What is Static Liquefaction Failure of Loose Fill Slopes?

What is Static Liquefaction Failure of Loose Fill Slopes? Charles W. W. Ng The Hong Kong University of Science and Technology, Hong Kong SAR

ABSTRACT: Static liquefaction failure of soil slopes has often been reported in literature. It appears that some researchers and engineers use different criteria to define and describe static liquefaction and they refer to different failure mechanisms. What is static liquefaction? How is it triggered? How can we identify and define static liquefaction failures? Does a strain-softening material necessarily mean static liquefaction? These are not all easy questions to answer and some of them may be even controversial. Based on some centrifuge model and triaxial element tests, suggested answers to some of these questions are explored, discussed and verified in this paper. 1 INTRODUCTION Slope failures occur in many parts of the world. A slope will become unstable when its shear resistance is smaller than any external driving shear stress, which may be induced by mechanical and hydraulic means such as rainfall, earthquake, vibration and seepage. Alternatively, a slope will also become unstable if its shear resistance is deteriorated and reduced due to weathering and any other mechanisms such as static liquefaction. Very often the terminology ³static liquefaction¶ is used to describe soil slope failures and reported in literature. However, it is evident that different researchers and engineers may refer to different failure mechanisms. Some use debris mobility (travel angle or run out distance) to judge whether a slope failure is caused by liquefaction or not. Clearly there is no direct relationship between liquefaction and mobility. For instance, a level ground can liquefy (at zero/small effective stress under seismic loading) with zero run out distance. On the contrary, a steel ball can run down a bare slope to reach a very long travel distance and this is nothing to do with liquefaction or not (Ng 2007). What is static liquefaction? How is it triggered? What is the effective stress at failure, if the slope is fully saturated initially such as undersea slopes? How can we identify and define static liquefaction failures? Does a strain-softening material necessarily mean static liquefaction? Is there any difference between slide failure and flow failure? What is the role of hydrofracture? How the angle of a slope affects the so-called static liquefaction? Is there any difference between fluidization and liquefaction? Will

static liquefaction occur in unsaturated soil slopes? How does the angle of a slope affect the potential of static liquefaction? Is there any relationship between the so-called static liquefaction failure and run out distance? Can soil nails be used to stabilize any loose fill slopes? Some of these questions have not been well understood and addressed and some of them may be even controversial. In this paper, some selected issues from above are investigated via laboratory triaxial element tests and centrifuge model tests on loose fill slopes using gap-graded Leighton Buzzard (LB) sand and completely decomposed granite (CDG), which is a well-graded silty sand. Observed key failure mechanisms of static liquefaction in the LB sand and non-liquefied slides of CDG fill slopes are identified and discussed, mainly following on the papers by Ng (2005, 2007 & 2008). 2 CLARIFICATION OF SOME TERMINOLOGIES RELATING TO STATIC LIQUEFACTION Figure 1 shows some typical results from monotonic triaxial tests on saturated, anisotropically consolidated sand specimens (Ng 2008). As shown in Fig. 1a, a very loose sand specimen, A, exhibits a peak undrained shear strength at a relatively small shear VWUDLQ DQG WKHQ ³FROODSVHV´ to much smaller shear strength at large strains. This behaviour is often causally referred to ³OLTXHIDcWLRQ´RU³flow liquefacWLRQ´ by many researchers and engineers. No matter ZKHWKHULWLVFDOOHG³IORZOLTXHIDFWLRQ´RU³OLTXHIDcWLRQ´ WKH WHUPLQRORJ\ WR GHVFULEH WKH PDWHULDO Ee-

3.1 Model material Centrifuge model tests were carried out at the GCF of HKUST (Ng et al. 2002a, Ng et al. 2006a) to investigate the failure mechanisms of static liquefaction of loose fill slopes subjected to rainfall, a rising ground water table and dynamic earthquake loadings (Zhang 2006, Zhang et al. 2006, Ng 2007). Leighton Buzzard (LB) Fraction E fine sand was selected as the fill material for the model tests. Fig. 2 shows the gap-graded particle size distribution of LB sand. D10 and D50 of the sand were 125 Pm and 150Pm, respectively. Following BSI (1990), the maximum and minimum void ratios of the LB sand were found to be 1.008 and 0.667, respectively (Cai 2001). The estimated saturated coefficient of permeability was 1.6 u 10-4 m/s. LB sand was chosen because of its pronounced strain-softening characteristics with its high liquefaction potential, LP, i.e., a substantial strength reduction in shear strength when it is subjected to undrained shearing (see Fig. 3a). The results from four loose specimens with different initial void ratios (eo) shown in the figure are obtained from isotropically consolidated undrained compression triaxial tests. The loose sand clearly shows pronounced strain-softening behaviour and substantial strength reduction in the deviator stress and shear strain (qHq) space and contractive responses in the mean effective stress (pc) and deviator stress (q) space, i.e. pc decreases continuously as q increases until the peak state is attained (see Fig. 3b), where pc and q are

Deviator stress

(a)

