MULTI-SPAN LARGE BRIDGES
PROCEEDINGS OF THE INTERNATIONAL CONFERENCE ON MULTI-SPAN LARGE BRIDGES, 1–3 JULY 2015, PORTO, PORTUGAL
Multi-Span Large Bridges Editors
Pedro Pacheco & Filipe Magalhães
Faculty of Engineering, University of Porto, Portugal
Organized by:
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Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
Table of contents
Preface Acknowledgements Committees
XV XVII XIX
Keynote Lectures General presentation of the Keynote Lectures
3
Large viaducts, some executions a few ideas J. Manterola
9
Design and construction of sea-crossing bridges – A review N. Hussain
17
Viaducts with progressively erected decks J. Strasky
27
Betwixt and between Portus and Cale A. Adão da Fonseca
37
The Octavio Frias de Oliveira and Anita Garibaldi cable-stayed bridges C.F. Ribeiro
51
Multi-span extradosed bridges A. Kasuga
67
Multi-span large bridges – interaction between design and construction A.F. Bæksted
83
Recent achievements in the design and construction of multi-span cable supported bridges in China A. Chen, R. Ma & X. Zhang Multi-span large decks – the organic prestressing impact P. Pacheco
93 103
Experts, Experiences & Landmark projects Crossing of Bjørnafjorden – Floating bridge B. Villoria, J.B. Wielgosz & S.M. Johannesen
127
Rion-Antirion Bridge – Challenging earthquakes E. Joly, P. Moine & A. Pecker
135
Innovative erection methods of steel cable-stayed bridges M. de Miranda
143
Viaduct over river Ulla in the Spanish Atlantic high speed railway line: An outstanding composite steel-concrete truss bridge F. Millanes, L. Matute & M. Ortega V
151
Juscelino Kubitschek Bridge, Brasília, Brazil F.B. de Barros & J. de Freitas Simões
159
Bridge over the Cádiz Bay, Spain J. Manterola, A. Martínez, J.A. Navarro, S. Criado, S. Fuente, M.A. Gil, L. Blanco, G. Osborne, M. Escamilla & J.M. Domínguez
167
Baluarte Bridge executive project G.R. Argüelles
173
Queensferry Crossing: Role of concrete in the design and execution of the project P. Curran
179
New Pumarejo Bridge over the river Magdalena in Barranquilla, Colombia J. Manterola, J. Muñoz-Rojas, S. Fernández, J.A. Navarro & S. Fuente
187
Delivering the Padma Multipurpose Bridge project, Bangladesh W.K. Wheeler & C.J. Tolley
193
The tied arch bridge of the Saale-Elster-Viaduct W. Eilzer, R. Jung, T. Mansperger & K. Humpf
201
Construction and design features of the bridge over the Danube River, Bulgaria J. Manterola, A. Martínez, J.A. Navarro, J.L. Alvárez & J.I.D. de Argote
209
TUNeIT – Towards a global World E. Siviero, A.B. Amara, M. Guarascio, G. Bella, M. Zucconi, A. Adão da Fonseca & K. Slimi
215
The Russky Bridge: Pylons design approach optimization L.V. Miklashevich & V.E. Rusanov
223
Bridge across the Waschmühl Valley, Kaiserslautern, Germany: A harmonic symbiosis between a historic monument and a new innovative bridge K. Humpf, V. Angelmaier & W. Eilzer
231
Viaduct over river Deba in the “Y-Basque” high speed railway line in the north of Spain F. Millanes, M. Ortega, P. Solera, H. Figueiredo & J. Ugarte
239
Structural solutions and construction methods for the main crossing of the Mersey Gateway Bridge Project G.D. Moir, S.H. Jang, J. Seo & P. Sanders
247
Design of the long-span footbridge over the Bug River in Niemirów J. Biliszczuk, J. Onysyk, W. Barcik, P. Prabucki, K. Ste˛pie´n, J. Szczepa´nski, R. Toczkiewicz, A. Tukendorf, K. Tukendorf & P. Wo´zny Design and proof checking of foundation, substructure and superstructure of Rail cum Road Bridge at Munger, Bihar, India H.M. Farook & G.S. Babu
257
263
Multi-span bridge bypass over the Dziwna Strait J. Hołowaty
271
Large multi-span bridges built in recent years in Poland J. Biliszczuk, J. Onysyk, P. Prabucki & R. Toczkiewicz
277
Kassuende Bridge over Zambezi River in Tete, Mozambique T. Mendonça, V. Brito & M. Monteiro
285
VI
Multi-span bridge crossings for improved road access to Szczecin sea port J. Hołowaty
293
Armado Guebuza Bridge over Zambezi River in Caia, Mozambique T. Mendonça, V. Brito & M. Almeida
301
Design and construction of a long-span continuous fin-back bridge Y. Lu, M. Fu, X. He & C. Zhou
309
Pinhal Interior Motorway Concession – IC3 – Section Condeixa – Coimbra – Special engineering structures – Construction processes T. Nogueira, A. Hipólito & N. Amaro
317
Viaduct Araranguá – The alternative design of viaduct of 1661.59 meters in the BR-101/SC Brazil I.C. Santos & F.P.S. Nunes
325
Design and construction of flyovers in Outer Ring Road, Delhi K. Ganesh & V. Shanmugham
331
Haramain high speed railway line J.M.G. Parejo, M.T. Serrano, M.M. Cañueto, M.B. García & F.J.M. López
337
Design and construction of viaduct to Mumbai International Airport P.G. Venkatram & K. Ganesh
345
Meriç Bridge: Construction and quality control S. Uluöz, S. Düzbasan, T. Uluöz, E. Yakıt & U. Akyazı
351
Design and construction of elevated viaduct at Nashik, India K. Ganesh & P. Murali
357
Conceptual design Development of a submerged floating tube bridge for crossing of the Bjørnafjord M. Reiso, T.H. Søreide, S. Fossbakken, A.S. Brandtsegg, S.A. Haugerud, A. Nestegård, J.H. Sekse & A. Minoretti
365
Three span floating suspension bridge crossing the Bjørnafjord J. Veie & S.H. Holtberget
373
Long railway viaducts with special spans: Part 1. Arch construction by balanced cantilever with auxiliary cables J. Manterola, A. Martínez, B. Martín, J.A. Navarro, M.A. Gil, S. Fuente & L. Blanco
381
Long railway viaducts with special spans: Part 2. Arch construction by tilting J. Manterola, J. Muñoz-Rojas, A. Martínez & S. Fernández
389
Long railway viaducts with special spans: Part 3. Precast girders J. Manterola & A. Martínez
395
Four spans continuous cable stayed bridges without extra cables J. Romo
401
Particular design features for a long span cable-stayed bridge over the Harbour of Port Louis, Mauritius J. Jungwirth, J. Casper & A. Baumhauer
409
A study on vehicular live load design based on actual vehicular load for a multi-span large cable-stayed bridge H. Sugiyama, H. Kanaji, H. Watanabe & O. Aketa
417
VII
Comparison of variants for New Peljesac Bridge in Croatia J. Radic, Z. Savor, M. Srbic & M. Pipenbaher
427
Gebze–Orhangazi–Izmir Motorway, Izmit Bay Suspension Bridge N. Güngör & F. Zeybek
435
Construction of cable-stayed bridge over the Drava River on Corridor Vc, Croatia P. Sesar, M.M. Buhin, D. Bani´c & S. Kralj
443
Strait crossing of the Thermaikos Gulf with a mixed long-span bridge and subsea tunnel system M. Malindretou-Vika & P. Spyridis
451
Segmental prestressed concrete multispan large bridges V. Barata, J.P. Cruz & P. Pereira
459
Experience of some long multi-span bridges in Queensland, Australia (Part 1) J.A. Hart & E. Kittoli
467
Experience of some long multi-span bridges in Queensland, Australia (Part 2) J.A. Hart & E. Kittoli
475
Multi-span bridges: The first Chilean experience and future challenges M.A. Valenzuela, M. Márquez & I. Vallejo
483
Optimization of cable weight in multi-span cable-stayed bridges. Application to the Forth Replacement Crossing A. Baldomir, E. Tembrás & S. Hernández
491
Design parameters of suspension bridges: Updates of state of art and its application on multi-span typology I. Vallejo, M.A. Valenzuela & M. Márquez
499
Comparative study of prestressing consumptions in 7 different constructive methods for 75 m multi-span box girders A. Ferreira, B. Lima, F. Lopes & P. Pacheco
507
KaTembe Bridge over Espírito Santo Estuary, in Maputo T. Mendonça, V. Brito & M. Monteiro
513
The South Approach Viaduct of Izmit Bay Crossing Project N. Güngör
521
Effect of hangers disposal on the steel consumption for bowstring arch bridges M. Daraban & I.R. R˘ac˘anel
527
Strategy for durability of structural concrete in Mega-Sealinks in tropical sea-waters V.K. Raina Project Westgate – Lekki Beltway Bridge, Lagos, Nigeria C.M. Bednarski & A. Adão da Fonseca
533 549
Innovative construction methods High productivity in bridge construction – the OPS effect P. Pacheco, H. Coelho, A. Resende, D. Carvalho & I. Soares
559
FlexiArch-Stress Ribbon combination for multi-span pedestrian bridges A.E. Long, D. McPolin, S. Nanukuttan, A. Gupta & D. Robb
567
VIII
Balanced lift method for the construction of bridges with two spans S. Foremniak, W. Weiss & J. Kollegger An innovative system of precast segmental span-by-span construction for span lengths of above 100 m J. Muñoz-Rojas, S. Fernández, C. Iglesias, P. Pacheco, H. Coelho & A. Resende
575
583
Launching of fully welded steel long span bridges: Bogibeel bridge A.K. Mathur, S.S. Shukla & J. Gupta
591
Swivel lowering operation of the viaduct over the River Tera F.J.M. López, M.B. García, M.M. Cañueto, J.M.G. Parejo & M.T. Serrano
599
Deck forces of a cable-stayed bridge – “Analysis of the construction and the in-service phases” P. Almeida & R.C. Barros
607
Building the decks of the world’s largest high speed train arch bridges with movable scaffolding systems A.A. Póvoas
615
The Patani Bridge (Nigeria): Innovative construction methods P. Stellati & L. Marenzi
625
Innovative spliced girder method for multi span bridges I.Z. Stern
633
Innovative formwork systems in bridge construction – Case studies A. Preuer, M. Kamleithner, M. Mihal & C. Beer
641
Prestressed I-beams made of ultra-high performance concrete for construction of railway bridges ˇ P. Tej, J. Kolísko, P. Bouška, M. Vokáˇc & J. Cech Preliminary assessment of wind actions in large span MSS A. Resende, H. Coelho & P. Pacheco
649 655
Cabriel River Viaduct in Cofrentes (Valencia, Spain) bypass at N-330. Construction design J.F.M. Soriano, J.I.C. Vázquez & B.D. Santana
663
Segmental precast technology for multi-span bridges (production, transportation and launching) V.N. Heggade
673
Construction of Panipat Elevated Expressway on NH-1 on BOT basis P.N.S.S. Sastry
701
Mold for full span method M. Kye
707
Special foundations and geotechnical site investigations Offshore pile driving foundations monitored by PDA® Test at Puente Nigale M. Rojas, I. Miquilena & A. Souza Ceira bridge foundations: Combined Micropile and Footing Foundations (CMFF). Static load tests J.M.S. Cruz, M.S. Neves & S. Gil IX
715
721
Tresfjord Bridge – Foundation of main span on 40 m caisson on soil seabed K.B. Dahl, L. Toverud & D.E. Brekke
729
Chiapas Bridge G.R. Argüelles
737
Life cycle Life-cycle costs of bridge bearings – Key considerations for bridge designers and owners T. Spuler, N. Meng & G. Moor
743
Application of the Monte-Carlo method to calculate the life-cycle costs of bridges C. Hofstadler & M. Kummer
751
Selective use of non-corrosive rebar to increase concrete durability A.E.C. Borderon
759
Monitoring, maintenance and management Dynamic characterization and continuous dynamic monitoring of long span bridges E. Caetano, A. Cunha, C. Moutinho & F. Magalhães Investigation and countermeasures for fatigue cracks that emerged on the finger joint of the cable-stayed bridge T. Kosugi, M. Takahashi, Y. Nakamura & H. Dobashi Management of the Severn Bridge Suspension Bridge C.R. Hendy, C. Mundell & D. Bishop Surveillance of continuous precast concrete bridge decks supported by monitoring-based techniques H. Sousa, C. Sousa, A.S. Neves, J. Figueiras & J. Bento Implementation of a B-WIM system in a centenary steel truss bridge F. Cavadas, B.J.A. Costa & J. Figueiras A novel inspection method for orthotropic steel decks using phased array ultrasonic testing T. Makita, H. Sakai, T. Suzuki & N. Yagi Self-evaluating smart expansion joints of multi-span and long bridges K. Islami & N. Meng
771
781 789
799 807
815 823
Evaluation of fatigue crack formation in cantilever brackets of a multi-span railway steel box girder bridge L.R.T. Melo, R.M. Teixeira, A.P. da Conceição Neto & T.N. Bittencourt
831
Investigations of post tensioned bridges with critical prestressing steel regarding hydrogen induced cracking (HIC) A.W. Gutsch & M. Walther
841
Fatigue management of the midland links steel box girder decks C.R. Hendy & S. Chakrabarti Improved structural health monitoring strategies for better management of civil infrastructure systems J. Winkler, C.R. Hendy & P. Waterfall X
847
855
Assessment of thermal actions in the steel box girder of the Millau Viaduct L. Defaucheux, H. Desprets, Z. Hajar, C. Servant & M. Virlogeux
863
Delayed deformations of concrete structures: The Savines bridge and the Cheviré bridge J.-P. Sellin, J.-F. Barthélémy, G. Bondonet, B. Cauvin & J.-M. Torrenti
871
Laser scanner in identification of pathological manifestations in concrete S. Pavi, P. Gorkos, F. Bordin, M. Veronez & M. Kulakowski
879
Management of the M4 Elevated Section substructures C.R. Hendy, C.T. Brock, A.D.J. Nicholls & S. El-Belbol
887
Using data mining and numerical simulations for on-line monitoring of long span bridges J. Santos, P. Silveira, C. Crémona, A. Orcesi & L. Calado Monitoring based assessment of fatigue resistance of 40 year old pc bridges H. Weiher & K. Runtemund Maintenance method for cable-stayed and extradosed bridge with composite main girder H. Sakai
895 903
911
Construction control of a long-span single pylon cable-stayed bridge C. Liu, L.J. Sun, Y.S. Ni & D. Xu
919
Effect of cable corrosion on the structural response of cable-stayed bridges O.A. Olamigoke, G.A.R. Parke & M. Imam
927
Fatigue analysis of cable anchorages on cable-stayed bridges N.A.M. Khairussaleh, G.A.R. Parke & M. Imam
937
Monitored-based methodology to predict the initiation of corrosion in RC structures E.A. Tantele, R.A. Votsis & T. Onoufriou
947
Analysis of indicators in concrete production decrease in Distrito Federal – DF: problem notes and solutions R.S. Simões, H.R. Filho, C.D.U. Palacio, M.T.M. Carvalho & S. da Silva Araújo
955
Incidents and accidents Structural performance of cable-stayed footbridges to the loss of cable(s) O.A. Olamigoke, G.A.R. Parke & M. Imam
967
Causes of the bridge falsework collapse near Levoˇca in Slovakia P. Paulík, J. Halvoník, V. Benko & L’. Fillo
975
New materials and special devices Lightweight concrete for long-span bridges R.W. Castrodale
985
Cable stayed footbridge with the deck made of UHPC J.L. Vítek & M. Kalný
993
Non-destructive measurements to evaluate fiber dispersion and content in UHPFRC reinforcement layers S. Nunes, F. Ribeiro, A. Carvalho, M. Pimentel, E. Brühwiler & M. Bastien-Masse XI
1001
New test methods for stay cable systems A.W. Gutsch, M. Laube & T. Nolte
1009
Mechanical properties and explosive spalling behavior of the recycled steel fiber reinforced ultra-high-performance concrete G.F. Peng, J. Yang, Q.Q. Long, X.J. Niu & Y.X. Shi
1019
Extreme loads Characteristic of traffic loading response for multi-span large bridge J.Y. Zhou, X. Ruan & C.C. Caprani Multivariate probabilistic seismic demand analysis of steel-concrete composite bridges under near-fault pulse-like ground motions Y. Liu, D.G. Lu & F. Paolacci
1029
1037
Rehabilitation Innovative rehabilitation of large bridges – the Indian way P.Y. Manjure
1049
Widening of San Timoteo and Canero viaducts F.J.M. López, M.B. García, M.M. Cañueto, J.M.G. Parejo & M.T. Serrano
1057
Impregnation technique provides corrosion protection to grouted post-tensioning tendons D. Whitmore, I. Lasa & L. Haixue
1065
Assessment of epistemic uncertainties in the resistance of RC columns confined by CFRP J.R. Ferreira & S.M.C. Diniz
1073
Expansion joint renewal – Solutions that minimise impacts on the bridge’s structure, its users and its owner’s finances G. Moor, N. Meng & T. Spuler
1081
Safety and serviceability An efficient methodology for fatigue damage assessment of critical details on a long span composite railway bridge C.M.C. Albuquerque, A.L.L. Silva, A.M.P. de Jesus & R. Calçada
1091
Cyclic behavior of continuous railway viaducts made with U-shaped precast concrete girders C. Sousa, R. Calçada & A.S. Neves
1099
Concrete box girder bridge assessment – a stiffness adaptation approach G. Schreppers, A. de Boer & D. Begg
1107
Residual bridges bearing capacity analysis during service period subject to safety variability L.V. Miklashevich, L.A. Chernyshova & V.E. Rusanov
1115
Alkali-Silica Reaction, ASR – Review on how to deal with ASR in concrete structures J. Custódio, A.B. Ribeiro & A.S. Silva
1121
XII
Structural analysis Non-linear ULS analysis of long-span reinforced concrete arches to EN 1992 J. Nebreda & F. Millanes
1129
Stressing sequence of steel cable-stayed bridges built by cantilevering A. Recupero, M. Calvo, M.F. Granata & M. Arici
1137
Dynamic analysis for fatigue assessment of reinforced concrete slabs in railway viaducts J. Malveiro, C. Sousa, R. Calçada & D. Ribeiro
1143
Optimized bridge deck design using a genetic algorithm B. Lima & A. Ferreira
1151
Stiffened flanges used in steel box girder bridges P.S. Ferreira & F. Virtuoso
1163
Finite Element Modeling of the Fatih Sultan Mehmet Suspension Bridge S.A. Kilic, H.J. Raatschen, B. Körfgen, A. Astaneh-Asl & N.M. Apaydin
1169
Numerical simulation of wind pressure of a continuous fin-back bridge M. Fu, X. Li, Z. Nie & Z. Tang
1175
Modeling bridge construction phasing by the balanced cantilever method – “Comparison between predicted and real camber values” L.G. Castro, R. Bastos & R.C. Barros
1181
Fatigue analysis induced by vibrations in stay-cables subjected to along wind turbulence component I. Failla, A. Recupero, G. Ricciardi & F. Saitta
1189
Rational and practical method for camber control in bridges built by successive segments R.N. Oyamada, H. Ishitani, R.A. Oshiro & A.M.L. Cardoso
1197
Application of nonlinear FEM to evaluate load bearing capacities – Capability and limitations S. Kattenstedt & R. Maurer
1203
Thermal analysis of a fin-back bridge under sudden drop in temperature F. Tian, Y. Lu & P. Zhu
1211
Comparison study of aeroelastic analysis of a pylon of the Mersey Gateway Bridge with its 2D/3D wind tunnel tests S.B. Kim, J. Rees, J.Y. Chung, S.H. Jang, G.D. Moir & J.H. Seo
1217
Numerical models used to simulate the “in situ” testing of a bridge on A1 motorway in Romania I.R. R˘ac˘anel
1223
Wind effects analysis on cable stayed bridges decks V.D. Urdareanu & I.R. R˘ac˘anel
1231
“Beam sectional analysis” an innovative technique for analysis of bridge superstructure K. Kashefi & A.H. Sheikh
1239
Large bridge in pergola for high velocity trains in Spain C. Jurado
1247
XIII
Advances of external prestressing tendons in multi-span curved box-girder bridges Y. Shen, T.Y. Song & G.P. Li
1255
Reinvestigation of post–tensioned bridge over Bitlis river B.D. Öztürk, E. Löker, E. Ökte & E. Talıblı
1263
A graphic exact method for analyzing hyperstatic spatial pergolaes A.G. Lacort
1271
Investigation on stability problems as a second order theory problem for piers with practically infinite bending stiffness V. Karatzas, G. Karydis, E.K. Roussou & T. Konstantakopoulos
1279
Suspension cables bridge and arches L.M. Laginha
1287
Author index
1295
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Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
Preface
Throughout the last decades, the increasing development of the urban metropolis and the importance of establishing fundamental infrastructure networks promoted the development of important infrastructure projects worldwide and several multi-span large bridges have been erected. Some are sea-crossing bridges, some are long viaducts and others include both sea-links and large viaducts. Moreover, due to their significant benefits to society, multi-span large bridges are being studied for potential execution in several countries over the next decades. Some are already underway. There are definitely several problems/solutions which are common to the wider field of bridge engineering and certainly many lessons emerge from there. However, the approach to multi-span large bridges comprises a specific knowledge in design, construction and managing points of view, where the scale – large – the repetition property – multi-span – and the frequent inclusion of main bridges – demand integrated solutions to satisfy society’s requests. The aim of the Multi-Span Large Bridges International Conference is to aggregate experts and experiences in a global meeting where the outputs – interaction and documents – are to effect an increase in the performance and knowledge of all participants. There is also the purpose of sharing state-of-the-art achievements not necessarily originating in this specific field of bridge engineering, but which clearly demonstrate a potential for application in multi-span large bridges. The increasing demand on safety, time and economic issues represents a challenge to all subcommunities of bridge engineers. In the particular case of the Multi-Span Large Bridges experts community, the knowledge is actually somehow limited to a restricted number of companies and entities in dispersed countries – sometimes only with singular, but important, experiences. That is the main goal of the conference: merge knowledge of disperse experiences. To this end, worldwide prestigious bridge engineers, from different regions of the world, were invited to share their experiences. Nearly 150 contributions from designers, constructors, members of academia and researchers were selected within an intensive undertaking of the conference’s Scientific Committee, who reviewed papers of more than 400 authors. These contributions were organized in 13 thematic sessions, as follows: • • • • • • • • • • • • •
Experts, Experiences & Landmark projects Conceptual design Innovative construction methods Special foundations and geotechnical site investigations Life cycle Monitoring, maintenance and management Incidents and accidents Durability New materials and special devices Extreme loads Rehabilitation Safety and serviceability Structural analysis
All the papers were peer reviewed by the Scientific Committee which, itself, also comprises distinguished bridge engineers and academics from more than 30 countries. Multi-Span Large Bridges is already a fact for the bridge engineering worldwide community – more than 50 countries are actively involved. An important part of the shared knowledge is in this book. Pedro Pacheco & Filipe Magalhães XV
Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
Acknowledgements
Multi-Span Large Bridges International Conference and this book are the result of an intensive contribution of a plethora of members of the bridge engineering worldwide community. Thus, the Editors are deeply grateful: To their colleagues of the Organizing Committee who shared their knowledge, their experience and their time, to prepare this conference; To the Invited Keynote Speakers who, with generosity, found the time to share their knowledge, experience and vision; To the Scientific Committee Members who are responsible for the quality of the conference and of the book; To the Technical Committee Members whose work is to be done after this book edition and before the conference, but who already found time and availability for cooperation; Of course, the Editors deeply thank the Authors whose projects, research works and experiences are the core of this book. The Editors and the Organizing Committee wish to thank the International Co-Sponsors who gave so much support to the Organization and to the quality of the event.
and the main Sponsors, whose contribution is of major importance to provide the necessary means for the organization. A grateful word to all the remaining Sponsors whose names will be published on the conference website.
Finally, a special acknowledgement of the Conference Secretariat: fundamental for this event.
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Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
Committees
ORGANIZING COMMITTEE Pedro Pacheco, University of Porto, FEUP – Chairman Filipe Magalhães, University of Porto, FEUP – Co-Chair João Almeida, University of Lisbon, IST – Co-Chair Manuel Pipa, National Laboratory for Civil Engineering, LNEC Paulo Cruz, University of Minho, EAUM Rui Calçada, University of Porto, FEUP António André, University of Algarve, UA INTERNATIONAL SCIENTIFIC COMMITTEE Raimundo Delgado, Professor, FEUP – Honorary Chairman Pedro Pacheco, FEUP – Chairman Filipe Magalhães, FEUP – Co-Chair Adrian Long, United Kingdom Airong Chen, China Alan O’Connor, Ireland Alvaro Viviescas, Colombia Amin Ghali, Canada Aníbal Costa, Portugal António Adão da Fonseca, Portugal Antônio Laranjeiras, Brazil Antonio Martinez-Cutilhas, Spain António Reis, Portugal Atorod Azizinamin, USA Azlan Bin Adnan, Malaysia Bratilslav Stipanic, Serbia Catão Francisco Ribeiro, Brazil Christian Cremona, France Christos T. Georgakis, Denmark Dan Frangopol, USA Elsa Caetano, Portugal Erhan Karaesmen, Turkey Fernando Branco, Portugal Fernando Sima Brum, Uruguay Francisco Milanes Mato, Spain Galo Valdebenito, Chile Gordon Clark, United Kingdom György Balázs, Hungary Hanz Rudolf Ganz, Switzerland Helmut Wenzel, Austria Hugo Corres Pireti, Spain
Jan Biliszczuk, Poland Jan Vitek, Czech Republic Jens Sandager Jensen, Denmark Jin-Guang Teng, China Jiri Strasky, Czech Republic Joan Ramon Casas Rius, Spain João Almeida, Portugal João Almeida Fernandes, Portugal João Pires da Fonseca, Portugal Joaquim Figueiras, Portugal John Anderson, South Africa John O. Sobanjo, USA Juan Sobrino, Spain Júlio Appleton, Portugal Jung-hwi Noh, South Korea Ken Wheeler, Australia Luís Oliveira Santos, Portugal Makoto Kawakami, Japan Manuel Jara Díaz, Mexico Manuel Pipa, Portugal Marco Rosignoli, USA Mario Petrangeli, Italy Mauricio Lustgarten, Venezuela Michael Daebritz, Germany Miguel Angel Astiz, Spain Mike Schlaich, Germany Milan Kalny, Czech Republic Mirek Olmer, USA
XIX
Mourad Bakhoum, Egypt Murat Dicleli, Turkey Nikolaos Malakatas, Greece Paulo Cruz, Portugal Paulo Lopes Pinto, Portugal Paulo Silveira, Portugal Peter Paulik, Slovakia Rui Calçada, Portugal Rui Faria, Portugal Santinho Horta, Portugal Serge Montens, France
Stein Atle Haugerud, Norway Thomas Vogel, Switzerland Tiago Abecassis, Portugal Tor Ole Olsen, Norway Victor Barata, Portugal Vinay Gupta, India Vincent de Ville de Goyet, Belgium Virindra Kumar Raina, India Walter Dilger, Canada Wolfgang Eilzer, Germany Yozo Fujino, Japan
TECHNICAL COMMITTEE Ademir Santos, Brazil Alípio Ferreira, Portugal António Fonseca, Portugal António Hipólito, Portugal António Souza, Brazil Campos e Matos, Portugal Chris Hendy, United Kingdom David Ramos, Portugal Elbasha Nuri Mohamed, Libya Eneo Palazzi, Brazil Erik Andersen, Denmark Filemon Botto de Barros, Brazil Guy Fremont, France Gustavo Rocha Arguelles, Mexico Hugo Coelho, Portugal Javier Muños-Rojas, Spain Joaquín Arellano Casanova, Mexico Jorge Fandiño, Colombia
José Carlos Clemente, Portugal José Hemilio Herrero, Spain Luís Afonso, Portugal Luis Gustavo Zanin, Brazil Luis Matute, Spain Mihai Predescu, Romania Nelson Vila Pouca, Portugal Paulo Barros, Portugal Pedro Borges, Portugal Pedro Moás, Portugal Pedro Morujão, Brazil Raja Rizwan Hussain, Saudi Arabia Renan Ribeiro Setubal Gomes, Brazil Renato Bastos, Portugal Tiago Mendonça, Portugal Venkatram PG, India Ziad Hajar, France
SECRETARIAT Brigitte Rouquet Manuel Carvalho
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Keynote Lectures
Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
General presentation of the Keynote Lectures
Since an early stage of the Conference preparation, it was established, as a priority, that some of the most prestigious and experimented Bridge Engineers worldwide could share their knowledge and vision in the Conference. The invitation criteria comprised the intention of benefiting of different experiences, from diverse countries. The 10 invited Key Note Speakers are presented in the following paragraphs in a sequence that respects the antiquity of their activity in Bridge Engineering, with no further criteria. The very summarized presentations don’t need additional words – they talk by themselves. After this introduction of the Invited Key Note Speakers, manuscripts are presented for 8 of the Key Note Lectures. An additional Lecture of one of the Editors is added.
JAVIER MATEROLA ARMISEN Carlos Fernández Casado, S.L. Spain
– Education: M.Sc.C.E., 1962, Technical University of Madrid; Ph. D., 1964, Technical University of Madrid. – Memberships: Association of Spanish Civil Engineers (CICCP); Spanish Association of Concrete (ACHE); International Association for Shell and Spatial Structures (IASS); American Concrete Institute (ACI); International Association for Bridge and Structural Engineering (IABSE); Royal Academy of Fine Arts of San Fernando (Spain). – Resume of Experience: Since 1966 Mr. Manterola has been one of the leading engineers in Carlos Fernández Casado, S.L. and has specialised on bridge design. He is one of the founders of the company. – Projects Undertaken: Author of more than 200 works, mainly bridges. – Last Distinctions: Medal Féderation Internationale de la Précontrainte (F.I.P.), 1996; National Award for Civil Engineering, 2001; IABSE Award (Spain), 2002; IABSE International Award of Merit, 2006; Civil Engineering Foundation Award, Galicia 2010; Puente Alcantara Award, 2010; Gold Medal of Fine Arts Circle of Madrid, 2010; Honorary member of the Italian Association of Concrete Armato and Presforzado, 2011; Outstanding Engineer Award, College of Civil Engineering and Ports of Madrid, 2013.
3
NAEEM HUSSAIN Director, ARUP Hong Kong, China
Naeem Hussain graduated from the West Pakistan University of Engineering and Technology in 1962 and joined Kenchington Little and Partners now WSP in East Pakistan as a structural engineer, before transferring to their London office in 1964. In London he studied architecture at the Architectural Association School of Architecture and then went on to obtain a Masters degree in Concrete Structures at Imperial College. He is Fellow of the IABSE, Fellow of the IStructE, Fellow of the ICE, Fellow of the HKIE, and Fellow of the HKEng. He is currently an Arup Fellow, Director and Arup’s Global Leader for Bridges. Amongst the notable bridges that he has designed are Oresund Crossing between Denmark and Sweden, and Stonecutters Bridge in Hong Kong. He is the concept designer for the new Forth Bridge in Scotland currently under construction and the designer for the 30 km Brunei Muara–Temburong Sea Crossing in Brunei. JIRI STRASKY Technical Director Strasky, Husty and Partners Czech Republic
M.Sc. and Ph.D. from the Technical University of Brno, Czechoslovakia, DSc. from the Czech Academy of Science. Professional Engineer in 7 states of the USA and in the Czech and Slovak Republic. Professor of concrete structures at Brno University of Technology & Technical director of the design office Strasky, Husty and Partners, Brno, Czech Republic and Greenbrae, California, USA. Expertise in concrete and steel bridge design and construction. Experience in practically all structural systems – suspension, cable-stayed, stress-ribbon arch, cantilever and technologies – span by span and cantilever erection, launching and lifting of steel or concrete structures. Experience with elastic and plastic design of bridges built in severe seismic areas of California, Oregon and Taiwan. MICHEL VIRLOGEUX Michel Virlogeux Consultant SARL France
Graduated from the École Polytechnique in 1967 and from the École Nationale des Ponts et Chaussées in 1970. From 1970 to 1973 he served in Tunisia on road projects and at the same 4
time gained his Engineering Doctorate from the Pierre et Marie Curie University. In January 1974 he joined the Bridge Department of SETRA. In 1980 he became Head of the Large Concrete Bridge Division, and in 1987 of the large Bridge Division, Steel and Concrete. During twenty years he designed more than 100 bridges. In 1995 he set up as independent consulting engineer; his major achievements include his participation in the construction of the ‘Second Tagus Crossing’, the Vasco da Gama Bridge in Lisbon and the design of the Normandy Bridge and Millau Viaduct in France. Several of his bridges have received architectural awards. Since 1977 Dr Virlogeux has been a part-time professor of structural analysis at the prestigious École Nationale des Ponts et Chaussées and at the “Centre des Hautes Études du Béton Armé et Précontraint” in Paris. He also has been very active in technical associations. Dr. Virlogeux received the inaugural IABSE Prize in Venice in 1983 and has received many other international awards. He was appointed an International Fellow of the Royal Academy of Engineering in 2012. ANTÓNIO ADÃO DA FONSECA Faculty of Engineering, University of Porto Portugal
– Civil Engineering Diploma in 1971 – University of Porto (FEUP), Portugal. – PhD in Structural Engineering in 1980 – Imperial College, University of London, United Kingdom. – Fellow and Specialist in Structural Engineering of “Ordem dos Engenheiros” – Portugal. – Professor of Bridges at FEUP, until 2010. – Member of the Council of Ethics of University of Porto. – Research Fellow of Engineering Research Centre of Sustainable and Innovative Bridges, Fujian University, China, from 2013. – National President of Civil Engineering College of “Ordem dos Engenheiros”, in 1995–1999. – President of ECCE – European Council of Civil Engineers, in 1998–2002. – Vice-President of APEE – Portuguese Structural Engineering Association” – Portuguese Collective Member of IABSE. – President and Technical Director at “AFA – Consultores de Engenharia”, in 1985–2005. – Designer of bridges and special structures at “Adão da Fonseca – Engenheiros Consultores” (Portugal), from 2006, and at “ADEAM – Engenharia e Consultoria (Brazil)”, from 2012. CATÃO FRANCISCO RIBEIRO ENESCIL – Engenharia de Projetos, Ltda Brazil
Acknowledged as one of the most respected Brazilian engineers specialized in the field of bridges and viaducts, he is the technical manager at ENESCIL (Design Engineering) since 1976, immediately after receiving his professional degree in Civil Engineering from the Polytechnic School of the University of São Paulo. 5
Winner of three editions of the Prize “Gerdau Talent of Structural Engineering” (2008, 2010 and 2014) and of two honourable mentions in the same Prize (2011 and 2013), he displays in his curriculum the executive design of the Cable-stayed Bridge for the Roberto Marinho Avenue Roadway Complex (in São Paulo), the Cable-stayed Bridge over the Guama River (in Belém), the João Isidoro França Cable-stayed Bridge (over the Poti River in Teresina), the Negro River (in Manaus), the Cable-stayed Bridge over the Guanabara Bay (in Rio de Janeiro), the Cable-stayed Bridge with a viewing deck across the Piaui river (in Sergipe), and of the Basic Design for the 12.5 kilometres-long Cable-stayed Bridge linking the city of Salvador to Itaparica Island in the Bay of All the Saints (Bahia), the latter carried out in partnership with COWI International, amongst many others. His design work is guided by the quest for economy, construction simplicity and great durability, allied to a bold aesthetics, always integrated to the environment. Until now, he has participated in more than 3000 projects of special engineering works of art, with the most varied structural concepts, from cable-stayed works to works with pre-fabricated beams and slabs. WOLFGANG EILZER Leonhardt, Andrä und Partner Beratende Ingenieure VBI AG Germany
Wolfgang Eilzer graduated as Diplom-Ingenieur (M.Sc.) Structural Engineering from the University of Stuttgart in 1982. The same year he started his career with Leonhardt, Andrä und Partner. He has extensive experience within bridge engineering from numerous bridge projects worldwide, but especially from the transportation projects after the German reunification. In 1991 he established a branch office in Dresden, in 2001 he was appointed executive director, in 2013, when LAP changed its legal form to a corporation, Chief Executive Officer of Leonhardt, Andrä und Partner Beratende Ingenieure VBI AG. Wolfgang Eilzer was awarded with the Structural Award 2008 for the Roadway Bridge across the Lockwitz Valley, the Structural Award 2014 for the Elbe Bridge Schönebeck and with the German Bridge Award 2010 for the Design of the Elbe Bridge Mühlberg. Numerous bridges that have been designed under his responsibility won prizes in design competitions and have been awarded with national and international engineering awards and prizes. He is amongst others registered with the German Chamber of Engineers and with the German Association of Consulting Engineers. AKIO KASUGA Sumitomo Mitsui Construction Japan
– – – –
Born in 1957, Graduated in Civil Engineering (1980) from Kyusyu University. Working for Sumitomo Mitsui Construction since 1980. Technical Director & Bridge Designer. Design & Construction of more than 200 Bridges. 6
– – – – – –
Visiting Scholar of The University of Texas at Austin in USA (1989–1990). R&D of Stay Cable Damper, Optimization and New Bridge Systems. Technical Adviser of St. Croix River Crossing in USA (2011–2012). fib Presidium Member (2015–2018). fib Awards for Outstanding Structure 2006. Premier Prix of Tropfy Freyssinet 2013.
ARNE FREDERIKSEN BAEKSTED COWI Denmark
Mr. A. Frederiksen Baeksted has an all round experience in bridge engineering covering different sizes and types of bridge structures and all of the design stages within bridge design. Project management of bridge projects and integrated road/rail and bridge design including also other value engineering aspects. Leadership of teams capturing integration of requirements in tender documents, design intentions and adaptation of Contractor driven requests and preferred methods and equipment are one of his focal points. AIRONG CHEN Department of Bridge Engineering, Tongji University Shanghai, P.R. China
In 1983, he got his B.S. of Bridge Engineering from Tongji University. In 1989, he got his M.S. of Bridge Engineering from Xi’an Highway Institute. In 1993, he got his PhD of Bridge Engineering from Tongji University. He served president of Chinese Group of international association of Bridge Maintenance and safety (IABMAS), vice chairman of China Institute of Highway Bridge and Structural Engineering Branch. His research interests focus on life cycle design theories of bridge engineering, structural performance of bridge engineering under extreme events, wind-resistance and CFD technique study of bridge engineering, bridge forms and Topology Optimization. He was in charge of numerous major national scientific research projects and professional design guidelines. He has published 7 books and more than 200 journal papers. And he was awarded many national prizes for Progress in Science and Technology and many provincial prizes for progress in Science and Technology.
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Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
Large viaducts, some executions a few ideas J. Manterola Carlos Fernández Casado S.L., Spain
ABSTRACT: In large viaducts it is crucial to be spot-on regarding the bridge typology chosen as well as the building procedure, two facts that generally go hand in hand. The typology also depends on the characteristics intrinsic to the site, whether it is a river, a sea, a great valley, etc.
1 BRIDGE OVER THE BAY OF CÁDIZ The overall viaduct length totals 3,082 m (Fig. 1) which may be divided into four stretches. 1.1 The approach viaduct on the Cadiz side This stretch has an overall length of 570 m, and a 5% longitudinal slope necessary to reach the 69 m clearance gauge on the navigation canal. The 34.3 m wide deck has a constant cross section all along the bridge, made up of a 3.0 deep trapezoid in the centre, 10 m wide at the lower end. In this area the cross section is composite, with a concrete deck and steel trapezoid and webs. The typical span in this area is 75.0 m (Fig. 2). The whole access from the Cadiz side is carried out by incrementally launching the steel structure, assisted by a small cable staying at the front end of the first span. 1.2 The demountable stretch This stretch responds to a request by Navantia shipyards to allow for the passage of 100 m tall vessels unable to pass through the main bridge’s 69 m clearance. The demounting operations, however, are likely to happen few times throughout the bridge existence.