B Strain hardening Dilation

Strain softening

C Strain hardening Limited liquefaction

Liquefaction Strain softening

A Axial strain

Deviator stress

3 INVESTIGATION OF THE FAILURE MECHANISM OF LIQUEFIED FLOW IN SAND FILL SLOPES BY CENTRIFUGE TESTS

equal to (V 1c  2V 3c ) /3 and (V1c  V 3c ) , respectively. After the peak state, q drops (the soil collapses) with a large deformation develops until the quasi-steady state (a shear strain of about 15%) or the critical state (shear strain = 30%) is reached. The critical state friction angle (Icc) of the sand is 30° (Cai 2001). Following the approach proposed by Lade (1992), the angle of instability (Icins) determined for the sand is 18.6°. It is well-known that Icins is dependent on void ratio and stress level (Chu & Leong 2002). For engineering assessment and design of remedial work for loose fill slopes, it may be reasonable to assume this angle is a constant as the first approximation.

Excess pore pressure

haviour observed in the laboratory is rather confusing and, strictly speaking, incorrect. Would it be clearer and more precise to describe the material behaviour of the loose specimen, A, and a dense VSHFLPHQ % DV ³VWUDLQ-VRIWHQLQJ´ DQG ³VWUDLQKDUGHQLQJ´UHVSHctively, in the deviator stress-axial strain space (see Fig. 1a)? In the mean effective stress-deviator stress space (see Fig. 1b), would it be more precise to use the terms ³XQGUDLQHG VWUHQJWK UHGXFWLRQ´ RU VR-called collapse (Sladen et al.  DQG³XQGUDLQHGVWUHQJWKLQFUHDVH´WRGescribe the strength changes of specimen A and specimen B, respectively? Of course, it is well-recognised that a reduction and an increase in undrained shear strength are caused by the respective tendency of sample contraction and dilation, leading to a respective increase and a reduction in pore water pressure ('u) for specimens A and B during undrained shearing (see relationship between 'u and axial strain in Figure 1c). It must be pointed out that these are just material element behaviour that does not necessarily capture and represent the global behaviour of an entire fill slope or an earth structure.

B Dilation

(b)

Undrained strength increase Phase transC due to dilative formation tendency point Limited liquefaction contractive tendency A Liquefaction Undrained strength reduction due to contractive tendency Mean effective stress Contractive tendency Liquefaction (c) A

Axial strain Limited li- C quefaction contractive tendency B

Dilation Dilative tendency

Figure 1. Liquefaction, limited liquefaction, and dilation in monotonic loading tests (modified from Castro 1969, Kramer 1996).

3.2 Model package and test procedures Figure 4 shows an instrumented 29.4o loose sand fill slope model together with the locations of the pore water pressure transducers (PPTs) (Zhang & Ng 2003, Ng 2008, Ng et al. 2009). The model slope was prepared by moist tamping. The initial relative compaction was 68%. The body of the sand slope was instrumented with seven PPTs and arrays of surface markers were installed for image analysis of soil movements. Linear variable differential transformers (LVDTs) and a la-

ser sensor were mounted at the crest of the slope to monitor its settlement. LB-Sand SKW-CDG CKL-CDG BH-CDG WTS-CDG

60 40 20 0 0.001

LVDT

0.01

0.1 Particle size (mm)

1

10 Drainage board

Figure 2. Particle size distributions of LB sand and CDG. 800

Temporary reservoir

(a)

700

Model container

Inlet hole PPT7

Reflector y

PPT5

x

q (kPa)

PPT6

PPT4

PPT2 PPT1

600 500 e0=0.973

Model scale

LVDT & Laser sensor

305

80

29.4

Percentage finer (%)

100

implies that the slope was vulnerable to instability, which could lead to liquefaction (see Fig. 3). At 60 g, the 18 m-height (prototype) slope was destabilised by rising ground water from the bottom of the model (Zhang 2006, Ng et al. 2009). The loose sand slope liquefied statically and flowed rapidly (see Fig. 5b), i.e., it followed a process in which the loose slope was sheared under undrained conditions, lost its undrained shear strength as a result of the induced high pore water pressure (see Fig. 6) and then flew like a liquid, called ³OLTXHILHGIORZ´

Liquefaction potential (LP)

Outlet hole PPT3

Sand 1130,7

1130.7

400

Figure 4. Centrifuge model of a loose sand fill slope subjected to rising ground water table at 60 g (Zhang & Ng 2003).

e0=0.970

300

Quasi-steady state Quasi-steady state

200 e0=0.983

100

e0=0.992

0 0

10

20

30

40

Hq (%)

700

(b)

500

300

100

Figure 3. Contractive behaviour of loose LB sand under consolidated undrained tests (a) in the Hq - q and (b) in pc - q planes (modified from Zhang 2006, data from Cai 2001).

3.3 Observed static liquefaction mechanism Although the initial angle of the loose slope was prepared at 29.4o at 1 g, the slope was densified to 80% of the maximum relative compaction due to self-weight compaction at 60 g. The slope angle was therefore flattened to 24o (see Fig. 5a), which is steeper than the angle of instability of 18.6°. This

Figure 6 shows the measured rapid increases in the excessive pore water pressure ratio ('u/Vcv) within about 25 seconds (prototype) at failure at a number of locations in the slope during the test. The maximum measured 'u/Vcv was about 0.6, which would be much higher if a properly scaled viscous pore fluid were used to reduce the rate of dissipation of excess pore pressure in the centrifuge. This means that the slope would liquefy much more easily. As shown in Figure 5b, the completely liquefied slope inclines at about 4o to 7o to the horizontal after the test. The observed fluidization from in-flight video cameras and the significant rise in excessive pore water pressures during the test clearly demonstrated the static liquefaction of the loose sand fill slope. It should be noted that measurements of sudden and significant rise of excessive pore water pressures are essenWLDOWR³SURYH´RUYHULI\WKHRFFXUUHQFHRIVWDWLF liquefaction of loose fill slopes if no video recording is available. The liquefaction of the loose sand slope was believed to be initially triggered by seepage forces in the test (Ng et al. 2009). It is obvious that soil nails cannot be used to stabilize a loose sand fill slope which has a high liquefaction potential (see Fig. 3a). Figure 7 shows five postulated zones, Z1-Z5, representing the sequence of the failure and liquefaction process of the slope (Ng et al. 2009). Z1 is a failure region de-stabilised by the loss of its toe due to the seepage force in the gully (drained failure). The soil mass at the toe of Z1 slid with the soil at the gully head to trigger the failure of Z2. The soil mass in Z2

collapsed rapidly (undrained) which was then followed by the collapse of Z3 (undrained) without inducing obvious deformation in the lower part. The collapses of Z2 and Z3 were due to the strainsoftening associated with the significant strength reduction (i.e. high liquefaction potential) of the loose LB sand as illustrated in Figure 3. The rapid undrained collapses of Z2 and Z3 were evident from the measured large excess positive pore pressures at PPT7 (see Fig. 6).

Subsequently, Z4 also collapsed as a result of the strain-softening associated with the significant strength reduction (high liquefaction potential) of the loose LB sand (see Fig. 3). The dotted line drawn between Z4 and Z5 in Fig. 7 represents the upper boundary of the stable region (Z5), monitored by markers and the small excess pore pressures at PPT1 and PPT2 (see Fig. 6) during the liquefaction process. Based on the observed mechanism, it is fair to suggest that soil nails cannot be used to stop any liquefied flow of loose sand fill slopes. However, the use of soil nails can reduce the magnitude of any excessive positive pore water pressure generated in a loose sand slope, minimize the chance of liquefaction and reduce damages after liquefaction (Zhang et al. 2006). Gully erosion

B 20

Water flow Sand movement

Gully head

A Water

Z1 Z2

15

PPT7

10

B¶

Z3

PPT5 PPT4

PPT6

PPT2 PPT1

5

Slope profile before failure

Final slope profile

Z4 PPT3

Z5

A¶ 0 0

5

10

15

20

25

30

35

40

45

50

55

60

65

(m)

Figure 7. Postulated failure zones during the liquefaction of slope SG30 (Ng et al. 2009).

4 OBSERVED EXCESSIVE SETTLEMENTS OF THICK LOOSE CDG FILL SLOPES IN CENTRIFUGE 4.1 Monotonic and cyclic behaviour of CDG from Beacon Hill (BH) Figure 5. Slope profile in a loose sand fill test (a) before rising ground water table; (b) after static liquefaction (Zhang & Ng 2003). 1.0 Slope failure

0.6

PPT7

0.4

PPT4 PPT5

'

Excess pore pressure ratio ('uw/Vv )

0.8

0.2 0.0 -0.2

PPT2 PPT1 PPT6

-0.4 -0.6

PPT3

-0.8 -1.0 37.8 38.2 38.6 39.0 39.4 39.8 40.2 40.6 41.0 41.4 41.8 42.2 42.6 43.0 Duration (min)

Figure 6. Measured sudden and substantial increases in pore water pressure at seven locations inside the slope (Zhang & Ng 2003).

Prior to centrifuge model tests, a series of undrained monotonic and cyclic triaxial tests on normally consolidated CDG specimens (70 mm in diameter and 140 mm in height) were performed to assist in the interpretation of centrifuge tests on loose CDG fill slopes. Figure 2 shows the particle size distribution of the well-graded CDG samples obtained from Cha Kwo Ling (CKL), Kowloon. In the figure, the wellgraded CDG taken from Beacon Hill (BH) is also included for comparison. The mean particle size, D50, of the CDG from CKL is 1.18 mm and the sample contains about 15% fines content. According to the British Standard, BS1377 (1990), CDG can be classified as well-graded silty sand. The triaxial specimens tested (Fig. 8) were prepared by moist tamping at the optimum moisture content (Ng et al. 2004a). The initial relative compaction of the specimens was 70% before saturation. Enlarged lubricated end platens were used in the tests to reduce the end constraints on the soil specimens. In the undrained monotonic triaxial compres-