Figure 1.
General view of the bridge.
9
Figure 2.
Figure 3.
10
Figure 4.
Figure 5.
The demountable stretch has a 150 m long span and its cross section is based on the standard one with the following difference: the central trapezoidal area varies from the 3 m depth of the typical cross section to 8 m in depth in the middle (Fig. 3). The piers are also the same along the whole viaduct excepting the approach viaduct from the Puerto Real side where the shape of a double trapezium joined by the wider side is more pronounced. Here the highest piers are 10 m wide at the base, 4.5 m wide at the “waist” and then widen again to form 10.5 m wide capitals. 1.3 The cable-stayed stretch The cable-stayed stretch is 1,180 m long. This is the stretch spanning the navigable canal that runs from the Cabeziela Pier towards Cadiz. Due to the conditions imposed by the maritime Authority the bridge’s main pier had to be built 70 m into the Pier (Fig. 4). 11
Figure 6.
Figure 7.
Figure 8.
The cross section is composite, steel and concrete, where concrete is present through precast slabs. The span distribution along the cable-stayed stretch is 120 m + 200 m + 540 m + 200 m + 120 m. The length of 200 m on the direct compensation span was adopted instead of smaller lengths that produce greater stiffness in the main span but in turn have an ill effect in the middle of the Bay. The towers, whose vertical design leaves a grommet for the deck to pass through, are 180 m high and their cross section is a double trapezium joined by the long base. The dimensions at the base are 94 m × 9 m. There are 176 stay cables. 1.4 The Puerto Real stretch The access to the main bridge from Puerto Real, the side opposite Cadiz, has the same cross section only in this case it is wholly made up of prestressed concrete. The spans remain 75 m long, with the exception of the final area where they turn into 40 m (Fig. 5). 12
Figure 9.
This stretch presents two particularities. The shape of the piers changes at a certain point and opens up to make way for the longitudinal road traffic carriageway in the bridge axis. The opening is archived by simply separating the two trapeziums forming the pier. The second particularity of this stretch lies in the deck. The concrete deck is carried out in two phases: the first includes the building of the central, 10 m box girder, while in the second phase the transverse cantilevers that complete the cross section are built.
2 BRIDE OVER THE TAGUS RIVER The High-Speed Railway Line between Madrid and Extremadura crosses the Tagus River at its partial outlet into the Alcántara Reservoir. The layout runs 60 m above ground (Fig. 6). The bridge overall length totals 1,488 m. with a 60 m long typical span and the Reservoir crossing resulting in a 324 m span. The deck width is 14.00 m. It accommodates the double railway line and is composed of a 4.00 m deep box girder, 5.00 m wide at the lower slab and 6.5 m wide at the upper slab that forms part of the deck (Fig. 7). The cross section is identical on either side of the bridge. The arch has a box girder cross section of a variable depth, 4.00 m at the abutment and 3.50 m at midspan. Its width varies linearly from 12.00 m at the abutment to 6.00 m at midspan (Fig. 8). The piers connecting the deck to the ground have a variable height ranging from 9.6 m to 7.5 m. The thickness is constant amounting to 3.00 m while the depth varies ranging from 5.00 m at the top to 3.2 m at the “waist” to 5.5 m at the bottom. All piers are deduced from this basic model depending on their height. 13
Figure 10.
Figure 11.
The deck is built using a 60 m span scaffolding truss (Fig. 9). The arch is carried out applying the free cantilever method, cable staying the cantilevers from the piers founded over the arch abutment and a 54 m high superposed steel tower, with 9 pairs of cables coming out of the tower itself, −14 pairs of cables altogether, counting those coming out of the piers (Fig. 10).
3 THE VIDIN (BULGARIA) – CALAFAT (ROMANIA) BRIDGE OVER THE DANUBE The crossing of the River Danube presents at this point two clearly differentiated areas. The first one lies between Vidin and an islet situated in the center of the river (non-navigable part) and the other lies between the islet and Calafat (navigable part). A 150 m horizontal clearance was required on the navigable part, thus determining that the configuration of the bridge be ordered as follows: 124 m + 3 × 180 + 115 m. 14
Figure 12.
Figure 13.
On the non-navigable area 80 m spans were adopted, repeated along 612 m. As it disembarks on the Vidin side, the railway originates a series of 40 m long spans, split away from the road deck due to its smaller longitudinal slope (Fig. 11). The transverse cross section of the entire viaduct, excepting the mentioned final railway spans, is constant. It consists of a central box-girder of a 4.5 m depth along which runs the railway, plus two cantilevers on either side supported by inclined struts resting upon the central box girder. The overall width thus totals 31.35 m (Fig. 12). The construction on the non-navigable part of the river was carried out using precast segments to build the central box girder. The strutted lateral cantilevers were built subsequently by means of a form traveler rolling along the completed box girder (Fig. 13). 15
Figure 14.
The 180 m span over the navigable part of the river keeps the same cross section assisted by an extradorsed cable staying system hanging from small-height towers erected over the piers (Fig. 14).
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Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
Design and construction of sea-crossing bridges – A review N. Hussain Arup Fellow and Director, Hong Kong
ABSTRACT: The paper describes the author’s personal experience and involvement in the design of large sea-crossing bridges. The approach to design has been based on construction methods that allow fast-track quality construction with use of large pre-fabricated bridge elements in both steel and concrete. 1 INTRODUCTION 1.1 Need for large sea-crossing bridges Infrastructure plays a key and vital role in the economic development and well-being of a region or country. In many parts of the world waterways whether wide rivers, bays and estuaries have meant large detours and/or use of ferries thus severely hindering the movement of people and goods and limiting economic development. This has led to the bridging of these waterways with road and rail bridges and since the 1990’s several large sea-crossing bridges have been built mainly in the Far-East. The methods of construction have varied for these crossings but the driver has invariably been fast construction with pre-fabricated elements. Another feature of the crossings has been bridging of navigation channels associated with long waterway crossings which has led to the development of long span cable supported bridges. 1.2 Development and features of crossings The features and development of the crossings which also depends upon its location is described with reference to Oresund Crossing Denmark-Sweden, Shenzhen Crossing Hong Kong, Incheon Bridge Korea, Hong Kong Zhuhai Macao Bridge, Queensferry Bridge Scotland and Brunei Temburong Bridge, all of which are sea-crossing bridges. The primary feature of all these bridges is that foundation and substructure construction can proceed in parallel with precast or prefabricated superstructure elements off-site in factory conditions, thus appreciably shortening the construction period. Design for ship impact is also an important feature of the navigation channel bridges, that are invariably associated with long sea crossings, and various protection methods have been devised to safeguard the bridge towers adjacent to the navigation channels. 2 ORESUND CROSSING DENMARK-SWEDEN 2.1 Location and general features Oresund Crossing across the Danish Straits connects Copenhagen in Denmark with Malmo in Sweden and was opened in July 2000. The Danish Straits are of special importance because they provide the only natural connection between the Baltic and the open seas. The straits also function as hydraulic links and are profoundly important for the maintenance of water quality and survival of marine life within the Baltic. Any scheme for the crossing had to ensure that obstruction of water flow was as little as possible. The total length of the link is 16 km and comprises of an 4 km immersed tunnel, 4 km of artificial island and 8 km of bridge. The final alignment is shown in (Fig. 1). 17
Figure 1.
Figure 3.
Figure 2.
Figure 4.
2.2 Alignment and standard spans The Treaty specified an alignment which was simply a straight line from the artificial island south of Saltholm to the landfall in Sweden. We proposed a S-curve alignment for the bridge to give users of the Link more interesting views (Fig. 2). The bridge reference design in the treaty comprised of single level bridge structures carrying two tracks of high speed rail and a dual 2-lane with hardshoulder road. In the ARUP competetion winning design the road and rail was separated with the road above and rail below. With this arrangement the most economical structural solution was to use steel trusses with diagonals connecting the upper and lower decks. These trusses are uniform throughout the bridge, but modified at the cable-stayed main spans so that every other diagonal has the same direction as the cables. The 20 m bay length of the truss is constant along the bridge and imposes a modular discipline on all the spans. The deep composite girders lead naturally to longer spans, which have environmental as well as visual and construction advantages. Longer spans meant less obstruction to water flow to meet the limit to blocking set by the environmental authorities in both countries at 0.5%. We were aware of the special floating heavy lift ‘the Swanen’ which can lift a 7200 t payload out of which 6000 t can be a structural element. The water depth along the alignment is shallow and foundation bedrock is also at shallow depth. Pad foundations could be used and hence precast cellular foundations were designed which could be lifted into place by the Swanen. The Swanen was also able to lift precast hollow columns and whole 140 m long bridge spans into place (Figs 3, 4). The chosen construction method meant that both the substructure and superstructure could comprise of large concrete and steel bridge elements that could be precast or prefabricated in controlled factory type environment with good quality control thus ensuring a long life durable structure in an aggressive marine environment. 2.3 Navigation span ARUP proposed a single navigation span of 490 m over Flintrannan instead of the 330 m and 290 m spans over Flintrannan and Trindelrannan specified in the Treaty. A truss sufficiently deep 18
to accommodate the railway is naturally stiff enough to act as a deck for a cable-stayed span considerably longer than that required by the brief, so the opportunity was taken to provide only one navigation span at Flintrannan. The inherent stiffness of the truss deck was also a factor in choosing a harp configuration for the cables. The live load moments in a slender deck are sensitive to the vertical stiffness of the cable system, which strongly suggests a fan arrangement. This does not apply to the truss deck. Its repeating geometry has also a natural affinity with the harp, which can be emphasised by adjusting the angles of the diagonals to match those of the cables. The harp system has a visual formality, particularly apparent when cable planes are vertical, and the towers were designed to express this. The effect is further enhanced because each cable plane is supported by independent towers unconnected above deck level and was a major visual feature of the bridge. 2.4 Ship impact protection As the water was relatively shallow, approximately 12 m at the navigation span and for aesthetic purposes the ship protection to the towers is provided by submerged artificial earth mounds. 3 SHENZHEN CROSSING AND DEEP BAY LINK HONG KONG 3.1 Location and general features The Shenzhen Crossing and Deep Bay Link (DBL) connects Yuen Long Highway and the future Route 10 in Hong Kong with Shenzhen in China (Fig. 5). DBL is approximately 5.4 km long and Shenzhen Crossing across Deep Bay is approximately 5.0 km long. The design and construction of the bridge whilst being of fast-track nature also had to take into account the minimisation of ecological damage. Following the concept ARUP had developed for the Oresund Crossing between Denmark and Sweden, ARUP proposed an S-shaped horizontal alignment for the bridge as shown on Figure 8. 3.2 Marine viaduct The marine viaduct is formed in modules of bridges. Each bridge consists of 8 continuous span concrete boxes with 75 m internal spans and 70 m end spans (Fig. 6). The constant depth viaducts were built by the precast segmental balanced cantilever technique using a combination of external and internal prestressing cables. The precast segments were constructed in China and delivered to the bridge site by barge. Precasting allowed quality construction of the segments. 3.3 Navigation spans There are two navigation channels in the bay, and each of them is bridged with a single inclined tower cable stay bridge. One cable stay bridge is in Chinese waters and the other in Hong Kong waters. The inclination of the towers are deliberately inclined towards each other to indicate amity between the Shenzhen and Hong Kong people (Fig. 7). The deck of the navigation channel bridges are orthotropic steel box girders. Almost the full length of the main was fabricated off-site, brought to site by barge and lifted by strand jacks into place. 3.4 Ship impact protection The foundations of the towers are large diameter bored piles. Piled dolphins are used for protecting the towers (Fig. 7). 19
Figure 5. Alignment.
Figure 6.
Figure 7.
4 INCHEON BRIDGE KOREA 4.1 Location and general features Incheon Bridge is a 12.3 km long sea crossing in South Korea. It connects the new Incheon International Airport on Yeongjong island to Songdo (New City). The majority of the length of the bridge is constructed as low level viaduct structures with pretensioned precast 50 m long concrete box girder spans. Where the alignment rises to cross the navigation channel, precast segmental balanced cantilever approach bridges with 145 m spans link the viaducts to the cable stayed bridge which provides the 800 m long navigation span (Fig. 8). 4.2 Low level marine viaducts The low level viaducts consist of 50 m spans and 250 m long five span bridge units. The soffit of the bridge is typically 4.5 m above H.H.W.L. and the substructure generally consists of pile bents with pile caps only adopted in deeper water. The 50 m spans are pre-tensioned and precast in a single pour in the contractor’s specially constructed casting yard. The spans are then erected using the Full Span Launching Method (FSLM). Since much of the viaduct is in shallow water and tidal flats which are inaccessible by floating cranes a self launching overhead gantry system was used to erect the deck (Fig. 9). However, the end of the viaduct is in deeper water and so each 1350 t precast span is lifted by floating crane onto multi-wheel transporter units which then deliver the span to the erection front. 4.3 Navigation span bridge The cable stayed bridge is a 1480 m long structure with an 800 m main span. Two planes of stay cables support a 33.4 m wide orthotropic steel box girder. The pylon is a reinforced concrete hollow section in a diamond configuration which provides torsional stability to the main span and minimises the size of foundation which must be protected from ship impacts. 20
Figure 8.
Incheon bridge.
Figure 9. Viaduct and launching gantry.
4.4 Ship impact protection Ship impact protection is provided in the form of circular sheet piled dolphins filled with crushed rock and tied together with a reinforced concrete cap. The dolphins were designed to provide both deterministic and probabilistic protection, the former being to stop a 100,000 DWT design vessel travelling at 4.5 m/s directly towards the cable stayed bridge pylon and the latter being to reduce the annual collapse frequency to less than 1 in 10,000 when considering a distribution of design vessels heading towards any point on the bridge axis in any direction. The dolphins work by dissipating energy through various mechanisms; crushing of the ships bow, local deformation of the dolphin, passive resistance of the soil and friction between ship and dolphin. A reliable way to estimate impact dissipation in soil structures is through testing of a physical model in a centrifuge which allows earth pressures to be correctly modelled at a reduced scale. However, due to the time and expense required for centrifugal model testing it is preferred to use the results to calibrate a non-linear finite element analysis which will then allow analysis of different configurations. This method, which had previously been adopted for Stonecutters Bridge (Lee & Peiris 2004), was followed for the design of the Incheon Bridge ship impact protection. 5 HONG KONG ZHUHAI MACAO BRIDGE 5.1 Location and general features The alignment of the bridge is shown in (Fig. 10). Currently the delta is bridged approximately 50 m upstream of the mouth of the delta and the journey time between Hong Kong and Zhuhai/Macao is about 4 hours by road and 1 hour by fast ferries. The new link will reduce the journey time to approximately 30 minutes and more importantly provide a safe and fast link to Hong Kong International Airport and Hong Kong Shipping Container Terminals. The link with a total length of 42 km will have one of the longest sea-crossing bridges in the world. Generally the orientation of the alignment has been kept normal to the water flow in order to minimize obstruction to flow of water and horizontal curves have been introduced to provide interesting views of the main bridge as seen by the driver and occupants of the vehicles. The alignment starts from the Boundary Crossing Facilities opposite Zhuhai & Macao, runs in open waters and ends at the Boundary Crossing Facilities at the north-east tip of Hong Kong Airport. The general arrangement of the link in mainland waters comprises of: 75 m short span viaducts approximately 7 km in length; 110 m long span viaducts approximately 14 km in length; Jiuzhou Navigation Bridge approximately 500 m in length; Jianghai Navigation Bridge approximately 700 m in length; Qing Zhou Navigation Bridge approximately 900 m in length; Approximately 5 km of immersed tunnel with two artificial islands. The Hong Kong section is approximately 12 km long and generally comprises of viaducts with spans ranging from 70 m–180 m. 5.2 Marine viaducts A number of long sea and river crossings have recently been constructed in China such as Sutong Bridge, Donghai Bridge, Hangzhou Bridge. In all of these bridges two separate prestressed concrete 21
Figure 10.
Figure 11.
HKZMB components.
Figure 12.
Figure 13.
box girder decks have been used, each supported by single column piers under each deck. For this bridge, from environmental considerations, it was decided to use a single column piers to support either two separate decks or a single wide deck. The reason for this is to provide the least obstruction to water flow specially as the water flow is not always normal to the alignment of the bridge. To further minimize obstruction to water flow, the piles are going to be buried in the sea-bed. 5.2.1 Short span viaducts The construction and whole life cost of the structure is dependent upon the cost of site establishment and preliminaries such as fabrication and assembly yards, transportation, availability and cost of large floating cranes, launching gantries, maintenance costs etc. The exercise showed that whilst the precast concrete boxes were possibly the cheapest, the single wide composite box with a span of 75 m is the optimum solution (Fig. 11). 5.2.2 Long span viaducts Cost analyses of concrete, composite and orthotropic steel boxes showed that the single wide orthotropic steel box girder with a span of 110m is the optimum solution taking into consideration quality construction, construction equipment, and construction period. (Fig. 12). 5.2.3 Foundations The substructure comprises of piles, pile-caps and pier columns with large diameter bored piles, buried pile-caps using precast housing as permanent formwork for the pile-caps and precast hollow pier columns. This construction method limits insitu concrete construction to piles and pile-caps, thus helping in minimising environmental impact and shortening the construction period. 22
Figure 14. Options to stabilize the internal towers n a multi-span cable-stayed bridge.
Figure 15.
5.3 Navigation channel bridges There are three navigation channel bridges. Jizhou Bridge near Zhuhai with a cable stayed main span of 260 m is visually the most interesting design with a distinctive central sail tower and composite deck, Figure 13. 6 FORTH REPLACEMENT CROSSING (QUEENSFERRY BRIDGE) 6.1 Location and general features The Forth Replacement Crossing is currently being built across the Firth of Forth to maintain and improve reliability of a vital transport link in Scotland. The total length of the new bridge, including approach viaducts, is approximately 2.7 km. The cable stayed section will include two 650 m spans to cross the two major navigation channels – the Forth Deep Water channel and the Rosyth Navigation channel. Beamer Rock divides the two channels, and forms the location for the central tower. The cable-stay bridge is a unique 3-tower cable stayed bridge with a pair of 650 m main spans across and overlapping stay cables in the middle of the main spans to stabilize the central tower. With two main spans required over the navigation channels, a major challenge the design had to address was the stability of the internal tower. As the internal tower is not connected to a stiff back span structure, out of balance live loading on only one of the main spans causes a significant sway of the tower resulting in large deflections of the tower and deck and large bending moments in the tower. This issue is well known for multi-span cable stayed bridges, and there are a number of configurations which can be adopted to stabilise the internal tower. The simplest of these is to adopt a very stiff deck, or very stiff towers. Other configurations are shown in Figure 14 in the following order: Provide anchor piers; Tie the top of the towers with stabilising horizontal cables; Use sloping stabilizing cables from the top of each internal tower to the junction of the deck with the adjacent towers Use overlapping stay cables. The option to extend the length of the stay cable fans beyond mid-span, so that they overlap in the central region of each span was investigated in detail to ascertain if this configuration provided a good solution. Parametric studies were carried out to investigate the ability of the overlapping cables to provide the stiffness required. As the length of the overlapping zone is increased, the system becomes stiffer, and the bending moments in the tower and deck reduce. The arrangement adopted in the final scheme is to overlap the stay cables over approximately 25% of the main span. (Fig. 15). 6.2 Ship impact protection The southern main span crosses the Forth Deepwater Channel, the main access to the upstream ports (Fig. 16). The northern main span crosses the approach into Rosyth port. A quantitative 23
Figure 16.
Figure 17.
Ship impact – workflow.
marine collision risk assessment (Carter et al. 2010), based primarily on Eurocode, was carried out to determine the appropriate design impact forces for the foundations and substructures. This risk assessment, based primarily on BS EN 1991-1-7 (2006), forms the backbone for determining the actual ship impact forces on the structure (Fig. 17). The navigational conditions in the vicinity of the bridge are complex, with bends in the navigation channels and significant obstructions, not least of which is the existing Forth Rail Bridge as shown in Figure 16. The holistic model of AASHTO (2009) was not adequate to address this and a semi-holistic model was developed following the principles of Eurocode and taking account of specific features of the site. The semi-holistic model considers: vessel aberrancy at any point on the transit paths in the vicinity of the bridge leading to a large number of aberrancy scenarios (defined solely by the point of aberrancy); post-aberrancy behaviour of the vessel in a holistic manner without attempting to explicitly track the path and velocity of the vessel taking into account specific human, mechanical and metocean factors. 7 TEMBURONG LINK BRUNEI 7.1 Location and general features The new 30 km Cadangan Projek Jambatan Temburong (Temburong Bridge Project) in Brunei will connect the relatively isolated district of Temburong with the more developed Brunei-Muara district (Fig. 18). link will comprise 14,6 km long marine viaducts, two cable stayed bridges across 24
Figure 18.
navigation channels in Brunei Bay, 12 km of low height short span viaduct across the Temburong peat swamp forest and a small area of mangroves, and approximately 3,6 km road in Brunei-Muara district where 3 lengths of tunnels are required as well as at-grade roads and viaduct ramps to link with the existing road network. The key challenges include very soft ground conditions, shallow waters, difficult access, lack of local raw materials and the required fast-track programme. 7.2 Marine viaducts Brunei Bay is shallow except at the two navigation channels. There are mud flats in the Bay, as well as mangroves near the landing point on Temburong side and other islands. The ground is very soft, with typically 20 to 30 m of soft marine clay. The shallow waters, very soft ground conditions and the fast track programme pose significant challenges to both design and construction. The viaducts will be in the form of twin post-tensioned concrete single-cell, box girders with e 50 m spans. Typical modules are formed by 6 spans of continuous deck. These standard modules have typical box cell with 3.8 m wide soffit (Fig. 19). The box girders will be post-tensioned with a combination of internal and external tendons. To achieve the 45-month programme, the viaducts were designed for construction by precast segmental span-by-span erection method. Construction by precast full-span launching method is also permitted. Continuity between separate spans is achieved by a 200 mm wide cast in-situ stitch. The successful contractor has chosen to use the full-span launching method and erecting two spans at the same time with a specially designed gantry which will be a world first. Due to the shallow water, dredging will be carried out to gain access to carry out piling works. Three types of piles are used: (i) majority of the piles will be 1 m diameter concrete spun piles, which are precast circular hollow sections with high density concrete achieved by spinning during the manufacturing process; (ii) steel tubular piles of 1 m diameter are adopted for locations where hard mudstone is at shallow depth and (iii) steel tubular piles of 1,6 m diameters are adopted at piers subject to higher ship impact forces near the navigation channels. One of the key drawbacks of the use of concrete spun piles is that they have limited capacity to resist bending and can behave in a brittle manner therefore high damping rubber bearings have been used for all the bearings of the marine viaducts. Not only this reduces the seismic force on the foundations, it also reduces the sensitivity of the design to the ground conditions. 7.3 Navigation span bridges There are two navigation channels in Brunei Bay, namely Brunei Channel and Eastern Channel. Each channel will be crossed by a cable stayed bridge. The Brunei Channel Bridge will be a gateway between the Brunei-Muara and the Temburong districts and hence designed as an iconic bridge. 25
Figure 19.
Figure 20.
Eastern Channel Bridge.
The Eastern Channel Bridge follows a similar form. To cater for future navigation needs of trade and navy vessels, Brunei Channel Bridge will have a main span of 145 m with the soffit level at about 23 m. The Eastern Channel Bridge crosses an international waterway with a main span of 260 m and the soffit level at about 34 m. Both bridges have a concrete ladder beam deck which comprises two edge girders and cross beams between the edge girders, with concrete slab spanning between the cross beams. Foundations will be 2,2 m diameter concrete bored piles (Fig. 20). 7.4 Ship impact design Apart from the two navigation channels, the rest of Brunei Bay is very shallow. Current vessel movements are very scarce in the Bay, with the largest vessel recorded being 36 m long. Despite that there is no known plan for future port development, a larger 80 m long rivertrade vessel was selected as the design vessel to safeguard future development opportunities. The bridges are designed for ship impact in accordance with BS EN1991-7 (2009). As there is generally no specific rule on the probability based analysis in this standard, the methodology in AASHTO Guide Specification and Commentary for Vessel Collision Design of Highway Bridges (2006) was adopted to carry out the probability based risk assessment. A large number of vessels was assumed conservatively such that the design ship impact forces are essentially established in a deterministic way. For piers further away from the navigation channels, vessel collision associated with drifting vessels at the speed of water current has been considered for robustness. 8 CONCLUSIONS Sea-crossing bridges have now been constructed upto 46km in length and longer crossings are in planning stages or being considered. Many of these bridges will be in deep water and across shipping lanes. Methods of foundation construction will be a development of off-shore oil platforms and environmental considerations of spillage from oil carrying vessels hitting the towers will be a major concern. It is inevitable that longer spans will be used and design and protection for ship impact will no doubt require further research and innovative approach. The use of composites will lead to lighter decks and longer spans with large prefabricated bridge units. REFERENCES Lee D.M. & Peiris N. 2004. Modelling of Ship Impact on a Bridge Foundation, IABSE Symposium Shanghai 2004: Metropolitan Habitats and Infrastructure, IABSE Report, Volume 88. Carter M., Kite S., Hussain N. 2010. Ship Impact Studies of the Forth Replacement Crossing, IABSE Symposium Venice 2010: Large Structures and Infrastructures for Environmentally Constrained and Urbanised Areas, September 2010, IABSE Report: A-0315. BS EN 1991-1-7:2006, “Eurocode 1. Actions on structures. General actions. Accidental actions”, British Standards Institution, September 2006. American Association of State Highway and Transport Officials, 2009. “Guide Specification and Commentary for Vessel Collision Design of Highway Bridges”, 2nd Edition.
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Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
Viaducts with progressively erected decks J. Strasky Brno University of Technology & Strasky, Husty and partners, Brno, The Czech Republic
ABSTRACT: Two types of bridges with progressively erected decks are described in terms of the architectural and structural solution and technology of their erection. The first type is formed by a concrete spine box girder that is additionally widened by precast struts and cast-in-place overhangs. The bridges have spans from 42 to 69 m; their width is from 25 to 31 m. The second type is formed by a steel trough that is composite with a concrete deck slab supported by steel struts. The deck slab is cast in a formwork formed by precast members. The bridges have spans from 45 to 66 m; their width is from 25 m.
1 INTRODUCTION Bridge structures formed by concrete or steel box girders with large overhangs supported by struts represent an optimum solution for crossings of deep valleys – see Figure 1c. These bridges are esthetically pleasing and structurally very efficient. Their economy can be enhanced by a progressive erection of their decks. This is illustrated on a construction of several viaducts built in the Slovak and Czech Republic.
Figure 1. Typical arrangement of long viaducts.
27
Figure 2. Precast struts formed by: a) single bars, b) a truss, c) slabs.
Figure 3. Shear flow in the: a) one cell box girder, b) three cell box girder, c) pseudo three cell box girder.
Figure 4. Viaduct across the Hostovsky Creek: Completed bridge and its erection.
The supporting struts can be formed by single bars, by a truss, or by slabs (see Figure 2). The truss and slab struts, taken together with the top slab, form a pseudo three-cell box girder – see Figure 3c. These struts significantly contribute to the bending and torsional resistance of the structure.
2 CONCRETE VIADUCTS Thirty year ago a cable stayed bridge across the River Elbe near a city of Podebrady was erected. Its 31.80 m wide deck is formed by a spine girder with large overhangs supported by not mutually connected precast slab struts. The one cell box girder assembled from precast segments was constructed first, then the struts were erected and the overhangs were cast in simple formwork that was supported by these struts. After that similar arrangement was used in a construction of the Vrsovice cable stayed bridge built in Prague and in several bridges designed by others. Recently similar approach has been used in construction of several long viaducts that have been built in Slovakia – see Figure 4. These bridges have span lengths up to 69 m (see Fig. 5), their widths are up to 28.70 m (see Figures 6 and 7). The spine girder is cast span by-span in a formwork suspended on a special overhead gantry with ‘organic’ prestressing system (OPS) that eliminates 28
Figure 5.
Concrete viaducts: elevations.
Figure 6.
Concrete viaducts: Structural arrangement of the deck.
deflection of the gantry. To reduce the self-weight of the spine girder as much as possible, the girder is very narrow. Therefore the transverse projection of the overhangs up to 11.00 m. It is evident that this solution requires not only a careful analysis and detailing, but also an experience contractor. Also the construction and service of these bridges have to be carefully checked and monitored. 2.1 Recent structures The first structure of this kind is a 975 m long viaduct across the Hostovsky Creek built on the Expressway R1 near a city of Nitra – see Figures 4 and 5 – R1 205. The bridge of the width of 25.66 m has 17 spans of lengths of 33.0 + 42.0 + 45.0 + 48.0 + 9 × 69.0 + 48.0 + 45.0 + 42.0 + 33.0 m. The depth of the girder varies from 4.00 to 2.60 m. The bridge was opened to traffic in September 2011. 29
Figure 7.
Cross section of the bridge D1 216.
Two bridges of similar arrangement are being built on a motorway D1 section Fricovce – Svinia near a city of Presov – see Figure 5 – D1 202 and 203. These bridges built across the Lazny and Stefanovsky Creek have total lengths of 269 m and 182 m; typical span length is 45 m. Both motorway directions are carried by one bridge of the total width of 29.5 m. The depth of the girder is 2.60 m. Since the bridge decks are frame connected with H shaped piers, the bridges form semi-integral structural systems. Another two bridges of similar arrangement are also being built on a motorway D1 section Janovce – Jablonov near a city of Levoca – see Figure 5 – D1 216 and 217. The bridges across the Lodina and Doliansky Creek Valley have total lengths of 367 m and 414 m; typical span length is 65 m. Both motorway directions are carried by one bridge of the total width of 28.70 m. The depth of the girder is from 4.00 to 2.60 m. Since the bridge decks are hinge connected with twin piers, the bridges form semi-integral structural systems. The spine girder of all bridges is progressively cast span–by-span in a formwork suspended on the overhead gantries. The girders are cast with short overhanging cantilevers. The decks of all bridges are longitudinally prestressed by internal bonded tendons situated within the basic cross section and by external non-bonded tendons situated inside the central box. The bonded tendons are coupled in each construction joint. External cables are anchored at pier diaphragms and are deviated at pier and span deviators. In the transverse direction the deck slab is prestressed by tendon formed by 4 strands lead at flat ducts situated at distance of 1.50 m. During erection the struts are suspended on two prestressing bars anchored at outer cantilevers of the basic cross section. The struts of a nominal width of 3.00 or 2.50 m are supported by short bottom corbels of the box girder – see Figure 8 and 10. The cast-in-place deck slab was cast in the formwork supported by already erected precast struts – see Figure 9. After the transverse prestressing of the deck slab is applied, the longitudinal external cables are post-tensioned. 2.2 Static analysis The structural solution was developed on a basis of very detailed static and dynamic analyses. The structures were analyzed by a MIDAS program system. The bridges were modeled as 3D structures formed by beam elements and a 3D structure assembled of shell and solid elements. A detailed 30
Figure 8.
Suspension of precast struts.
Figure 9.
Formwork of the overhangs.
Figure 11.
Figure 10. Connection of the struts with the box girder.
Progressive erection of the deck.
time dependent analysis of the progressively erected structure (see Figure 11) was also performed. Also a strut and tie models were used for checking of important details. The design of the first structure – the Viaduct across the Hostovsky Creek Valley – had been performed according to the Slovakia’s standards (STN) valid at the time of the design. After that all other viaducts have been designed according to Eurocode. All aspects of the analysis are presented in (Novotny et al. 2012). The detailed analysis has confirmed that although the outer struts are not mutually connected, they contribute to resistance of the structure both in the bending and torsion. Since the function of the bridge depends on perfect connection of the struts with the box girders’ corbels that depends on perfect workmanship, following conservative approach was accepted. The spine girder was designed without contribution of the struts both for bending and torsion. On the other hand the precast struts were designed for the stresses that originate on condition of the perfect connection of the struts with box girders’ corbels. This approach guarantees that the struts will not be damaged during service of the bridge. 31
Figure 12.
Long term measurement.
2.3 Bridge monitoring during construction and service The first structure, the Viaduct across the Hostovsky Creek Valley (R1 205), has been carefully monitored during construction, in depth loading tests and during service. For monitoring of concrete stresses, strain gauges were placed in four sections. Two sections were in span 6 and two sections were in span 7. In both spans one section is situated at mid-span and one section situated at distance 0.5 from the pier table. The gauges were connected with Data Taker situated inside the bridge box. After casting of the structural members in which strain gauges are placed, the measurement was performed. After that the measurement were done before and after post-tensioning of bonded tendons, striping of the formwork, erection of the struts, casting the overhangs, post-tensioning of the external cables, applying of the additional dead load, loading test and bridge opening. During service of the bridge the measurement is performed twice a year. Figure 12 presents calculated and measured values from which a good agreement of results is evident. The strain measurement was also performed during the static loading tests for three loading stages. The spans 5 and 7 were loaded by 16 Trucks of average weight of 32.05 tons situated symmetrically to the bridge axis that created maximum positive bending moments. Span 6 was loaded by 8 trucks situated only on the left bridge side that created maximum torsion. The measured and calculated values were in a good agreement. 3 COMPOSITE VIADUCTS After successful completion of the bridge across the Lochkov Valley Creek on the Prague Ring Road, Czech Republic, which deck is formed by a by a steel trough that is composite with a cast-inplace deck slab, two other long composite viaduct are being built on the highway I/11 in the North Moravia, Czech Republic – see Figure 13. However, to simplify their construction, their deck slab is cast in a formwork formed by precast members – see Figures 14 and 17. 32
Figure 13.
Construction of the Bridge across the Kremlica Creek Valley.
Figure 14.
Composite viaducts: Structural arrangement of the deck.
3.1 Recent structures The first viaduct, the Bridge across the Hrabynka Creek Valley (Bridge 206) of the total length of 330.0 m, is formed a continuous girder of 6 spans of lengths from 39.0 to 66.0 m, the second one, the Bridge across the Kremlice Creek Valley (Bridge 207) of the total length of 528.0 m, is formed by a continuous girder of 11 spans of lengths from 33.0 to 57.0 m – see Figure 15. While the Bridge 206 has a straight axis, the Bridge 207 is in a plan curvature with radius of 900 m that transfers by a transition curve into the straight axis. Both directions of the highway are carried by a bridge decks formed by a steel girder and a 25.5 m wide composite deck slab – see Figure 16. The steel girders of the trough cross section assembled of top and bottom flanges and inclined webs are supplemented by central stringer and two edge stringers. While the central stringer has I cross section, the edge stringers have V shape with smooth surface that simplify the bridge maintenance. At distance of 3.0 m the stringers are supported by diagonal pipes attached to the girder’s bottom corners. The shape of the structure is secured by top transverse ties anchored at the top flanges and at the edge and central stringers – see Figure 17. The deck slab is composite of precast slab members and additionally cast deck slab. 33
Figure 15.
Composite viaducts: elevations.
Figure 16.
Composite viaducts: cross section.
Figure 17.
Composite viaducts: Progressive assembly of the deck.
34
Figure 18.
Precast formwork.
Figure 19.
Buckling of the reinforcing bar.
Figure 20.
Lunching of the Bridge 206.
Figure 21.
Lunching of the Bridge 207.
The precast members of thickness of 100 mm are stiffened by steel trusses welded from reinforcing bars – see Figures 18 and 19. Their function both, for erection and service load, was verified by loading tests done at a Brno University of Technology (Klusacek et al. 2010). 3.2 Construction of the bridges Both bridges are incrementally assembled beyond the abutments and consequently launched into a design position. The steel structure of the Bridge 206 is assembled of 20 segments of lengths from 13.0 to 21.3 m. The steel structure was incrementally launched with precast members; only a part of the structure of the length of 66 m beyond the launching nose was formed by the steel section – see Figure 20. When launching was completed, remaining precast members were erected and the deck slab was progressively cast. The steel structure of the Bridge 207 is assembled of 25 segments of lengths from 13.0 to 29.0 m. Due to the complex bridge geometry the steel structure is assembled from two parts and it was launched from both abutments – see Figures 21 and 22. The division is situated at the fifth span at the point of zero bending moment approximately in the place, where the strait bridge axis follows 35
Figure 22.
Lunching of the Bridge 207.
the transition curve. At first, the part of the steel structure close to the abutment 1 was incrementally assembled and launched, and then the part of the structure close to abutment 12 was assembled and launched. Due to the variable plan curvature the launched structure was temporarily supported by pier transverse steel girders that allowed a transverse movement of the deck. To reduce the weight of the launched structure, the steel structure was launched without precast members. After connection of both parts the precast members were progressively erected and the deck slab was cast. 4 CONCLUSIONS The described structures were erected without any significant problems. They proved to be elastically pleasing and cost effective. ACKNOWLEDGMENTS The bridges were designed by a design firm Strasky, Husty and partners, Brno, Czech Republic. The general contractors of the bridges in Slovakia are Eurovia SK, Kosice, and Strabag, Bratislava, Slovakia. The general contractors of the bridges in the Czech Republic are Eurovia CS, Prague and SKANSKA, Prague, Czech Republic. REFERENCES Novotny, P., Juchelková, P., Jurik, M., Pawelczak, M. 2012. Bridges with progressively erected cross section. Design of Concrete Structures Using Eurocodes. Third International Workshop – Vienna. Klusacek, L., Necas, R., Dvorak, T., Strasky, J. 2010. Zkoušky prefabrikovaných filigránových panel˚u mostovky pˇred a po spˇražení – most pˇres údolí potoka Hrabynka. 17. Betonarske dny, Hradec Kralove.
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Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
Betwixt and between Portus and Cale A. Adão da Fonseca Faculty of Engineering, University of Porto, Portugal Adao da Fonseca – Engenheiros Consultores, Lda., Portugal ADEAM – Engenharia e Consultoria, Ltda., Brazil
ABSTRACT: “Portus” and “Cale” – the cradle of PORTUGAL. The river “Durus” – providing perfect environments for exceptional wines and, when approaching its destiny in the Ocean, an inspiration for bridge designers and a challenge for bridge builders. There are not many settings in the World with so many outstanding bridges. Undisputable, there is no other scenery in the World with four spectacular arch bridges. All are amazing bridges and World records, if only just for a few years. Over the ultimate stretch of the river Douro, where slow moving visits of the Ocean tide come up and go down twice a day, seven bridges were built and five more are pending into the future.
1 AN HISTORIC SITE A Celtic hamlet, a “Gall”, which means themselves, developed in the south bank of river “Dwr”, which means water. The Romans called them “Gale” and “Durus”. “Gale” or “Cale”, for, in Classic Latin, letters “g” and “c” are not well differentiated. Life with the river was easier in the south side, but the greater depth of the river in the north side provided the best local for the “Portus”. Arrival of the Moors in the VII Century forced the population to move to the north side, and “Portus” became more important than “Gale”. Together, “Portus and Gale” soon took the leadership of an active and prosperous region that, in 1147, turned into independent Portugal. Now in the XXI Century, Porto is a city of less than half a million inhabitants facing the more populated city of Gaia across the river Douro. The entire metropolitan area claims more than one and a half million inhabitants. Porto is the cultural and economic centre of Northern Portugal, but Gaia nurses the cellars of Port, the supreme wine, and profits from vintage views over the river and the steep slopes of the medieval quarters of Porto. In 1996, UNESCO has declared the area a World Heritage Site.
2 AN INSPIRATION FOR OUTSTANDING BRIDGES For many centuries, the steep granite slopes of the riverbanks have witnessed life on both sides of the river with few commercial contacts. Crossing a deep river at least 160 m wide, every winter with water sometimes speeding at close to 10 m/s, could be quite an impossible task. On the XIV Century, calm waters allowed for a temporary bridge of boats to be installed for the urgent crossing of the river by Portuguese troops on their way to the city of Guimaraes, where a siege by Castilian troops had to be repelled. Other temporary boat bridges were set up in the following centuries, but it was only in 1806 that a semi-permanent bridge boat was open for all to use. Unfortunately, three years later, hundreds of Porto inhabitants drowned in the river when the boat bridge collapsed under the weight and confusion of the population fleeing away from the Napoleon Army approaching from the north. 37
Figure 1.
Suspension footbridge (view towards west).
Figure 2.
Suspension footbridge (view towards east).
Two more boat bridges were set up before a permanent suspension bridge came into operation in 1843 across the 160 m narrowest section of the river and above its maximum flood water level (Figs. 1 and 2). The design and construction procedure devised by Stanislas Bigot for this bridge was in line with similar ones built around the world, but its sway movement under moderate winds and corrosion of its iron components meant this bridge was short lived, and it was demolished in 1888. In the meantime, the railway from Lisbon had arrived in Gaia and crossing the river Douro was a priority. An international competition was organised and the solution presented by Gustave Eiffel, together with his partner Théophile Seyrig, was chosen and built in less than two years (Figs. 3 and 4), and open to traffic in 4 November 1877. Besides this world record at the time of its construction, Gustave Eiffel designed also de double deck (lower level road and upper level railway) Viana do Castelo Bridge, 90 km north of Porto, and 38
Figure 3.
“Maria Pia” Railway Bridge under construction (view towards west).
Figure 4.
“Maria Pia” Railway Bridge (view towards east).
other smaller bridges before returning to France, where a few years later he built the very similar railway Garabit Viaduct over the River Truyère, with the arch spanning 165 meters. The last quarter of the XIX Century was a period of industrialization in the north of Portugal and Porto was then a town in great prosperity. The pedestrian suspension bridge was displaying corrosion and it was not able to fulfil the requirements of an approaching motorised society. Therefore, another international competition was organised and the double deck arch bridge (Figs. 5 and 6) designed by Théophile Seyrig was open to traffic in 31 October 1886, only nine years after the Maria Pia Bridge. This 128 years old beautiful bridge is presently one of the ex-libris of Porto and dominates de World Heritage Site. Almost 80 years passed for another bridge to be built in Porto. This bridge was open to traffic on 22nd June 1963 and carries a motorway at the height of 68 m over a reinforced concrete arch spanning 270 m across the river (Figs. 7 and 8). A world record at the time of its construction, the design by Edgar Cardoso of this beautiful bridge near the mouth of the river has received wide acclamation. The bridge designed by Gustave Eiffel was 114 years old when the railway lines were transferred, on 24th June 1991, to a new portal frame bridge (Figs. 9 and 10) designed also by Edgar Cardoso. This outstanding bridge has a central span of 250 m and runs almost parallel to the Eiffel’s bridge. 39
Figure 5. “Luiz I” Road and Tram Bridge under construction (for about one year, side by side with the Suspension Bridge).
Figure 6.
“Luiz I” Bridge – Road lower deck and presently Light Metro upper deck.
More recently, a new motorway east of Porto crosses the river Douro at a section with shallow margins, over twin multi-span bridges with two central spans of 150 m (Figs. 11 and 12), designed by Antonio Reis. These twin viaducts are an efficient design and they were well built. However, it is regretted that the Portuguese Road Authority did not pick up one of the more interesting alternatives also submitted by the same Designer. With some extra cost and vision, the river Douro could boast one more outstanding “work of art”. Most of the XX Century was a dormant period for Porto, but new life and new ambitions rose as the last Century was coming to a close. Porto and Gaia became more interdependent with a population of several hundred thousands crossing daily the river Douro, and a brand new light Metro system went into construction. A straight line in between the administrative centres of the two cities goes precisely along the upper deck of the Luiz I Bridge. Of course, this deck was chosen to be used by the Metro lines uniting those two centres, and a not too far new road bridge had to be constructed. An international “design and built” competition was set up and I have had 40
Figure 7. Arrabida Motorway Bridge under construction.
Figure 8. Arrabida Motorway Bridge.
the pleasure of creating and designing this winning bridge together with Fernández Ordóñez and Francisco Millanes. A bridge inspired in Maillart’s works-of-art, with an extremely shallow and thin arch “flying” 280 m over the river with a rise of 25 m (Figs. 13 and 14). Start of construction took place in January 2000 and opening to traffic happened on 30 March 2003. This bridge is named after the Portuguese Infant Dom Henrique, no doubt one of the most remarkable sons of Porto and Portugal, who has lead the European adventure of meeting other civilisations around the Globe. Navigation by tall ships is possible up to the Luiz I bridge and the riverbanks are becoming major leisure and tourist areas. Obviously, the lower deck of that bridge cannot be closed for road traffic and its narrow sidewalks are uncomfortable and even dangerous for pedestrians. Then, Porto and Gaia Municipalities got together just before summer 2000 and asked for a “XXI Century” pedestrian 41
Figure 9.
Figure 10.
“St. John” Railway Bridge under construction.
“St. John” Railway Bridge.
bridge to be studied for the exact location of the XIX Century suspension bridge. Obviously, this location would not affect the navigation of tall ships. The challenge was to conceive a truly state-of-art footbridge that would enhance the beauty and character of the World Heritage Site, but the very powerful presence of the “Luiz I” Bridge seemed to raise an impossible task. Since power is best dealt with by not facing it head on, lightness would have to be a main feature of the new bridge, but with no “flying” cables. In addition, a true state-of-art structural material was sought to be used. A splendid solution of a slender single stainless steel arch spanning 156 m with a shallowness ratio of 1:13 (Figs. 15, 16 and 17) that raises the arch slightly above the Luiz I bridge lower deck, in order to secure the “reading” of the latter, was conceived. Several photomontages of the bridge were displayed for public scrutiny and it has attracted and immense appraisal. Construction tender took place in 2001, but higher than expected budget and 42
Figure 11.
“Freixo” Motorway Bridge (view towards west).
Figure 12.
“Freixo” Motorway Bridge (view towards south).
Figure 13.
“Infant Dom Henrique” Road Bridge (in three construction phases).
43
Figure 14.
“Infant Dom Henrique” Road Bridge.
Figure 15.
Footbridge arch (model).
some opposition against having two bridges so close to each other, lead to the cancellation of the project. Although not unanimously, the political will of integrating the two cities of Porto and Gaia runs strong, and for it more crossings over the river Douro are required and being planned. In 2002, the need to have a light metro line crossing the river in the west sides of both cities, together with the need to alleviate the Arrabida Bridge from interurban traffic, lead to the invitation for a multifunction road/metro bridge to be designed along the planned alignment of the metro line coming from the “Casa da Musica” metro station. A powerful box-beam for the deck was designed to take the metro lines inside the box and to provide the required lanes for urban traffic on the top slab (Fig. 18). The suitable height of 7.5 meters for the box-beam meant a span of about 160 meters could be accomplished with a continuous beam of constant height deck, with that span to be complemented with other spans to surpass the over 300 meters wide river and the distances required to achieve the upper grounds of both cities (Fig. 19). Five years were to pass over the cancellation of the stainless steel footbridge project for the will to build a footbridge serving the increasing commercial and leisure activity on both riversides 44
Figure 16.
Upstream view of the Footbridge, in front of Luiz I Bridge (photomontage).
Figure 17.
Downstream view of the Footbridge, behind Luiz I Bridge (photomontage).
(“ribeiras do rio”) was taken up again by the Municipality of Gaia. This time the decision was for the footbridge to be located at the centre of the two “ribeiras”, although there the river widens to about 300 meters. A suspension bridge with two catenaries but one single mast near the Gaia bank, and curved into the flat platform on that same side (Figs. 20 and 21), was designed. This time, the pedestrian crossing got unanimous support but the project is in stand-by, waiting for a more favourable political environment and for a better public financial state of affairs. In 2013, again by initiative of the Gaia Municipality, two more road bridges were studied to meet the so much demanded low level crossings. In the east side of the cities, a 180 meters span network bowstring (Figs. 22 and 23) was designed with differentiated road, pedestrian and cycling lanes. In the west side of the cities, a 180 meters span cable-stayed bridge (Figs. 24 and 25) was designed, also with differentiated road, pedestrian and cycling lanes. 45
Figure 18.
Golgotha Road and Metro Bridge (cross-section).
Figure 19.
Golgotha Road and Metro Bridge (photomontage).
Figure 20.
Ribeiras Footbridge model.
46
Figure 21.
Ribeiras Footbridge (photomontage with view towards west).
Figure 22. Areinho Road Bridge (elevation and cross-sections).
3 EPILOGUE Betwixt and between Porto and Gaia, over a noble river dams do not completely tame, stands a unique set of Bridges whose design and construction raised immense challenges to the art and the skill of designers and builders. Out of the 12 presented bridges, the 8 listed below were (at the time of its construction), are or would be (if built today) world records in their spans (in respect to their function and structural material/structural typology). – – – – –
Maria Pia Bridge – railway iron arch bridge Luis I Bridge – double deck iron arch bridge Arrabida Bridge – reinforced concrete arch bridge St. John’s Bridge – portal railway bridge Infant Dom Henrique Bridge – bridge with slender arch stabilized by rigid deck 47
Figure 23. Areinho Road Bridge (photomontage).
Figure 24.
Massarelos Road Bridge (elevation and cross-sections).
Figure 25.
Massarelos Road Bridge.
48
– Stainless steel Footbridge – stainless steel arch bridge – Ribeiras Footbridge – single mast suspension footbridge – Areinho Bridge – network bowstring bridge In all 12 bridges, aesthetics was a corollary of structural efficiency.
The river Douro meets its destiny From east to west – 1. Areinho Bridge – 2. St. John’ Bridge – 3. Maria Pia Bridge – 4. Infant Dom Henrique Bridge – 5. Luiz I Bridge – 6. St. Antony Bridge (definitely cancelled) – 7. Massarelos Bridge – 8. Golghota Bridge – 9. Arrabida Bridge (bridges 1, 6, 7 and 8 are photomontages)
49
Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
The Octavio Frias de Oliveira and Anita Garibaldi cable-stayed bridges C.F. Ribeiro Enescil Eng. Projetos Lda., São Paulo, Brazil
ABSTRACT: The Octavio Frias de Oliveira Cable-stayed Bridge is in the Rela Parque Roadway Complex and cross the Pinheiros River at the end of Jornalist Robert Marinho Avenue, in São Paulo, Brazil. It is constituted by two curved ramps giving direct access between the Journalist Roberto Marinho Avenue and the expressway of the Nações Unidas Avenue. All such curved access lanes extend themselves and cross over the Luiz Carlos Berrini Avenue. The way encountered to overcome all of interference was to create four foundation pile caps on which the towers start to rise to the pylon for the central support of the structure. The pylon was started with 12 meter tall pillars, with a variable, rectangular section. Above this first level, two rectangular hollow cross section pillars begin to rise. After 11.4 meters of height, the extremities of these pillars are again connected longitudinally by a reinforced concrete platform, constituted by two beams and a slab. From the 23.4 meter two inclined towers rise, with a 1:3 ratio and a 57.6-meters height, the extremities of which converge towards the same point. These towers have a section with rectangular hollow cross section walls, with a width of 40 cm. At the 81 meter level, the extremities of these two towers meet, creating a cellular link with 9 meters of height (the knot of the "X"). The number of strands inside each stay-cable varies between 10 to 25. Due to the variable geometry of the complex, characterized by inclinated towers and overlapping decks, with a constant radius of curvature, an unique and innovative spatial arrangement was obtained. The cable-stayed decks present a curvature in plan with 16- meter width and a constant radius, with 140-meter spans to cross the Nações Unidas Avenue and the CPTM trains, and 150-meter spans across the Pinheiros River. The cable-stayed bridge over the Laranjeiras Channel, officially named "Anita Garibaldi Bridge", is localized in the city of Laguna, State of Santa Catarina (SC), and is included in the Duplication of the BR-101/South Highway project. The work has an extension of 2.830 meters, being constituted by 49 approach spans with 50 meters in length and three cable-stayed spans each side span with 100-meter and a central, 200meter span. The approach spans of the bridge are built with typical, 50-meter spans, with 3.65-meter long segments, each of these segments with 80-tons. The precast concrete brackets supported by the box girder lower slabs allow the typical cross-section of the bridge to have a total width of 24.10 meters. The foundations of the piers of those spans are built with concrete steel shearing piles, with a 2.50-meter diameter in the soil shaft and 2.30-meter diameter in the rock shaft. The 49 approach spans are all simply supported spans, prestressed and made up by bonded segments. This system is only possible due to the technological resources of the BERD Company, where a launching gantry of the LG50 model with OPS (Organic Prestressing System) is used. It is a cable-stayed section with a total length of 400 meters made up by two side spans of 100 meters and a 200-meter central span. The side spans have 24 segments and the central span has a total of 48 segments, all of which with a 3.30-meter length and a weight of 91 tons each. The key segment has a 2.0-meter length and is casted in place. 51
There is a single pylon at each pier, with a hollow, rectangular and variable cross-section in the first 16 meters of height. The pylons are built totally vertical without suffering any change in direction despite the fact that the deck is curved in plan.
1 INTRODUCTION The Octavio Frias de Oliveira Cable-stayed Bridge is integrated to the process of the restructuring of the road system of the Municipality of Sao Paulo. It is in the Rela Parque Roadway Complex and cross the Pinheiros River at the end of Jornalist Robert Marinho Avenue. It is constituted by two curved ramps giving direct access between the Journalist Roberto Marinho Avenue and the expressway of the Nações Unidas Avenue The cable-stayed decks present a curvature in plan with a constant radius, with 140-meter spans to cross the Nacoes Unidas Avenue and the CPTM trains, and 150-meter spans across the Pinheiros River. The geometry of the "X"-shaped tower makes this Bridge as a new post-card for the city of Sao Paulo. The cable-stayed bridge over the Laranjeiras Channel, officially named "Anita Garibaldi Bridge", is localized in the city of Laguna, State of Santa Catarina (SC), and is included in the Duplication of the BR-101/South Highway project. This bridge will allow to increase and modernize the highway, improving the connection between the productive hubs and the region’s seaside harbors. The investment, according to initial estimates, reaches R$ 600 millions and the bridge is built by three construction companies: Construbase, M. Martins and Camargo Correa. The work has an extension of 2.830 meters, being constituted by 49 approach spans with 50 meters in length and three cable-stayed spans each side span with 100-meter and a central, 200meter span. Of the 52 spans which constitute the bridge, only 11 approach spans are built over the ground, while the rest are built over the water.
2 THE REAL PARQUE COMPLEX The Octavio Frias de Oliveira Cable-stayed Bridge, which is part of the Real Parque (Royal Park) Roadway Complex, is the 13th bridge crossing the Pinheiros River and it is integrated to the process of the restructuring of the road system of the Municipality of Sao Paulo, currently saturated by more than seven million vehicles. With the objective of creating new access alternatives between the Pinheiros Riverbank Avenue and the Journalist Roberto Marinho Avenue, this engineering work aims at reducing traffic in the crossing of the latter with the Luiz Carlos Berrini Avenue, as well as the traffic on the Morumbi Bridge. Added to this localized improvement, this complex also aims at alleviating pressure on the Bandeirantes avenue, which will occur after the extension of the Journalist Roberto Marinho Avenue, all the way to the Imigrantes Highway. Within this ambitious policy of improving local traffic, the roadway lay-out of this enterprise could not be simple. It is constituted by two curved ramps giving direct access between the Journalist Roberto Marinho Avenue and the expressway of the Nações Unidas Avenue. All such curved access lanes extend themselves and cross over the Luiz Carlos Berrini Avenue. To make this roadway lay-out the most efficient as possible, there was the need to overlay the two curved, cable-stayed decks which cross the Pinheiros River, thus imposing the innovative geometry of the "X"-shaped tower. This new post-card for the city of Sao Paulo adds up to 2.910 linear meters of elevated roads, of which 580 meters of their overall extension are cable-stayed.
Figure 1.
Roadway study.
Figure 2.
Schematic plan of the foundation.
2.1 Foundations To match the planned roadway lay-out, it was necessary to position the pylon of the bridge on the right riverbank of the Pinheiros River, where the Journalist Roberto Marinho Avenue begins. However, in this location there is a series of forms of interference which represent a severe hindrance. The underground concrete service gallery, which houses the CTEEP three transmission lines with 345 kVA, runs parallel to the right riverbank, as well as the railway line of the CPTM (Paulista Company of Metropolitan Trains) -. In addition to these two kinds of interference, there is also the adduction channel of the pumping station of the Aguas Estaiadas stream, buried exactly under the axis of the pylon. The way encountered to overcome these forms of interference was to create four foundation pile caps on which the towers start to rise to the pylon for the central support of the structure. Each one of these blocks is supported by 28 excavated piles, with a 130 cm. diameter (reduced to 120 cm. when embedded in rock) and 10 root-piles with a 41 cm. diameter (reduced to 31 cm. when embedded in rock). 53
Figure 3.
Main Pier.
For this construction work 2.880 cubic meters of concrete and 184 tons of CA-50 steel were used. In its cable-stayed portion, the foundations for side pier, facing the Journalist Roberto Marinho Avenue, were built with 11 excavated piles, with 110 cm. of diameter (reduced to 100 cm. when embedded in rock)..With regards to the side pier located on the left riverbank, due to the presence of transmission lines which made it impossible to use large-size equipment, 50 large root-piles were built, with a diameter of 41 cm. (reduced to 31 cm. when embedded in rock). For the construction of the foundations of these piers, in each extremity, 490 cubic meters of concrete were consumed. 2.2 Main Piert The pylon was started with 12 meter tall pillars, with a variable, rectangular section. In this position, the pillars are cross-linked transversally by a ribbed slab and longitudinally by a concrete prestressed platform constituted by two beams and one slab. From this platform the deck of the journalist Roberto Marinho Avenue, seeking the Pinheiros Riverbank Avenue, begins. Above this first level, two rectangular hollow cross section pillars (walls with a 40 cm and a 60 cm width) begin to rise. After 11.4 meters of height, the extremities of these pillars are again connected longitudinally by a reinfored concrete platform, constituted by two beams and a slab. From this platform the deck of the Pinheiros Riverbank Avenue, in the direction of the Journalist Roberto Marinho Avenue, is born. From the 23.4 meter (where the reference is the side of the block) two inclined towers rise, with a 1:3 ratio and a 57.6-meters height, the extremities of which converge towards the same point. These towers have a section with rectangular hollow cross section walls, with a width of 40 cm. At the 81 meter level, the extremities of these two towers meet, creating a cellular link with 9 meters of height (the knot of the "X"). From this point upwards, two other towers with an inclination ratio of 1:4.5 (this proportion was necessary to minimize the torsion effects in the pylon due to permanent loads) in a way that its extremities draw away, giving the "X"-shape to the pylon. In these 42 meter-long parts, the 54
Figure 4.
Details of the central support.
anchoring devices of the stays are located. the walls of their rectangular hollow cross section are pre-stressed so as to assure that no traction tension may exist due to the horizontal component of the stays’ force. Lastly, a beam with a cellular section was created at the 107 meter level with the objective of blocking these towers, in order to reduce the flexion and torsion moments. Above this part of the pylon, already at the 132-meter level, an architectural detail with 6-meter of height, is used to convey to the pylon a better aesthetics and lightness. In short, to execute this 138-meter high pylon, 5.600 cubic meters of concrete and 1.220 tons of CA-50 steel, in addition to 63 tons of CP190-RB steel. 2.3 Side Piers The side piers are constituted by massive pillars with a rectangular section and rounded-off corners, which support a cross-beam with the purpose of making a monolithic link between the pillar and the stayed deck, besides serving as a support for the pre-fabricated concrete girders of the approach spans ramp. 2.4 Stay-cables The stay-cable corresponds to the system used in suspension bridges to transfer directly the active loads between the deck and the pylon. Each stay-cable is constituted by a set of parallel strands. These strands, in turn, are formed by seven galvanized steel wires, helically distributed, protected by an individual HDPE (high-density polyethylene) sheath filled with wax. In addition to the individual protection of the strand, there is also a HDPE tube, yellow in color, which wraps all the strands which make up the same stay-cables. The number of strands inside 55
Figure 5.
Details of the side piers.
each stay-cable varies between 10 (stay-cables nearer to the pylon) to 25 (stay-cables nearer to the side piers). Due to the variable geometry of the complex, characterized by inclinated towers and overlapping decks, with a constant radius of curvature, an unique and innovative spatial arrangement was obtained. In the cable-stayed spans over the Pinheiros River, once the decks are positioned on the opposite side to the towers that support them, there is an interlacement of stay-cables that creates a special configuration in their geometric disposition which has not yet been verified in any other cable-stayed bridge in the world. All the stay-cabling of this work consumed approximately 375 thousand meters of strands, equal to 462 tons of steel.
2.5 Deck The cable-stayed decks present a curvature in plan with a constant radius (equal to 275.1 meters measured axially), with 140-meter spans to cross the Nacoes Unidas Avenue and the CPTM trains, and 150-meter spans across the Pinheiros River. The 16- meter width of the deck is constituted by two 1.5-meter longitudinal girders (where the anchors of the stay-cables may be found), two 85 cm.-wide sidewalks, two 40 cm-wide concrete barrier and a 10.5-meter carriage way (on a 48 cm.- thick slab). Built with the cantilever technique, the longitudinal structural members were prestressed with penpons positioned in its upper side to avoid excessive traction tensions during construction phases. Another prestressing operation during the deck’s construction was the one constituted by cables located near the inner edge of the deck, with the objective of overcoming the flexion moments in plan (horizontal plane), originated from the horizontal vectorial component of the force of the stay-cables, which does not align with the longitudinal axis of the deck. A transversal prestressing of the slab was also carried out during the deck’s construction, so that the slab may support loading without cracking. The only prestressing carried out after the conclusion of the deck’s construction refers to the continuity tendons, mostly distributed near the lower surface of the girders to assure that the section 56
Figure 6. View at top of the pylon.
Figure 7.
Bridge deck.
57
Figure 8.
Cross section of the deck.
should not be submitted to tensions greater than those stipulated by the Brazilian regulations, during the service phases, that is, under the action of mobile loadings. There is no bearing support equipment on the site, and the connections between deck and pillars were made monolithically fixed). For the construction of the 590-meter deck the following were used: 6.400 tons of concrete and 885 tons of CA-50 steel, as well as 420 tons of CP190-RB steel. 2.6 Engineering and Architecture It is important to emphasize that this bridge complex was completely conceived by engineers, especially and initially by the professionals of ’Enescil Engenharia de Projetos Ltda’ and, later, by the teams of ‘ANTW Engenharia de Projetos Ltda’, ‘Antranig Muradian Ltda’, and ‘Geometrica Engenharia de Projetos Ltda’, who approved the calculations and confirmed the initial design of the work as feasible, functional, safe and economic for its site. After the definition of the engineering solution for the work, the architectural office ‘Valente e Valente Arquitetos’ was responsible for complementing the design, softening contours, defining friezes, choosing the color of the stay-cables and the form of the guard-rail, amongst other activities. Shortly, the architecture and conception of the work, especially the "X"-shaped-pylon, was the exclusive work of the engineers. The architect’s team was responsible for the complementary, although important, task of beautifying and giving finishings to the work. 3 ANITA GARIBALDI BRIDGE The cable-stayed bridge over the Laranjeiras Channel, officially named "Anita Garibaldi Bridge", is localized in the city of Laguna, State of Santa Catarina (SC), and is included in the Duplication of the BR-101/South Highway project. With the current daily capacity limited to just 20 thousand vehicles, this work will allow to increase and modernize the highway, improving the connection between the productive hubs and the region’s seaside harbors, for the well-bring of the regional and national economies, as well as those of other Mercosul countries. The investment, according to initial estimates, reaches R$ 600 million and the bridge is built by three construction companies: Construbase, M. Martins and Camargo Correa. The work has an extension of 2.830 meters, being constituted by 49 approach spans with 50 meters in length and three cable-stayed spans each side span with 100-meter and a central, 200-meter span. Of the 52 spans which constitute the bridge, only 11 approach spans are built over the ground, while the rest are built over the water. 3.1 Approach spans of the Bridge The approach spans of the bridge are built with typical, 50-meter spans, with 3.65-meter long segments, each of these segments with 80-tons. The precast concrete brackets supported by the 58
Figure 9.
Figure 10.
Bridge location.
Longitudinal layout and plan of the Bridge.
box girder lower slabs allow the typical cross-section of the bridge to have a total width of 24.10 meters. The foundations of the piers of this spans are built with concrete steel shearing piles, with a 2.50-meter diameter in the soil shaft and 2.30-meter diameter in the rock shaft. The pillars in the approach spans form a continuity with the foundations, with their starting points located above the steel sheath and ending points in the corbels and cross-beams, where the plinths for the super-structure are located. The 49 approach spans are all simply supported spans, prestressed and made up by bonded segments. This system is only possible due to the technological resources of the BERD Company, where an launching gantrie of the LG50 model with OPS (Organic Prestressing System) is used. The construction of the spans using this truss had initially an estimated time-schedule of one week for the building of each span; however, during the course of construction, shorter periods of four to five days per span were achieved. 3.2 Construction sequence of the LG50 BERD truss 3.3 Images of the construction sequence of the Bridge After the construction of the spans by means of the launching gantry (launching, bonding and prestressing of the segments), the launching of the precast concrete brackets by means of cranes was started on both sides of the deck, fixed to the box girder through high-resistance rods. 59
Figure 11.
Cross section of the simply supported spans.
Figure 12.
Support of box section at the simply supported spans.
Figure 13.
3D Model of the LG50 Launching Gantry by BERD.
60
Figure 14.
Launching Gantry model LG50 - BERD.
Figure 15. LG50 Model launching gantry. Preparation for lifting the segments.
Figure 16. gantry.
Lifting and suspension of the first 7 segments. Placing of the counterweight for the launching
The lateral slabs are constituted by precast plates supported by the precast concrete bracket and a thin concrete surface with a medium thickness of 14 centimeters, adding up to a 25-centimeter final slab. Since these slabs have an 7.55 meter width, a transversal-prestressing is used to guarantee the safety of the slabs and a firm linkage with the box girder. 61
Figure 17.
Suspension, placing and bonding of the 7 remaining segments.
Figure 18. Placing and bonding of the first suspended segments. Launching of the gantry for the next span.
Figure 19.
Images of the construction sequence of the Bridge.
3.4 Cable-stayed spans It is a cable-stayed section with a total length of 400 meters made up by two side spans of 100 meters and a 200-meter central span. The side spans have 24 segments and the central span has a total of 48 segments, all of which with a 3.30-meter length and a weight of 91 tons each. The key segment has a 2.0-meter length and is casted in place, together with the key segments of the side span. The cross-section of the cable-stayed spans is slightly different than the approach spans crosssection, the box girder being a little larger, the box girder slabs increasing their thickness, while diagonal ties are placed in the cable-stayed segments in addition to a groove in the middle of the 62
Figure 20.
Precast overhang bracket.
Figure 21.
Pre-stress transverse tendons.
Figure 22.
Longitudinal elevation of the cable stayed spans.
upper slab, between rigid barriers. These ties are prestressed with high resistance rods to help to counter the concentraced force of the cable and to redistribute the forces to the webs. In the side piers we find columns with the same diameter as the piles and cross-beams at their tops. These cross-beams are quite different of the other cross-beams and corbels, since they establish a continuity with the key segment of the cable-stayed side spans. The foundations of the pylons are 20 piles excavated in soil and rock, with a diameter of 2.50 meters in the soil shaft and 2.30 meter in the rock shaft. Above the piles stepped pile caps were built, composed by a lower layer, 3.0 meter high, and an upper layer, with a 3.5-meter height. Above the foundations of the pylons we find two walls, 1.3 meter thick, for greater flexibility with regards to the effects of creep and shrinkage, where we have the pier segment, with a total length of 39.6 meters and a 26.22-meter width. There is a single pylon at each pier, with a hollow, rectangular and variable cross-section in the first 16 meters of height. The pylons are built totally vertical without suffering any change in direction despite the fact that the deck is curved in plan. 63
Figure 23.
Cross section of the curved cable stayed spans.
Figure 24.
Piles caps of the pylon.
The definition of the vertical pylon forced the use of transversal cable-stays, 4 cable-stays for each pylon, to confront the transversal forces in the pylon. These cable-stays begin at the extremities of the prestressed cross beams of the pier segment. Two temporary cable-stays for each pylon to help with their construction, reducing the wind forces and confronting the horizontal share of the cable-stays due to the curvature of the bridge. The pylon receives a horizontal prestressing in the longitudinal direction of the construction work, near the blisters of the cable-stays to counteract the concentrated forces of the cables. The cable-stays are in a semi-fan arrangement with a single plane, varying the quantity of strands between 50 (first prestressed segment near the pylon) and 72 (last segment near the key segment), with 281 thousand meters of strands in the cable-stayed section, equivalent to 366 tons of steel. 64
Figure 25.
Pier segment.
Figure 26. Transverse stay cables.
65
Figure 27.
Prestress tendons at the pylon cross section.
Figure 28.
Numbers of strands on each stay.
The construction of the deck of the bridge using a traveler, cable-stayed with prefabricated and bonded segments, was carried out by means of high-resistance rods to guarantee the bonding of the segments. The greatest concern in the sizing of the cable-stayed spans with bonded segments is with the limitation of the traction tensions between segments, allowing only compression tensions on both upper and lower fibers of the deck.
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Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
Multi-span extradosed bridges A. Kasuga Sumitomo Mitsui Construction, Tokyo, Japan
ABSTRACT: Some examples of multi-span extradosed bridges in Japan are introduced together with their structural features. Furthermore, a new solution to extend the span length of extradosed bridges to around 500 m is proposed, using a new technology called the butterfly web. The structural behavior of a multi-span long extradosed bridge, including the results of wind tunnel tests, is discussed.
1 INTRODUCTION A large number of extradosed bridges have been constructed around the world since the theory was presented by Mathivat in 1988 (Mathivat 1988), and the Odawara Blueway Bridge was constructed in 1994. Japan is the country with the largest number of bridges constructed (Kasuga 2006). The most distinguishing structural feature of extradosed bridges is that the designer can freely choose the stiffness balance between stay cables and main girder. This principle has been adopted in the Japanese standards (JPCI 2000), providing a continuum spanning the design of girder and cablestayed bridges. Depending on the stiffness balance between stay cables and main girder, the design can be performed with a safety factor of 1.67 for the stay cables (limiting cable tension to 0.6fpu ,) in spite of being a cable supported structure. This is the value used for internal or external cables. The factor of safety can also be varied linearly, allowing a limit value for cable tension ranging between 0.4fpu , the value specified for cable-stayed bridges, and 0.6fpu . The span length of extradosed bridges built to date has been in the range from 100 to 250 m. Even for a composite bridge using steel girders, 275 m is the maximum span length that has been achieved so far. Innovation in cable-stayed bridges and extradosed bridges consists of finding ways to make the main girders lighter. Recently, there has been remarkable progress in composite structures that permit lighter main girders, including the successful construction of many corrugated steel web bridges and composite truss bridges. Further progress is now being made in Japan with the construction of bridges using a new form of composite structure, the butterfly web (Ashizuka et al. 2012). Butterfly web structures use panels of steel or fiber reinforced concrete as an alternative to a double warren truss, and make it possible to achieve a similar level of lightness to corrugated steel web bridges (Fig. 1). Furthermore, it is a highly durable member since there are no reinforcing bars placed in the 15 cm thick butterfly web as the tensile side is reinforced with pretensioning steel. There are now a number of extradosed bridges constructed with main girders that use these sorts of composite structures. Furthermore, a method to further extend the applicable span length of extradosed bridges up to the range of 500 m is presented. To achieve this, the butterfly web is used. With the use of butterfly webs, the total weight practically stays the same even if the depth of the main girder is increased. This means that the main girder stiffness can be considerably increased. Hence, the main tower can be shorter than cable-stayed bridge towers at the 500 m span length range (Kasuga et al. 2015). Here, an actual long-span cable-stayed bridge is compared to the case in which it is designed as an extradosed bridge. One issue that has to be taken into account with such a bridge is its wind-resistant stability. Consequently, the effect of the butterfly web, which has openings in 67
Figure 1. Butterfly web bridge (Takubogawa Bridge).
Figure 2.
Ikuchi Bridge
Figure 3.
Nhat Tan Bridge.
the webs of a deep girder, was verified through wind tunnel testing. The results of the study of the long-span extradosed bridge are shown below. 2 LONG SPAN CABLE SUPPORTED BRIDGES 2.1 Composite Girder Construction Techniques Composite bridges are structures that combine steel and concrete, selecting the material to suit the characteristics required for each part of the bridge. In Japan, where there are frequent earthquakes, reducing girder weight brings a range of benefits, and this has been a driving force behind the remarkable pace of development over the past twenty years. Reducing the weight of the superstructure is the key to innovation in long span bridges supported by cables. The reason is that this approach enables stiffness to be increased without increasing girder weight. The hybrid girder is a well-used approach for composite bridges. In Japan, long span cable supported structures with hybrid girders are used, for example, in the Ikuchi Bridge (Fig. 2) and in the Nhat Tan Bridge (Fig. 3). The corrugated steel web bridge dates back to 1965 when Shimada proposed that corrugated steel plate may be used in the web of main girders (Shimada 1965). However, it was a long time before this marvelous concept was realized in an actual bridge. That was in 1984, when a composite bridge using this approach was completed in France, far away from the birthplace of the idea. By replacing the concrete web with corrugated steel plate, a corrugated steel web bridge enables the weight of the main girder to be reduced by around 10 to 15%. An example of corrugated steel 68
Figure 4.
Himi Bridge.
Figure 5.
SBS Link Way Bridge.
Figure 6.
Fudo Ohashi Bridge
Figure 7.
New concept of composite truss bridge.
webs used in cable supported structures includes the Himi Bridge (Fig. 4) (Maeda et al. 2002). The Himi Bridge, an extradosed bridge, is particularly interesting as it was the world’s first use of a corrugated steel web in a cable supported structure. One form of composite truss bridge that permits substantial weight reduction is the space truss structure. A cable-stayed structure using the space truss can achieve weights that are around 40% less than conventional concrete girders (Muller 1990). The only actual bridge using this structure is a very small pedestrian bridge in Singapore (Fig. 5), but in Japan, there is one example of the extradosed bridge utilizing composite truss techniques but replacing just the web with a steel pipe truss, as can be seen in Figure 6. In addition, as a means of reducing axial forces acting on the truss nodes, it is possible to use a composite structure combining a concrete box girder with a space truss as shown in Figure 7. Table 1 shows that selecting an appropriate mix of concrete and steel trusses in combination with a cable supported structure enables the maximum axial force acting on the trusses to be controlled to within 1500–2000 kN. The concrete-steel composition is a parameter to be considered when designing an optimum structure. 2.2 Tay Cable Design Japan was also the location for the world’s first extradosed bridge, the Odawara Blueway Bridge constructed in 1994, which marked the debut of a structural form that could cover span lengths for which cable-stayed bridges are uneconomic. This represented a major innovation in thinking about the design of stays. Cable-stayed bridges and external-cable structures could be conceived as points on the same continuum when determining the maximum allowable stress for stays. Taking this approach enables the use of stays with allowable stress in a continuous range between 0.4fpu and 0.6fpu . Moreover, the allowable stress can be determined according to the range of live load stresses 69
Table 1. Maximum axial force in various composite truss structures.
in the cables. Instead of the conventional approach of stipulating a single value of allowable stress for the whole structure, the structure could be designed with different values for each stays. This design method was formalized in Japan’s specifications for highway bridges, enabling engineers to produce more rational designs for stays (Fig. 8). 2.3 Ibi River Bridge – An example of multi-span extradosed bridge The Ibi River Bridge is a five-span continuous extradosed bridge which was completed in 2001 (Fig. 9). The longest span length is 271.5 m long and was built with precast segments of up to 400 tons (Fig. 10). It is a hybrid structure made with 100 m of steel girders at the span center. The two-point bearing supports on top of the piers serve to increase the overall stiffness of the bridge, while the shape of the main towers provides additional stiffness in the longitudinal direction of the bridge. Since it is an extradosed bridge, the stiffness of the main girder is high. However, the maximum stay cable stress variation due to live load is 112 N/mm2 because it is multi-span. A limit value of 0.6fpu is used for the stays since prefabricated cables (DINA) were used. One characteristic feature of an extradosed bridge in comparison with a cable-stayed bridge is its ultimate limit state behavior (Kutusina et al. 2002). Since the increase in stay cable tension due 70
Figure 8. Allowable stress versus stress change owing to live loads.
Figure 9.
Ibi River Bridge.
Figure 10.
400 tons precast segment.
to the ultimate load starts from 0.6fpu , the stay cables yield before the ultimate limit state is reached (Fig. 11). In contrast, the stay cables of a cable-stayed bridge starts from 0.4fpu and do not yield. In an extradosed bridge, the construction error distribution at the time of introduction of the stay cable forces produces a small error distribution in the bridge because it is the stay that cables yield, which may lead to a smaller factor of safety at ultimate load like external tendons (Woelfel 1990). Extradosed bridges inherently have higher main girder stiffness than cable-stayed bridges, which makes it possible to raise their overall stiffness even with multiple spans. This suggests potential for exploring the possibility of ranges beyond the commonly stated applicable span length of 250 m for extradosed bridges.
3 BUTTERFLY WEB BRIDGE 3.1 Outline of Butterfly Web The butterfly web (Kasuga et al 2010) is a new structure with butterfly-shaped web members having the following characteristics. (1) The web is configured with butterfly-shaped panels placed independently and not joined continuously. The shape limits the orientation of compression and tension in the panel due to shear forces, meaning that the structure is similar to a double warren truss (Fig. 12). 71
Figure 11.
Behavior of extradosed cables up to ULS.
(2) The butterfly web uses 80 MPa steel fibre reinforced concrete, and has prestressing steel oriented in the direction that tensile forces act (Fig. 13), limiting the occurrence of cracks. It does not use steel reinforcements, relying instead on steel fibers and prestressing to achieve the required strength. (3) Transmission of shear forces between the butterfly web and deck slabs is achieved by the joint between the slab concrete and dowels embedded in the panel. Many corrugated steel web bridges and steel truss web bridges have been built. These bridges had rational structures and excellent structural characteristics, but at the same time, they required complex machining of steel members, on-site welding, or other special skills for fabrication or construction. In contrast, as the butterfly web is a precast product, all that is needed to construct a girder is to combine the web with the slabs on site. The prestressing steel oriented in the same direction as the tensile forces in the web is pre-tensioned at the factory, so there is no need to work on the butterfly web at the construction site. The potential weight reduction of the main girder is similar to that of a corrugated steel web bridge, achieving about a 10% to 15% reduction compared to a conventional box girder section. Consequently, the length of segments that can be constructed using a form traveler can be 50% longer because of light weight of the girder. A butterfly web bridge, which uses butterfly-shaped panels instead of a double warren truss, is a new structure that has both the corrugated steel web bridge’s advantage of being able to simplify the joints with the concrete slabs, and the truss bridge’s advantage of not needing on-site work to make joints between the butterfly-shaped panels that carry the shear forces. 3.2 Innovative Bridge Project One of these solutions for long span extradosed bridges is being used in an innovative project currently under construction. The Mukogawa Bridge, shown in Figure 14 and Figure 15 is an extradosed bridge using butterfly web technology. This is a 5-span continuous rigid frame bridge with a span length of 100 m. The tallest piers are 81.2 m, and they were designed for rapid construction. The cross-section incorporates four butterfly webs, and the extradosed cables are located in the center of the cross-section. The main girder is constructed by free cantilevering, with individual segments having a length of 6.0m and incorporating two butterfly web panels parallel to the longitudinal direction. After setting panels, the concrete deck is cast in place. The reduction in superstructure weight achieved enables a substantial reduction in pier thickness and the size of foundations. The cables for the Mukogawa Bridge use ultra high-strength strands (Kido & Hoshino 2010) that are 30% stronger than regular strands. 72
Figure 12.
Behavior of butterfly web bridge.
Figure 14.
Mukogawa Bridge.
Figure 15.
General view of Mukogawa Bridge.
Figure 13.
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Butterfly web panel.
Table 2. Comparison of material quantities.