4.2 Centrifuge modelling of loose CDG fill slope due to rainfall infiltration (Take et al. 2004) With the support of the Geotechnical Engineering Office (GEO) of the Civil Engineering and Development Department of HKSAR, collaborative and complementary centrifuge model tests were carried out at the University of Cambridge and HKUST. Bulk samples of CDG taken from BH were delivered to the two universities for centrifuge model tests. Figure 9 shows an initially 45o loose fill model slope. The model was constructed by moist-tamping with only a minimal compaction effort. To reduce particle size effects, the fill material was first sieved to remove all particles in excess of 5 mm in diameter. To simulate effects of rainfall infiltration by con-

trolling water contents (or moisture), the fill slope was installed in an atmospheric chamber, which was sealed from the external environment (Take et al. 2004). It is well known that suction is related to the water contents in soil pores (Ng & Menzies 2007). In Take et al¶s test, positive and negative pore water pressures were measured using a network of miniature PPTs and pore pressure tension transducers (PPTT), respectively, at each of locations indicated by open circles in the figure. The deformation of the model fill slopes was captured by PIV (White et al. 2003) at 60 g. 300

CU050

CU100

CU200

CU300

CU400

(a)

Shear stress, q (kpa)

250 Mins=1.12 (I'ins=28.2o)

M=1.54 (I'=37.8o)

200

Critical state line

150 Instability line

100

e=0.78 e=0.82 e=0.85

50 e=0.94 e=1.05

0 0

50

100 150 200 250 300 350 Effective mean normal stress, p' (kPa)

400

450

300

CU050

CU100

CU200

CU300

CU400

(b)

250

Shear stress, q (kpa)

sion tests, the soil specimens were consolidated isotropically to different initial mean effective stresses before shearing. Figure 8a shows the effective stress paths of five isotropically consolidated undrained compression tests with the initial pc ranging from 50 kPa to 400 kPa (corresponding to void ratios varying from 1.05 to 0.78). The effective stress path of each loosely compacted specimen is characterized by its initial increasing q with decreasing pc, due to an increase in pore water pressure during undrained shearing resulting from the contractive tendency of the soil. After a peak is reached, q decreases with a further reduction in pc until the critical state (M=1.54, Iƍ o) is reached, illustrating the unstable nature of the specimen. By joining the stress origin and the peak of each stress path, an instability line (Lade 1992) can be identified with its slope equal to 1.12, corresponding to Iƍins=28.2o. Strainsoftening behaviour with very small liquefaction potential but without any phase transformation phenomenon was noted in these tests (see Fig. 8b). During the cyclic tests (Ng et al. 2004b), a cyclic deviator stress of equal magnitude in compression and extension was applied to the specimens. Figure 8c shows a typical result of CDG (e=0.821) from a cyclic triaxial test with a cyclic stress ratio (CSR) of 0.1, where CSR is defined as the single amplitude cyclic shear stress (ıd) divided by twice the initial effective confining pressure (ı 3), i.e. CSR=ıdı 3). In the test, pc decreased monotonically but the rate of the pore water pressure build-up decreased as the number of cycles increased, due to the relatively low CSR. Eventually, the pore water pressure ceased to develop further as the contractive and dilative tendency of the soil specimen balanced out. The total deviator strain developed was less than 0.2% at the end of the test. On the other hand, for a cyclic test on CDG with CSR=0.15 (e=0.821) as shown in Figure 8d, the pore water pressure accumulated continuously and resulted in a continuous decrease in pc, illustrating a typical cyclic mobility phenomenon (Castro 1969).

e=0.78, Vc=400kPa

200

e=0.82, Vc=300kPa

150

e=0.85, Vc=200kPa

100

e=0.94, Vc=100kPa

50

e=1.05, Vc=50kPa

0 0

5

10

15 20 Axial strain, Ha (%)

25

30

(c)

(d)

Figure 8. Triaxial tests on loose CDG: (a) Stress paths of static triaxial tests; (b) Stress-strain relationships of static triaxial tests; (c) Cyclic triaxial test with CSR = 0.1; (d) Cyclic triaxial test with CSR = 0.15.

Figure 10 illustrates the changes of the initially moist-tamped structure of the model fill at the crest during the test. At 1 g, the very loose soil had an initially very open structure (see Fig. 10a), which consisted of large voids supported by capillary suction. One such void is circled in the figure. At 60 g, many of these macro-voids were observed to collapse (Fig. 10b). However, not all the voids collapsed. In particular, the voids at low stress levels (i.e. shallow depths) such as the highlighted void in Figure 10a simply settled along with the fill. The observations of the collapse and the mechanisms shown in these two figures cannot be easily obtained from the field or numerical analyses even with large-strain formulations. After the initial self-weight consolidation, the fill slope was subjected to the equivalent of six weekly periods of rainfall infiltration in centrifuge. A significant portion of the soil suction was destroyed very rapidly at the shallow location after the arrival of rainfall on the slope surface (Take et al. 2004). The loose model fill responded immediately to this loss of surface tension by collapsing the macro-voids

that had survived self-weight consolidation (Fig. 10c). Although the slope was suffered from excessive settlement, no flow slide and no liquefaction were observed in the test. This finding is consistent with the test in CDG reported by Ng et al. (2002b).

Figure 9. Model geometry of CDG fill slope (Take et al. 2004).

Figure 10. The observed changes of soil structure of the crest region due to rainfall infiltration (Take et al. 2004).