Concrete Rebar Prestressing steel Stay cables
Girder Pylon
m3
m3 ton ton ton
CSB
EDB
48900 18500 15960 259 6030
43700 14400 13410 261 6300
4 STRUCTURAL BEHAVIOR OF MULTIPLE SPAN EXTRA DOSED BRIDGE The models used for this comparative study are a cable-stayed bridge and an extradosed bridge composed of five continuous spans with a central span of 500 m (Fig. 16). In order to raise the overall stiffness of the multi-span cable supported structure, overlapping cables are distributed at the middle of the span, and the stiffness of the bridge piers and main towers is increased. Moreover, highly stiff butterfly webs with a 6.0 m depth are used for the extradosed bridge. Using these two structures, the main girder stresses at the serviceability limit state and stay cable stress variations, as well as other variables such as the behavior during earthquakes, were compared, and the structural feasibility of the proposed continuous long span extradosed bridge is verified. 4.1 Service limit state With regard to the limit stresses, concrete stress at dead load is 0.4f ’ck = 18 N/mm2 on the compression side and the full prestress on the tension side. At design load, concrete stress is 2/3 0.6f ’ck = 27 N/mm2 on the compression side and −0.5f ’ck for 50% live load on the tension side; for 100% live load, cracking is allowed, with the amount of reinforcement set such that the reinforcement tensile stress is 0.6fsy or less. The bending moment and stress diagrams of the main girder at 50% live load are shown in Figs. 17 to 20. From these results, it was confirmed that the structure is sufficiently sound even with the extradosed bridge configuration. Figure 21 shows a comparison of the stress variations in stay cables due to live loads. In the JPCI code of Japan, up to 0.6fpu , the limit value for stay cable tension, is allowed for road bridges as long as the stay cable stress variation is 70 N/mm2 or less. However, it was confirmed that almost all of the cable stress variation exceed 70 N/mm2 except some cables of the side span because multi-span structures are very flexible. Table 2 shows a comparison of the material quantities for the two configurations. The amount of concrete for the extradosed bridge using butterfly webs is lighter, despite having a deeper girder. Moreover, the material required for the pylons was reduced. However, there is no big difference of the weight of stay cables in the multi-span structures. 4.2 Seismic resistance In verifying dynamic behavior under seismic loading, it is interesting to examine the differences between cable-stayed bridges with their tall main towers and flexible girders, and extradosed bridges with short main towers and stiff main girders. Seismic response analysis was performed using the wave of Japanese seismic specifications (Fig. 22) for soft soil sites. The natural periods for each of these bridge types are shown in Table 3, and the maximum bending moments of girders and pylons are shown in Figs. 23 and 24. The response value of the extradosed bridge girder is larger than that of the cable-stayed bridge, but the bending moment of pylons has no large difference. For long span bridges, adoption of an extradosed bridge with a light main girder instead of a cable-stayed bridge was found to be advantageous in regions prone to earthquakes. 74
Figure 16.
500 m five-span bridge.
Table 3. Natural period.
CSB EBD
1st
2nd
3rd
7.92 8.00
5.14 5.38
4.17 3.99
75
Figure 17.
Girder bending moment (CSB).
Figure 18.
Girder bending moment (EDB).
Figure 19.
Girder bending stress (CSB).
Figure 20.
Girder bending stress (EDB).
Figure 21.
Stay cable stress variations due to live load.
Figure 22. Wave of Japanese seismic specifications.
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Figure 23.
Girder bending moment.
Figure 24.
Figure 25.
Girder configurations of butterfly web.
Figure 26.
Girder configurations of conventional web.
Pylon bending moment.
5 WIND TUNNEL TEST When bridge structures are stretched to very long spans, one concern is the more noticeable aerodynamic vibration as the entire structural system becomes more flexible. Therefore, for this chapter, we conducted wind tunnel tests to examine the wind-resistant stability of the main girder in the 500 m span extradosed bridge described in the previous chapter. Note however that a wider bridge deck was used as the main girder section, as shown in Figure 25. Furthermore, in order to study the effects on wind-resistant stability of the web openings in the butterfly web structure, a comparative study was conducted through wind tunnel testing on another model with the main girder’s butterfly web structure openings covered (Fig. 26). 5.1 Test outline Modal analysis was performed for the bridge to set the test conditions. The results are shown in Table 4, and the wind tunnel test specifications are provided in Table 5. To tailor to the test facility, a scale of 1/100 was used. A logarithmic decrement of about 0.03 is usually used for cable-stayed and 77
Figure 27. Test outline and setup. Table 4. Results of eigenvalue analysis. Vibration mode
Natural frequency
Girder, heaving 1st mode Girder, torsion 1st mode
0.139 Hz 0.682 Hz
other bridges to take the damping ratio under aerodynamic vibration into account, but the damping ratio in this test was minimized as much as possible, so that the aerodynamic properties induced by the shape of the butterfly web structure itself can be compared to those of the conventional box girder section. Wind-resistant stability for the actual bridge was determined from excitation forces (damping factor) for the various vibrations exhibited. Wind tunnel testing was performed on two-dimensional rigid body spring models. The tests are outlined in Figure 27. 5.2 Test Results (1) Heaving vibration For the closed section box girder studied for comparison, the single-degree-of-freedom heaving vibration test results are given in Figure 28 (smooth flow, angle of attack + 3 deg). The figure shows the test results when the model was excited with a 5 m/sec full scale wind velocity and then left to freely vibrate. It also shows the results when the model was left to freely vibrate without adding excitation (initial amplitude of vibration is 0), which confirms the presence of limited amplitude vibration that is apparently vortex-induced vibration, with a double amplitude of about 600 mm for the actual bridge. However, for subsequent wind velocities, there were no vibrations for the tests without added excitation. For the tests with added excitation, unsteady amplitude increases as the wind velocity increases. Although peak amplitudes were large for full-scale wind velocities over 65 m/sec in particular, these were not considered as divergent vibrations such as galloping as these large vibrations did not occur in a stable form. 78
Table 5. Wind tunnel test specifications. Item
Actual bridge
Test mode
Girder width (m) Girder height (m) Mass per unit length (kg/m) Scale factor Wind angle (degree) Heaving Frequency (Hz) Wind speed magnification Logarithmic decrement Scruton number Torsion Frequency (Hz) Wind speed magnification Scruton number
30.0 6.0 6.353×104
0.30 0.06 6.353 1/100 −3, 0, +3 2.237
Figure 28. Flexural vibration test results of conventional web.
0.139 6.257 (average) 0.030 91.94 0.682 184.02
Figure 29. fly web.
0.00402 (average) 12.28 (average) 6.426 (average) 10.582 (average) 39.47 (average)
Flexural vibration test results of butter-
Limited amplitude vibrations similar to those for the closed section model with a 6 m/sec full scale bridge wind velocity were not found when the same tests were performed on the butterfly web section model. The subsequent major vibrations were also not found (Fig. 29). These results suggest that, with the use of the butterfly web structure, heaving vortex-induced vibration was suppressed by the more complex air flow induced by the provision of openings for approximately 30% of the web. (2) Torsional vibration The single-degree-of-freedom torsional vibration test results for the closed section are given in Figure 30. From this figure, it can be seen that limited torsional vibration with double amplitude of about 2 degrees was exhibited at a 25 m/sec full scale wind velocity. Moreover, torsional vibrations developed from 60 m/sec full scale wind velocity, with the rotational displacement increasing as the wind velocity increases. Since the aspect ratio B/D, where B and D are the characteristic width and the height respectively, of the section is in the relatively large range of about 5.0, these divergent vibrations are most likely to be torsional flutter. Torsional vibration test results for the butterfly web structure are given in Figure 31. Similar to the closed section box girder, limited amplitude vibration was exhibited at a 25 m/sec full scale wind velocity, although the peak value is small with double amplitude of a little over 1 degree. Moreover, divergent vibrations at high wind speed range were exhibited at 90 m/sec wind velocity, shifting to the range that is not an issue in practical terms. In other words, the presence of openings due to the use of the butterfly web structure can be expected to enhance aerodynamic stability with regard to torsional vibration, as well as with regard to heaving vibration. 79
Figure 30. Torsional vibration test results of conventional web.
Figure 31. Torsional vibration test results of butterfly web.
Figure 32. Excitation force (2.4 m/sec model wind velocity).
Figure 33. Excitation force (6.0 m/sec model wind velocity).
VS
amplitude
VS
amplitude
Here, the aerodynamic vibrational state of the structure is determined by the relative magnitudes of the aerodynamic excitation force and the inherent damping force of the structure (structural damping). In other words, vibration increases when the aerodynamic excitation force exceeds structural damping; conversely, vibration attenuates when the aerodynamic excitation force is less than structural damping. Based on this, the limited vibrations occurring in the butterfly web structure at the neighborhood of 25 m/sec full scale wind velocity (2.4 m/sec model wind velocity), which is apparently torsional vortex-induced vibration, is investigated, and the relationship between the excitation force (logarithmic decrement) and amplitude as the vibration progresses are presented in Figs. 32 and 33. Note that since the logarithmic decrement of the test model is 0.00647 (on average), which corresponds to its structural damping, the vibration will increase when an aerodynamic excitation force higher than this value is applied, while the vibration will decrease when only a smaller value of aerodynamic excitation force is applied. In Figs. 32 and 33, the vertical axis represents the sum of the model structural damping and aerodynamic excitation force. The region showing a logarithmic decrement smaller than 0.00647 is the region where aerodynamic excitation force may be said to be acting on the model. Figure 32 shows the test result for the model starting at rest, and where the torsional amplitude gradually increased from 0 until it reached 1.3 degrees double amplitude and assumed steady-state vibration. In other words, for the region where the amplitude is less than the 1.3 degrees double amplitude, the logarithmic decrement is shown as negative, and the vibration may be said to be 80
increasing because the aerodynamic excitation force is greater than the structural damping. Furthermore, Figure 33 shows the test result for the model starting from a 4 degree double amplitude excitation, and where a positive damping factor is shown immediately after the test started with gradually decreasing amplitude until it reaches steady state at 1.3 degrees double amplitude. However, this test was conducted with a logarithmic decrement of 0.006, which is much smaller than the value for the actual bridge, in order to understand the aerodynamic vibration properties induced by the shape of the butterfly web structure. If the damping factor for the actual bridge, which is a logarithmic decrement of about 0.03, is taken into account, this is equivalent to adding more damping with a logarithmic decrement of 0.024 to the test results. In other words, since it can be regarded as converging to the amplitude for the logarithmic decrement in the test minus 0.024, the amplitude of the actual bridge due to vortex-induced vibration can be inferred to be about 0.5 degrees double amplitude. 6 CONCLUSIONS This paper proposes a design for extradosed bridges with span lengths of the 500 m range, and examines their structural characteristics on the basis of a comparison with cable-stayed bridges. Studies revealed that a long-span extradosed bridge using butterfly webs is structurally feasible, and that it has characteristics such as those shown below. (1) Use of butterfly webs enables a lighter main girder, making it possible to increase the depth of the girder without increasing weight. This counteracts the larger bending moment of the main girder resulting from shorter main towers, and as there is less stress variation of the stay cables due to live load , it also becomes possible to reduce the factor of safety for the stay cables. The quantities of stay cables is similar to that of a cable-stayed bridge. (2) For seismic behavior, the extradosed bridge with a lighter main girder due to the butterfly web has smaller bending moment responses than those of a cable-stayed bridge. This allows a smaller substructure and provides other characteristics that are beneficial in earthquake prone regions. (3) For the proposed extradosed bridge, in spite of having a deep main girder, openings in the web due to the butterfly web design increase wind-resistant stability to some extent. Galloping is eliminated, and torsional vibration tests showed that divergent vibrations were limited to a wind speed range that is not an issue in practical terms. This paper has shown that a long-span extradosed bridge using butterfly webs has many advantages over a conventional cable-stayed bridge in terms of both structure and economy. As a result of the proposed design, the potential for extradosed bridges is dramatically expanded from the 250 m level that has been the maximum span for such bridges to date. Mathivat’s introduction of the extradosed bridge concept in 1988 greatly increased the freedom of design for cable-supported structures. By fusing new materials and new structures with his concept, this proposal expands design freedom once again. REFERENCES Ashizuka, K., et al. 2012. Construction of Butterfly Web Bridge. fib Symposium Stockholm: 545–548. JPCI. November 2000. Specifications for Design and Construction of Prestressed Concrete Cable-Stayed Bridges and Extra dosed Bridges: Japan Prestressed Concrete Institute. Kasuga, A. 2006. Extradosed Bridges in Japan. Structural Concrete Journal of fib: No.3, 91–103. Kasuga, A., et al. 2010. Study of a bridge with a new structural system using ultra high strength fiber reinforced concrete. Proceedings of 3nd fib Congress: Washington DC. Kasuga, A., et al. 2015. Study on 500m span extradosed bridges. fib Symposium Copenhagen. Kido, T. & Hoshino, Y. 2010. Recent Development of External Tendon System Using Ultra High Strength Prestressing Strand with Epoxy Coating. 3rd fib International Congress: Washington DC.
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Kutsuna, Y., et al. 2002. Non-linear behavior of the Ibi River Bridge under ultimate loads. 1st fib Congress 2002 Session2: 93-100. Osaka. Maeda Y., et al. 2002. Design and Construction of the Himi Bridge – Extradosed Bridge with Corrugated Steel Web. 1st fib Congress 2002 Session1: 95–100. Osaka Mathivat, J. 1988/2. Recent Development in Prestressed Concrete Bridges. FIP notes: 15–21. Muller, J. 1990/12. Les ouvrages d’art autoroutiers. Travaux: 88–94. Shimada S. Dec. 1965. Shear Strength of Steel Plate Girders with Folded Web Plate (Ripple Web Girders). Journals of the Japan Society of Civil Engineers No.124: 1–10. (in Japanese) Woelfel, E. 1990. External versus internal bonded prestressing. SP-120:External Prestressing in Bridges: 425–436. ACI
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Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
Multi-span large bridges – interaction between design and construction A.F. Bæksted COWI A/S, Copenhagen, Denmark
ABSTRACT: The design adopted on the new link across Lake Maracaibo carries both road and railway traffic. Due to the time schedule and phasing of the project as well as important aspects of construction, it was deemed advantageous to design and construct two parallel bridges with independent superstructures for road and rail, respectively, but with common foundation. This has led to a one-level superstructure type, utilizing concrete as the preferred construction material. This is based on availability of materials and tradition for maintenance. Present note will present some key features of the design in the project and try to extract what governs the design and construction of the project. A few comparisons with other fixed links are made. 1 PUENTE NIGALE – MOST IMPORTANT FEATURES AND PHASES 1.1 Overall scheme of the project Construction of the second fixed link across the Maracaibo Lake in Venezuela – the Puente Nigale – started in 2012. The first link – Puente General Rafael Urdaneta – is more than 50 years old and very much trafficked. The total length of the new link is around 11 km from coast to coast and consists of approximately 1.3 km short or medium span bridges at the coastlines, 4.5 km low level bridges, 4.8 km elevated bridges and two parallel 910 m cable stayed bridges with 430 m main span across the navigation channel in precast segmental construction. A dual two lane highway incl. emergency lanes alongside a double track railway line shall expand the infrastructures’ capability in transporting passenger and freight trains and vehicular traffic from the western to the eastern shore of the lake. The first forecasts and planning for a new fixed link dates more than 15 years back and a number of sketch layouts and studies were carried out over the years.
Figure 1. The project is located at the strait at the northern end of Lake Maracaibo.
83
1.2 Project during 2008–2010 The project broadly comprised multiple span low level approach bridges, an artificial island and a submerged tunnel under the navigation channel in the lake. The low level bridges entailed 75 m spans and was composed of separate bridges for road and rail. The substructure of the low level bridges was intended made by bored cast-in-place piles and in-situ cast pile caps. The superstructures for both rail and road was designed as prefabricated post-tension concrete girders. Erection of the girders was intended to be conducted from barges subsequent from load out at a nearby construction yard. 1.3 Project in 2012 After appraising the difficulties of the prevailing ground conditions at the navigation location invoking on the feasibility of the proposed submerged tunnel a re-setting of this part of the project was suggested. The tunnel was abandoned and conceptual study made which led to a high bridge over the navigation channel. A clearance of 50 m was introduced. 1.4 Project 2012 to date Thorough studies were made to prepare what was deemed an optimal alignment for the combined rail and road corridor. In particular for a bridge with cable systems it was important that the bridge designers had influence on the overall geometry to ensure suitable and viable curvature and symmetry. Thus, focal points should concentrate on where the combined advantages are the greatest. This results in a desire of providing symmetric horizontal and vertical alignment, centered around the mid point of the main bridge. During manufacturing and construction the merits of such a design aspect should not be underestimated. After in-depth assessments of ship collision risk the length of the navigation span was determined to 430 m. This was however also impacted by many other parameters of which a few but most important can be mentioned: structural articulation, skew angle between navigation channel and bridge alignment, construction method of main bridge, foundation arrangements including concept, constraints from available materials for concrete, specific design aspects of a main span with precast segmental and environmental aspects. For phasing, see Figure 2.
Figure 2. Layout and phasing of the project, 2015. Legend: WSSB – West Short Span Bridge, LLB – Low Level Bridge, EB – Elevated Bridges, MB – Main Bridge, EMSB – East Medium Span Bridge.
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Figure 3.
Rendering of the main scheme of the project, 2015. Seen towards West.
1.5 Primary parties involved The owner of the project is the Ministry of Popular Power for Land Transport, Venezuela. The Contractor is Constructora Norberto Odebrecht S.A [Odebrecht] who is the executing party of the contract with the owner. The main consultant to Odebrecht in the design and build contract is COWI A/S, Lustgarten y Asociados, C.A. as detailer of the design, and the independent checker is T.Y. Lin International. 2 EXCERPT OF BASIS AND GUIDES TO INTERACTION 2.1 Design code The design code is AASHTO LRFD, however augmented on the loading side for combined road and railway loading. For this purpose inspiration from Eurocode EN 1991-2 has been adapted. 2.2 Bathymetry and climatic conditions At most of the 11 km between the shores of Lake Maracaibo in the alignment the water depth is greater than 3–4 m. Shallower water is naturally encountered near shore straddling a few hundred meters from the shore line. The maximum currents during normal operations (1 year return period) are less than 1.1 m/s with typical current speeds of 0.6 m/s. In the 100 year return period the current speeds are 0.2–0.3 m/s higher. Temperatures are rather constant throughout the year compared to many other locations in the world and typical average daily temperature is around 30 deg. The wave climate is quite favorable for seaborne operations too as the significant wave height even for a 100 year return period hardly exceeds 1.5 m and for the 1 year event the value is around 0.9 m. It shall be noted that if global sea rise is included the wave heights naturally becomes a little larger, in particular in the shallow water. 2.3 Preparation of design basis and design statements Preparation and use of a project specific design basis has been essential. Several volumes for General, Road, Rail, Bridge and Geotechnical are prepared to capture project specific basis notwithstanding being enforced by legislation, codes, owner requirements or from other parties. A number of design statements have been prepared to specifically address items where the basis design code (AASHTO and in a few instances Eurocode) is not concise or is open for choices. Those items are explained and directions are given in these documents. 85
Figure 4.
Short Span Bridges (left) and Low Level Bridges (right).
2.4 Availability of materials and equipment A constraint on the design and construction is e.g. cement and to some extent aggregates. The concrete mix design exposed to an aggressive environment during the service life are thus carefully trimmed to the various primary exposure zones: submerged, splash/spray and atmospheric. Other examples on project defining parameters are the possibility for obtaining permission for rental of large scale equipment such as driving hammers fitted to the scale of project and barges for erection of the structural members. In broader terms the local conditions – technical, tradition and preferences, political, considerations to neighbors and other stakeholders, wishes and requirements from the Owner – impose and mold the solutions. This is at the same time an enchanting challenge as it may give rise to some frustrations in short lapse of time. 2.5 Mutual understanding in and between the project teams – project charter Noteworthy for the designers are to adhere to local and project specific traditions and requirements in all aspects including what might often be contemplated by designers as being of little or no importance. What could appear to be simple logic may not be so easy to have in the mindset. This could be project specific setup of drawings which are tailor-made to either procurement team, to production team or to the construction team. What is in focus in one team may not necessarily be in focus in the other and information provided to one team may be utterly superfluous to another. The importance of feedback from planning, procurement, fabrication and construction teams cannot be underestimated. Having said this it can never be achieved that everyone know all about everything. But as ever important is the fine art of communication between the parties involved in a project and the larger the project is the more important a common designation and understanding is. The communication part is all about people and how people interact. 3 MAIN MOTIVATIONS FOR PROJECT LAYOUT 3.1 Short span bridges At the western shore 500 m of short span bridges are designed and under construction. There are a number of reasons for this concept with 25 m spans. See Figure 4 (left). Important drivers are the availability of equipment, reasonably early commencement of the project, involvement of local sub-contractors and vendors, setting up production facilities for primary construction materials in particular concrete and steel piles. This gives not only valuable experience with the not-so-often used design with combined load of rail and road but probably more importantly evolvement of knowledge for the various 86
Figure 5. Transition from Short Span Bridge, West and Low Level Bridge.
processes: design, design check, procurement and production of materials, preparation of access facilities and construction plant, adaptation to local legislation for e.g. import of materials and plant, feedback from production and construction to design and alterations to design. The learning process and gained experience through this part of the project constitutes a good basis for the remaining parts. 3.2 Low level and elevated bridges The majority of the bridge structure is featured by this concept where a module of 75 m span length was re-affirmed as the preferred. The articulation chosen applies isolation of the superstructure from the substructure. It was during the design considered to use a monolithic connection but in particular the seismic load would in such case have been overridingly governing the design and it was deemed to have un-desired complication for construction. By isolation of the superstructure the horizontal seismic load is more in line with the applied ship impact loads from vessels and barges. The substructure consists of driven, vertical 1.8 m steel piles with a sheet thickness of 25 to 32 mm. Pile caps entails a prefabricated outer shell consisting of several uniform pieces to allow for easy lifting operations and a load transferring and carrying reinforced core. A delicate detail is the structural connection between the steel piles and the pile caps. This is facilitated by an insitu cast reinforced concrete plug in which shear rings ensure transfer of shear and flexure to the concrete plug. One of the difficulties are the position of the shear rings which are to be forecasted from the encountered – and varying soil conditions – during the driving operation. An example of feedback from site to be responded to is a design-projected shift from high grade steel to lower grade and simultaneously a shift in sheet thickness of the sheet. Due to unforeseen soil conditions encountered in a specific location an alteration to the design is developed and hence implemented on site. Piers, segmented into elements of less than 400 tons and pier heads with a weight of maximum 1000 tons are all prefabricated and only stitches between elements are made in-situ. The superstructures for both rail and road have been considered constructed by both full span construction (FSC) erected by barges and by in-situ construction on a movable scaffolding system (MSS). With FSC the primary challenges are the heavy lift of up to approximately 4500 tons for the road girder elements and 2600 tons for the rail girders. A concern worth noting is the fact that the heaviest elements conjure with the highest lifts. Even though the climatic conditions may not appear too onerous they potentially pose a risk which infer on the decision for the construction method. Contrary to FSC the MSS method eliminates the very heavy lifting operations. The method has some merits in the construction of the most shallow water areas. The main risks may be the logistics 87
Figure 6. Transition from Low Level Bridge to Elevated Bridge, seen towards West.
in transportation of work force and materials to site location on a daily basis. The construction schedule should be able to include for a slack subsequent to a delay at a single pier/foundation to eliminate costly delays or workarounds. For all structural components a close interaction between the design, production logistics and facility and the construction methods has been exercised. In the continuation of the project execution the feedback will for certain principles and details give rise to some alterations. 3.3 Main bridge The main bridge is a cable stayed bridge with a main span of 430 m, side spans of 162.5 m and 77.5 m. The main bridge features two superstructures with central cable planes. The pylons, anchor piers and transition piers for the road and rail bridges sit on common foundations. The layout with anchor piers in the side spans has proven to be efficient. It leads to uniform stay cable forces in the side spans, minimizes moments in the pylons and avoids complications from particularly heavy cables that would also require extra space at the anchorages in the pylon. Tie-down systems at the anchor piers are designed to avoid uplift in the bearings under Service conditions. The articulation chosen adapts a monolithic connection between pylon and superstructures. During the design this concept has displayed the most optimal utilization of materials against the loads applied on the structure. This part of the project is the one that yields the highest demand to the compressive strength of the concrete and also the highest demand on the foundation piles. This reasoned with the high load intensity from dead load and live load on the superstructure and the design governing loads from ship impact and seismic load on the substructure. The substructure are in-situ cast solid pile cap resting on driven vertical 2.0 m steel piles. A major challenge for the substructure is the rather deep temporary cofferdam structures that is necessary for the construction of the pile caps that have bottom at level −14 m to Mean Seal Level. Conversely to the remaining parts of the link, the pile caps of the main bridge and a few for elevated viaduct spans are located below sea level. This is to reduce the risk of ship impact as well as the possible force eccentricities, and to reduce demands on the piles in case of ship impact as well as the impact during seismic event. The tip level of the driven piles is estimated to level circa −65 m. Use of caisson foundations has also been studied, which would be very efficient both for ship impact and seismic events, but has been omitted due to construction considerations. Substantial yard facilities is established for fabrication of large numbers of precast I beams for the short-span bridges as well as precast elements for piers and pile caps and (possibly) full-span prefabricated box girders of the low level and elevated bridges. Having such yard facilities available 88
Figure 7.
Main Bridge with Elevated Bridge, East in the background.
and planned for makes it particularly attractive to use prefabrication also for the main bridge, which is anyway advantageous in terms of erection time and achievable quality. Thus it was chosen to use precast segmental construction for the girders of the main bridge and with 430 m main span this certainly is in the upper range of what has been recorded in the past for a railway girder. The superstructure construction based on precast segmental method is the most demanding for the post tensioned concrete where the limitation set forth is a compressive cylinder strength of 45 MPa. The latter has been thoroughly discussed during the design and in turn selected to ensure as high compliance with the requirements to concrete quality during construction as acceptable and desired. From a pure design perspective surely a somewhat higher strength such as 60 MPa in many ways would have been desirable. That said, the selected value grinds out very well respect for the fine balance between design, available materials, procurement, production and execution. This is where one of the finest dilemmas in design and build contracts really is displayed. In the end a close corporation between the contractor and the designer, spiced up with valuable contribution from the independent checker, yields the best value available for money. The longitudinal design of the girder is very much governed by the segment joints. The verifications that particularly govern the required concrete dimensions and the required amount of post tensioning are: – concrete stress limitations in Service Limit State, – verification for vertical shear and torsion in Strength Limit State, and – deformation limitations in Service Limit State At the commencement of the design an assessment of the precast segmental method versus in-situ construction was conducted and also here the pro’s and con’s were evaluated including the various parameters displayed above. 3.4 Geotechnical conditions The regional geology is dominated by the Maracaibo Basin, which is an intermountain basin lying between the Perija and Merida Andes in the northwest corner of Venezuela. The main part of the basin floor is covered by the waters of Lake Maracaibo and recent deposits. Along the alignment, the sedimentary deposits sampled during the borehole campaign consist mainly of recent (holocene) material resting un-conformably on the “El Milagro” formation of inferred Pleistocene age. The recent cover is generally less than 5 m thick and consists predominantly of very fine grained, loosely deposited, dark greenish-gray muds. The upper clay layers are generally soft and of low plasticity, whereas normally consolidated clay with high plasticity is found at greater depth. At greater depth medium dense to very dense sand and over-consolidated clay with low plasticity are found in the assumed Pleistocene unit. The very dense sand and over-consolidated clay superposes bedrock, consisting of claystone and siltstone with interbeds of sandstone, which are also referred to the Pleistocene unit. The surface of the bedrock is found at various depths bellow surface/seabed. The minimum depth is 8 m, but at several boreholes the surface of bedrock has not been found at depths up to 70 m. 89
Figure 8. Main Bridge – image from above deck level featuring two separate decks, each having a central cable plane.
Figure 9.
Stratigraphy in the alignment (Legend; green: clay, yellow: sand, magenta/blue: bedrock).
The essence of the above potpourri of stratigraphy and geology moreover unveils a very large variety of geotechnical conditions, that interacts with the structural behavior of the structure and to some extent also the loading as a function of soil-structure interaction. Thus, it is of the utmost interest to develop a foundation concept that simultaneously conform to design requirements and is flexible against encountered geotechnical conditions on each foundation location; the latter explored full scale during driving of the foundation piles. For the short span bridge the design and execution on site was particularly challenging as the geotechnical site investigations in some locations were carried out simultaneously with the construction (driving) of the upper part of the piles. A smooth transfer of driving records from test piles on site at each individual pile cap combined with a clear method for handling the results in turn made an instruction to execute the final part of the driving, e.g. the uppermost 6 m, in an expedite manner. This methodology of feedback from site and subsequent adaptation in the geotechnical design is even more important in the construction of the sea-based bridge parts.
4 EXAMPLES FROM OTHER PROJECTS 4.1 Qatar Bahrain Causeway A 40 km combined rail and road link, tendered as a design and build contract, with around 25 km bridges was the core of this project (2008–10). It was essential with the preparation of a concise design basis from Employers Requirements. In this project the soil conditions – and to some extent references from nearby King Fahd Causeway – led two parallel bridges with a substructure concept comprising precast monopiles installed in drilled shafts and with the void between soil strata and piles grouted. 90
Figure 10.
Pile driving rig working side by side of rig executing CPT tests.
A decisive factor for the layout was a requirement for continuous welded rail and thus with a strong incentive for prefabrication the approximately 400 viaduct spans (low and elevated), each was 50 m and based on full span erection. All, except deck structures of the two main bridges are constructed in concrete. The integrated planning and design – simultaneous with posted liaison officers from the independent checker – were conducted from common offices close to the owner’s premises. The value of the common office during a very intense design period was un-measurable. 4.2 Storstrømmen, Denmark The project, a 4 km combined rail and road bridge crossing a strait, is tendered as a design and build contract during 2015. An initiative to observe and adapt design and construction aspects from the market has been taken through a technical dialogue prior to the design and build tender and this has fertilized the common perception of the projects scheme. This process involving the owner of the project and eight contractor consortiums has unfolded a number of characteristics that the owner could contemplate to benefit from and to include in their entirety or partly in the project specifications. Some of the most important views from the contractors has been: – – – – – –
formulate a consistent design basis, provide a certain degree of freedom to the design and construction, but do not open up for everything as allowed variables, make room for various construction methods if possible, if possible allow for a certain slack in the time schedule, and introduce a certain risk sharing for climatic and geotechnical conditions and uncertainties
4.3 Sheikh Jaber Al Ahmad Causeway, Kuwait An approximately 25 km highway link crossing predominantly shallow water. The project was tendered in a design and build contract and is currently progressing on site. Also in this project the substructure concept for viaduct spans entail precast monopoles installed in drilled shafts. Again a strong incentive for prefabrication of the viaduct spans (low and elevated) and subsequently full span erection was selected by the contractor. 4.4 Mersey Gateway, United Kingdom The centrepiece of the project is a new 2.3 km-long, six-lane toll bridge over the River Mersey of which the 1.0 km main bridge is a central part. The design features a post tensioned concrete box supported on a single line of central stays and hence an array of three mono-towers form the iconic features of the crossing, with the central tower being the shorter. The approach span are constructed 91
by means of a Movable Scaffolding System (MSS) and construction effects are accommodated for in the design. The substructure for the approaches a made from bored in situ cast concrete piles and pile caps whereas the main bridge is supported on pad foundations. All construction work is based on in situ construction. A significant asset for the team has been to collectively gauge on the implications on design depending on construction method and vice versa in the sense that design choices naturally impact on the construction including plant and equipment. 4.5 Queensferry Crossing, Scotland Towering over the Forth alongside the iconic 125 year old railway bridge the construction of the new cable stayed road bridge is well under way. The 40 m wide composite bridge deck is sustained by a central stay plane and three towers featuring two 650 m main spans and about 550 m approach spans on the southern side of Forth. The superstructure sits on concrete caisson and piers. All of the works are conducted in a design and built contract. Some of the characteristics – let aside the long spans and wide use of prefabrication – are the lessons learned in the tendering process and the post award phase. It was appreciated by the bidders that the Employer – Transport Scotland – had a clearly defined framework for the scheme presented. A very comprehensive competitive dialogue was planned for and the framework directed the bidders well and yet allowed for ingenuity. Post award all parties except the independent checker were gathered in the same office premises. To date this is by all parties evaluated as being a primary asset in achieving close collaborations and mutual understanding of the project charter. 4.6 Pulau Muara, Brunei This project is a 2.7 km highway link crossing to connect to new urban area. The project is a design and build contract and consist of two approach bridges 0.9 km and 1.4 km and a central 0.4 km main bridge. The superstructure construction was subject to intensive assessment in the tender group prior to selection of the preferred method. Selection was made from precast segmental method (PSM – either free cantilever or span by span), free cantilever method (FCM) and incremental launching (ILM) and various combinations of the method. It was decided that precast segmental construction was the optimal for the group. One of the fundamental again was the very close liaison in the group in developing the concept which was best fitted to geometrical constraints, competencies, experience, price, allowance for utilities and risk profile. 5 CHARACTERISTIC FEATURES From the Puente Nigale project the signposts will be: – – – –
structure layout adapted to functionality and equipment, plant, equipment, materials available and tradition masterminds the design, have focus on communication and exercise the fine art of it, roll up the sleeves and get the job done in a transparent and honest project atmosphere. What are the main learnings for comparison between projects in design and built contracts?
– – – – – – –
give a certain degree of freedom under clear constraints, define success criteria, high degree of prefabrication will almost always be desired for parts of the scheme, planning – communication – execution – feedback: essentials, gauge on preferences, equipment and risk assessment, be aware that parties works most efficient when situated in the same physical office, if you are acting as consultant: exercise skills in all thinkable construction methods, and be minded for change management and act accordingly. 92
Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
Recent achievements in the design and construction of multi-span cable supported bridges in China A. Chen & R. Ma Department of Bridge Engineering, Tongji University, Shanghai, China
X. Zhang CCCC Highway Consultants CO., Ltd., Beijing China
ABSTRACT: In recent years, many multi-span cable supported bridges have been designed and built in China. Among them, there are Taizhou Bridge, the largest multi-span suspension bridge, and Jiashao Bridge, the largest multi-span cable-stayed bridge. Taizhou Bridge, spanning over Yangtze River in Jiangsu Province, is a three-pylon two-span suspension bridge with a span arrangement of 390 + 1080 + 1080 + 390 m. It has a steel middle pylon and two concrete side pylons and steel bridge decks. Jiashao Bridge, spanning over Hangzhou Bay in Zhejiang Province, is a six-pylon cable-stayed bridge with a span arrangement of 70 + 200 + 5 × 428 + 200 + 70 m. It is the largest multi-span cable-stayed bridge in the world with the widest bridge deck of 55.6 m. In this paper, some key design and construction technologies of the two bridges are presented. The following topics regarding the design and construction of multi-span cable supported bridge are discussed: for Taizhou Bridge, (1) vehicle load of multi-span suspension bridges, (2) the anti-slide problem between the main cable and the saddle on the top of the middle pylon, (3) the construction of the longest main cable; for Jiashao Bridge, (4) the structural system of the six-pylon cable-stayed bridge, (5) the solution to release temperature stress for long bridge deck, (6) the maintenance equipment for multi-span twin-deck steel girder, and (7) the construction of bridge deck. These studies offer significant advances in understanding and predicting the behavior and the construction of those cable supported bridges with multiple spans.
1 INTRODUCTION In recent years, many long span bridge have been built in China. And a new trend can be found that multi-span cable supported bridges (MSCSB) have come into our lives as a new bridge family (Virlogeux, 1999). In China, many multi-span cable stayed bridges and multi-span suspension bridges have been built, which are listed in Table 1. During the design and construction of those MSCSBs, there are many technical problems in need of settlement. In this paper, two bridges are introduced, which are Taizhou Bridge, the largest multi-span suspension bridge(MSSB), and Jiashao Bridge, the longest multi-span cable-stayed bridge. Some key design and construction technologies of the two bridges are presented. The following topics regarding the design and construction of multi-span cable supported bridge will be covered are discussed: for Taizhou Bridge, (1) vehicle load of multi-span suspension bridges, (2) the anti-slide problem between the main cable and the saddle on the top of the middle pylon, (3) the construction of the longest main cable; for Jiashao Bridge, (4) the structural system of the six-pylon cable-stayed bridge, (5) the solution to release temperature stress for long bridge deck, (6) the maintenance equipment for multi-span twin deck steel girder, and (7) the construction of bridge deck. These studies have led to significant advances in understanding and predicting the behavior and the construction of those cable supported bridges with multiple spans. 93
Table 1. Some large span MSCSBs in China. Bridge Name
Span Arrangement (m)
Bridge type
Erqi Bridge in Wuhan, Hubei Province Ting Kau Bridge in Hong Kong Jiashao Bridge in Zhejiang Province Yiling Bridge in Yichang, Hubei Province Dongtinghu Bridge, Hunan Province Taizhou Bridge in Jiangsu Province Ma’anshan Bridge in An’hui Province Yinwuzhou Bridge in Hubei Province
90 + 160 + 2×616 + 160 + 90 127 + 448 + 475 + 127 70 + 200 + 5×428 + 200 + 70 38 + 38.5 + 43.5 + 2×348 + 43.5 + 38.5 + 38 130 + 2 × 310 + 130 390 + 2 × 1080 + 390 360 + 2 × 1080 + 360 225 + 2 × 850 + 225
cable stayed bridge cable stayed bridge cable stayed bridge cable stayed bridge cable stayed bridge suspension bridge suspension bridge suspension bridge
Figure 1. Taizhou Bridge.
Figure 2.
Jiashao Bridge.
Taizhou Bridge, spanning over Yangtze River in Jiangsu Province, is a three-pylon two-span suspension bridge with a span arrangement of 390 + 1080 + 1080 + 390 m. It has a steel middle pylon, two concrete side pylons and steel bridge decks. In 2012, the bridge was open to traffic. Jiashao Bridge, spanning over Hangzhou Bay in Zhejiang Province, is a six-pylon cable-stayed bridge with a span arrangement of 70 + 200 + 5 × 428 + 200 + 70 m. It is the largest multi-span cable-stayed bridge in the world with the widest bridge deck of 55.6 m. In 2013, the bridge was open to traffic. 2 VEHICLE LOAD FOR MSCSB In the design of MSCSB, the traditional approach for vehicle load determination by referring to the related codes and guidelines is not suitable any more due to the exceedingly long span of these bridges (Guo, 2012). For the MSCSBs, dead load will be the major load because of their large scale. Under this circumstance, normal heavy vehicles have no significant influence on the whole performance of these bridges. However, some load parameter, such as the constitution of vehicle types, the space headway, the waiting time for accidents, have more obvious influence on the whole 94
Figure 3.
Load distribution schemes for MSSB.
Figure 4.
Establishing process of vehicle load model.
Table 2. Design vehicle load model for Taizhou Bridge in ULS. Loading distribution
single span
two spans
Standard value of lane loading Reduction factor (daily volume 100
B-C C C* C* C* C* A
C A A-B A-B A-B A B-C
C A A A-B A A-B A-B
A A-B A-B A A A A-B-C
>100
A
B-C
B
A-B
>100 >100
A A
B-C B-C
B-C C
A-B A
Legend: A – Good; B – acceptable; C – critical; * – with identified potential progress.
by the construction method. In other words, most probably, in several cases, the optimal span may be “rejected” not by design criteria, but due to construction method restrictions or inadequacy. 1.1 Key issues for the constructive method selection Each construction method has its own merits and, most probably, each method has a preferred field of application. Surely the “best” decision should be taken in a case by case basis. The adoption of a construction method is part of a “triangular decision” of conceptual design which always comprises Structural System (& Cross Section) – Material – Constructive Method. Several well-known factors and several well-known conditions are to be considered in each case. Among all factors, there are four that have a systematic importance: Technical Validation, Cost, Time and Durability. Let us make an exercise of qualitative evaluation of the most common constructive methods used for multi-span large viaducts in the beginning of the present century (see Table 1). The observation of Table 1 allows to stablish a main conclusion: in the beginning of the XXI century there were several construction methods that were unsuitable for the construction of 70– 100 m multi-span viaducts due to technological limitations. In the following point a little deeper approach on the reasons which justify this conclusion is made for the particular case of, in situ, span by span construction with Movable Scaffolding Systems (MSS). Afterwards, similar conclusions are drawn for precast segmental, span by span construction.
2 PART 2 – WHY LARGE MOVABLE SCAFFOLDING SYSTEMS (LMSS) WERE NOT USED BEFORE: TECHNOLOGICAL CHALLENGES A Few years ago, there were 4 main reasons why Large Movable Scaffolding Systems (LMSS), i.e., movable scaffolding systems for the range 70 m to 100 m, were not used. • • • •
Excessive Deflections Excessive Weight of Equipment Doubts on the Stability of large spans MSS related with the Wind action and effects Productivity and Construction Time 105
Other possible reasons could be identified, as logistics, concrete pouring control (due to duration and volume), or off-shore restrictions, but surely reasonable solutions to overcome them could be developed, if the formers were overcome. 2.1 Excessive deflections in former MSS technology for 70 to 100 m span The deflection of a MSS is one of its main operational characteristics. Recently, the larger span MSS (60 to 70 m span) were characterized by mid-span deflections (during concrete pouring operation) which were near the limits of acceptable operational values. And it was known that for larger spans that limit would be overcome. Indeed, the common values of mid-span deflections (D) during concrete pouring operation, for most common MSS were within the range L/1000 < D < L/400. For the span range 70 m < L < 100 m, that would represent absolute values within the interval 70 mm < D < 250 mm (Vasques de Carvalho 2008; Pacheco et al. 2011). It is easy to verify that, even applying pre-cambers with a 75% efficiency (meaning that the difference of measured deflections and theoretical ones would be less than 25% of the elastic deflection), that would imply that the potential geometric deviations could be easily greater than 30 mm, or even greater than 50 mm (depending on the span length), i.e., above common geometric acceptable tolerances. Additionally, such amount of deformations could induce structural problems on the erected decks, as cracks, if the concrete pouring operation was long (what is expectable for such spans) and/or if more than one stage of concrete operations occur, where the concrete of first stages could become with damage (Pacheco et al.). According to previous studies, for LMSS, good results are achieved if the mid-span deflection limit is L/1000 (Pacheco et al. 2011). Thus, to reasonably increase the MSS span range, new technologies were to be developed. 2.2 Excessive weight of former MSS technology for 70 to 100 m span The weight of an MSS may have impact on the deck design and also on the piers design. Additionally, it is known that the design of the deck may significantly depend on the constructive joint location (eventually between at L/4 or L/5 from the pier. Associated with this issue it is also relevant the location of MSS back support (typically on the deck cantilever). Let us refer Figure 3, where the registration of the weight of 23 different conventional MSS are pointed and where a linear approximation of MSS conventional weight is drawn. Very simplified previous studies, which equalize the deck flexural moment over the last pier with deck, for the maximum constructive loading scenario and for bridge service loading (the same pier with the complete deck), allow to obtain grossly approximate values of what would be approximately a neutral weight for the deck design (Pacheco et al. 2011). In Figure 3, a simplified indicative linear projection of “MSS neutral Weight” is also drawn. In Figure 3, it is, then, possible to observe that above approximately 65 m of span, the weight of conventional MSS tends to be conditioning for the deck design, and that tendency clearly increases with the span length. It should be clear that this analysis is simplified and no conclusions are to be established in an exact or even nearly exact basis. Just for example, the deck weight obviously depends on the width of the deck (Pacheco et al. 2011). Nevertheless, this exercise allows to obtain general conclusions. In what concerns to piers design, any approximation, even simplified, would be much difficult to achieve. Indeed, the piers height, the structural design, the specific actions of the local, etc., could have important influence in each case. Nevertheless, if conventional MSS were applied for 70 m to 100 m spans, Bridge Designers should expect service horizontal forces values on the top of the piers that could easily reach values from 75 ton to 150 ton, this applies, neglecting longitudinal slope effects. Regarding this matter, the MSS sliding devices specification is also an important issue (Pacheco et al. 2011). 106
Figure 3.
Conventional MSS weight/span relation – versus – MSS neutral weight.
This very simplified exercise, most probably frequently done by several Bridge Engineers, in an intuition/common sense basis, shows another reason why a few years ago MSS construction, for 70 m to 100 m spans, was not an option. 2.3 Former doubts on stability LMSS with wind effect The study of wind action and wind effects in MSS involves some particularities where importance increases in the particular case of LMSS. In both cases, for MSS and LMSS, wind actions are to be substantially different in launching operation stages and in equipment stationary stages. This results from: (1) the duration of the stages/operations and adequate return period; (2) common predicted limitations in the MSS Operational Manuals and (3) due to changing in support conditions, changing in span distribution, changing in mass values and in mass distribution and, finally, the evolution on the structure location, during the equipment launching stage (Resende et al. 2015). The wind actions in these type of equipment are not included in known wind action codes, then it is necessary to create specific rules and criteria for this type of structures, with, as explained before, very particular characteristics. Nevertheless, in the particular case of MSS, there is specific and useful bibliography (almost with normative significance) that treats this disparity between situations, using different wind velocities for each situation (SEOPAN 2007). But, previous studies (Rosignoli 2013, Pacheco et al. 2015) show that natural frequencies of MSS during launching stage, as expectable, decrease with the span increase. Although such operation is to be conditioned by actual winds measured during the operation, if natural frequencies are too low, there is not an adequate domain of involved phenomena. This fact induces doubts on the applicability of previous knowledge to the study of stability of LMSS with wind effect. Indeed, if in LMSS natural frequencies are below the former typical values of natural frequencies for MSS (Pacheco et al. 2011), as dynamic phenomena may occur, additional research is to be developed for LMSS. This issue is another fundamental point which needed further research in order to validate LMSS technology. 2.4 Construction time and productivity of MSS technology for large spans The last point which demanded further technological development was the productivity. As a matter of fact, productivity is obviously a key issue for all intervenients in a Bridge Project but, particularly important for the Constructors. 107
In the beginning of the XXI century there is plethora of experiences referring to cruise MSS deck erection productivities nearby the value 3 km/year, for the span range 40 m to 60 m. But, according to known relates, that value could decrease down to 2 km/year (or lower values) of cruise deck erection for 70 m span decks. And, eventually, the expectative for larger spans was not better. For the decision makers, these values of productivity were not compatible with the industrialized needs of multi-span viaducts, which could be near from 4 km/year achieved with other methods. Again technological development was demanded. 2.5 Intercalary conclusion: the need of relevant technological developments in LMSS technology – for 70 m to 100 m span range It is known, that the construction of decks of bridges and viaducts with several spans with movable scaffolding systems (MSS) may be a very efficient and competitive constructive method. In particular, this method is among the most economic constructive methods (if not the most) in what concerns to materials consumption. Recent comparative studies show that the PC, cast in situ, span by span construction for span range 70 m to 100 m (with LMSS) may be even more efficient in terms of material consumption (Morim 2008; Lopes 2015). In fact, in this range, surprising economical results may be achieved if the number of spans is high and/or if the costs of piers and foundations are relatively high. Moreover, the use of LMSS may represent very significant cost reductions if the accesses to the front line of site are difficult – for example, high piers or water access, because this may imply for most significant costs of elevation equipment and/or site facilities. Finally, the span by span construction also ensures important advantages as the perfect continuity of the deck. Thus, the presentation of previous points may stablish a first conclusion: in the beginning of the XXI century beginning there was a clear need of relevant technological developments in MSS technology for 70 m to 100 m span range – meaning the development of LMSS (Pacheco et al. 2011). In the following Part of the present paper, relevant recent achievements, mainly related with technological progress in the construction of PC, span by span, in situ and segmental precast, multi-span viaducts are presented: the Organic Prestressing technology. 3 PART 3: THE ORGANIC PRESTRESSING (OPS) TECHNOLOGY Organic prestressing system (OPS) is a concept inspired by the behavior of an organic structure found in nature: the muscle. It is nothing more than an active control prestressing system, whose objective is to reduce deformations and/or stresses due to live loading (Pacheco 2002). Still in early XX century, Freyssinet and Zetlin mentioned the possibility of strengthening structures with active cables (Falcó et al. 1990). They did not continue studies in this field probably because the technological context then was unhelpful. After 1970, several applications on active control of structures involving active cables were developed, mainly for vibrations control (Soong 1990). In the last years of the XX century, a Research Project on Organic Prestressing was initiate at FEUP (Pacheco et al. 2002, 2007). After the first numerical simulations it was concluded that the concept of organic prestressing would be particularly useful to apply on large bridge equipment structures, due to the importance of live load and deformations, where OPS is particularly efficient (André et al. 2006). Figure 4 summarizes the main steps of the project development, since R&D stage at FEUP, to several first key applications, in particular in the deck erection of multi-span large bridges and viaducts. 3.1 OPS technology brief description OPS consists of known, unrelated technologies. The main elements are (a) the actuator in the organic anchorage, (b) the unbonded cables, (c) the sensors and (d) the electronic controller in the 108
Figure 4.
Historic synopsis of OPS technology development and main achievements.
girder control unit (as in Figures 5 and 6). All of them have been used before with reliable results, but not in the present combination (Pacheco et al. 2007). An effective and simple control system was achieved, in which using mid-span deflection as main control variable (input) is feasible and simple to control geometry and indirectly to control internal forces. Great part of the OPS software (clearly the major part) comprises safety complementary codes, as the control code, itself, is most simple (Pacheco et al. 2010). Present software, with 2 109
Figure 5.
Main components of OPS technology for underslung MSS.
intercalary upgrades has been used in the construction of several kilometers of bridges and viaducts, worldwide, without any incident. The symmetric algorithm controls the bridge deck prestressing stage (reverse process). In both stages software filters are used to oversee vibrations. Indeed, this control algorithm is valid for static control. Thus, to avoid control instability, unit-step changes performed by the actuator (output) must not depend on vibrations. More than one technique may be used to achieve such a filtering procedure. One solution consists of defining time increments as the computed average of a convenient number of consecutive mid-span deflection measures, during an adequate analysis period (Pacheco et al. 2010). The first OPS concept was developed for underslung equipment. Main components, in a similar application, may be observed in Figure 5. The OPS bowstring concept was developed between 2007 and 2009 (Pacheco et al. 2009). Again reproducing another elementary static classical concept – see Figure 6 – this solution brought important features. Indeed, the bowstring concept, is obviously much suitable for overhead equipment which comprise some important advantages in some type of applications. Mainly, (1) overhead MSS are more adequate for bridges with important plan curvature, (2) are frequently easier to assemble (behind abutments) and most important, (3) have more stability in launching stage (Pacheco et al. 2008). OPS Bowstring concept is the solution already developed and applied in Large Movable Scaffolding systems (LMSS), presented later. More recently, another classical static solution was used to develop the Cable Stayed OPS concept, as in Figure 6. This system is particularly useful for double span Equipment, taking advantage of the nearly symmetrical structure and loading. In this equipment pouring operation is to be performed in a symmetrical progression. 3.2 Some experiences worldwide, using OPS technology (span by span, cast in situ) Important knowledge arises from each application. The success of each experience results from an adequate work comprising an integral bridge design-equipment-operation approach. In fact, the interaction with the 3 corresponding intervenient players – bridge designer, equipment supplier, operation crew – is fundamental. Any unilateral approach of any intervenient may represent an apparent benefit in one perspective, but may be harmful to the global solution. In Figure 8 some experiences in span by span, cast in situ, construction using OPS technology are shown. 110
Figure 6.
Main components and layout of the OPS bowstring concept.
Figure 7.
Main components of OPS technology for underslung MSS.
3.3 OPS application in Precast Segmental construction Since 2013, the OPS technology is also applied in Precast Segmental Construction, in launching gantry equipment (Pacheco et al. 2014). Main features are described later in detail – for a particular case study – but again the deflection control is one of the most important features, in this particular case, with very relevant impact on productivity, as it will be seen (Sousa 2014). 111
Figure 8.
Some projects with movable scaffolding systems with OPS.
As a matter of fact, the technological challenges of OPS technology application in large span launching gantries are simpler to overcome than in movable scaffolding systems. Indeed, from the four technological challenges for large span technology identified in Part 2, there are 3 that are naturally overcome: (1) Excessive Weight of Equipment, (2) Doubts on the Stability of large spans MSS with Wind and (3) Productivity and Construction Time. The non-existence of formwork in the travelling mass, explains why in large launching gantries either the excessive weight of the equipment – due to relevant mass reduction – either the doubts on the Stability with Wind action – due to relevant wind exposure area – are not so restrictive issues as in MSS technology. The industrialized segmental precast solution, explains by itself why the productivityconstruction time is not an issue, in large launching gantries technology. This explains why, in the present state of art, there are functional and efficient launching gantries, for precast segmental cantilever construction, that reach the span range 100 m to 120 m, with success. In the case of span by span, segmental precast construction, with launching gantries, the technological challenge – excessive deflections – is not overcome, with current technology, for spans larger than 40–50 m. 112
Figure 9. Precast segmental, span by span, launching gantry up to 90 m (in tender stage).
Figure 10. Precast segmental, hybrid span by span – cantilever, construction, 100 m span (courtesy Javier Muñoz-Rojas/CFC- SL).
Again, the OPS technology becomes particularly useful. Indeed, there are already mature equipment solutions for OPS strengthened launching gantries for 70 m to 90 m span range – see figure 9 – for precast segmental, span by span construction. Several projects are being developed worldwide with this technology and one was already object of a public governmental tender in Colombia (INVIAS 2014). Still on tender stage, but a mature equipment. The importance of deflections control overcome in precast segmental, span by span construction, allow new limits of construction and allow the development of other creative solutions, as the case of precast segmental, hybrid span by span – cantilever construction recently developed for a tender multi-span viaduct with 100 m span (Muñoz-Rojas 2015). 3.4 Other OPS applications studies – still in research In parallel with previous developments, other R&D projects on OPS technology applications are being developed, namely and for example on the Constructive Process of Arches and Cable Stayed bridges, as per Figure 11. In the case of the application of Constructive Cables in Arch Bridges Erection, by means of fast and safe implementation of stress/release matrix of stress variations it is possible and feasible to implement a good approximation to the optimum level of stresses in each construction stage. Simultaneously it is fast and safe to implement a geometry control plan (Cunha 2014). In the case of the application of Constructive Cables in Cable Stayed Bridges Erection, by means of fast and safe implementation of stays stress during the correspondent segment pouring, 113
Figure 11. Ongoing R&D studies on constructive active cables using OPS technology: Left – active cables for arch bridges erection; Right – active cables for cable stayed deck erection. Table 2. Main features of OPS technology. Mid-span deflection reduction: for unslung MSS, above 90%; for overhead MSS, above 80%; Mid-span deflection reduction for launching gantries (span by span): may be above 50%; Lighter equipment units are achieved (steel quantity may reduce above 30% for large spans); Higher load capacity of equipment units is ensured; Continuous monitoring of the scaffolding structure enables higher safety levels. Ability to automatically predefine deflections makes the equipment more efficient. Much simpler steel connections are achieved (maximum tensions are substantially reduced).
it is possible to significantly reduce the negative moments in the deck during constructive stages. Simultaneously it is easier to implement an adequate geometry control plan (Almeida 2013). 3.5 Summary on organic prestressing main features Through the following years important features were observed (Pacheco et al. 2007). Moreover, the following indirect advantages are achieved: (a) greater versatility of the scaffolding equipment (may be used for different spans with slight changes); (b) easier transportation; (c) easier on site assemblage of the scaffolding equipment and finally (d) reduction of space needs to store equipment. 4 PART 4: HOW LARGE MOVABLE SCAFFOLDING SYSTEMS (LMSS) WITH OPS ARE OVERCOMING FORMER TECHNOLOGICAL CHALLENGES In Part 2, four technological challenges to enable the use of Large Movable Scaffolding Systems (LMSS) for 70–100 m span range, were identified (1) Excessive Deflections, (2) Excessive Weight of Equipment, (3) Doubts on the Stability of large spans former MSS with Wind, and (4) Productivity and Construction Time. In the present Part, solutions to overcome such challenges are described. 114
Figure 12. Mid-span deflection data during a concrete pouring operation in a LMSS strengthened with OPS technology.
4.1 Overcoming excessive deflections in former MSS technology for 70 m to 100 m span, with OPS technology As mentioned before, organic prestressing is nothing more than an active control prestressing system, whose objective is to reduce deformations and/or stresses due to live loading. It is a simple and feasible technology, reflecting a simple concept. To understand its functioning, nothing better than referring to the control algorithm. During the concrete pouring operation, the OPS loading mode is “on”. Every time that a predefined value of mid-span deflection is measured and reached, the controller gives an order to the hydraulic jack (in the organic anchorage) to increase the prestressing level, thus reducing/compensating such mid-span deflection. Such functioning can be better understood in Figure 12 where the mid-span deflection of LMSS strengthened with OPS technology is presented. In this Figure 12 it can be observed both the mid-span deflection (left vertical axis) and the OPS hydraulic jack stroke (right vertical axis) during approximately the last 4 hours of a concrete pouring operation. The maximum mid-span deflection (neglecting vibrations) is about 25 mm. As the span of the bridge (Rio Cabriel, Spain) is 70 m, that give us an approximate mid span deflection of L/3000. In the particular case of underslung MSS strengthened with OPS technology, as described in previous publications (Pacheco et al. 2014) that value may be reduced to L/10.000 or even less. In all MSS and LMSS applications with OPS technology, maximum mid-span deflections observed are always about L/ 2000 and in some cases much less (Pacheco et al. 2011). Meaning that the technological need of mid span deflections bellow L/1000 is clearly overcome. This OPS technology’s characteristic, if not the most, is certainly among the most important OPS features and contributions for the actual state of art in bridge construction. 4.2 Overcoming excessive weight of former MSS technology, with OPS technology A similar exercise as in the previous point may be done to understand the reduction of equipment weight that OPS technology enables. Of course, the static effect of organic prestressing is nothing different from the static effect of conventional prestressing. Meaning that, when OPS is applied, a very important compensation of flexural moments is achieved. In fact, each one of the 3 mentioned statics concepts presented in 3.1, may be used, accordingly, to simultaneously (and indirectly) control the stress levels of the MSS or LMSS main girders. And, if stresses are reduced, the main girder profiles are reduced, and, obviously the weight of the equipment reduces. 115
Figure 13.
MSS weight (with and without OPS)/Span relation – versus – MSS neutral weight.
Several studies published before (Pacheco et al.) allow to conclude that for conventional MSS span ranges (until 70 m span) that weight reduction may be above 30%. And that relative difference increases with the span length. For large spans (70 m to 100 m) that comparison is not possible as conventional MSS for multispans larger than 72 m are not known by the authors. Let us consider previous Figure 13 and represent, in the same picture, a simplified projection of OPS-LMSS weights. Again it is to be clear that this exercise is simplified and no exact values are to be established, but also again it allows clear general and important conclusions. Analyzing Figure 13, it is possible to observe that the point of interception between the “MSS Neutral weight” and the “LMSS weight (with OPS)” lines is clearly above the former 65 m and approximately near from 90 m. It should be understood, that even, for example, for 100 m spans, the cast in situ, span by span, construction is to be still a competitive solution. What results from presented curves is that the constructive method becomes conditioning for the deck design, but that is no different from what is a normal case for any construction method: most probably much less conditioning. Presently there is an application of a LMSS with OPS for 90 m span (presented later). This equipment traveling mass weight is about 1250 ton, which reasonably accomplishes predicted “LMSS (with OPS)” line. This technological former restriction on MSS weights is also overcome. 4.3 Overcoming doubts on wind effect assessment on MSS for large spans To overcome any doubts on wind effect assessment on LMSS, additional research is being developed following previous preliminary studies on critical issues related to wind action on LMSS. It is also important to understand that the wind effects on LMSS – either the static, and/or dynamic ones – are to be evaluated in a probabilistic basis, as the wind forces and their effects on the LMSS structure cannot be formulated on definite mathematical functions of time (Andre et al. 2015; Pacheco et al. 2011). At present, a robust methodology to overcome this issue comprises a set of 6 measures which are to be implemented together: – Previous studies on local (near the site) wind actions are to be obligatory; – The wind velocities are to be carefully chosen by the LMSS designer in order to obtain sufficiently small probabilities of occurrence, ensuring a safe operation of the LMSS (Resende et al. 2015) and without causing excessive operational restrictions; 116
Figure 14.
Simplified planning for a 50 spans viaduct over water, with 6 construction strategies.
– During the Conceptual design of LMSS, measures are to be taken in order to ensure adequate natural frequencies, in particular in maximum cantilever configuration during LMSS launching operation (Pacheco et al. 2015); – To develop the Operational Manual by Versions, upgraded in first launching operations: with incremental operational wind velocities limits; – Continuous monitoring of wind action is to be always applied (already common); – In first types of each LMSS, monitoring of structural response are to be implemented; Present strategy is already being implemented in a 90 m LMSS. This strategy is a robust solution to overcome the previous technological/scientific and mostly normative need on the assessment on wind effects on LMSS.
4.4 Upgrade in construction time and productivity of LMSS technology Recent works on productivity on, span by span, cast in situ, and segmental precast emerging technologies have been recently published, comprising possible peak productivities of 3 km to 4 km/year of deck/equipment unit (Pacheco et al. 2014). 117
Figure 15. Anita Garibaldi Bridge, Santa Catarina, Brazil (Courtesy of Consórcio de Laguna and Enescil).
Additionally these methods, in particular, span by span, cast in place, enable to verify, that, with this technology, the decks erection may begin in a much earlier stage of construction, thus, even if the “Deck” task line in the chronogram is longer, the total construction period may be shorter. In Figure 14 a simplified planning exercise is made for an exemplificative multi-span viaduct with 50 spans over water. 3 different construction Methods are considered: (1) Full Segment; (2) Precast segments, span by span; (3) In situ construction, span by span. For all, 2 different rhythms are also considered, meaning 6 simplified chronograms are presented. Main conclusion is that, under certain circumstances, both cast in situ and precast segmental, span by span, construction methods, may achieve similar levels of productivity, or even in some 118
Figure 16.
LMSS – Viaduct across Hostovsky Creek, Slovakia (courtesy of Jiri Strasky/SHP).
cases higher than other fast construction methods. Surely, this should be analyzed in a case by case basis, but the productivity of mentioned span by span methods is not an obstacle anymore. 5 PART5: MULTI-SPAN DECKS AND LMSS EXPERIENCES WITH OPS TECHNOLOGY In the last few years a growing number of projects were developed comprising equipment with OPS technology. Several of them are multi-span large bridges or viaducts. Some are already concluded as per Figures 15, 16 and 17. The Anita Garibaldi bridge over “Canal das Laranjeiras” in Laguna, Santa Catarina State (Brazil) – see Figure 15 – is a cable stayed concrete bridge with a total length of 2830 m. It is the first cable stayed bridge built in Brazil with plan curvature. The bridge is inserted on BR-101, the main road for land transportation between Brazil and other southern American countries. The existing road was visibly under-dimensioned for present traffic. The need to accelerate bridge opening led to the choice of segmental construction process. A launching gantry with OPS technology was chosen for the construction of the East Viaduct and also the West Viaduct, after a disassembly, transportation and re-assembly process. On both viaducts, a total of 43 spans of 50 m were built with this technology. The Viaduct across Hostovsky, in Slovakia, (Figure 16) is a multi-span viaduct characterized by a truly clean design with a coherent integration in the landscape. 119
Figure 17. Bridge over Corgo river, Portugal, general view, cross section and lateral view (courtesy of LCW, Soares da Costa and FCC).
Figure 18.
Movable scaffolding system for 70 m span (BERD) (courtesy of Jiri Strasky/SHP).
120
Figure 19. High speed railway viaduct (1451 m extension); Turkey; Cast in Situ, span by span, current span 90 m; (A) Lateral view; (B) Cross section; (C) 3D model of 90 m span LMSS (under construction).
The cross-section of the viaduct, comprises a box girder and slab struts, which taken together with the top slab, form a pseudo three-cell box girder. During the construction stage the struts were supported bellow in salient elements in the boxgirder and, in the top, they were supported by prestressing bars. For that reason in the erection of the box girder (with LMSS) it was fundamental to achieve a rigorous geometric control, as such struts were precast elements. Such geometric control was performed by OPS, perfectly achieving the desired accuracy. Mid span deflection was lower than L/2500. The LMSS in figure 16 was the first equipment type with OPS bowstring concept and was simultaneously the first equipment with OPS technology working in the 70 m to 100 m span range – in the lower limit (first application in Rio Cabriel, Spain, with a current span of 70 m). It is to be enhance that the construction of these types of decks in viaducts with several spans with movable scaffolding systems may be a very efficient and competitive constructive method. Recent studies show that this technology has very significant advantages in terms of material consumption, namely in prestressing consumption, when compared with other methods for similar spans (Francisco et al. 2015; Menn 1990). 121
Figure 20.
LMSS (BERD).
Finally, it is verified that such material consumption reduction, may be so relevant, that it may have a not neglectable impact in global sustainability (Pacheco et al. 2009). Bridge over Corgo river, in Vila Real (north of Portugal) – see Figure 16 – is a concrete bridge with a total length of 2796 m, including approach viaducts and the 552 m long cable stayed main bridge. By the time of conclusion, this was the 2nd highest cable stayed bridge in Europe. The East viaduct, with a total length of 1278 m comprising 22 spans with a maximum length of 60 m was built with MSS with OPS technology. In this application, the deck was erected in 3 stages. First the U section, then the box was completed and, at last the wings were erected by a complementary equipment. The maximum deformation observed in stage 2 – which was critical for the U section – was below 5 mm, due to OPS control. This feature was extremely important to avoid concrete cracks in first stage concrete. The same concept adopted before (see Figures 18 and 20) is now adopted for LMSS with 90 m spans. Several projects are being developed worldwide with this system. One equipment, under manufacturing stage is to be applied few months after the present publication, for the construction of 4 viaducts. One of them is shown in Figure 19.
6 ONE CONCLUSION Most probably, OPS technology is still in its infancy, but so far and surely, the OPS technology has already brought new limits to some of the most common methods for multi-span large decks (see Figure 21). In particular, it is now possible to increase in situ, span by span methods span range up to 100 m span. The first application for 90 m span is already ongoing (under construction). Several other projects (still in design stage) are already being developed worldwide within the new span range limits given by the OPS impact. 122
Figure 21. Actual span ranges for the most common constructive methods/bridge types: the OPS impact.
Certainly, there are now less restrictions for decisions makers, and most probably, in several cases, this span range increase will contribute for the adoption of the most rational and competitive solution, when such method is the most adequate. Mainly, this represents another degree of freedom for bridge designers, for constructors and for project owners. Further achievements – too soon to be mentioned as conclusions– only with facts will be (or not) an effective part of the Bridge Engineering state of art. ACKNOWLEDGMENTS The author wishes to deeply thank all the BERD team, both in R&D and in these Projects, specially to Hugo Coelho, Pedro Borges, António Guerra, Teresa Oliveira, André Resende, Diogo Carvalho, Igor Soares, António André and of course Diogo Moura and Sir David Ramos. Truly, they are co-authors of present paper. Not less, not more. The author also wishes to thank his colleague at FEUP, in the discipline of Bridges, Filipe Magalhães: for all his suggestions, comments and so positive team working. Also to FEUP colleagues, through all these years, in particular to his Masters Adão da Fonseca, Raimundo Delgado and of course Joaquim Sarmento. The Author also wishes to thank all the Designers and Constructors whom the author has the privilege to work with and with whom has learn so much. REFERENCES Almeida, P. 2013, Study on Forces in Cable Stayed Bridges during constructive stage, in Portuguese, MSc Thesis, FEUP. André, A., Pacheco, P. and António Adão da Fonseca, 2006, Experimental study of a launching gantry reduced scale model strengthened with organic prestressing, Structural Engineering International, Journal of the IABSE, 16, 49–52. Cunha, J. 2014, Study on Internal Forces and Geometry Control on Arches Bridges erected with constructive Active Cables, MSc Thesis, FEUP. Falcó, X., Aparicio, A.C., Barbat, A.H. and Rodellar J., 1990, Control activo de puentes sometidos a cargas de tráfico, Centro Internacional de Métodos Numéricos en Ingineria, Barcelona, 1990. INVIAS, 2014, Puente Pumarejo, Volumen VIII Estudio y Diseno de Estructuras – http://www.invias.gov.co/ index.php/informacion-institucional/2033-puente-sobre-el-rio-magdalena-en-barranquilla.
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Konstantinidis, D. and Maravas, A., 2003, Egnatia Motorway concrete bridges statistics, Proceedings of the 31st ASECAP Study, Portoroz, Slovenia, May 18–21, pp. 92–109. Lima, B., Ferreira, A., Lopes, F., Pacheco, P., Resende, Lima, B. and Ferreira, A. 2015, Comparative study of prestressing consumptions in 7 different constructive methods for 75 m multi-span box girders, Proceedings of the Multi-Span Large Bridges Conference, Porto 2015, CRC Press. Menn, C., 1990, Prestressed Concrete Bridges, 1990, Birkhauser. Morim, M. 2008, Study on a viaduct with 90 m span, erected by span by span, in situ, construction method, in Portuguese, MSc Thesis, FEUP, 2008. Munoz-Rojas, J.F.C., Fernández, S., Iglesias, C., Resende, A., Coelho, H. and Pacheco, P. 2015, An Innovative system of Span-by-span construction with precast segments for span lengths above a 100m, Porto, Proceedings of Multi-span Large Bridges 2015 International Conference, CRC Press. Pacheco, P. 2008, Key Note Lecture on Movable scaffolding systems strengthened with organic prestressing, Proceedings of International Conference CCC 2008, Porto. Pacheco, P. and Adão da Fonseca, 2002, A. Organic Prestressing, Journal of Structural Engineering, ASCE, pp. 400–405. Pacheco, P., André, A., Borges P. and Oliveira, T. 2010, Automation robustness of scaffolding systems strengthened with organic prestressing, Automation in Construction, Elsevier(5), Vol. 19, No. 1. pp. 1–10. Pacheco, P., Adão da Fonseca, A., Resende, A. and Campos, R. 2009, Sustainability in bridge construction processes; Clean Technologies and Environmental Policy, Springer(4), Volume 12, Issue 1, Page 75–82. Pacheco, P., Coelho, H., Borges, P., Guerra, A. 2011. Technical Challenges of Large Movable Scaffolding Systems, Structural Engineering International (IABSE), Vol. 21, Number. 4. pp. 450–455. Pacheco, P., Coelho, H., Resende, A. and Soares I. 2014. High productivity in bridge construction – the OPS effect. 9th International Conference on Short and Medium Span Bridges, Calgary, Alberta, Canada. Pacheco, P., Guerra, A., Borges, P. and Coelho, H., 2007, A scaffolding system strengthened with organic prestressing – the first of a new generation of structures, in Structural Engineering International, Journal of the International Association for Bridge and Structural Engineering, Vol. 17, Number 4, November 2007, pp. 314–321(8). Resende, A., Coelho, H. and Pacheco, P., 2015, Preliminary Assessment of Wind Actions in large span MSS, Porto, Proceedings of Multi-span Large Bridges 2015 International Conference, CRC Press. Rosignoli, Marco, 2013, Bridge Construction Equipment, Thomas Telford. SEOPAN, 2007, Confederación de la Construcion, 2007, Manual of Self launching scaffoldings, in spanish, 1a ed., CNC, Madrid. Soong, T.T. Active Structural Control: Theory and Practice, Longman Scientific and Technical: New York, 1990. Sousa, P. 2013, Study on geometric control of span by span precast segmental brdges of medium large spans, MSc Thesis, FEUP. Vasques de Carvalho, D., 2008, Study of the application of the prestressing application stage in decks constructed span by span – deformation of the scaffoldings effects, in Portuguese, MSc Thesis, FEUP.
BIBIOGRAPHY AFONSO B., Mobile Equipments for Bridge Construction, in portuguese, MSc Thesis, IST, Lisbon, 2007. CEN, BS EN 13670: Execution of concrete structures, 2009. EUROCODE 1: 2005, Actions on structures – Part 3: Actions induced by cranes and machinery. EUROCODE 1: Actions on structures – Part 3: Actions induced by cranes and machinery, 2005. EURONORM 12811-12: Temporary Works equipment – Part 1: Scaffolds, 2003. Lima, B. & Ferreira, A. 2015. Optimized bridge deck design using a genetic algorithm, Proceedings of the Multi-Span Large Bridges Conference, CRC Press. Mathivat J., The cantilever construction of prestressed concrete bridges, 1st Spanish edition, EDT, S. A., Barcelona, 1980. Pacheco, P. Auto-adjustable prestressing, 2004, PCT Patent, pct/pt2004/011, WO2004/109018, Gazette OMPI. Schlub Peter, 1981, Formwork Launching Girders, IABSE Periodica 4, IABSE SURVEYS s-18/81, 1981. Schlaich, J., Scheef, H., 1982, Concrete Box–girder Bridge. International Association for Bridge and Structural Engineering. Valter, Vásquez, J., Domínguez Santana, B., Viaducto Río Cabriel – Análisis Dinámico Pilas, Report, Number 074.08.P23/IN-005.2, PAVASAL, Valencia, February 2009.
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Experts, Experiences & Landmark projects
Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
Crossing of Bjørnafjorden – Floating bridge B. Villoria, J.B. Wielgosz & S.M. Johannesen The Norwegian Road Public Administration, Norway
ABSTRACT: The aim of the present paper is to provide a description of the ongoing work regarding the feasibility studies that are currently being performed. Different alternatives are being discussed. Environmental loads, constructability of the structure and the risk of ship impact can be listed among the decisive aspects of the design.
1 INTRODUCTION The Norwegian Public Road Administration (NPRA) has initiated one of the most ambitious and ground-breaking large scaled infrastructure programs whose objective is to connect Kristiansand to Trondheim without ferry crossing. The crossing of Bjørnafjorden constitutes one of the most challenging parts of the project. Different solutions are currently being investigated. The present paper deals exclusively with the feasibility study that is being conducted regarding a floating solution whose combined constraints in terms of depth of the seabed and total length would be unprecedented. The NPRA is being assisted by engineering consultancies and research institutions with extensive experience with large bridge projects. The purpose of the feasibility study is to provide the Norwegian authorities with the means to reach a conclusion as to the constructability and cost effectiveness of a floating solution. This paper summarizes the main steps of the analysis that is being performed as well as the difficulties that need to be overcome. The response of the structure to wave loads, the modelling of ship impact loads or the determination of appropriate construction methods can be listed among the central issues that need to be addressed. 2 DESCRIPTION OF THE DIFFERENT ALTERNATIVES 2.1 Concept The Crossing is envisaged as a Multi-Span Structure resting on floating elements referred to as Pontoons. Some of the Pontoons would be anchored to the seabed by means of mooring lines whose properties and arrangement remain to be determined. The main objective of the feasibility study is to develop a floating bridge concept for the crossing of Bjørnafjorden that will remain valid regardless of the final position of the navigation channel. Different alternatives are currently being investigated with regards to the type of high bridge, the position of the navigation channel and the curvature of the bridge. The various alternatives are discussed below. 2.2 Navigation channel located at the South of Bjørnafjorden 2.2.1 Alternative 1A: Curved bridge – wide navigation channel The navigation channel is assumed to be 400 m wide and is placed outside the south island. The bridge deck consists of two box girders and follows a horizontal curve with a radius of 5000 m. 127
Figure 1. Wide navigation channel – South.
The passage over the navigation channel would be supported by means of a cable-stayed whose pylon will be placed on the island. The bridge deck would be fixed in its southern extremity but simply supported by the pylon, allowing the deck to move sideways. 2.2.2 Alternative 2A: Curved bridge – narrow navigation channel This scenario requires the blasting of the island located at the south of Bjørnafjorden. The width of the navigation channel would then be 200 m. The bridge deck would consist of two box girders placed in a horizontal curve with an assumed radius of 5000 m. The end supports are assumed to be hinged. The road elevation gradually increases so as to reach the desire clearance. In the considered alternative the passage over the Channel would be supported by a cable-stayed bridge whose pylon would be positioned ashore. 2.2.3 Alternative 3A: Side anchored straight bridge – narrow navigation channel In this scenario, a side-anchored straight solution is considered. It is assumed that the bridge deck consists of one box girder and that the pylon is positioned ashore. Similarly to Alternative 2A a cable-stayed is envisaged over the navigation channel and its pylon is placed ashore. 2.2.4 Alternative 4A: Side anchored straight bridge – wide navigation channel Like alternative 3A, the bridge is assumed to be straight and side-anchored. Its deck consists of one box girder. The pylon of the cable-stayed high bridge rests on the island allowing for the navigation channel to be placed outside the island. The bridge deck is restrained laterally by the pylon, which entails that the latter has to be designed accordingly. 2.3 Navigation channel located in the middle of Bjørnafjorden 2.3.1 Alternative 1B: Curved bridge – network arch The navigation channel is positioned in the middle of the fjord. The high bridge is assumed to be a network arch and the bridge deck follows a curve similar to Alternative 1A. The bridge deck consists of two box girders connected with transverse elements maintaining a constant spacing of 75 m. The height of the arch is initially set as 80 m and it supports a span of 400 m. The bridge deck is hinged at both ends allowing exclusively rotations about the vertical axis. 2.3.2 Alternative 2B: Straight bridge – network arch Alternative 2B differs from Alternative 1B in that the bridge is assumed to be straight and side anchored. The bridge decks consists one unique girder. The pontoons supporting the network arch will be side-anchored. In addition two pontoons, located at a quarter and three quarters of the total length of the bridge, will be side-anchored. 128
Figure 2. Navigation channel – middle of the fjord.
2.4 Pontoon Geometry It has been determined that the displacements in heave of the pontoons shall be restrained below a certain threshold set as Linfl /350, where Linfl is defined as the length of the influence line regarding heave motions of the pontoons. For combined roll and heave from traffic the freeboard change at the edge of the pontoons is limited to 1.0 m. The roll of the pontoon due to the action of 1 year static wind has been restricted to 0.5 degree. The criteria listed above have been used to determine target values for the stiffness in roll and heave of the pontoons. The final dimensions of the pontoons will be chosen in order to provide the structure with sufficient stiffness. 2.5 Mooring system As described in the above paragraphs, the straight bridge alternatives have to be side-anchored. The introduction of mooring lines into the system induces an additional risk and a certain number of uncertainties. The number of mooring lines as well as their dimensions remains to be further investigated.