18.900

Prototype Scale

e 0.600 Unit in metre

Scale

6.240

To complement the rainfall infiltration tests carried out at Cambridge, a series of centrifuge model tests on loose CDG fill slopes with and without soil nails was subjected to rising ground water at HKUST (Ng et al. 2002b, Zhang 2006). The CDG fill material used for the tests in Hong Kong was also from BH. A model slope was initially prepared to incline at 45o to the horizontal and the initial relative compaction of the fill was less than 80%. At 60 g, a 300 mm high model slope was equivalent to an 18 m high prototype slope. Figure 11 shows the measured displacement vectors of a 45o unreinforced loose CDG fill slope destabilised by the rise of the ground water. Excessive settlement was measured but no sign of liquefied flow or slide of the slope was observed

during and after the test. This was probably because of the small liquefaction potential of the CDG (Ng et al. 2004a).

24.240

4.3 Response of loose fill slope subjected to rising ground water in centrifuge (Ng et al. 2002b)

Figure 11. Displacement vectors in unreinforced slope (CG45) (Ng 2007).

4.4 Response of loose CDG fill slopes to earthquake loading in centrifuge 4.4.1 Centrifuge model and test procedures (Ng et al. 2004b, Ng 2007) To further investigate the possibility of flow liquefaction of loose CDG fill slopes, uni-axial and biaxial dynamic centrifuge tests were carried out using soil samples taken from BH (Ng et al. 2004b). The model CDG fill slopes were subjected to shaking ranging from 0.08 g to 0.28 g (prototype) in the centrifuge at HKUST. All the models were essentially the same in geometrical layout and made of loose CDG with the same initial dry density. Figure 12 shows a typical model slope (6 m in prototype) initially inclined at 30o to the horizontal with its instrumentation. A rigid rectangular model box was used to contain the CDG samples compacted to an initial dry density of about 1.4 g/cm3 (or 77% of relative compaction). Five pairs of miniature accelerometers were installed in the slope. Each pair was arranged to measure soil accelerations in two horizontal directions (i.e., X- and Y-directions). Four miniature pore pressure transducers were installed in the soil near the accelerometers to record pore water pressures during shaking. On top of the slope, three LVDTs were mounted to measure the crest settlement, and one LVDT and one laser sensor (LS) were used to measure horizontal movement of the crest. To simulate the correct dissipation rate of excessive pore pressures in the centrifuge tests, sodium carboxy methylcellulose (CMC) powder was mixed with distilled deionized water to form the properly scaled viscous pore fluid and to saturate the loose CDG model slopes. After model preparation, the speed of the centrifuge was increased to 38 g. Once steady state pore pressure condition was reached at all transducers, a windowed 50 Hz (1.3 Hz prototype), 0.5 s (19 s prototype) duration sinusoidal waveform was then applied (Ng et al. 2004b). After triggering each earthquake, the centrifuge acceleration was maintained long enough to allow the dissipation of any excess pore pressure. Due to page limits, only some results from one biaxial shaking test are discussed here. Other details of all the tests are presented in Ng et al. (2004b). 140

LVDT-v3

LVDT-v2

LVDT-v1

LVDT-h1

LS-h1

660

30

ACC-T-X,Y,Z

150

X PPT4 ACC4-Y ACC4-X ACC3-Y PPT3 ACC5-Y ACC3-X ACC2-Y PPT2 ACC2-X ACC5-X ACC1-X ACC1-Y PPT1

Z

712

Figure 12. Configuration of the model slope and instrumentation (Ng et al. 2004b).

4.4.2 Measured responses of the loose CDG fill slope subjected to bi-axial shaking (M2D-0.3) (Ng et al. 2004b) Figure 13 shows some measured horizontal acceleration time histories in the X- and Y-directions together with their normalized amplitudes in the Fourier domain. In the biaxial shaking test, the base input accelerations (recorded by ACC-T-X & ACCT-Y as shown in the figure) were 11.26 g (0.28 g prototype) and 7.77 g (0.19 g prototype) in the X± direction and Y-direction, respectively. The windowed sinusoid waveform applied in the Y-direction lagged the X-direction input signal by 90°. Recorded by the accelerometer near the crest, the peak acceleration in the X-direction increased by 45% at ACC4-X, higher than that measured in a corresponding uni-axial shaking test (Ng et al. 2004b). A similar trend of variations in the acceleration was also found in the Y-direction. The normalized spectral amplitudes of acceleration at the predominant frequency of 50 Hz decreased by about 9% in the Xdirection but increased by about 4% in the Ydirection in the upper portion of the embankment.

Figure 13. Seismic acceleration history and Fourier amplitude spectrum (M2D-0.3) (extracted from Ng et al. 2004b).

Figure 14 shows the time history of the excess pore pressure ratios along the height of the model embankment during shaking. Peak acceleration occurred at about 0.25 s after the start of shaking. The maximum pore pressure ratio occurred at about 0.33 s at each of the three transducers (PPT1, PPT2 & PPT4). PPT1 and PPT2 recorded about the same maximum pore pressure ratio of 0.87, whereas PPT4 registered the smallest of 0.75. These measured values were less than the theoretical value of 1.0 for liquefaction, even though the pore fluid was correctly scaled in the test. The excess pore pressures dissipated to zero at about 12 s (6.8 minutes in prototype) after the start of shaking. Figure 15 is a photograph of the model taken after the completion of a shaking test. The deformation profile for the slope was similar in both the uni-axial

and bi-axial shaking tests. The observed profile of the deformed slope clearly illustrates that no liquefied flow and non-liquefied slide took place during the shaking. The significant difference between the observed physical test results from the loose LB sand and CDG fill slopes may be attributed to the difference in fine contents, gradation and liquefaction potential of the two materials (see Fig. 3).

most of the existing fill slopes formed before 1977 in Hong Kong.

1.0 1.0 PPT1

PPT2 (Z=100mm)

PPT2

0.5

'u / Vcv

PPT1 (Z=145mm)

0.0

0.5

PPT4

-0.5 0.0

0.2

0.4 0.6 Time (s)

0.8

1.0

0.0 PPT4 (Z=10mm)

Figure 16. General view of the slope (from Tang & Lee 2003). -0.5 0

5

10

15

Time (s)

Figure 14. Measured excess pore-water pressure ratios in biaxial shaking test M2D-0.3 (Ng et al. 2004b).

Ground water level LVDT LVDT

Laser sensor Laser sensor

Laser sensor

Laser sensor

Original liquid surface

Figure 15. A typical profile of a loose fill slope after shaking (Ng et al. 2004b, Ng 2007).

5 OBSERVED EXCESS SETTLEMENT OF CDG FILL SLOPE IN THE FIELD Tang & Lee (2003) reported a large-scale field trial on a partly reinforced 33o loose CDG fill slope (see Fig. 16). The bulk fill material was taken from BH. The height and width were 4.75 m and 9 m, respectively. It was constructed by the end-tipping method and resulted in a loose state with an initial dry density ranging from 70% to 75% of the maximum dry density. It was considered that the stress state of this slope would represent reasonably well to that of

Two rows of grouted nails were installed at a grid of 1.5 m x 1.5 m at an inclination of 20o from the horizontal. Holes of 100 mm diameter with two different lengths (8 m and 6 m) were drilled. A 25 mm diameter steel ribbed bar was inserted into each hole and the hole was filled with grout from the bottom up using a plastic hose. In order to destabilize the slope, a water re-charge system was used. This re-charge system comprised crest recharge trench, buried piping system and surface sprinkler and they were installed separately so that a rise in groundwater table in combination with a rainfall event could be simulated in the field. To increase destabilising forces, 1 m x 1 m x 0.6 m concrete blocks were stacked up to 3m high at the central area of the slope crest. They imposed a surcharge loading of 72 kPa. The slope was heavily instrumented. Details of the instrumentation are described by Tang & Lee (2003). After water was being recharged into the slope through the piping system, it was observed that the deformation of the slope increased rapidly. The total deformations at the crest and mid-slope were 139 mm and 33 mm, respectively. Due to the large deformation, the surcharged blocks tilted and collapsed (Fig. 17). The settlement-induced toppling failure of the blocks was restricted at the crest zone and the slope remained intact. No sign of static liquefaction and flowslide was observed in this largescale field test. The observed excessive settlements and large measured nail forces in the field are similar to those measured in the centrifuge model tests, as shown in Figure 18 (Ng et al. 2002b, Ng 2008).

Figure 17. General view of slope after failure (from Tang & Lee 2003)

Figure 18 compares the displacement vectors of the loose CDG fill slopes obtained from two centrifuge tests, one without and one with soil nails. The soil nails were installed in-flight at 60 g and it can be seen that the soil nails substantially reduced soil movements by at least a factor of 5. No sign of static liquefaction of the slopes was observed during and after the tests. Similar findings are also reported by Take et al. (2004) from independent centrifuge model tests using the same loose CDG fill at Cambridge University and by Tang & Lee (2003) from large-scale field tests conducted at Hong Kong University.

loose CDG fill slopes. Centrifuge model tests were commissioned to investigate possible failure mechanisms of loose fill slopes. Figure 19 shows an instrumented centrifuge model created to study the potential static liquefaction of a loose shallow CDG fill slope subjected to a rising ground water table. The particle size distribution of the CDG used is denoted as WTS in Figure 2. The initial fill density was 66%. This model was used to simulate a 1.5 m thick, 24 m high layered fill slope when tested at 60 g. In addition to laser sensors (LSs) installed for monitoring soil surface movements, PPTs were installed to measure excess pore water pressures during the tests. Effects of layering were considered by tilting the model container during model preparation. The slope was destabilised by downward seepage created by a hydraulic gradient, which was controlled by the water level inside the upstream temporary reservoir and the conditions of the outlet hole located downstream (see Fig. 19). Two failures were induced in the test. PPT Unit: mm Model box Upstream drainage board

Coarse soil

PPT1 PPT2 PPT3 Loose CDG (WTS)

Downstream drainage board

LS3 PPT4

Downstream temporary reservior reservoir

LS2

Upstream temporary reservior reservoir

PPT5 LS1

Wood block

Coarse soil

PPT6

Inlet hole

PPT7

PPT B

Outlet hole

PPT8 PPT9 PPT C

Figure 19. Model package of an instrumented shallow fill slope (Ng et al. 2007).