3 WAVE AND WIND LOADS 3.1 Methodology At the time of writing, the main focus had been placed on the development of a reliable methodology to capture the response of such a complex solution to wave and wind loads. The present paragraph describes the assumptions and simplifications that have been made. It is possible that the final conclusions alter significantly. The two main software packages that have been used are Novaframe, developed by Aas-Jakobsen (Norway) and Orcaflex, developed by Orcina. The purpose of the different models is to predict the response of the structure under the dynamic environmental loads it is exposed to (static and turbulent wind, current, first and second order waves). Important simplifications have been made in order to facilitate the comparison and the calibration of the key parameters. Preliminary evaluations of wind and wave loads are based on the data collected by the weather stations located in the vicinity of the fjord. The most relevant station is located at open sea, outside Bjørnafjorden. Local wind and waves are then evaluated by extrapolation of the available data. 129
Weather buoys have now been positioned at relevant locations in the fjord and will collect data throughout the entire lifecycle of the project increasing the statistical confidence in the prediction of extreme values. 3.2 Hydrodynamic loads 3.2.1 Characterization of the expected wave loads In order to characterize the hydrodynamic loads the bridge will undergo, it is required to define the following parameters: – – – –
Wave Spectrum. Directional spreading spectrum Spatial variation of the above parameters Current parameters
The chosen wave spectrum could be two-peaked in order to combine the effects of wind and swell components. Second order effects shall as well be accounted for. 3.2.2 Methodology In the model developed in Orcaflex, the bridge deck is modelled as a continuous beam element, with adequate stiffness and mass properties, rigidly connected to the pontoons. The high bridge is also incorporated. The effect of the mooring system is included in the model by means of simplified horizontal springs. The latter simplification, as mentioned in paragraph 3.1, has been made in order to make it possible to compare the findings obtained in Orcaflex with those obtained in Novaframe. In a later stage of the project, a more detailed analysis the mooring lines will be performed. The pontoons are modelled as “Vessels” whose hydrodynamic properties are determined from diffraction and potential theory. Another software, WAMIT, is used to analyze the behavior of the pontoons in the presence of waves. It is based on linear and second-order potential theory. The panel method is used to determine the velocity potential as well as the fluid pressure on the outer surface of the pontoons. Sets of Response Amplitude Operators (RAOs) can then be exported from WAMIT into Orcaflex. 3.3 Wind loads The wind dynamic response of the structure is based on procedures for random variables and stochastic processes. The wind load and the response of the structure can be subdivided into two terms: – A mean wind load and a mean structural response due to constant (or static) wind velocity taken as the mean wind velocity during the assumed stationary period (usually taken as a 10 minutes period). – A fluctuating wind load and a fluctuating structural response characterized by the turbulence intensity of the wind and the standard deviation of the response. The expected extreme structural response rmax during the assumed stationary period can be expressed as follows:
where µr = mean value of structural response(static wind); σ = Standard deviation of response (Fluctuating wind load); and kp = peak factor of the fluctuating structural response. The structural response of the considered structure to the static component of wind can be evaluated independently from its response to dynamic wind. The global response to wind loads can finally be obtained by summation of the static and dynamic components. 130
As a first approach, the guidelines from NS-EN1991-1-4 have been used to evaluate the effects from wind. The design group is currently trying to elaborate a reliable methodology to describe the distribution of the wind along the bridge. 3.4 Modal analysis 3.4.1 Calibration of the hydrodynamic stiffness of the pontoons The restoring coefficient corresponding to roll and pitch movements of the pontoon can be evaluated as follows:
where V = Volume of the considered pontoon; zG = Elevation of the centre of gravity; zB = Elevation of the centre of buoyancy and Awp = Area of the water plane. Equation (2) can be rewritten:
where zG,i refers to the elevation of the centre of gravity of the different elements forming the bridge. For every displacement of the pontoons a new equilibrium needs to be found. It is described by equation (4):
where t refers to the considered time step. In order to take into account the variation of the buoyancy forces acting on the pontoons, their hydrostatic stiffness is defined as follows in Orcaflex:
where C44 is the modified hydrodynamic stiffness used in Orcaflex from which is deducted the weight of the elements connected to the pontoons (the pontoon itself corresponds to element i = 1 in equation (5)). 3.4.2 Findings Initially the bridge was modelled in Novaframe and Orcaflex without high bridge so that the comparison of the eigenmodes obtained in both softwares would be facilitated. The influence of added mass was not accounted for in this preliminary model. The same side-anchoring is assumed in the different models discussed in the present paragraph. The ten first Eigen periods can be found in Table 1. Similar eigen forms and eigen periods are obtained in both programs. In a subsequent stage, the high bridge was incorporated in the model. Table 2 gives the eigen periods of the first ten modes. Table 2 shows that the correspondence between the two softwares in terms of eigen forms and eigen periods remains valid after a high bridge has been added to the models. Finally the added mass generated by the pontoons is added to the models. Table 3 shows the influence of added mass. 131
Table 1. Model without high bridge – no added mass. Mode
1
2
3
4
5
6
7
8
9
10
Eigen Periods [s] Orcaflex Eigen Periods [s] Novaframe
36
35
34.7
32.5
25.7
24.6
20.3
16.8
16.2
15.3
36.6
35.2
34.9
32.6
25.6
24.5
21.4
18.3
16.1
15
Table 2. Model with high bridge – no added mass. Mode
1
2
3
4
5
6
7
8
9
10
Eigen Periods [s] Orcaflex Eigen Periods [s] Novaframe
48.6
35.3
35.1
34.4
26.9
26
17.4
17.2
15.1
14.6
50.1
35.2
34.9
33.7
26.4
25.7
17.4
17.1
15
14.1
Table 3. Model with high bridge – with added mass. Mode
1
2
3
4
5
6
7
8
9
10
Eigen Periods [s] Orcaflex Eigen Periods [s] Novaframe
57.2
43.7
39.1
39
30.1
29.8
22.2
20.5
20.4
16.7
55.7
40.0
39.1
38.3
33.0
30.4
22.4
20.5
18.9
17.5
4 SHIP IMPACT 4.1 Context A ship impact analysis has been initiated. Its purpose is to provide a basis for the determination of potential ship impact scenarios. Probabilities of impact as well as a levels of impact energy are associated to each of the considered impact scenarios. Ship collision and the resulting impact energy could prove dimensioning for the floating bridge. The risk assessment regarding ship collisions has been carried out by SSPA Sweden AB. The results from this analysis will play a decisive role in the choice of the final bridge layout.
4.2 Probabilistic approach The existing ship traffic and past trends were analyzed in order to establish a prognosis of the ship traffic for 2070 in terms of number of ships, size and speed. The prevailing traffic following a north-south route together with the traffic entering the area of the planned crossing represent a risk of collision with the planned bridge. Traffic prediction will greatly influence the choice of design ship. An established American calculation model issued by American bridge design guidelines (AASHTO, 2009) was used as a starting point for Monte Carlo simulations. Collision energy was calculated and systematically simulated based on the design ships and validated against the AASHTO methodology. 132
Figure 3.
(a) USFOS model and (b) Response in Orcaflex.
The output of the risk assessment described above can be summarized by the two following aspects: – Identification of a design ship – Identification of events whose occurrence probability exceeds a probability criterion of 10−4 per year. These scenarios cannot be ignored and further analysis has to be carried out. SSPA uses a probabilistic approach based on Monte Carlo simulations to reflect the real traffic situation. The modelled traffic pattern is based on long term real Automatic Identification System (AIS) recordings from the area and predicted route alterations resulting from the construction of the planned bridge. A large number of impact simulations has been performed, covering a large variety of types of failures. The most frequent root cause for collision accidents is human error, Other causes are technical failure (loss of propulsion or failure in steering system) or a combination of technical and human failure. The risk of collision could be reduced by introducing traffic regulations in terms of speed and category of vessels that would be allowed to enter the fjord area. Energy dissipation mechanisms would also be an option. At this stage, the benefit of risk mitigating measures has not been examined. 4.3 Bridge response The design group has used simplified models to develop a better understanding of the expected response of the floating bridge under a ship impact scenario. An assembly of springs has been defined in USFOS where the contribution of the mooring system, the bridge deck as well as the water plane stiffness and viscous effects are accounted for. The ship impact is modelled by means of an impulse load. The response of the structure is then exported into Orcaflex in the form of time series. The loads undergone by the structure can then be studied. It appears from the first models that have been established (Figure 3a and Figure 3b) that the impact energy is dissipated throughout the entire bridge. 5 CONSTRUCTION STAGE Regardless of the solution the design group will opt for, the construction phase entails a large number of uncertainties. Given the large scale of the structure and its innovative nature, very few projects can be used as references. 133
The structural capacity of the different bridge elements will have to be investigated during the temporary phases. Initially the following assumptions are made: – The pontoons will be produced in Norway at a location yet to be determined. – Bridge elements of 30 to 40 m will be transported to the construction site and welded. – Lifting vessels equipped with dynamic positioning systems will used to lift the bridge deck sections to the desired elevation. Installation aids such as bumpers guides can be considered. Temporary mooring lines could constitute a good supplement to installation vessels. The construction process will have to follow strict weather restrictions throughout the entire process which could result in significant additional costs. 6 CONCLUSIONS It has not yet been determined whether the construction of a floating bridge was a viable solution for the crossing of Bjørnafjorden. The topics discussed in the present paper represent crucial aspects of the feasibility study. A floating bridge concept applied to a crossing of this scale represent an opportunity for the parties involved in the project to further develop their knowledge and understanding of the floating bridge technology through close collaboration between the offshore and bridge construction industry. REFERENCES Statens Vegvesen. 2011. Håndbok V499, Bruprosjektering. SINTEF. Bridge across Bjørnafjorden Meteocean conditions NORSOK STANDARD N-001. 2010. Rev. 7, Juni. DET NORSKE VERITAS. 2010. Offshore Standard DNV-OS-E301, Position Mooring, October. DET NORSKE VERITAS. 2010. Offshore Standard DNV-OS-E302, Offshore Mooring Chain, October. DET NORSKE VERITAS. 2010. Offshore Standard DNV-RP-C205, Environmental Conditions and Environmental loads (October). AASHTO 2010. Guide Specifications and Commentary for Vessel Collision Design of Highway Bridges.
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Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
Rion-Antirion Bridge – Challenging earthquakes E. Joly & P. Moine VINCI Construction Grands Projets, France
A. Pecker Géodynamique et Structure, France
ABSTRACT: The Greek bridge at the edge of the Gulf of Corinth is a vital part of the transEuropean road network, it links the Peloponnesus to the mainland in a high seismicity area; it has been built from 1998 to 2004. Its innovative design, from the isolated shallow foundations to the 2252 m continuous deck, is still to date an inspiration for projects around the world.
1 INTRODUCTION The Rion-Antirion Bridge, now called Charilaos Tripoukis Bridge, is a five-span cabled-stayed bridge; it still remains a remarkable feat of engineering 11 years after it was commissioned in August 2004. The huge technical challenges involved in its construction enables the VINCI Group to make pioneer progress in terms of designing and building such structures. The paper will focus on its specific design features. The description of the very severe site conditions and of the rationales behind the design, will allow understanding how all the various engineering’s capabilities, from the geotechnical aspects to the cable-stays, together with outstanding construction means, have helped realize an exceptional structure. 2 MAIN CHARACTERISTICS The main bridge has a total length of 2252 m, 3 main spans of 560 m, for a total overpass of 2883 m with the access viaducts. The seabed presents fairly steep slopes on each side and a long horizontal plateau at a depth of 60 to 70 m. Towers are 230 m high. The deck is a composite steel-concrete structure, 27.20 m wide, consisting of a concrete slab, 25 to 35 cm thick, connected to twin longitudinal steel I-girders, 2.20 m high, braced every 4 m by transverse cross beams (Fig. 2). It is continuous over its total length of 2252 m, with expansion joints at both ends, and is fully suspended by 8 sets of 23 pairs of cables.
Figure 1.
Bridge elevation.
135
Figure 2.
Deck details.
3 SITE DATA An exceptional combination of severe site conditions: – High water depth, up to 65 m, – Deep soil strata of weak alluviums: No bedrock has been encountered during soil investigations down to a depth of 100 m. Possible liquefaction identified at North shore, – Heavy navigation, with possibility of ship collision on piers (180 000 t at 30 km/h), – Very windy conditions: reference wind 50 m/s (180 km/h) at deck level, – Active tectonic fault along the Gulf of Corinth and as consequences: • Tectonic movements: up to +/−2 m in any direction, at any location • Strong seismic activity (the 2000 years return period earthquake has a magnitude of 7 and a peak ground acceleration of 0.48 g at sea bed level). The conjunction of all these unfavorable conditions has led to unusual conceptual problems.
4 DESIGN CONCEPTS To deal with all these severe unusual conditions, innovative concepts were developed for several key subjects with the leading idea that finally the deck, where users stand, needs to be the most isolated as possible from the soil where the earthquake shaking comes from. The first concept was for the foundation condition. The idea was to have shallow foundations for the pylons, with a possible sliding on the seabed thanks to a specific gravel layer. To avoid soil ruptures, it has been strengthened at each foundation with about 200 steel pipes, 2.0 m diameter, 30 m long, at a mesh 7 m × 7 m. There are moreover many voids in the piers foundation, which, thanks to buyoancy, allow compensating for 50% of the vertical loads. This innovative foundation system behaves hence efficiently as a seismic energy dissipation device. The second concept was to design a continuous deck, longitudinally fully suspended to the cablestays system. This contributes highly to isolate the deck, making it insensitive to any permanent movements of a pier under seismic or tectonic effects. Transversally, some sets of dampers (4 per pylon, capacity in the range of 3500 kN each) and fuses (10000 kN capacity at each pylon) damp efficiently the lateral movements of the deck and avoid pounding with the pylon legs (Fig. 5). 136
Figure 3.
Response spectra.
Finally, the high stiffness of continuous pylon, thanks to their four pylon legs and a composite steel-concrete structure, allows easily withstanding the loads from the ship impacts and guaranteeing the overall stability. Sophisticated analyses tools, ANSYS for the dynamic analyses, and GP3D (in-house software) for the static, construction and geometry analyses have been used to account for all non-linear effects, large displacements, hysteretic and non-reversible behaviors, sliding capacities, cable-stays, etc. and also Soil Structure Interactions.
5 CONSTRUCTION Although construction methods remain those commonly used for offshore structures, the particularly difficult site conditions with high depth water, and the huge dimensions of pylons have needed quite exceptional means. Hence from the soil to the final deck erection, the main steps of the works were: – Dredging and steel pipes driving works to reinforce the soil, and laying of a 8000 m2 gravel bed interface, 3m thick, at each foundation location, This was a major marine operation which necessitated special equipment and procedures. A tension-leg barge was used with active vertical anchorages to reach the required stability. – A “dry dock” at seashore (230 m × 100 m, and 14 m deep) has been arranged to build easily, 2 by 2, the four footings or the four pylons. The Dry dock had an unusual closure system: The first circular footing was built behind the protection of a dyke; but once towed out, the second foundation, for which the erection had already started, was floated to the front place and used as a dock gate (Fig. 7). 137
Figure 4. Typical pylon.
– Footing erections (conic parts) were continued in a “Wet dock” with enough water depth, where the base footings have been towed. – Pylon bases are ballasted and immersed at their final locations, before ending to erect on site remaining elements of the pylons. – Deck erection of precast segments (12 long) from a floating barge by the balanced cantilever method. – Stiching of the cantilevers. 6 BRIDGE IN OPERATION Over these first eleven years of operation, VINCI has also been able to develop unique expertise in terms of monitoring and maintenance. During this period, actually several events, such as the 2008 Achaia-Ilia earthquake in the area, demonstrated the effectiveness of the processes settled and the merits of close collaboration between designers, construction team, partners, and the teams in charge of operation. 138
Figure 5.
Deck dampers/fuse at Pylon.
Figure 6.
Dredging works.
These actions and analyses of the data collected thanks to many sensors during all these events have improved knowledge of bridge behavior. They are also the core of the structural health monitoring system developed by VINCI and greeted enthusiastically by the scientific community which has resulted in development of numerous tools for handling and analyzing events that may have an impact on large structures. 139
Figure 7.
Dry dock.
Figure 8. Wet dock.
The implementation of the Bridge Management system follows: Goal: Monitor seismic and wind events, data processing to help taking quickly the good decision to preserve both users’ safety and bridge integrity. Means: Full set of sensors 140
Figure 9.
Figure 10.
Erection of Pylons on site.
Deck erection.
Automated Data processing with predefined criteria which allow to classify the extent of the event and the decisions to take: – Inspection of the bridge and its equipment, – Closure of the traffic.
7 CONCLUSIONS It has been shown how the specific design features of the Rion-Antirion bridge have made it possible to build a bridge able to withstand all the severe conditions of the site, with huge and innovative construction methods. It has of course needed many steps and sometimes to change of options but this successful adventure is the result of a perfect combination of different expertise from the geotechnical aspects, to the calculation capabilities, the wind engineering or cable-stays technology. 141
Figure 11.
Cantilevers stitching.
Figure 12.
Sensors location.
The bridge, opened in 2004 for the Olympic Games, is still to date a reference and a source of inspiration for many projects all over the world. REFERENCES Combault J, Pecker A, Teyssandier JP & Tourtois JM, 2005. Rion-Antirion Bridge, Greece – Concept, Design and Construction. Structural Engineering International 1/2005, 22–27. Pecker A, 2003. A seismic Foundation Design Process, Lessons learned from two major projects: The Vasco da Gama and the Rion-Antirion Bridges. La Jolla, California. ACI International Conference on Seismic Bridge Design and Retrofit. Teyssandier JP, Combault J & Morand P, 2000. The Rion-Antirion Bridge Design and Construction. Auckland, New Zealand. 12th World Conference on Earthquake Engineering. Teyssandier JP, de Maublanc G, Morand P & Tourtois JM, 2002. The Rion-Antirion Bridge. Osaka. LaTechnique Française du béton. Teyssandier JP, de Maublanc G, Tourtois JM, Pecker A & Morand P, 2004. Le Pont de Rion-Antirion. Revue “Travaux” n◦ 809.
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Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
Innovative erection methods of steel cable-stayed bridges M. de Miranda Studio de Miranda Associati, Milano, Italia – Rio de Janeiro, Brasil
ABSTRACT: This paper introduces three non-conventional construction methods of steel cablestayed bridges designed by the same engineer and author of this paper, over the past few years. Each method results as an adaptation to different design requirements and constraints. – In the first case, the bridge has a 390 m main span girder erected in preassembled segments by a cantilever method, on a river, without the use of barges. – In the second case, has transversally slanted mast and a curved deck erected by pre-assembling of macro segments on the sea, by using a barge as erection yard and means of transport and putting in place. – In the third case the bridge, which also presents a single slanted mast, with a central plane of stay cables, the structure was erected by a combination of longitudinal launchings and rotation, lifting the 75 m mast, without the use of cranes. The project’s presentation will show the methods kinematics as well as the real job site erection operations. A discussion of the three systems, together with a comparison with respect to more conventional systems, closes the report, addressing further developments of construction methods of cable stayed bridges. 1 INTRODUCTION Construction methods of bridges, even if formalized in a finite number, are practically infinite, since they have to adapt to specific sites, available equipment, workers experience and to different bridge design (Miranda, 2004). Cable-Stayed bridges present a great variability in design and, in turn, require great diversity in construction methods. Sometimes, more than one construction method is possible and the choice of the better one is done by considering all present restraints together with the available resources. This paper wants to briefly show, by discussing real cases, three different ways of solving and interpreting the subject of a proper construction method of cable stayed bridge for which construction, for various reasons, was not simple. All the presented bridges were designed, together with their construction method, in the last ten years by the author. In all cases, their design was influenced by the chosen construction method, contained in their conceptual design phase. This condition, of great importance, allowed the work to go on smoothly during the construction of each bridge. 2 BUILDING A LONG SPAN BY PURE CANTILEVER: THE HIGUAMO BRIDGE 2.1 Bridge concept This bridge crosses the Rio Higuamo near San Pedro de Macoris, in the Dominican Republic. The site area is prone to tropical hurricanes with design wind speed of the order of 67 m/s (240 km/h) at deck level. Details can be found in Miranda (2007) & Miranda (2008). A single span across the river was required due to poor soil conditions in the river bed. The most suitable structure was found to be a cable stayed bridge with a main central span 390 m long 143
made by composite steel concrete construction and short side spans made by cast-in-place solid reinforced concrete girders. The design concept was to build a light central span counterweighted by heavy side spans of simple and economical construction; the side spans had short lengths since the deck had a low height from the ground. This layout prevented any risk in the unloading of back cables and also induced small stress ranges in these cables, and good fatigue performance. A composite steel concrete structure for the main span was found to be the most convenient choice, when compared with a totally concrete option, which would have required a higher quantity of stay cables and of counterweight mass. And it was also a more cost effective option when compared to a solution totally made of steel, because this would have required a more expensive deck making it more difficult to stabilize against wind because of its lightness. The central span section was made by a top slab 25 m wide, supported by two lateral plate girders and transverse trussed beams spaced at 4.10 m. Stay cables were spaced 16.40 m. With reference to the choice of a deck cross section, the following procedure was defined. The adopted structural cross section is effective from the point of view of manufacturing, structural weight and cost. Its aerodynamic behavior, however, performs less well than a box streamlined section, and, according to preliminary analysis, would require some improvement in order to ensure the required safety against aerodynamic instability for the design wind speed. As a consequence, in the preliminary design stage, it was decided to adopt λ shaped towers in order to increase the torsional frequency and consequently the critical speed, and to provide lateral triangular fairings in order to improve the aerodynamic behavior. 144
In order to define possible alternative features for the aerodynamic measures to be checked by wind tunnel tests, some different options were analyzed in order to choose the better one. The performed wind tunnel tests confirmed the initial decisions and defined the numerical values of critical speeds. 2.2 Bridge construction The provided construction method was performed according to the following sequence: – In the first place, the building foundations piers and abutments and concrete tower were placed then, starting from the abutments, the building of the side concrete spans by traditional scaffoldings; – The second step, consisted in erecting the steel top parts of the towers, in segments, with a tower crane; – The third step, consisted in installing the steel girder segments with the use of special cranes running on the deck; This operation was carried out by means of special equipment designed by DMA – Studio de Miranda Associati, that was able to lift the steel segment, previously tilted and transported by special trailers just below the gantry, shift it up to the end of the cantilever; then tilt it again, lower it to the final position and finally joint it to the already erected segment by bolted connections; – After first tensioning a pair of cables, the concrete slab was built by placing precast thin slabs, placing reinforcements, and pouring concrete; – At the end of the first curing the second tensioning of stay cables was applied. The construction of the deck progressed according to expectations. After a learning phase, the rate of one segment was completed in a week giving an average progress of around 2.0 m/day. This construction method, conceived in 1999 and realized in 2008, proved to be efficient and simple, and was applied in our recent project for the crossing of Paraná river in Foz do Iguazù between Brazil and Paraguay, were proved to be the most cost effective design solution. 3 BUILDING A CURVED DECK ON THE WATER: THE VENICE HARBOUR BRIDGE 3.1 Bridge description The bridge, crosses the Commercial Harbor of Venice, it presents a curved deck, steel-concrete composite structure, a single layer of stay cables displayed on a conical layout, and a slanted single lateral mast. It was completed in 2007. The bridge details can be found in Miranda (2010). 145
The construction method was based on the erection of the deck’s structure by floating and lifting of large macro-elements. The girder system was designed according to the concept of deck structure protected by a stainless steel enclosure. The crossing has a total length of 466 m, including the cable-stayed bridge and two approach viaducts. As for the other two bridges De Miranda Associates developed, and was responsible for: final design of the bridge, the process of construction, and the construction engineering. Overall, the bridge provides a continuous deck of six spans, each of (42.80 + 105 + 126 + 30 + 42 + 42 = 387.80 m), and 23.70 m wide, set out on a plan to follow a curve of radius 175 m, and a pair of transitional isostatic girders spanning 38.80 and 42.80 m. The structural depth of the deck is 2.15 m. The two largest spans are suspended by cable stays anchored along the central axis of the deck. The cables are anchored to the top of a single-column tower in pre-stressed reinforced concrete, transversally inclined, and structurally independent of the deck. The deck structure is of composite steel-concrete type, with a top slab poured on precast panels and transversely pre-stressed in the pylon area; and at the bottom slab at the bearing points. The steel part consists of a central box beam, two longitudinal perimeter beams, and a series of transversal beams that support the concrete deck slab. A system of horizontal cross-bracing on the underside of these beams, provides stiffness and torsional strength to the deck structure, forming a very wide, torsionally rigid virtual box beam. The tower, surmounted by the steel anchoring structures for the cable stays, is located outside the deck, on the inner side of the curvature, and is inclined towards the outside. Triangular in section, of variable width, it is also eccentrically pre-compressed. Its inclination is such that its central axis coincides with the resultant of the forces in the cable stays, under permanent loading conditions. 3.2 Construction process Clearly, this condition cannot occur during construction, unless all the permanent loads have been imposed. For this reason, the phases of construction were calibrated, to avoid excessive longitudinal and transversal flexing of the tower. In this bridge, in fact, unlike bridges whose structure lies along one or more parallel planes, the balance of forces on the tower must be simultaneously guaranteed in both directions. Maintaining this balance was thus a rather complex task, since this requirement, which in itself is difficult to meet, had to be reconciled with other static and geometric requirements as follows: – maintaining the strength of the box beam and the cable stays, in the various construction phases; – ensuring that the designed internal forces in the deck, the cable stays, and the tower, would be achieved after completion of all the construction phases; – acquisition of the planned design geometry of the deck and the tower, after completion of construction. Construction of the bridge was carried out by means of the following general sequencing. In the first phase all substructures as well as piers and inclined tower were built; this was done during fabrication of the deck’s steel structure. Then, the erection in the canal bed of a pair of temporary steel column supports took place; they were designed to act as intermediate support for the beam assemblies. At this point, due to the lack of any adequate alternative, assembly of the cable-supported spans was carried out on a floating barge, on which two macro-elements were mounted in successive phases, each of maximum length 92 m and weight 720 t. These were curved beam assemblies of high longitudinal slenderness: the thickness of each beam assembly was about 1/48 of the span. The erection in situ of the deck structure occurred by lifting by more than 10 meters the deck macro-segment of 100m length by means of a system of jacks and modular shims up to the final position. 146
The barge was then towed to the location where the beam assembly was to be engaged, where it was lowered so that it was supported and restrained by the temporary column supports, and by the previously erected deck structures. This operation, which was particularly sensitive due to the large size, slenderness, and weight of the components being manoeuvred, their longitudinal and transversal curvatures, the large deformations occurring during the load transfer phases, the influences of the tides, and variations in temperature, was completed successfully on time.then it was lowered to the temporary supports, linked by temporary hinges to the already erected segments and then welded to create continuity. At the end of the steel deck erection, the installation of the cable stays and, in parallel, construction of the concrete deck slabs completed the bridge structure. Removal of the temporary column supports and final tensioning were the completing operations. Within this sequence, the final phases involved a series of operations of particular interest and some complexity but, being all sequence and procedures well defined in a Detailed Method Statement, they were not major problems. In the more easily accessible zones, assembly and placement of the deck structures was carried out on site using temporary column supports and conventional methods. 4 LIFTING A 540 t MAST WITHOUT CRANES: THE CURITIBA BRIDGE 4.1 Bridge summary In the town of Curitiba, State of Paranà, Brasil, a steel cable stayed bridge was recently completed. The bridge has a 225 m length and a 129 m main span, with two side spans of 70 m and 29 m. 147
The deck has steel-concrete composite structure: a multi-cellular box girder forms the main element and a sequence of side transverse beams give support to the concrete slab. Deck width is 23.60 m. All joints of longitudinal structures are welded; joints of transverse internal structures are bolted. Stay cables are all contained in a unique central plane, and are parallel, forming a “harp” shaped configuration, with only three main strong visual directions of: side cables, central cables and mast. The mast, full steel construction, has a height of 75 m and is inclined backwards. It has a box girder structure, trapezoidal cross sections and orthotropic plate type. A bridge description can be found in Miranda (2014). Bridge Design and Construction Engineering were developed also in this case by DMA – De Miranda Associati. 4.2 Erection method The bridge construction had to be executed in a short time span, with a small amount of space available, being it in an urban area and having the traffic free to run around and below the erecting girders. These were the constraints that defined the erection method. The structure erection took place according to the phases described below: – The steel deck was assembled on a temporary embankment created on the first side span, so occupying only a small space outside the bridge foothold path. It was assembled by a gantry crane and, as soon as completed a segment, launched longitudinally by means of long stroke jacks, roller, launching nose, and all typical launching equipment, optimized only for this bridge. The advancement was created with steps for about 40 m. – After the completion of the assembling and launching of the full deck and after lowering it on its supports, the construction of the concrete slab took place. – In the meantime, behind the abutment, the mast was assembled horizontally, in good working conditions, by using the cited gantry crane. Finally, four rollers used to move the deck were dismounted and placed below the Tower, so as to give it a rolling support. The tower was then ready for its longitudinal launching. In this phase a challenging task had to be carried on: How to erect a 540 t mast of a cable-stayed bridge without using cranes. A seventy-five meter high steel tower, slim lined and inclined, whose weight was the equivalent of 400 medium size cars, had to be hoisted to its final position without additional help. The erection of a heavy inclined steel tower can usually be done in two ways: – By erecting it in place, segment by segment, in vertical, using cranes and welding the segments on site, by means of traditional scaffolding. – Or, by welding the tower on the ground in a horizontal position and later lifting it by using one or two large crane sections. 148
But in this case, the first option would have been difficult, expensive and time-consuming; and the second option was not applicable since there was no room for placing such large cranes. An innovative solution was devised: to hoist the tower by rotation using a set of strand-jacks and a special hoisting boom. – The tower was assembled, as said, in a horizontal position behind the abutment, then launched forward, passing above the already built deck and up to engage the bottom hinges installed in the hoisting structure. – Then the auxiliary structures was fitted, i.e. the hoisting boom and a set of longitudinal and transverse bracings and it was lifted by a rotating movement. It was done by using a system of parallel, computer driven, hydraulic jacks designed for the scope. The movement was fluid, without problems, and the operation was completed in around 6 hours. – This operation, even if acrobatic and suggestive, was performed in an unhoped-for smooth manner to everyone’s full satisfaction. – With the tower in place, after the welding of its base, the stay cable installation and tensioning, by symmetric and tuned phase, took place. All construction phases were described in a Detailed Method Statement and all construction equipment was specially designed within the Construction Engineering, and every operation statement and structural piece was ready just on time to start construction. The steel structure assembling and erection had a duration or nine months. The bridge was completed in 18 months and was opened to traffic in April 2014. 149
5 CONCLUSIONS Cable Stayed Bridges can present very different layouts and have to accommodate different site conditions; so their construction, often requiring an innovation effort, can present objective difficulties and can follow very different construction processes. Given this great variability and difficulty, Steel CSB has many players involved during their construction: – – – – –
The Main Contractor, in general directly responsible for the concrete parts of the bridge; The Steel Fabricator, usually, but not always, responsible for the erection of the bridge as well; The Steelwork Erector; The Stay-Cables supplier; The bearings and special devices supplier.
Each of these players, has a specific role in the bridge global design but also have specific financial interest, own ideas, specialists and time schedules, that at times can create small or large conflicts, misunderstandings, delays or claims. A coordination of all these players, taking into account the whole bridge concept and structural behaviour, is a must; and when this coordination can be carried on, with a tight supervision by the bridge designer, it becomes the premise for the success of the project. The case presented in this report, shows a set of completely different erection methods which are: – By total cantilevering of pre-assembled large segments using specially designed equipments, without using barges; – By erection of macro elements with extensive use of barges and jacks; – By multiple longitudinal launchings, with an autogenously lifting of a heavy tower, by rotation, without external aid. The common element of these methods, that mark a difference with more traditional processes, consists on the assembling close or over the bridge in construction of large elements, weighting even hundred of tons, and conceiving the best way of putting them in place. Although there is always need to solve the many problems that always occur during the construction of a bridge, the work in all these cases was fluid and successful. A reason for this, in the opinion of the writer, can be found in the fact that for all these three projects, the erection method was conceived, from the beginning, together with the bridge structural design and detailing, and was all managed by the designer of the bridge. By adopting this approach it was possible to get the integration of all design and construction issues, and avoid reasons for conflicts, creating instead opportunities for synergies that are necessary in order to build bridges in a safe, proper and cost effective way. REFERENCES de Miranda, M. 2004. “Metodi di costruzione di ponti in acciaio” (Steel bridges construction methods), Rivista Strade & Autostrade n. 4/2004. de Miranda, M., 2007. “Puente Mauricio Báez Sobre el río Higuamo, San Pedro de Macorís” – Archivos des Arquitectura Antillana – AAA 027 – Luglio 2007. de Miranda, M., 2008. “Il ponte sul rio Higuamo”, Il Giornale dell’Ingegnere n. 5 – 15 Marzo. de Miranda, M., 2010. “Construction of Cable-Stayed Bridge in the Commercial Porto of Venice, Italy”, Structural Engineering International n. 1/2010. de Miranda, M., 2014. “Um viaduto estaiado em Curitiba”, Congresso Sulamericano de Estruturas metálicas – Construmetal 2014 – Rio de Janeiro 2–4 Settembre.
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Viaduct over river Ulla in the Spanish Atlantic high speed railway line: An outstanding composite steel-concrete truss bridge F. Millanes, L. Matute & M. Ortega IDEAM, S.A., Madrid, Spain
ABSTRACT: Viaduct over the river Ulla constitutes the most important intervention in the High Speed Atlantic Railway Line between A Coruña and Pontevedra, in Galicia. The viaduct is 1620 m long with a span distribution of 50 + 80 + 3 × 120 + 225 + 240 + 225 + 3 × 120 + 80 m, which means that the main span is about 20% longer than the current world record, the Nantenbach bridge over river Main in Germany (208 m long span). The main spans are resolved with a double truss depth-varying deck, with a depth ranging from 8.5 m at the midspan to 17.25 m at the section over the piers. The article describes the structural conception of the bridge and the three different constructive processes used for erecting the deck: erection of the 5 central spans by 4 equilibrated cantilevers, launching of the four lateral spans nears abutment 1 and the complete lifting of the three lateral spans near abutment 2.
1 DESCRIPTION OF THE BRIDGE Viaduct over river Ulla constitutes the boldest and high-profile undertaking in the High Speed Atlantic Railway Line in Galicia, in northwestern Spain. Its location, close to the firth of Ulla, a landscape of outstanding natural beauty and strong environmental constraints, led to the Spanish Railway Administration to hold a competition of ideas among the most renown Spanish structural engineering companies. The alternative presented by IDEAM was finally awarded and this paper describes the project and construction of the Viaduct. The project constraints focused especially on the following aspects (Millanes F. et al. 2008): – The outstanding nature of the project, which required serious consideration of the aesthetic qualities and viaduct integration. – The reduction of the number of piers in the water course, minimizing the impact on the riversides. – The erection procedures, being suitable to the works scale, had to be kept as independent as possible from the inlet’s course, to avoid environmental damage as much as possible. – Visual transparency and minimal bridge interference with the surrounding landscape. All these factors guided the solution to a haunched steel-concrete composite lattice, with double composite action at the hogging zone, three main spans 225 + 240 + 225 meters long, and several 120 m long approach spans. The main span is 20% longer than the current world record, the Nantenbach bridge in Germany, with a main span of 208 m. The resulting viaduct is 1620 m long with a span distribution of 50 + 80 + 3×120 + 225 + 240 + 225 + 3 × 120 + 80 meters (Fig. 1). The structural solution of a steel lattice with double steel-and-concrete composite action adequately solves the previously stated conditions (Millanes F. et al. 2014). The deck was designed as a haunched lattice in the five main spans (Fig. 1), their depth ranging from 9.15 m to 17.90 m, and as a 9.15 m constant-depth lattice in the approach spans. The four central piers, with architectonic shapes (Fig. 1), are rigidly connected to the lattice deck creating composite frames which bestow the required stiffness upon the three central spans 151
Figure 1. View of Viaduct over river Ulla.
Figure 2.
Lateral view of the main span of Viaduct over river Ulla.
in order to withstand the stresses arising from loads acting on alternate spans within the stringent deformation limitations established by Eurocodes for High Speed Railway bridges. The lateral piers P-5 and P-8 (Fig. 2) were designed with a lighter cross-section consisting of two separate concrete shafts embedded in the deck and in the foundation. This allowed to preserve some degree of stiffness against alternate loads as well as the necessary flexibility to allow for temperature and shrinkage imposed displacements. The structure’s design, preserving the structural orthodoxy, placed special care on the integration of shapes and geometry between the concrete piers and the deck’s steel lattice. The smooth depth variation along the deck, with an upward concavity, confers a serene appearance over the Ulla river’s course. The colour choice, pearly grey for concrete and green for the lattice, enhances the effect. Piers distinctly show a double typology. Firstly, the four main piers are rigidly connected to the deck, configuring a frame which increases the structure’s stiffness and improves its behavior regarding horizontal forces. These calyx-shaped piers are formed by a trapezium head 17.5 m high and 11.00 m to 16.80 m wide, and a shaft 8.00 m wide, growing with a 1H:25V slope in piers P6 and P7 (Fig. 2) and a 1H:50V slope in piers P5 and P8 (Fig. 15). The average height of the piers, measured up to the lattices’ lower member, is about 42 m (60 m up to the crowning point). 152
Figure 3. Sagging cross sections (left figure) and hogging cross section in the variable depth central zone (right figure).
The stiffness of these piers was optimized in order to restrain deck rotations at the pier section but preventing that bending moments taken by the pier itself (and then transmitted to the foundations) from becoming a decisive design constraint. In this way, piers P5 and P8 (Fig. 2), at the sides of the 225 m spans, were designed with two detached shafts from base to head, in order to avoid the excessive bending moments arising from two main sources: the disproportion of a 225 m span next to a 120 m span, and those produced by the temperature and shrinkage displacements, larger than in central piers due to their greater distance to the neutral displacement point. Lateral side span piers P1 to P4 and P9 to P11 are of a more conventional design. Their hollowbox cross section, with a 0.30 m thick wall and a 3.50 m × 8.50 m cross-section at the top, varies in depth both transversely and longwise. The pier height ranges from 52 m to less than 20 m. The main spans are designed with a double haunched lattice deck, with a total depth ranging from 9.15 m at the midspan section (Fig. 3 left figure) to 17.90 m at the pier section (Fig. 3 right figure). The lattices, modulated in 15 m long segments, are 6 m apart, measured between the upper flange midpoints, featuring a 1H/17.5V outward slope. The adjacent spans giving access to the haunched main ones were designed with constant depth. Both the upper and lower members’ cross-sections are parallelogram-shaped girders, 0.80 m wide, 1.00 m deep the upper chord and 1.20 m deep the lower one. Diagonal members are also parallelograms (0.8 m wide and 1.00 m deep). The upper member features a box-like head embedded in the concrete slab lodging the connection, allowing a shear transference closer to the center of gravity of the composite upper chord and minimizing the appearance of local forces and moments in the connections. The steel grade is S-355-J2+N and K2+N for the approach spans and S-460-M and ML for the three main spans. The upper slab thickness is 0.46 m at the center line and 0.25 m over the steel upper chords. The slab, made of cast-in-place C35/45 concrete, is poured on precast concrete slabs. 153
Figure 4. Aerial view of the construction of the foundations of the three central piers.
Along the hogging zone, a C50/60 bottom concrete slab is arranged between members, thus allowing for double composite action. The thickness of this lower slab ranges from 0.30 m to 1.10 m. Along the sagging zone, the deck’s lower face is visually closed using thin precast concrete slabs, with no structural role but to create a path to allow for extremely easy inspection and maintenance operations.
2 DESCRIPTION OF THE CONSTRUCTIVE PROCESS The bridge has been built in the period of 2009–2014 by the JV composed by Dragados and Tecsa. The chosen procedure to construct the viaduct has conjugated minimal river affection (always reversible) and erection means suitable to the bridge magnitude. The foundation of the piers P-5, P-6 and P-7 are located on the river (Fig. 4), and they have being built with a huge double enclosing sheet pilling circular wall (the exterior one has a diameter of 68 m and the interior one of 48 m) to allow the drain construction of the piles and the pile cap in P-5 and P-6, or the shallow foundation of the pier P-7. So as to access to the foundation of the 3 piers located on the river, a provisional steel access bridge has been built, supported in temporary driven piles placed each 6 m. The construction of this provisional access (Fig. 4) was carried out respecting the natural course of the river avoiding any possible contamination or affection to the protected local fauna. That simplifies the works with road access from both sides of the river, avoiding the need of boat special resources. Once the foundations have been completed the piers are erected by means of a climbing formwork. When the shaft of the piers P5 to P8 is finished, the zero steel segment (with “W” shape) was assembled in horizontal position at the bottom of the piers, and once both side truss segments are finished, they were lifted and fixed in their position on site over the piers. Each of these huge segments weights around 375 t each, and their dimensions are 35 m length per 17.5 m depth (Fig. 5). Completed the concrete part of the head of the piers, the steel truss of the central spans with variable depth will be simultaneously erected by a successive cantilever method, from the pier section to the closing segment at midspan. This method ensures independent work at the bridge from the marshes, river and surrounding vegetation. The welding of the different elements that constitute one segment: nodes, chords, diagonals and transverse bracings, are done at one of the three steel workshops located near the edges of the river. Once the segment is completely finished, it is transported to the pier base in modules that measure 15 m long by means of a special platform with multiple axes accessing. 154
Figure 5. View of the erection of the zero “W” steel segment over one of the central piers.