Figure 18. Comparisons of measured soil displacements without (CG45) and with soil nails (CGN45) in two centrifuge tests using CDG loose fill at 60 g (dimensions in metres at prototype scale) (Ng et al. 2002b).

6 OBSERVED NON-LIQUEFIED SLIDE MECHANISMS OF SHALLOW CDG FILL SLOPES IN CENTRIFUGE 6.1 Destabilisation of loose shallow CDG fill slopes near the crest (Ng et al. 2007) The Housing Department of HKSAR has been actively looking for innovative methods to preserve the environment by minimizing the need for felling trees when improving the stability of existing shallow

Figures 20 and 21 show the occurrence of a nonliquefied slide and the measured excessive pore water pressure during two failures, respectively. The slide was initiated near the crest. Based on the observed failure mechanisms and the small excessive pore water pressures measured, it was concluded that non-liquefied slide of loose shallow CDG fills slopes could occur but static liquefaction was very unlikely to happen in the slopes. 6.2 Destabilisation of loose shallow CDG fill slope at the toe in centrifuge (Take et al. 2004) Take et al. (2004) also carried out centrifuge model tests to investigate the possible slide-flow failure mechanism of a loose thin CDG fill layer. The CDG used was taken from Beacon Hill. Figure 22 shows the geometry adopted. The slope angle was 33o. At 30 g, the model corresponded to a fill slope of 9 m in height, with a vertical depth of fill of 3 m. The chosen soil profile for the model fill also represents an idealized case of layering in which the CDG fill material has been sieved and separated into its coarse

and fine fractions and placed one on top of the other to form a layered backfill. The layer ends blindly at the toe of the slope to generate elevated transient pore pressures (Take et al. 2004). This ensures that the rate of arrival of the seepage water at the toe greatly exceeds that of the leakage, thereby ensuring a more rapid local transient build up of pore water pressures in this region than would have existed in the absence of layering. In this experiment, the impermeable bedrock layer was modelled by a solid wooden block, the top surface of which was coated with varnished coarse decomposed granite to ensure a high interface friction angle. The density of the fill material in the first layered slope model was very loose, with an approximate relative compaction of 77%. After preparation, the model fill slope was installed on the centrifuge and slowly brought to the testing acceleration of 30 g.

of the fill slope. As intended, the rate of water transfer into the toe region exceeded the seepage velocity through the model fill, causing a transient increase in the pore water pressure at the toe. The local pore water pressure was observed to increase at a nearly constant rate reaching a maximum value of 16 kPa at point B in Figure 23a. As this seepage front progressed towards the toe, the slope was slowly creeping (Fig. 23b). After time B, the slope mass is observed to accelerate (points B-C on Figure 23b). By analysing images captured by PIV (White et al 2003) at the onset of more rapid failure, it is found that the toe accelerated horizontally with an average velocity of approximately 6 mm/s (Fig. 24). The observed displacement field over this time interval indicates that the surface of the model fill moved down-slope at a slower velocity. When the fill material finally came to rest, it formed a low-angle run-out. This failure mechanism differs from that of the slope destabilised by downward seepage in the test for the Housing Department in which the slope was not blinded hydraulically at the toe (see Fig. 19). The initiation of the non-liquefied slides differed in these two slopes.

Figure 20. Top view of the model showing a non-liquefied slide (Ng et al. 2007).

Figure 22. A slide-to-flow landslide triggering mechanism model (Take et al. 2004).

Figure 21. Variations in the measured pore water pressure at the crest (PPT2) and at the toe (PPT7) of the slope with time (Ng et al. 2007).

Figure 23a shows the arrival of the transient pore water at the toe of the slope. Once the line source of seepage water was activated, the high transmissivity of the coarse layer quickly delivered water to the toe

Figure 23. Observed behaviour of slide-to-flow models (Take et al. 2004).

pressure required to initiate failure (see Fig. 23a), but it made the failure more brittle (Take et al. 2004). Based on the non-liquefied slides observed in both loose and dense CDG shallow fill slopes, it is evident that soil nails VKRXOGEHDEOHWR³QDLOGRZQ´ these non-liquefied slides since the CDG still possess sufficiently large shear strength after the peak (see Fig. 8b).

Figure 24. Displacement field prior to final acceleration of loose fill model (Take et al. 2004).