Figure 6.
Schematic view of the Cantilever progress in the central spans 5 of variable depth.
Once the segment has arrived to the pier base, a gantry crane picks up the module close to the pier shaft, translating it to its final position and lifting it to be welded in place (Fig. 6). The constant depth spans of both sides are built by different procedures due to the different inferior crossing conditions. The side near the abutment A1 has several local road crossing and a local railway crossing, so the constructive process is by launching in three different phases (Fig. 7). As there is not enough free space behind the bridge to prepare a launching yard as it would be conventional, due to a tunnel very near to the end of the abutment, the launching yard has been established between abutment A1 and pier 2 (50 + 80 m), over temporary props. Each 120 m span are assembled on site over the launching yard by welding the segments over temporary supports, and once finished it was launched. The second launching operation moved two complete spans of 120 + 120 m length, and finally the lateral side spans 1 and 2 (50 + 80 m) are welded on site over the temporary props by erecting each segment with the use of cranes (Fig. 7). The approaching spans on the side of the abutment A2, do not have the same restrictions as the ones on the side of the abutment A1. As there are no inferior interferences, a complete span lifting procedure has been designed. 155
Figure 7. View of the three phases of the constructive process of the lateral approaching spans near A-1.
Figure 8. Sequence of the lifting process of the lateral spans 10, 11 and 12. View of the lifting of span 12 and span 10.
The complete span are welded on site by assembling each segment propped on the ground (Fig. 8), and finally each span will be vertically lifted and welded to the previous one giving continuity to them. Once the assembly of the steelwork has been completed, the lower precast concrete plates were placed and the subsequent lower slab concrete casting done. The upper concrete slab is poured over precast concrete slab.
3 MANUFACTURE OF THE DECK’S METALLIC STRUCTURE The singularity and complexity of the bridge’s metallic structure, as well as the huge amount of structural steel, nearly 20,000 tons, made it necessary to divide the manufacture in 4 groups of workshops, 3 of which located in the north of Spain and 1 in Portugal. In order to handle, manufacture and ship the steel elements to the worksite, the deck’s trusses are broken down into the following simple elements (Fig. 9): upper nodes, upper chords, lower nodes, lower chords, horizontal struts and cross bracings. Once each module’s individual elements are ready, and prior to being shipped to the worksite, they are welded in larger subsets, depending on the case: node + chord or node + chord + node. The subsets and the rest of simple elements (diagonals, horizontal struts and bracings) are shipped and, later on, assembled in the on-site shops, thus making up the modules of the truss. The assembly of individual pieces and pre-welded subsets required setting up remarkable permanent facilities on both banks of the river. It was also necessary to arrange large storage and assembly yards both below the viaduct’s vertical projection between pier P-9 and abutment E-2 and in a nearby expanse with access to the river through a wharf. The outstanding size of the steel modules, 8.75 m high in the constant depth segments and up to 17.5 m high in the haunch spans, made it forceful to set up assembly workshops as large as actual steelwork plants, with over 20 m clearance, nothing to do with temporary facilities. The complexity of the metallic structure has required an important amount of beforehand work in studies and development of a series of very repetitive details in order to do the assembly drawings. 156
Figure 9.
Break-down into simple elements for shop manufacture.
This way, the assembly drawings solve every encounter, welding, transition and specific detail, avoiding future executing problems. The 131 drawings of the project, that define the metallic structure in detail, are developed in more than 5000 workshop drawings and nearly 18000 broken down drawings, defining with absolute accuracy each one of the plates, encounters and weldings of the bridge. This important engineering effort is essential to ensuring the correct design of all details, which have to accomplish very strict requirements related to fatigue resistance due to its being a composite bridge for high speed trains. REFERENCES Millanes F. et al. 2008. Viaduct over the Ulla river in the HSRL Eje Atlántico in Spain. An outstanding structure in the field of Composite Steel-Concrete HSRL Bridges. Proceedings of Eurosteel 2008. Graz, Austria. September 2008. Millanes F. et al. 2014. Viaduct over river Ulla: an outstanding composite (steel and concrete) high speed railway viaduct. Structural Engineering International, Vol. 24, Nr. 1/2014, published by IABSE, pg. 131–136, Feb-2014.
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Juscelino Kubitschek Bridge, Brasília, Brazil F. Botto de Barros & J. de Freitas Simões Projconsult–Engenharia de Projetos, Ltda., Brazil
ABSTRACT: With the overwhelming growth of Brazil’s capital city, Brasilia, the need to expand the outer city and providing the nation’s executive power center (Federal District) with better accesses, had become a major concern to the local authorities. This led to a redistribution of the city’s road network, in an attempt to mitigate the daily heavy traffic developed in the existing Costa e Silva bridge. In addition, the new Juscelino Kubitschek bridge was designed and built.
1 INTRODUCTION Named after one of the most important and representative presidents of the Brazilian republic, the Juscelino Kubitschek bridge rises in the Brasília city’s landscape, over the Paranoá lake, described by engineer Mário Villaverde as “a rock graciously skipping over the surface of a pond”. Notable for its structural excellence, this project by Projconsult Engenharia de Projetos Ltda. was entirely developed in Brazil, since its conceptual phase which integrated the tender, until later stages of the executive project, with detailing of all structural elements, execution methodologies and special equipment needs, such as special proppings, sliding framework, pile blocks’ building systems, etc. With Brasilia’s growth exceeding all estimates, the Brazilian capital was, in 2002, home to 2 million inhabitants. This number largely exceeds the five hundred thousand for which it was initially intended. More than half of this initial number was expected to work in the Federal District (Brazil’s center of the executive power), but to live in the city’s outer perimeter.
Figure 1.
Pilot plan of Brasília.
159
Figure 2.
Presidential Palace (“Palácio do Planalto”).
Figure 3.
Metropolitan Cathedral of Brasília (“Catedral Metropolitana de Brasília”).
This urban expansion forced the city’s authorities to enlarge, adapt and think of new solutions for the city’s road network. This redesign would include a new bridge, which would specially serve the South Individual Habitation Sector (SHIS) and the nearby cities of Paranoá and São Sebastião. The residential sector of Dom Bosco, São Bartolomeu and Jardim Botânico would also benefit from this new access. Studies made throughout the 1990 decade showed a significant growth in the number of vehicles daily crossing the Costa e Silva bridge and it’s accesses (from 32,000 in 1992 to 55,000 in 2000), which generated massive car traffic in all of the accesses to the Federal District. This traffic would later be mitigated with the implementation of a new access. In the Figure 1, the pilot plan of Brasília can be observed, with a red circle over the area where the bridge was implanted. The main goal of the J.K. bridge was to act as one of the main arteries on the fast growing city’s organic road network, establishing itself as a core part of the urban development. It should also integrate the city, not only as means to get into it, but should, as well be a harmonious extension of the aesthetics, environmental and landscaping context of the Federal district’s Pilot Plan. Culturally, the bridge ought to be an architectural and artistic beacon to unveil to all national and international users, the excellence of Brazilian engineering and architecture, therefore contributing to the national pride. In the Figures 2 and 3, some of the arquitectonical symbols of Brasília can be observed. With all these demands, the work shall be a reference in technical terms, meeting all the normative specifications regarding car traffic, urbanism and structural design, while safety and maintenance issues shouldn’t be neglected nor should the cost be too high. 2 THE CONCEPT AND THE PROJECT To ensure the necessary sculptural aspect, required by the public tender, the structure could not simply be composed of small spanned beam decks supported by columns. Also, the minimum height 160
Figure 4. Small spammed beam decks supported by columns (left); under arch bridge concept (center); concept of deck near arch mid-height point (right).
Figure 5. Askew arch (left); askew arch (center); effect of arch inside another arch (right).
for crossing of light ships should be verified. Nor would the light ship crossing item be verified. A big span would have to be a premise. The doubt remained nonetheless, on which system would be the best for the structure, while the options were: straight beams, arches, or three-dimensional trusses. The arches came as an obvious choice, because not only would they allow bigger spans, but would also better integrate the bridge within the city’s aesthetics, filled with various curve shaped structures and other arches. The deck would be placed near an arch at a mid-height point, leaving behind the under arch bridge concept, which would place the deck too high above the water, at a loss of a water mirror experience landscape (Fig. 4). With the diverse types of arch bridges configuration known to date (both sides parallel arches with trusses, mostly used in past century railway bridges, single central arch, with vertical cables, creating a lira-like effect and others) the option was to use a much less used cable stayed arch system, the askew arch. To ensure harmony in-between this asymmetric figure, the askew arch was repeated two more times throughout the bridge to give the driver an arch inside another arch effect. The follow-up arch would start right were its predecessor would end, creating the open tunnel sensation while driving through the plan curved deck, giving the whole system a unique motion (Fig. 5). Special illumination would have to be a big part of this concept for the structure to be as impressive by night as it is by day: direct illumination on the downside of the arches above the deck, on the bases of the arches and upwards on the cables, from their connection to the deck. Both the design and structural concept for the bridge, were idealized by Projconsult’s engineering team, headed by Eng. Mário Villaverde and coordinated by engineers Filemon Botto de Barros and Piotr Slawinsky. During the execution of the basic project, Projconsult had important assessment from the Danish COWI Engineering, a state-of-the art bridge and viaduct design company, and also included Brazil’s foremost personalities in the engineering scenario as consultants, on later stages of the project. Namely, Dirceu Veloso, PhD, Nelson Aoki, PhD, Eng. Carlos Alberto Fragelli and Eng. Ulisses Cordeiro. The project was reviewed and highly complimented by German’s Leonhardt Partners – AG and was executed by a consortium composed of Via Engenharia & Dragados and Usiminas Mecânica. The whole design phases and construction lasted 31 months and cost nearly US$120 Million. 3 THE BRIDGE 3.1 General geometry and building overview This 1200 m long heavy structure has a 24 m-wide deck which provides 6 motorway lanes, a bicycle lane, and lateral sidewalks for pedestrians. 161
Figure 6.
JK bridge.
The structure integrates 2 accesses composed of 10 spans between 45 and 48 m long divided by both banks and three central main spans 240 m long, which are sustained by cable stayed metallic arches. These 3 arches are slightly askew from the deck by 24.6 degrees in plan view, creating a “zigzag” effect, which makes it unique. The deck of the three central 240 m spans has an orthotropic steel box girder section and it was pre-fabricated in the banks and subsequently incrementally pushed to its final location over the V shaped arch-ends’ pillars and over temporary towers raised between these piers throughout the three 240 m long spans. The abutments of the three arches are made of reinforced concrete and are placed side-by-side with the also reinforced concrete 4 laminar v-shaped pillars. The pillars organically follow the shape of the arch abutments until the deck level is reached. After that height is reached, the arches grow in 20 ton box segments with variable cross section, from a trapezoid 6.50 m (widest base) × 6.50 m (height) to a 5.0 m (width) × 3.0 m (height) rectangle section when the top of the arch is reached. Those segments were transported from the fabrication point to the construction site by barges and were placed in their final position with mobile cranes with 400 ton capacity, placed over the pre-launched deck. In the Figure 6, the JK bridge can be observed, after conclusion. 3.2 Foundations The Federal district is located in the geological region of the Planalto Central, characterized by being a combination of layered non-saturated porous clay soils and diverse clayed silts. The multiplicity of results of the geological studies on the lake, determined a number of 7 borings for each of the foundation blocks should be made. The soil was an important decisional factor in the global positioning of the bridge, as three alternatives were examined before choosing the definite plan for the bridge’s axis. The maximum water depth of 22 m is reached at Pillar P6, below one of the arches’ junction, close to the Federal District margin of the Paranoá River. The foundations in which the arches merge are composed by underwater pile header concrete blocks, 26 m wide, 41 m long and 6 m high, from which groups of 88 vertical and battered (1 to 4 slope) Wirth steel cased piles, with 1.20 m diameter, depart, being 30 of those vertical and 58 battered (Fig. 7 – center). To ensure correct foundation conditions, and reach the 23 m deep hard soil level, it was established that the bigger piles should have a 54 m length. The execution of the underwater concrete pile heading blocks followed numerous and meticulous steps before completion. First, the lower slab was cast over the water level, around the piles’ steel cases, with all the formwork supported on them (Fig. 7 – left). Second, the casting of the block’s walls. After this step, several 200 ton jacks lowered the blocks’ shell to the final position acting the shell itself, as a cofferdam for the remaining casting. Holes with a slightly larger diameter than the piles’ were left on the lower slab, to act as guidance for the battered piles that would later be executed, for this jack-assisted lowering would not be possible it they were already in place. It is possible to observe the lowering sequence described in the Figure 8 below. 162
Figure 7. Completed metallic beams with already positioned 200 t jacks (left); foundation blocks, Pillar 6 (center); reinforcement of the block and dowels of the inclined pillars and concrete arches (right).
Figure 8. Lowering sequence of the Pillar P7 concrete caisson, to the final position, 6 meter below the water level (from left to right and top to bottom).
With a 20 m gap between the deck and the average water level, the regular water traffic of the region wasn’t affected. 3.3 Pillars The pillars in the accesses’ spans have a hollow diamond shaped (11.0 × 2.50 m) cross section and their design is mainly focused on vertical loads. With exceptions of the P4 and P9 pillars (the last ones right before the cable stayed decks and arch abutments), in which its core had to be partially filled with concrete to be able to sustain the horizontal force of 120 ton.f (1,176.80 kN) generated by the hydraulic jacks during the deck push to the cable stayed spans. As mentioned above, the V-shaped pillars follow the shape of the arch right off the abutments and their section is a 13.20 × 2.00 m rectangle. In each pile block, the two parts of the V pillar, are connected at deck level by pre-stressed concrete beams that support the horizontal tension effects, therefore closing the force polygon (Fig. 9). During the construction, temporary supports had to be built to ensure correct support to the pushed steel decks, which would be required to correctly accommodate the arches’ proppings. In the Figure 10 below, it is possible to see an overview of the pillars during construction. 3.4 Decks Two types of decks were designed for this bridge. A composite steel/concrete deck for the accesses with smaller spans (Fig. 11 – left) and orthotropic box girder deck for the main 3 spans (Fig. 11 – center). The steel framework of the decks on the accesses were incrementally pushed from each 163
Figure 9. Project design of foundation block B7 and pillars P7A and P7B (left); General view of the pillars P7A and P7B (right).
Figure 10.
Overview of the bridge’s construction.
Figure 11. Construction of the bridge decks: accesses decks (left), main decks (center) and decks being pushed (right).
bank to their final position, and were further completed with a 6 cm precast concrete layer and 26 cm concrete layer cast in loco (Fig. 11 – right). The main span decks, composed of orthotropic steel shell are made of steel plates from 12.5 mm to 16 mm. To prevent local buckling all the plates have internal longitudinal Y shaped stiffeners and there are transverse diaphragms every 4.0 m to provide the necessary torsional stiffness to the deck. The cables connect directly to the decks, through joists every 20 m with 450 ton capable mobile anchorages. Vertical tether anchorages were provided to prevent disconnections from the deck and the pillars, in case of the occurrence of one way only car traffic, generating eccentric loading on the deck. 3.5 Arches and stay cables The steel arches arise from the concrete abutments parallel to the V shaped pillars in 5 m long segments. Each one has internal and perimeter T shaped stiffeners and their weight is limited by the crane operation capability, 35 tons. The walls of these hollow box segments vary from 16 mm 164
Figure 12.
Rising and assembling of the modules of the steel arch.
Figure 13.
Supporting concluded. View of the final step of the assembly of the steel arch.
plates, closer to the abutments, to 12.5 mm closer to the top. The arch’s section varies from a 6.50 × 6.50 m trapezoid right off the abutments to a 5.0 × 3.0 m rectangle section on the top. In the Figure 12, it is possible to observe various phases of the placement of the modules that compose the steel arch. The arches’ special proppings were placed over the orthotropic deck, right above the temporary towers. In the Figure 13, the final step of the steel arch assembly can be observed. Special non-linear calculations were made regarding second order effects due to lateral deformation on the arch, which, when combined with the permanent compression, generate additional lateral flexure on the arch, and additional deformation and deflection, which generate more flexure. This iterative procedure, generally ends when the increment is lower than a pre-established value, for which the method is considered to converge. This calculation method often generates much bigger stresses than the linear analysis, for very long and slender steel structures. To support the decks, each arch accommodates 16 stay cables, being each pair 18 m apart in the arch and 20 m in the deck. To ensure the necessary clearing height for the vehicle crossing, the cables had to be connected to the deck, not just by the deck’s border, but small extensions to the joists had to be provided. 3.6 Accessory devices Several devices were used on this bridge, such as: diverse bearing devices (Fig. 14 – right), with multidirectional purposes, waterproof 200 mm expansion joints (Fig. 14 – left and center), pavement special multilane illumination and atmosphere electrical discharge devices, just to name a few. 165
3.7 Conclusion To undertake this wondrous work, several support building units had to be built to ensure proper support to the two main building sites, in each bank, including: – – – – –
Industrial units had also to be created to ensure quicker production of the steel pile sheets; A 40 m long quay with two 12.5 ton gantries; An exclusive building site to a 150 ton crane capable barge; An 1140 m long and 2 m wide pedestrian service bridge; Two 750 kVA substations and several other generators. The total support facilities area mounted up to an area of 10500 m2 . The following list, includes some of the equipment used in the construction:
– – – – – – –
3 trussed cranes over floating barges with 30, 70 and 120 ton capacities; 4 cranes with 2 ton capacity and 30 m range; 2 WIRTH drills with 1.10 m diameter; A Demag-44 hammer, for the steel sheets pile driving; 4 Putzmeister concrete pumps; 12 pressurized air chambers to pile excavation; A 365 cv bi-motor tug. Quantitatively speaking, the following facts should be underlined:
– 8200 linear meters of pile steel casing were used to execute the 12,170 m of piles – Overall, the volume of concrete used was 38.900 cubic meters and 16.580 tons of steel in which 12580 tons of corrosion resistant SAC 350 were used just for the steel structure, the rest being reinforcing steel bars. In the Figure 15 below, it is possible to observe the sight of the JK bridge at night.
Figure 14. Accessory devices – Expansion joints (left and center) and steel bearings of the arches spans (right).
Figure 15.
JK bridge.
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Bridge over the Cádiz Bay, Spain J. Manterola, A. Martínez, J.A. Navarro, S. Criado, S. Fuente, M.A. Gil, L. Blanco, G. Osborne, M. Escamilla & J.M. Domínguez Carlos Fernández Casado S.L., Spain
ABSTRACT: The bridge over the Cádiz Bay has a total length of 3157 m and crosses from Cádiz City to Puertorreal. The main bridge is a cable stayed solution with a main span of 540 m and approach spans of 200 m and a maximum vertical clearance of 70 m. It will be one of the longest cable-stayed bridges in Europe. The deck is a trapezoidal box girder 3.0 m deep in a composite construction steel concrete in the bridge over the bay and prestressed concrete in the bridge on the land side. The simply supported deck has a variable depth made of steel with an orthotropic deck. The total width varies from 33.20 to 34.20 m. The bridge will support 2 lanes carriageways for vehicles and 2 tracks carriageway for a tram. The pylon is a double Y shape reinforced concrete structure. The bridge is currently under construction. The main span will be built by free cantilever system with 20 m long segments. The approach span on the Cadiz side will be built by incrementally launched segments. The approach spans on the Puertorreal side will be constructed span by span with a centering. The simply supported 150 m span will be lifted from a barge. 1 INTRODUCTION The old aspiration of the city of Cádiz to build a new access that would reach the old town and the Port directly from Puerto Real was fulfill by the authorities of the Spanish Ministry of Public Works and Transport with the launching of the project of the Bridge over the Cádiz Bay. The Bay navigation canal by the Cabezuela – Puerto Real Quay is 400 m wide and 14 m deep. The port authorities enlarged this horizontal clearance up to 540 m to avoid occupying the Cabezuela Quay shore and to provide for easier ship maneuvering. The pier on the Cabezuela Quay side, is placed as far as 70 m into the quay, thus enabling easier loading and unloading operations from the service cranes. As for the vertical clearance the carriageway is placed at a formidable height of 69 m, which makes this bridge one of the world’s highest. However, at the insistence of Navantia whose factories are located within the Bay, the vertical clearance was set at 100 m for a horizontal clearance of 140 m. This circumstance made us design a bascule bridge that was later modified and replaced by a removable span 150 m long allowing for the possibility of the passage of an exceptionally high vessel. The bridge itself can be divided in four separate stretches, depending on their different functional characteristics. – The Approach Viaduct stretch on the Cadiz side corresponds to the access to the main stretch from Cadiz. Length 570.0 m. – The Removable Deck stretch. Length 150 m. – The Main Bridge stretch is the cable-stayed bridge spanning the Navigation Canal and its cablestayed side spans. Length 1,180 m. – The Approach Viaduct stretch on the Puerto Real side corresponds to the access to the main stretch from Puerto Real. Length 1,182 m. The bridge total length amounts to 3,082 m. It is by far the longest bridge in Spain and one of the longest in the world. The main stretch corresponds to the bridge spanning the navigation canal 167
Figure 1.
General view of the bridge and general elevation.
Figure 2.
Main bridge cross section.
and constitutes the bridge’s main reason for being. That is to provide the city of Cadiz with a new access, spanning the navigation canal, the main entrance to the Port, without interrupting the road traffic, despite the great volume of vessel traffic, not as the case of the Carranza bridge that has to be successively opened and closed thus producing the ensuing road traffic interruption. For the clearance needed the present-day technology recommends the use of a cable-stayed bridge. In this case the 540 m main span and each of the two 320 m long side spans hang by 176 stays from two pylons 180 m high. The bridge has been designed in an integral way. The expansion joints have been installed on both abutments and on the removable deck. All the deck is supported over sliding spherical bearings except the removable deck which I supported over elastomeric bearings. 2 THE DECK 2.1 Main span The deck is 34.3 m wide, divided into four 3.5 m wide traffic lanes, two in each direction, plus two tram tracks, and all the other elements needed to ensure the perfect bridge functionality: shoulders, railings, stay-cables sockets as well as windshields to protect traffic from the wind. The structure of the deck must be light-weight, aerodynamic and slender. It is therefore a composite steel concrete structure made of a 3.00 m deep box girder with perfectly rounded edges. The main span will be built by free cantilever system with 20 m long segments that will be assembled on the Cabezuela Quay and will then be floated to the bridge and lifted from a barge 168
Figure 3.
Main span erection.
using crane form travellers placed at the front end of the segments. Once lifted, the segments will we welded to the already built deck and to the cable-staying from the tower. The upper slab will then immediately be reinforced and concreted and the stay cables tensioned. 2.2 Removable deck This is a removable bridge, design to allow for the passage of vessels exceeding the 69 m height, the maximum possible vessel height permitting the passage under the main bridge. This circumstance is unlikely and will happen few times throughout the bridge life. The bridge type chosen in this case is a simply supported one with a variable cross section ranging from 3.0 m, over the supports, to 8.0 m at midspan. 2.3 Approach Viaducts The approach viaduct on the Cadiz side is 570.0 m long with 75 m spans and one 45.00 m end span on the city side. The width of this stretch amounts to 30.5 m and the longitudinal slope is 5%. The design is due to the fundamental idea of the bridge as a whole: a slender, aerodynamic deck, a composite steel-concrete structure. The construction of this viaduct is carried out by incrementally launching the steel structure together with a part of the concrete slab. The approach viaduct on the Puerto Real side has a 5% slope. Its total length is 1,182 m and its width 30.5 m. The whole stretch is made of prestressed concrete. This stretch can be divided into three sub-stretchs. The one on the side of the cable-stayed bridge is made up of three 75 m spans whose cross section is externally identical to that of the approach viaduct on the Cadiz side only in this case it is made of prestressed concrete. The piers are identical to those on the Cadiz side. The second sub-stretch is made up of the following spans: 75.0 + 68.0 + 4 × 62.0 + 54.0 m. The reason behind this second sub-stretch is the presence of axial traffic flow under the bridge and the access to the factories along this thoroughfare, which made it necessary to design portal-frame shaped piers with a 13.5 m clearance between the supports. The shape of all the elements is owed to the general design of the piers. In this case the double trapezoid general shape of the typical pier is divided in two trapezoids, one on each pier of the frame. The height of the piers varies from 13 m to 34 m. The deck is exactly the same as in the already described sub-stretch. The third sub-stretch by the abutment 2 on the Puerto Real side changes. Its typical span is 40 m while the span over the abutment is 32 m due to the fact that the pier height drops radically as the piers get closer to the abutment. The approach viaduct on the Puerto Real side is built span by span, with construction joints placed at each quarter-span, using a centering. 169
Figure 4.
Removable deck elevation, cross section and construction.
Figure 5. Approach viaduct Puerto Real cross section.
3 PYLONS 3.1 Elevation The cross section of the pylon corresponds to two trapezoids that follow the axis of the pylon anits braces, and vary in dimension in the direction of the cross section of the bridge keeping the external line from the lateral view. 3.2 Stay-cables anchoring structure on the tower Initially, the upper vertical part of the towers consisted of a prestressed concrete structure. The concrete of the inner cavity had blisters to hold the stay-cables upper anchorages, with transverse prestressing arranged at each level. However, in order to make the installation and execution of the towers easier while intending to keep the modifications of the outer shape and dimensions to the minimum the use of steel cases was proposed in the vertical mast. The function of these cases would be also part of the resistant cross section of the mast, which structural type would therefore be composite. and the of transverse prestressing could be eliminated. This metal structure would be placed in a first stage of the execution and would later be concreted. 170
Figure 6.
Pylons.
Figure 7.
Stay-cables view and steel cases fnite element model in the upper part of the pylon.
The feasibility of this new typology was studied by means of a tridimensional finite elements model. The tensional state obtained through this model, once the transverse prestressing is removed, proves the convenience of removing the concrete on the tower front faces from which the stay cables exit the tower, slightly increasing the thickness of the metal plates of the cases in this area. The steel cases on the towers are 2.70 m wide (perpendicular to the deck axis) and 5.40 m long (parallel to the bridge axis). A module or a steel case is defined for each level of stay cables on the tower. In order to provide the lodging for the upper anchors on the inside a system of steel metal beams were devised welded to the two parallel loads of the steel cases. The stay-cables load can thus reach the whole structure from the anchorages. These beams have a 2.70 m span and their cross section is made of a double C-shaped profiles. The different strength mechanisms arranged on the towers cannot be studied using a single structural analysis model, which makes it necessary to use different modelling depending on the element to be studied. Thus, the global study of the towers is carried out using a bar model elaborated for the bridge as a whole. This model gives us the distribution of stresses for the ELS and ELU load combinations. The dimensioning of the metal girders within the box girders is carried out using a bar model isolated for each level of stay-cables for the configuration of a simply supported girder, given the fact that these girders are welded only at the extremes to the metal plates of the mast box girders. Once the global safety of the different cross sections of the towers is guaranteed, the need arises to check the tension level of the different structural elements in the areas prone to undergo concentration of local stresses, such as the anchor plates, discontinuities on the metal beams, points 171
Figure 8.
Pier: Cross section and final view.
of union between metal elements and areas of connection between the vertical mast concrete and the steel box girders. To this end, a shell type finite elements model was performed in which all structural elements of the towers are modelled using plane elements with the combined properties of membrane and plates, representing the vertical mast, most particularly the transition area between the mast and the slanted shafts along 25.0 m. Finally, in order to study the local effects likely to appear in the concrete areas near their connection to the steel box girders of the towers, a finite elements model of the vertical mast was developed, in which the concrete is modelled using brick-type elements. This model enabled us to study and integrate local stresses that appear in the area of connection between concrete and steel due to the frame effect in both vertical and transversal planes. These stresses, once properly integrated, have been used to complete the dimensioning of the ELU of the connection as well as to define the reinforcements necessary for the transverse concrete reinforcement, not only in the areas with connecting studs but also on the sides of the composite cross section parallel to the bridge axis. 4 THE PIERS The piers are the same along the whole bridge: the shape varies in the same way in all of them regardless of their height that ranges from 8 m to 52.5 m. The double rhombus shape has a 10.5 m width at the base of the highest pier and 4.2 m at the “waist” which is where the pier head opens to end up regaining the 10.5 m width. The transverse dimension is 4 m at the center and 2.9 m at the edge. The surface is therefore warped. In the case of the cross section of the piers at Puerto Real approaching viaduct, the double trapezoid general shape of the typical pier is divided in two trapezoids, one on each pier of the frame. The height of the piers varies from 13 m to 34 m. 5 CONCLUSION The Bridge over the Cádiz Bay is an outstanding bridge mainly due to: – It follows an integral idea, with the same cross section throughout the remarkable 3088 m length, with the only expansion joints in the 2 abuments and the removable deck, thus holding most of the longitudinal forces in the main towers. – The 540 m cable stayed main span has a back span of 200 +120 m in order to improve its aesthetical appearance in a beautiful landmark as the Cadiz bay, configuring a maximum balanced cantilever of 218.5 m before reaching the first support after the main towers. – The 150 m length simple supported removable deck is a unique solution for a unique situation, again totally integrated in the whole design, allowing sea structures higher than 69 m to cross below the bridge. 172
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Baluarte Bridge executive project G.R. Argüelles Grupo Triada, Mexico City, Mexico
ABSTRACT: The Baluarte Bridge has a total length of 1,120 m, a main span of 520 m, and a height from the bottom of the Baluate River of 402.57 m, for which it was awarded with the Guinness World Records Certificate to the highest cable-stayed bridge in the world. The Project required site’s topographic-hydraulic, geological, geophysical, geotechnical, seismic risk and wind incidence studies, in order to develop the structural and constructive design of the bridge. 1 BACKGROUND The Mexican Communication and Transportation Ministry (Spanish acronym, SCT) is considering finishing and renewing 14 main roadways: longitudinal, North-South direction, from border to border; and transversal, from the Pacific Ocean to the Gulf of Mexico. The missing link to complete the Mazatlan-Matamoros Axis was the Durango-Mazatlan roadway, which has a 230 km length and closes the path between both cities, increasing travel safety and comfort. However, this segment was critical due to its high cost and technical difficulty, as it crosses the Sierra Madre Occidental over its sheerest section, known as “El Espinazo del Diablo” [The Devil’s Backbone]. It is precisely on kilometer 157 + 400 of the highway that the Baluarte Bridge (BB) is located, over the deep river canyon of the same name, that separates the States of Sinaloa and Durango. In 2003, the SCT contracted the professional services to develop the basic studies and the executive project; the latter was followed up during construction, since the beginning in 2007, until its end in 2012. 2 BASIC STUDIES 2.1 Topography Defined by the geometrical project data and the SCT reference terms; a topographical survey was performed measuring 1140 m along the roadway trace and 60 m up and down current. A flight was carried out at a 1:6000 scale, to view the site’s physical characteristics and stereoscopic paired color photographs were obtained. 2.2 Hydraulic and hydrological Semi-empirical methods, associated with a 100-year return period were used to determine the discharge of the river current, using rainfall information from the isohyets drawings for rainfall intensity-duration-return period. A discharge of 525 m3 /s was recommended for design. From the hydraulic section at the site the design water level (DWL) was obtained, associated with a 4.0 m/s speed, which was non-significant to the BB design, given the great height of the project level line over the riverbed. 2.3 Geological It started with the gathering and analysis of the site’s existing information. Then a field geological survey was performed, helped by three site helicopter over flights, due to the difficult accessing 173
to the cliffs. Aerial color photographs were available to perform a photo-geological study of both river margins. The main conclusions included: – Rocks on both river canyon slopes have the required resistance and quality for the structure support. However, the volcanic rocks in the cliffs presented pseudostratification with slight inclination towards the river. – A recommendation was made to modify the bridge trace, away from the North shelf of the Cerro de la Olla on the right margin, to be centered between the faults on the left margin. – Excavations to reach the finish grade for the footing must be performed with explosive control to avoid loosening of the rock near the canyon cliffs. 2.4 Geophysical Consisted on a resistivity and seismic refraction study, in both river margins, focused in the foundation zone for the piers that would support the bridge. Several underlying rock units were identified by the seismic and electrical characteristic properties measured. This information was compared to the geological and geotechnical studies profiles, to improve interpretation and modeling of the rocky strata. 2.5 Geotechnical Direct exploration and sampling works began once the working sites accesses were prepared, for drilling machinery and brigades. Six boreholes were performed on the river’s left margin, two on the main pylon of this margin and one for each remaining support. An additional eight borings were performed on the river’s right margin. Afterwards, another exploration campaign took place, in which nine borings were re-drilled, to reach convenient depths below the footings level grades, once they were determined. The results from the field and laboratory works were integrated together with the other studies, to define the geotechnical characteristics of the subsoil rocky layers and their profile modeling. Geotechnical integration confirmed that the rocky massifs in the margins are formed by competent rocks, adequate for supporting the bridge’s foundations in their final positions. The failure/service limit states were analyzed to determine the permissible load capacity between 45 and 110 t/m2 , and expected settlements not greater than 5 centimeters. Excavations to house the footings were projected with slopes 0:25:1 (horizontal:vertical), until reaching the final grade indicated in the project. 2.6 Wind incidence Wind maximum velocity and turbulence were estimated for service and construction conditions, at a site with rugged and narrow topography. A sustained wind velocity of 130.5 kph was determined and wind gusts of up to 192.6 kph, for 200-year return periods. The following actions were recommended to have a better knowledge of the wind at the site: – Physical measurements with installed instrumentation monitored during construction. – Wind tunnel measurements of scale models of the topographical influences, were carried out in Nantes, France, resulting in recommendations for the installation of wind deflectors in the center of the main cable-stayed span. 2.7 Seismic risk According to this study: – The BB is in the B risk seismic zone of Mexico (moderate seismicity), and foundations will be supported by Type I firm rocky terrain. – Recommended spectrum for the elastic design was provided, corresponding to the failure limit state and service and construction conditions. 174
3 EXECUTIVE PROJECT Once the basic studies were advanced or finished, the structural analysis and design stage proceeded, to generate the memories of the design calculations and executive project drawings. The structural solution better suited for the site’s topographical and geotechnical conditions was chosen, once the bridge’s trace position was optimized. Alternatives studied included: suspension bridge, arch bridge and cable-stayed bridge. The latter was chosen due to cost advantages and construction ease, considering the impossibility of placing intermediate supports in the deep river canyon. The resulting structure was complex, due to its dimensions and characteristics: a bridge with a cable-stayed main span of 520 m, total length of 1124 m and piers of up to 147.3 m in height, with a vertical distance of 402.57 m from the bottom of the river to the bridge deck, the highest in the world for its type. The design was based on the SCT standards and the AASHTO Standard Specifications for Highway Bridges, Edition XVI (1996); also, when applicable, in the AISC, AREA or European Regulations Specifications. For the structural analysis, combinations of gravitational loads (death load + live load), accidental loads: seismic and wind were carried out, in accordance with corresponding regulations, and information resulting from the specific studies. The superstructure, measuring 22.06 m of total width and 16.60 m of road width, was divided in 11 spans of 44, 68, 68, 70, 520, 54, 56, 72, 72, 60, and 40 m. The bridge-deck structural solution is a combination of prestressed concrete beams in the spans 1–2 to 4–5, 6–7 to 11–12 and both ends of clearing 5–6, and metallic beams at the center of the latter, with a length of 410 m. To support the 5–6 main span, a multiple cable-stayed system was used, with a total of 152 cables in two planes, the longest measuring 280 m. The bridge-deck was solved with a mixed superstructure, formed by steel longitudinal beams and cross-sectional bridge pieces spaced 4 m, on which the reinforced concrete road slab was casted. The metallic beams were mounted and attached together by screws and welding in 12 m lengths, supported by the cables to move forward over the canyon from piers 5 and 6, until they were joined at the center of the clearing. Piers 2 to 4 and 7 to 11 and abutment No. 12 consisted of two reinforced concrete pillars casted on site, of rectangular section, joined by horizontal crossbeams of the same material. The pylons for supports 5 and 6 are hollowed concrete, with an inverted “Y” shaped geometrical configuration, thus the cable anchorings in the pylons are joined to provide torsional rigidity to the bridge-deck for wind and seismic forces. Foundation for the abutments, piers and support pylons consisted of reinforced concrete footings of varying dimensions, supported on the rock at convenient depths. 4 FOLLOW-UP DURING CONSTRUCTION During the development of a project of this kind, the contractors that will be in charge of construction are unknown, thus the designer has to make assumptions regarding the procedures and equipment for the works to be carried out: – Installation process and trajectory of the different bridge and superstructure elements. – Weight and location of hoisting devices: fixed and mobile cranes. – Weight and location of auxiliary equipment: compressors, energy plants, etcetera. Given the structure complexity, it was required a high degree of precision and safety conditions for construction, thus leading to confirm or improve the former design criteria and hypothesis: – Bridge and superstructure components real weight and geometry. – Materials measured properties for concrete, structural and cable steel, etc. 175
– – – –
Observed geological conditions in the excavations. Mechanical properties of resistance and deformability of rocks supporting foundations. Thermal variation influence during the construction process. Dominant wind and seismic activity during construction.
Due to the particular characteristics of the bridge, SCT contracted services for advisory and follow-up to the project under construction, which include project modifications, clarifications and requirements during construction. The services rendered were: • Structure – Verification of geometry and constructive procedures for piers, pillars and superstructure, via precision topographical control. – Detailed computer analysis to check and adjust the calculation hypothesis in accordance with the measurements performed. – Defining the camber or countercamber level to be considered at each stage of deck construction. – Determining the initial tension applied when installing each cable, and their final tension, to achieve the project expected theoretical level line. – “As Built” Drawings. • Geology and Geotechnical – Verifying that the exposed rock in the walls and at the bottom of the foundation excavations is within the determined geological and geotechnical basis established in previous studies. – Checking that excavations are performed as specified, with adequate equipment and procedures, so that the rock conditions in the foundations and slopes are not altered. – Checking that foundation settlements are within the project foreseen levels. • Instrumentation – Monitoring of lateral displacements of the top of the slope in the margin of the Durango side, using collimation lines and measurements with high precision topographical equipment. – Monitoring of lateral displacements at depth from the Durango side, using three inclinometers installed down to 70 m of depth. • Wind ocurrence during construction. Winds were monitored via the installation of 12 meteorological stations, in four 40 m height towers. At each tower, three stations were placed at 10, 20, and 40 m heights over the surrounding ground, to record values measured every 30 minutes. Measurements were important, as on October 21 2009, Hurricane Rick lashed over the site, with winds ranging from 80 to 100 kph and wind gusts of up to 149 kph. After analyzing the records, the following was concluded: – Analyzed measurements cover close to a 4-year period, from February 2008 to February 2012. – Rick’s winds analyzed and modeled were determined to have no consequences on the BB design, according to the results of the wind incidence study. – From the turbulence analysis, hurricane wind parameters were determined and the recorded values were found to be within the recommended limits. • Seismic activity during construction. Analysis of the earthquakes that occurred close to the BB site resulted in the following: – 36 earthquakes were recorded with epicenters located at 300 km or less from the site. – Earthquakes had a minimum magnitude of 3.3 and maximum of 5.6 in the Richter Scale. – Neither of the two more important earthquakes that occurred during construction could be detected, nor did they have impacts on structure. – Results allowed for confirmation and validation of the seismic risk study recommendations.
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Figure 1.
Photography of Baluarte Bridge.
Figure 2.
Elevation Baluarte Bridge.
5 TERMINATION After four years from construction start, the last casting of the closing slab for the central span was performed on January 5, 2012. BB construction included that of 17 km of an internal roads net to access the different working sites.