(a)

(b)

(c)

(d)

6.3 Destabilisation of a dense shallow CDG fill slope at the toe in centrifuge (Take et al. 2004) Unlike the static liquefaction mechanism of loose sand fill slopes, the non-liquefied slide triggering mechanism is argued to be independent of soil density (Take et al. 2004, Ng 2007). In order to verify this hypothesis, the experiment was therefore repeated with a fill compacted to 95% maximum Proctor density while all other factors remained constant (Take et al. 2004). As before, seepage water was introduced to the crest of the model fill slope and it was quickly transmitted to the toe of the slope, building up localized transient pore water pressures at an identical rate as in the loose fill model (Fig. 23a). Since the slope material was dry, the position of the wetting front could be observed. This dense slope exhibited a much stiffer response to the build up of pore water pressures, with less than one half of the pre-failure displacements signalling the onset of failure (see Fig. 23b). Just before reaching the failure pore water pressure, the brittle fill material cracked and water rapidly entered the fill. As high-pressure water entered the crack, the acceleration of the slide increased. The extent to which this crack injected water into the fill material at time B is shown in Fig. 23a. After time B, the slope mass accelerated, although at a slower slide velocity than observed in the loose fill slope (points B-C in Fig. 23b). The subsequent behaviour of the model fill slope is laid out pictorially in the remainder of Fig. 25. As the toe continued to accelerate horizontally, the surface of the model fill accelerated towards the toe (Fig. 25b), with the velocity increasing to such a point that it exceeded the shutter speed of the camera (Fig. 25c). Eventually, the slope came to rest (Fig. 25d). Similarly to in the shallow loose fill slope, the landslide event triggered from localized transient pore water pressures formed a low-angle run-out. The densification of the fill slope slightly increased the pore water

Figure 25. Failure mechanism in the dense fill model (modified from Take et al. 2004).

7 CONCLUSIONS Both static and dynamic model tests on LB and CDG were carried out. In-flight rainfall infiltration, rising ground water and dynamic loadings were simulated. Based on the tests, it can be concluded that static liquefaction/fluidization of the loose LB sand fill slope due to a rising ground water table was successfully created in the centrifuge. The occurrence of liquefaction in sand was observed by in-flight video cameras and verified by the significant and sudden build-up of excessive positive pore water pressures measured at various locations in the slope. It is found that strain softening of the material is a necessary but not a sufficient condition to cause flow liquefaction. A trigger such as seepage force or additional loading is needed. No liquefied flow and slide was observed in thick loose CDG fill slopes when they were subjected to rising ground water tables, heavy rainfall infiltration and very strong bi-axial shaking. Only excessive soil settlements were induced. Consistency was found between centrifuge model tests and full-scale field trial of a loose CDG fill slope. The significant difference between the observed physical test results on the LB sand and CDG models may be attributed to the difference in the fine contents, gradation and liquefaction potential of the two materials.

Although static and dynamic liquefaction did not occur in loose CDG fill slopes because of &'*¶V small liquefaction potential, non-liquefied shallow slides were observed in both loose and dense shallow fill slopes. Depending on the boundary conditions, different initiations of non-liquefied shallow slides were captured in the centrifuge. The landslide event triggered by highly localized transient pore water pressures at the toe results in a low-angle runout in both shallow loose and dense CDG fill slopes. For improving the stability of loose fill slopes, it is vital to differentiate the potential differences between a liquefied flow and a non-liquefied slide. It is evident that a potentially non-liquefied slide can be stabilized by soil nails. 8 ACKNOWLEDGEMENTS The work presented here was supported by research grants DAG00/001.EG36 and HKUST3/CRF-SF/08 provided by HKUST. The author is grateful for research contracts provided by the Geotechnical Engineering Office of the Civil Engineering and Development Department and the Housing Department of the HKSAR. Moreover, the author thanks Dr Robin Zhou for checking and formatting the paper. 9 REFERENCES BSI. 1990. BS1377: Methods of tests for soils for civil engineering purposes. British Standards Institution, London. Cai, Z.Y. 2001. A comprehensive study of state-dependent dilatancy and its application in shear band formation analysis. PhD thesis, HKUST. Castro, G. 1969. Liquefaction of sands. Harvard Soil Mechanics Series 87, Harvard University, Massachusetts. Chu, J. & Leong, W.K. 2002. Effect of fines on instability behaviour of loose sand. Géotechnique 52(10): 751-755. Kramer, S.L. 1996. Geotechnical earthquake engineering, Prentice-Hall, Inc. New Jersey, USA. Lade, P.V. 1992. Static instability and liquefaction of loose fine sandy slopes. J. Geotech. Eng. ASCE, 118(1): 51±71. Ng, C.W.W. 2005. Invited Country report: Failure mechanisms and stabilisation of loose fill slopes in Hong Kong. Proc. International Seminar on Slope Disasters in Geomorphological/Geotechnical Engineering. Osaka. 71-84. Ng, C.W.W. 2007. Keynote paper: Liquefied flow and nonliquefied slide of loose fill slopes. Proc. 13th Asian Regional Conference on Soil Mechanics and Geotechnical Engineering, Kolkata. Vol. 2. Allied Publishers Private Ltd. Ng, C.W.W. 2008. Invited special lecture: Deformation and failure mechanisms of loose and dense fill slopes with and without soil nails. Proc. of 10th Int. Sym. On Landslides and Engineered Slopes. ;L¶DQ&KLQDVol. 1, 159-177. Ng, C.W.W & Chiu, C.F. 2003. Laboratory study of loose saturated and unsaturated decomposed granitic soil. J. Geotech. and Geoenviron. Eng., ASCE 129(6): 550-559. Ng, C.W.W. & Menzies, B. 2007. Advanced Unsaturated Soil Mechanics and Engineering. Publisher: Taylor & Francis. ISBN: 978-0-415-43679-3 (Hard copy). 687p.

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