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Queensferry Crossing: Role of concrete in the design and execution of the project Peter Curran Project Director, Forth Crossing Design JV, Ramboll, London, UK
ABSTRACT: Once constructed the Queensferry Crossing will be the longest spanning composite deck cable stayed bridge in the world. With its unique crossing stays form and record length it will be assured a place among the globe’s most notable bridge structures. The different structural forms and options studied during the tender process for the various elements of the bridge are outlined and in particular how future maintenance considerations influenced the selected design solutions is discussed. The normal way of ensuring the design service life for concrete structures in severe environment is by controlling the binder composition of the concrete, the maximum w/c ratio, the minimum cover thickness and applying crack width limitations. The Employer’s Requirements demand for minimum 120 years of intended working life for all permanent structures. Therefore, an extensive investigation has been carried out by the Contractor Joint Venture and the Design team in order to ensure this. The paper describes the durability requirements and the means by which these are fulfilled. The main focus is placed on the durability against chloride ingress and corrosion in the splash zone of the towers. The project’s progress during construction will be briefly presented along with how maintenance decisions were handled for this Design/Build project. Keywords:
Design, design/build, construction, cable stay bridge, maintenance, service life design.
1 INTRODUCTION The Queensferry Crossing is a new multi-main span cable stay bridge located across the Firth of Forth just west of the city of Edinburgh in Scotland. Transport Scotland is the Client and is administering the project, and it is the biggest transport infrastructure project in Scotland for a generation. The Queensferry Crossing is vital to the economy of Scotland and is delivering significant economic benefits with employment on site and additional benefits through the supply chain for Scottish businesses. When completed it will provide essential future links across the important cross-Forth transport corridor in the east of Scotland. It will be the longest bridge of its type in the world and incorporates many engineering innovations which have contributed to its efficiency. The Client’s procurement process delivered a successful Design and Build tender within the budget and schedule expectations. The Specimen Design developed by the Client provided a high level of detail that resulted in cost certainty for the Client during concept and tender stage and also allowed the potential contractors to develop early cost comparisons during the tender. The Forth Crossing Bridge Constructors (FCBC) Joint Venture is a consortium made up of: Dragados (Spain), Hochtief (Germany), American Bridge International (USA) & Morrison Construction (Scotland). It was prequalified to participate in a detailed and in depth dialogue process with the Client during which time options were studied for final inclusion in the tender outline design. The options studied during the tender process will be outlined along with future maintenance considerations that influenced the option selection at an early stage of the project. 179
Figure 1. A view of progress, looking north. February 2015.
FCBC made a successful bid for the design and build contract of £790 million at 2011 prices which was a significant saving on the initial estimated cost range of £0.9 billion to £1.2 billon. FCBC was awarded the contract to execute the project for the Client in April of 2011. The Contract requires a five year defects notification period which is longer than many standard “warranty” periods for construction projects. This quasi mini-maintenance period forces the contractor to consider and assess maintenance and operations risks more than typically. The Queensferry Crossing is a major construction project requiring resources from all over the globe to complete on time and within budget. Both unique and adaptations of standard construction techniques have been developed to deliver the project successfully. The project’s progress during construction will be briefly documented within the paper and presentation. The current status, as at February 2015 is illustrated in Figure 1. 2 DIALOGUE PROCESS, TENDER DESIGN AND FINAL TENDER In addition to the Client’s Employer’s Requirements, a Specimen Design was provided as a reference to the contractor. This offered a very good basis to start analysing the costs of the project. FCBC’s approach toward the tender was to put preliminary prices to the different construction options presented within this Specimen Design while also studying the Employer’s Requirements for potential areas of refinement and optimization. At a very early stage of the project FCBC met with their design and checking team to discuss the impacts of different solutions in concentrated workshops. After agreement was reached within FCBC, design and check teams, the design team provided further information regarding details and quantities for FCBC to perform better cost estimates on the viable solutions. If any alternatives were identified that appeared to be economically advantageous but differed significantly from the Specimen Design solution, FCBC approached the Client to discuss with them their viability as part of the dialogue process. FCBC implemented many revisions to the Client’s Specimen Design within the Tender Design ranging from approach pier shapes to network connection structure optimizations however two optimizations significantly affected the tender. The Specimen Design primarily adopted a precast pile cap solution to all marine foundations except for the Central Tower which utilized precast shell segments making up a direct footing. 180
Figure 2.
Cross section, typical composite box segment (Tender Design).
This well-developed reference was compared to in-situ poured pile cap solutions with optimized geometry as well as a completely different construction and design approach dealing with spread footings established by sunken caissons or sheet pile cofferdams. The approach adopted in the Tender Design was that of caisson and sheet pile cofferdam spread footing solutions. The chosen solution contributed several benefits to the tender including reduced programme, reduced direct costs and increased operational control by eliminating the need for heavy specialty equipment and subcontractors. Outside of these construction benefits, the spread footings realized improved soil-structure interaction and saw a positive impact from the increased loads from a composite deck superstructure when considering ship impact loads. Due to its unique design and construction technique, the solution was discussed heavily with the Client to determine if it was a viable solution that satisfied all of the Employer’s Requirements. In addition to the substructure, FCBC focused on optimizing the main crossing superstructure. FCBC has strong experience with both full orthotropic box and composite box cross sections, so exhaustive studies were undertaken for both options presented within the Specimen Design and allowed by the Employer’s Requirements. After intense study and discussion, the team came up with a similar conclusion to the Client’s Specimen Design study that the cost differential was neutral although there were different identified risks and opportunities with each option. An early option presented by the design team was a variant on the composite cross section that minimized the steel in the nose area. This was accomplished by employing a steeper web angle and developing a cantilevered edge similar to the south approach viaduct. Preliminary design quantities showed a potential saving on the structural steel of around 10% compared to the Specimen Design composite solution shown in Figure 2. While not directly compensated in the tender evaluation, the Client recognized that the selected composite deck reduced the need for monitoring of the orthotropic deck steel work and asphalt surfacing which are both known historically as maintenance problems due to fatigue and surfacing delamination. In addition, any reduction in steel surface area would result in eventual reductions in carbon foot print and maintenance costs for painting. 3 DETAILED DESIGN The operation and maintenance requirements for the bridge are a key consideration within the detailed design of the crossing and form an integral part of the design process. Some of the key aspects are described below. 3.1 Performance based design The design philosophy applied to the bridge was based on the Eurocode system EN 1992-1-1 (CEN, 2008) and the concrete technology is based on EN 206-1 (CEN, 2002) together with the British national standards BS 8500 series (BSI, 2006). Furthermore, Transport Scotland specifications for durable highway structures (TS IA 32, May 2007) apply. The Employer has required a design 181
service life (=intended working life) of minimum 120 years for all permanent components of the structure. However, the specifications mentioned above normally only cover up to 100 years of service life. Hence, a certain extrapolation is needed in order to comply with the Employer’s Requirements. Early in the project it was decided to apply slag cement in the concrete mix design for various reasons. First of all for durability reasons but also in order to avoid cooling pipes during execution, casting and curing of massive concrete sections, particularly within the foundations. The concrete used in the towers and the bridge deck, being the most severely exposed structures, is strength class C55/67 in order to obtain sufficient early age strength for climbing of formwork, post tensioning, etc. For other parts of the structure, including the foundations, concrete C32/37 is used in mass concrete for less severely exposed parts. The towers are directly founded on large spread footings. The North and South flanking towers are designed as cylindrical steel caissons, backfilled with underwater mass concrete with a reinforced concrete spread footing resting on top. At the Central Tower a rectangular cofferdam was created within which a reinforced concrete spread footing was constructed. The outer skins of the caissons were designed to be sacrificial over the lifetime of the structure. The three concrete towers are made from high performance concrete C55/67 and are heavily reinforced. The towers are being cast within climbing formwork in a 4 m casting height per lift. A total of 54 lifts are detailed for each tower. The towers are 14 m × 16 m at thier base reducing to 7.5 m × 5 m at the top. The wall thickness is typically 1.6 m at the base reducing to 0.87 m at the top, but locally 2.4 m in thickness at the ‘power joint’ providing a monolithic connection between tower and deck about which the bridge articulates. The environmental exposure classes for the towers are based on the concrete standards (EN 206-1 (CEN, 2002)), (BS 8500 (BSI, 2006)). The most severe environmental loads on the concrete are chlorides from the seawater in which the bridge stands, combined with freeze-thaw cycles. For the sections of the towers close to the roadway de-icing salt exposure is also of high importance. Comparing the Eurocode for structural concrete and EN 206-1 together with the British national rules it is clear that only up to 100 years service life is generally covered. Hence, the durability design for the bridge structure needs special attention and to be documented in order to demonstrate the required performance. The durability design employed by FCBC and the Design team was based on the following items: – An assessment of the national requirements normally applied to obtain 100 years service life was carried out. The purpose was to quantify the possible combinations of cement and binder, w/c ratios and cover thicknesses. – Numerical calculations of chloride ingress from seawater exposure were carried out to demonstrate the cover thickness requirements in order to obtain minimum 120 years service life. This exercise determines the minimum requirements for concrete quality and cover thickness. Numerical calculations were conducted only for the most severe environmental class XS3. For other exposure classes simplified calculations were made. – The material requirements in terms of chloride diffusion coefficient were documented by laboratory tests on the specific mix design to verify that the numerical calculation assumptions were fulfilled. This initial work was documented in a Service Life Verification Report (SLVR). The numerical calculations of service life and chloride ingress were based on a well-documented method reported in several scientific reports on the subject (Bamforth, 2004), (fib TG 5.6, 2006), (fib, 2012). The durability principles for the concrete are mainly based on passive measures where the material and the execution are chosen particularly for high performance rather than relying on active measures such as cathodic protection, etc. In the severely exposed splash zone of the towers the durability requirements need both calculations and laboratory testing. Furthermore, in these severely exposed locations where inspections are difficult to carry out stainless steel reinforcement is employed in the cover zone. 182
3.2 Instrumentation and monitoring A Structural Health Monitoring System (SHMS) will provide a complete real-time event driven system to monitor the variations of the following categories of loads, characteristics and responses of the Main Crossing. In general the system consists of a sensory system based on commercially available sensors and a data acquisition system which is similarly based on commercially available software, albeit made bespoke for the project. A cabling network system is provided which is based on high performance fibre optic cabling. Additionally a structural health data management system and a bridge response characteristics system both custom built for the project will be provided. 3.3 Protective measures The design of the structure integrates both active and passive systems for its protection, including for example, a bespoke lightning protection system (LPS). Lightning protection has been designed for the structure in accordance with BS EN 62305 which was issued in September 2006. Application of the risk based approach detailed in this standard does require a degree of interpretation for a cable supported bridge structure and a degree of pragmatism in particular with regards to the protection of the towers and cable stays. As part of the background research LPS provision on existing cable supported structures was assessed and found generally to be fairly minimal and although there were few recorded instances of any damage occurring there were some reports, in particular from Sunshine Skyway in Florida where concrete was reported as having been spalled from the top of the tower and on the RionAntirion Bridge in Greece where, as a result of a lightning strike and subsequent fire, a cable failed and collapsed onto the deck. In response to the Rion incident, which was exceptional, there have been a number of bridges when owners have endeavoured to protect the upper most cable with an auxiliary cable placed above the upper most stay which acts as a combined air termination and down conductor. Such an approach was considered but discounted because of the minimal risk of damage, the limited protection offered by such a solution, the impact on the cable climbing cradle and the additional maintenance burden which would be generated. The new British Standard also advises that an air termination network should typically be provided over the top 20% of a tower or mast greater than 60m tall. An air termination would ordinarily comprise a network of copper or stainless steel straps; the provision of such a layout over the top 20% of the towers would be unprecedented and a series of risk assessments were carried out to reduce this to a provision better aligned to international practice. 4 CONSTRUCTION PROGRESS With the Client’s dialogue period being a considerably lengthy period of time, the majority of the FCBC’s team was well developed and established when the contract was awarded in April of 2011. This gave FCBC the advantage of minimizing the startup planning and mobilization duration. The commencement of the work was able to start two months after signing the contract. The temporary works planning and design of the caisson foundations occured concurently with the Detailed Design which allowed for a fast track fabrication process. The fabrication of the caissons was peformed at a steel shipyard in Poland. Figure 3 shows the first shipment arriving on site in May of 2012 less than a year after the commencement of works. Preparatory excavations and dredging operations occured during the begining of 2012 allowing all of the caissons to be placed on the seabed floor. FCBC utilized a custom designed and built sheerleg crane to offload the caissons from semi-submersible ships and place them in their planned position for excavation. The complete excavation of the three deep water caissons to rock level has been successfully completed. 183
Figure 3.
Caissons arriving at site.
A large quantity of structural steel for the project has been supplied from China. The structural steel for the approaches to the cable stay bridge has been supplied from a UK fabricator. Both of these key suppliers as well as the Chinese fabricator of the tower anchor boxes were assigned during 2012. The single box cross section of the cable stay bridge started fabrication during september of 2012 with the final shipment to site expected in the summer of 2015. Clear and very visible progress has been made on the project throughout 2014 with the towers climbing and intial starter segment falsework being erected. This is set to continue in 2015 with the deck cantilevering process due to commence. The 6000 tonne steel elements of the approach viaduct have been fully assembled on site. Currently the south approach viaduct steelwork has been incrementally launched 450 metres with one final push launch required to take it to its final full length of 543 metres. The concrete deck will be added during 2015. With steelwork complete in the South Approach Viaduct the focus of construction has now shifted to the north shore. Steel for the North Approach Viaduct is being assembled on site into a single continuous section, before it is pushed out across the two north approach viaduct piers to become an integrated part of the deck fan from Queensferry Crossing’s north tower. FCBC will open the Queensferry Crossing to traffic in the year 2016. After bridge opening there is a series of traffic phases that finalize the tie in to the existing road network on the north and south ends of the project. Final turnover is marked by completion of the landscape planting. 5 CONCLUSION The Queensferry Crossing is a unique multi span cable stayed bridge and when complete it will be the longest composite bridge deck in the world. The chosen design solution results from a detailed tender process in which maintenance and operations are key considerations by the contractor, designer, and client. The detailed design utilises performance based design principles to achieve the required service life of bridge components and results in state of the art corrosion protection measures, including 184
Figure 4.
Starter segment falsework being assembled.
extensive dehumidification and lightning protection systems. In addition to active systems, inspection and maintenance strategies utilise robust access facilities and computational structural health monitoring systems, both of which help maintain the bridge well into the future. The construction process is well into the penultimate year and will be open to traffic in 2016, after which, the contract uniquely requires that a five year defects notification period begins. The notification period, while advantageous to developing robust solutions with consideration of maintenance and operations, poses procurement risks to the contractor and client which have been given careful consideration by both parties. REFERENCES Bamforth, P., 2004. Enhancing Reinforced Concrete Durability, The Concrete Society, London, TR-61, Concrete Society Technical Report No. 61. BS 8500, (BSI, 2006). Concrete – Complementary British Standard to BS EN 206-1. EN 1992-1-1, (CEN, 2008). Eurocode 2: Design of concrete structures. EN 206-1, (CEN, 2002). Concrete – Part 1: Specification, performance, production and conformity. fib TG 5.6, 2006. International Federation for Structural Concrete, Lausanne, Switzerland, ‘Model code for service life design’, fib Bulletin 34. fib, 2012. Model Code 2010 Volume 2. International Federation for Structural Concrete, Lausanne, Switzerland. fib Bulletin 66. Møller, P., 2012. Service Life Verification Report, Concrete, FCBC Construction JV. TS IA 32, 2007. Transport Scotland, Implementation of BS 8500-1, TS Interim Amendment no 23, May.
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Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
New Pumarejo Bridge over the river Magdalena in Barranquilla, Colombia J. Manterola, J. Muñoz-Rojas, S. Fernández, J.A. Navarro & S. Fuente Carlos Fernández Casado S.L., Spain
ABSTRACT: The overall length of the new bridge over the River Magdalena amounts to 4 km, distributed between the 1,865 m of the work’s main trunk and the 1,920 m of the different branches on the left riverbank. The main feature is the 830 m long cable-stayed bridge with a 380 m central span, required to comply with the new navigation requisites. The entire deck is made of prestressed precast concrete sections. With the aim of unifying the construction solutions for both the main bridge and approach viaducts as well as the secondary branches a precast segmental solution was chosen. 1 INTRODUCTION The new bridge over the Magdalena River is to replace the bridge designed by Eng. Riccardo Morandi in 1972. The purpose of the new bridge is to improve the river navigation conditions and allow the increase of the traffic flow at the crossing. This project is one of the priority actions undertaken by the INVIAS, the Road Department of the Ministry of Public Works of Colombia. The new navigation conditions require a channel over 300 m wide with no piers within, whereas the need for the passage of vessels makes it necessary to raise the alignment of the new bridge to up to 45 m in height. On the other hand, the expected, medium-term increase in vehicle traffic requires an extremely wide platform, able to accommodate 6 traffic lanes. Different areas may be distinguished along the work: – The 830 m long approach viaduct on the right riverbank up to the main navigation channel. – The main bridge crossing the navigation cannel, spanned by a 830 m long structure with a 380 m long central span. – The access viaduct on the left riverbank with three, 70 m spans, the last of which serves as the transition element connecting the different branches. – Finally, four branches that enable diverse links of the vehicle and pedestrian traffic with the city roads. As for the longitudinal configuration of the civil work and in spite of its great length, the deck is continuous from one abutment to the other and is fixed exclusively at the pylons of the cable-stayed bridge, liberating the remaining piers by using pot-type bearings.
Figure 1.
Elevation view of the approach spans on the right riverside.
187
Figure 2.
Functional and structural distribution of the deck.
The entire deck is made of prestressed concrete. In order to unify the construction procedures, in both the main trunk and the branches, the construction solution proposed was a prefabricated one applying match-cast segments in most of the work. 2 APPROACH VIADUCTS The approach viaduct on the right riverside is 830 m long, and is composed of thirteen spans whose distribution is 12 × 70 + 55 m. It has a curved plan with a 461 m radius. The deck width ranges from 35.10 m at the abutments to the 38.10 m at the connection with the cable-stayed area needed to accommodate double carriageways with three lanes each and footpaths on either edges. The deck cross section is a double-cell box precast girder that laterally extends into cantilevers supported over rectangular, concrete struts every 5 m executed in a second stage. Its depth is 3.65 m, the lower slab width amounts to 12.0 m, and its outer webs are slightly inclined in order to exceed 16.20 m in width. The variable width of the cross section is thus accommodated through these cantilevers, whereas the central trunk remains the same along the entire bridge. The approach viaduct on the left riverbank is composed of another three, 70 m long spans. The two first ones are identical to those of the other approach, and the third one is a special span with a variable plan that gradually widens in order to link up with the four branches into which the platform is divided and connect with the city road network. The piers are vertical elements made of reinforced concrete with a void, hexagonal cross section. Their longitudinal dimension is constant, 2.50 m, whereas transversely they are variable, widening at the head so as to hold the bearings within. 3 MAIN BRIDGE The solutions proposed for the crossing of the river’s main branch, situated next to the left margin and where the 300 m navigation channel lies is a symmetrical cable-stayed bridge with two towers and a 380 m central span, whose highest point above the water surface reaches 45 m. It is 830 m de long, and is composed of five spans whose distribution is 70-155-380-155-70, the three central ones hanging from as many central, cable-staying towers by means of double stay cables anchored in the cross section axis. The cable-staying configuration is a classical semi-harp with 40 m windows without cables in areas near the towers where the deck is rigidly supported. Intermediate piers are not arranged on the compensation span, there is only one pier at the extreme end. The proportion of this span in relation to the central one (0.40 of the main span) is the guarantee of an adequate performance of the cable-staying system as well as of the bending moments in the deck and the pylon under asymmetrical loads. In order to avoid high concentrated forces in a single stay, the retaining force is not provided by means of a single stay but it is distributed in the last four cables that are anchored on either side of the anchor pier, at the same time enabling a symmetrical configuration of the cable-staying with a better formal result. The vertical up-lift forces produced by the live loads acting on the main span are countered by concrete fillings within the box girder. The pylon and the deck have a stiff connection; that way the longitudinal action (earthquakes, braking actions, …) of the entire bridge are resisted by these. 188
Figure 3.
Elevation and general view of the cable-stayed stretch.
Figure 4.
Functional and structural cross sections of the cable-stayed stretch.
Figure 5.
Stay cables’ anchorages in the tower.
The deck cross section is practically identical to that of the approach spans, with a constant, 38.10 m width, increased from 35.10 m in order to make room for the cable-staying tower. The stay cables, anchored every 10 m in the bridge axis, are made up of 0.6 strands with a triple level protection: (wax filling, galvanized and high density polyethylene (HDPE) coating), with units ranging from 35 to 58 chords. The cables are anchored in the axis of the cross section solid triangular anchor blocks located above the intersections on either side of the central web. Transverse diaphragms are installed at these sections in order to transfer the loads from the outer webs to the central stays. Cables are anchored inside the pylon by means of steel box section structures connected to the pylon concrete walls. The anchorages are supported on transverse beams fixed to the steel box sections, that are inclined according to the angle of each stay cable. The vertical cable-staying towers are placed on the median strip while the cable-staying is arranged along the centre of the deck. Their overall height is 130.15 m and they are made up of a 40.95 m pier below the deck and a 89.20 m high mast above the deck. Such a disposition produces a very sleek and elegant bridge compelling us to design a deck which has torsional stiffness in order to counteract the effect of the eccentric load that cannot be countered by the cable-staying. The pier geometry below the deck is variable, similar to that of the piers on the approach viaducts. 4 BRANCHES The roads on the left margin are connected with the bridge through four branches with a curved plan layout. Three of these are intended for vehicle traffic (Barranquilla-Santa, Santa MartaBarranquilla, Santa Marta-Puerto) and one for pedestrian traffic. 189
Figure 6.
Elevation views and cross sections of the cable-staying tower.
Figure 7. Typical cross sections of the deck on different branches.
The smaller height of the deck above ground level in civil works such as this one enables their modulation with shorter spans. The typical span adopted in this case was a 40 m long one. The number of spans is variable depending on the branch. These are connected with the trunk by means of two transition spans, 50 and 60 m long, respectively. The cross section of the branches for vehicle traffic is made up of a single, prefabricated, central, 6.70 m wide box girder. To accommodate the required platforms the branches are completed with variable cantilevers, one of which is supported on struts. 5 CONSTRUCTION 5.1 Approach spans Diverse construction procedures were compared. All of these are suitable for the construction of prestressed concrete, 70 m long spans and their different modalities: cantilevered system or span-by-span construction, built both in-situ (using form travellers or movable falsework) and with prefabricated segments (lifting frames at the advance front, segment launching gantries, span-byspan launching gantries). A comparative study carried out in collaboration with the firm BERD showed that the most competitive construction procedure for this civil work was the span-by-span method with prefabricated segments, even in spite of the enormous loads to be resisted by the launching gantry. One element that proved beneficial for this decision was the extensive deck surface potentially suitable for prefabrication, given the fact that the approaches and the main bridge share the same cross section. Also, the branches could be standardized based on a typical cross section. This meant additional advantages: apart from the favorable execution schedule, we could also reuse the auxiliary equipment (the prefabrication yard, transport elements, segment launching gantries). The prefabricated segments are installed by means of heavy-duty launching girders with the help of a detailed study of BERD, which proved the feasibility of the proposal. 190
Figure 8.
Scheme of the prefabricated segments and the casts.
Figure 9. Launching gantry scheme developed by BERD for the span-by span construction.
Figure 10. Transverse form traveller for the side cantilevers constructione.
In order to reduce the weight of the suspended elements and facilitate their installation, in a first phase only the central box girder is prefabricated and put in place, whereas the cantilevers on struts are built in a second, in-situ phase utilizing transverse form travelers. As with the approach spans, the branches were also built applying the span-by-span method with prefabricated segments smaller in size than those used in the trunk. These segments were standardized, which allowed the use of the same cast and the same device in all the branches. 5.2 Cable-stayed stretch Cable-stayed deck it is built by symmetrical balanced cantilevers. Prefabricated segments are transported on barges and lifted by means of lifting frames at the advance front. The progression is symmetrical beginning at the cable-staying towers and connecting the subsequent segments to the previous ones by means of prestressing bars. The built-in union at the towers guarantees the stability during construction, eliminating the need for auxiliary “tie-down” elements, with a maximum lag of one, 5.0 m segment. The segments of this area are to be installed using the cantilevers already prefabricated in the yard due to the fact that the additional weight and size of these elements are not critical as is the case in the approach spans. 191
Figure 11.
Construction of the cable-stayed bridge by balanced cantilevers.
Figure 12. Aerial view.
6 CONCLUSIONS The new crossing over the Magdalena River will be the largest bridge built to date in Colombia and the one having the longest span. The technical complexity implied in the execution of a civil work of such a magnitude offers the chance to introduce into the country some of the most advanced and innovative systems currently offered by bridge engineering science for a construction that is costeffective, fast and safe. This civil work therefore has a two-fold interest since it can also showcase other works to be undertaken within the ambitious plan of infrastructure development implemented by the Government. The hallmark elements that summarize this civil work are: – Configuration of the civil work with the continuous deck along the entire structure of the bridge, the approach viaducts, the main bridge and the branches, unifying the construction procedures and granting a uniform and singular image to the whole. – Usage of a single cross section for the central trunk of the work with a multi-cell box girder, completed with cantilevers over struts. – Development of a structure of branches based on a central box girder identical for all the branches, completed with varied cantilevers. – Central span with a cable-stayed stretch of remarkable dimensions, destined to become a distinctive element of the city and the country alike. – Application of advanced, highly mechanized, bridge construction techniques (prefabrication, launching gantries, self-climbing formworks, … etc.). 192
Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
Delivering the Padma Multipurpose Bridge project, Bangladesh W.K. Wheeler AECOM, Sydney, Australia
C.J. Tolley AECOM, Auckland, New Zealand
ABSTRACT: The US $2.9 billion Padma Multipurpose Bridge Project comprises a new fixed crossing of the Padma River in Bangladesh incorporating a combined road and railway bridge 6.15 km long (41 × 150 m spans) with separate road and rail viaducts at each end, major river training works and 13.6 km of approach roads, minor bridges and bridge end facilities. The bridge, currently under construction, will be the longest in South Asia and provide substantial economic improvements and poverty alleviation to the south west region of the country. This paper charts AECOM’s delivery as lead consultant of this mega project for the Bangladesh BridgeAuthority from development of concepts, studies, detailed design and documentation and procurement assistance over a period of six years, culminating in the award of five major construction contracts.
1 INTRODUCTION 1.1 The project The detailed design of the Padma Multipurpose Bridge was delivered by a team of international and national consultants headed by AECOM under an Asian Development Bank Technical Assistance loan to the Bangladesh Bridge Authority (BBA). The team comprised AECOM, SMEC International, Northwest Hydraulic Consultants and ACE Consultants, with additional assistance from Aas Jakobsen and HR Wallingford. The Padma Bridge is a multipurpose bridge designed to carry four lanes of highway traffic, a single freight rail track, a high pressure gas main and various communication facilities. The bridge is on the Asian Highway Route A-1 and Trans-Asian Railway Route connecting the two regions of India, approximately 35 km southwest of Bangladesh’s capital Dhaka (see Figure 1). Figure 2 shows the location and general layout of the project which comprises a new bridge 6.15 km long across the Padma River, approach viaducts, major river training works and 13.6 km of approach roads and bridge end facilities, including toll plazas, service areas and offices. 1.2 The Padma River The Padma River is one of the three major rivers in Bangladesh. It is approximately 100 km long and flows in a south east direction from the confluence of the Jamuna (Brahmaputra) and Ganges Rivers. The mean flow is around 30,000 m3 /sec. The present general location of the river dates from about 1826 and the plan form changes from a single channel to multiple braided channels. The bed and bank materials of the Padma are generally fine sands that can be mobilized by quite low river flows. The north or left bank near the bridge site is somewhat erosion-resistant along substantial lengths due to clay soils, but the south or right bank consists generally of unconsolidated, noncohesive, fine-grained soils and has exhibited major erosional and depositional changes within periods of only a few years. 193
Figure 1.
Figure 2.
Project location.
Project layout.
2 DELIVERY 2.1 Scope and governance The scope of AECOM’s work included review of the outcomes of the previous studies; develop and confirm the design criteria for the detailed design; identify, commission and supervise necessary additional studies to provide input to the detailed design; update and expand on previous traffic surveys; prepare a comprehensive financial analysis and financing plan; undertake economic evaluation and confirm the economic viability of the project; address the social and environmental impacts arising from the project and develop the necessary safeguards to be employed; develop a procurement strategy to comply with the Multilateral Development Bank Guidelines – the proposed source of funding; and assist with the procurement of the civil works through prequalification of contractors, assistance during bid period and bid evaluation. AECOM established a project office in Dhaka in early 2009 with a core team of 8 full time international and 60 full time national staff. The detailed design was delivered from multiple offices with 44 short term international staff visiting Dhaka during the course of the project – Dhaka (overall project management, river training works, approach roadworks, bridge end facilities, safeguard compliance), Hong Kong (main bridge structural and geotechnical design), Vancouver (river training works design and physical modelling), Sydney (financial analysis, economic evaluation, main bridge viaducts), Melbourne (traffic and transport planning, road design, approach bridges), Auckland (procurement documentation), Toronto (environmental impact assessment), UK (bridge architecture, river training works numerical modelling) and Oslo (main bridge verification. AECOM established an internal Project Management Board with representatives from AECOM Australia, New Zealand and Hong Kong to review progress on a monthly basis and the AECOM’s US Audit Services Group separately reviewed the in-country operations. BBA engaged a Panel of Experts comprising internationally recognized specialists and local academics to meet with the AECOM team in Dhaka on a regular basis and review all technical design deliverables. BBA also engaged an Independent Checking Engineer (Flint & Neill/COWI) to verify the main bridge and river training works detailed design. The World Bank, Asian Development Bank, JICA and Islamic Development bank sent missions to review design progress during the course of the works with a view to facilitating the formal loan approval process. 2.2 Key challenges The key challenges of the project included the technical engineering issues of complex river training works in a river subject to substantial annual flooding and high seismicity, and construction of a major bridge with deep pile foundations in loose alluvial deposits subject to extreme scour depths. 194
Proper handling of social and environmental impacts arising from the project was required, including addressing land acquisition and resettlement impacts on affected people and the environmental impacts on regional hydrology and ecosystems. Additional challenges included coordinating with the various organizations involved in the project (our client (BBA), numerous other government agencies, independent reviewers, multiple potential financiers); managing a large team working in multiple locations and time zones; running a project office in Dhaka with intermittent communications and unreliable power supplies; ensuring the safety, security and well-being of a team of national staff and short and long-term expatriate personnel; and operating effectively in a country with significant cultural differences, ongoing political instability and widespread corruption. 3 PREVIOUS WORK AND ADDITIONAL STUDIES 3.1 Previous studies A considerable amount of work had been undertaken prior to AECOM’s commencement, primarily since completion of construction of the Jamuna Bridge in June 1998. This included the Prefeasibility Study, 2000 and the Feasibility Study, 2005. These documents were reviewed for their accuracy, completeness and relevance to the detailed design phase of the project. The objectives of the Prefeasibility Study were to determine the most suitable location for the Padma Bridge and to look at possible configurations for it. The Feasibility Study recommended a preliminary design comprising a prestressed concrete extradosed bridge with 180 m main spans, framed reinforced concrete piers, supported on 3.15 m diameter raked steel tubular piles driven to typical depths of 80m through silty sands. The river training works design proposed no artificial constriction of the river at the site with continuous revetment of the north bank for 6 km, continuous revetment of the unstable south bank for 4 km in addition to 6 km revetment upstream along the bank of the present secondary channel. This study provided the basis for the Government of Bangladesh (GoB) to proceed with the design and construction of the bridge. 3.2 Additional studies A total of 37 additional studies and surveys were undertaken in conjunction with the detailed design. These included traffic surveys; topographic surveys of the approach roads and river floodplain; bathymetric surveys of the river at various times before, during and after flood; river flow, scour and hydrological studies in conjunction with physical modelling; geotechnical investigations for the main bridge, approach roads and river training works; site specific environmental design parameters such as climate study and seismic; and a shipping study. The geotechnical investigations included 29 boreholes to a maximum depth of 150 m with insitu SPT tests, various field and laboratory tests to investigate shear strength parameters for the soils and determine mica content. However the poor performance of the drilling contractor led to termination of this contract and resulted in confirmation investigations in the central section of the river being included in the bridge construction contract. 4 DESIGN CRITERIA The preliminary design criteria developed for the Feasibility Study were reviewed and updated based on all the information available for the site. The main changes comprised: • The highway design loading was increased from the AASHTO loading to the British Bridge Code BS5400. The adoption of BS5400 was consistent with the designs for the Jamuna and Bhairob bridges in Bangladesh. • The railway design loading was increased to comply with the Indian Railways Dedicated freight Corridor loading with a 32.5 tonne axle loading. 195
• A site specific seismic hazard study was carried out by the Bangladesh University of Engineering and Technology (BUET) to determine suitable seismic parameters for use in the design. Two levels of seismic hazard were adopted. The bridge is required to remain operational for all traffic under an Operating Level Earthquake (OLE) with a 65% probability of exceedance within a return period of 100 years with a peak ground acceleration of 0.052 g. A Contingency Level Earthquake (CLE) with a return interval of 475 years and a 20% probability of being exceeded during the 100 year design life of the bridge was also considered with a peak ground acceleration of 0.144 g in the dense sand at a depth of 120 m. Any damage sustained from such an earthquake was required to be easily detectable and capable of repair without demolition or component replacement. • A detailed assessment of scour based on satellite images over the last 40 years and simple analytical methods was undertaken. Design general scour levels for the 100-year return interval were adopted as −34.8 m PWD in the middle of the river and −46.7 m PWD adjacent to the river bank. To these estimates were added the effects of local scour that occur from the obstruction to the flow caused by the bridge piles. Local scour was estimated from model test experimentation work to deepen the design general scour by a further 15 m for the proposed raking pile configuration. • A shipping study was undertaken to confirm the nominated vertical clearances and the design ship impact loading. Given the braided nature of the Padma River, it was decided that the number of principal navigational spans should be increased to provide the minimum vertical clearance over the central 4.8 km of the river. An additional allowance of 0.4 m vertical clearance was provided for the effects of future climate change. The load combinations given in BS 5400 Part 2 were generally followed, but this code does not adequately cover how to combine seismic loading, ship impact and scour of the foundations. The effect of scour was thus given special consideration as the nature of the Padma River is unique. Scour can occur over prolonged periods and when infill of scour holes later occurs, the material that fills the holes is loose and remains uncompacted for a long period after the event. The loose material is susceptible to liquefaction and cannot be relied upon during a seismic event. Hence scour with a 100-year return interval, combined with loading from a CLE seismic event, was adopted. In the case of ship impact, liquefaction of the infill material is not considered a problem and a lesser return period of 10 years for scour was adopted. Wheeler (2011) provides further detailed discussion on the development of the design criteria. 5 MAIN BRIDGE 5.1 Superstructure With the difficult foundation conditions at site, efforts were made to maximise the span length. Cast-in-place concrete box girders allow long spans to be achieved, but are slow to construct and their large mass exacerbates design for seismic loadings. Precast segmental construction is much quicker to construct and provides the added advantage of high levels of quality of workmanship associated with precast construction. The maximum span length, however, is limited by the available erection equipment and the maximum segment weight that can be handled. Similar superstructure solutions were used for the Jamuna, Bhairab and Paksey Bridges in Bangladesh with span lengths up to 110 m. The extension of this span length is possible through the use of extradosed cables, as proposed in the Feasibility Study (180 m spans). However review of the Feasibility Study design at the commencement of the detailed design phase indicated that the 180 m span length for the river spans needed to be reduced to satisfy the serviceability deflection and rotation performance requirements under the rail loading. This led to the adoption of a composite steel superstructure solution with greater span lengths. The resulting main bridge river crossing is in the form of a composite steel truss comprising 41 spans of 150 m (with typical modules of 6 spans) with two levels – a single railway track at the lower deck level and two 10.0 m wide highway carriageways at the upper deck level (refer Figure 3). Two 196
Figure 3.
Main spans superstructure.
Figure 4. Transition of river spans with approach viaducts.
main Warren truss planes, transversely spaced at 12 m, form the major structural component of the superstructure, with hollow steel box sections for top and bottom chords and for the diagonals. At the lower deck level, transverse lower cross beams at 18.75 m spacing connect the two bottom chords and form a platform for the railway track. At the upper deck level, a concrete deck comprising precast sections of approximately 22.0 m wide is made composite with the top chords. The trusses were envisaged to be prefabricated, transported to site in modules, assembled into full span lengths and erected one span at a time. Precast deck slab sections are then lifted into place and prestressed. The viaduct spans are separated into the approach road and the railway viaducts. With the main bridge as a two level structure, a complex arrangement of viaducts was required to separate the railway from the highway. There are a total of four viaducts supporting the highway, two on each side of the river. The approach road viaduct lengths range from 720 m to 875 m long and comprise 38 m spans of precast, pre-tensioned concrete ‘Super-T’ girders. There are two viaducts supporting the railway, one on each side of the river. The railway viaduct lengths range from 2.36 km to 2.96 km and they also comprise 38 m spans of precast, post-tensioned concrete I-girders. Figure 4 shows the transition arrangement to the river spans. 5.2 Substructure Piers for the main bridge comprise reinforced concrete columns supported by a deep pilecap and a group of six 3.0 m diameter steel tubular piles raking in symmetric pattern driven to founding levels at 114 m depth. Two types of piles were investigated in the design – large diameter raking steel tubular piles and large diameter cast in situ vertical concrete bored piles. The raking piles were found to be more efficient in resisting lateral loads resulting from seismic and ship impact effects. The superstructure is supported by friction pendulum seismic isolation bearings, which significantly reduce the seismic loading generated at the top of the piers and hence to the piles. Sham (2010a), Deery (2011) and Sham (2010b) provide a comprehensive description of the detailed design of the main bridge structural components. 197
6 RIVER TRAINING WORKS 6.1 Design approach The river training works are designed to protect three distinct areas from damage by the river. At the north bank they are designed to prevent possible outflanking and erosion of the viaducts and end facilities. At the south bank they are placed near the bridge and viaducts to prevent outflanking and erosion of the viaducts and end facilities and are also placed upstream of the bridge to protect the new approach road, approach road bridges, drainage structures and the two riverside resettlement villages. To achieve this, the design incorporated in-depth study of river morphology, numerical analyses, physical models tests, geotechnical investigations and slope stability analyses. 6.2 Outcomes After consideration of a number of alternative layouts, a layout fairly similar to that proposed in the Feasibility Study has been adopted. Rather than attempting to maintain the river in its present fairly straight alignment towards the bridge, this layout allows the main river to re-occupy the present minor south channel from time to time, permits the south bank training works to be constructed on top of the slightly more consolidated sediments of the present river bank, and leaves room for the river to adjust to possible future developments including climate change. The works consist of a continuous embankment and revetment that starts downstream of the bridge and continues upstream for a length of about 10 km, turning into the right bank of the south channel. This extension is designed to prevent outflanking of the south bridge abutment and eroding of floodplain land towards the south approach road. The work on the relatively stable north bank will consist of a relatively short length of embankment and revetment. The typical embankment or revetment cross-section has three main components: upper-slope wave protection using concrete blocks, a dredged underwater slope with erosion protection using rock riprap and geobags, and a launching apron at the toe of this slope using rock riprap or geobags. Because potential maximum scoured depths far exceed the feasible maximum depth of the dredged underwater slope, the launching apron is critical for long-term stability and new ones may need to be placed at lower levels as scour deepens. This will require a long-term capability for in-river operations that goes beyond normal maintenance requirements. Neill (2010) describes in detail the development of the river training works design. 7 SAFEGUARD COMPLIANCE 7.1 Social action plan The project covers three districts with an estimated 13,000 households (74,000 persons) affected by the project construction. Of the total affected households, about 4,000 required relocation prior to project construction. Four resettlement sites were identified for relocation of the affected households and these sites were constructed with all civic amenities for resettlement of the affected families. Safeguard documentation “packaging” was very important to demonstrate the full coverage of impacts. The social and resettlement safeguards were presented in an 11 volume Social Action Plan to ensure comprehensive coverage and to assist the review of compliance with the documentation requirements of all co-financiers. 7.2 Environmental action plan The project boundaries extend 15 km upstream and 7 km downstream of the main bridge, laterally 6 km from the river bank at Mawa towards Dhaka and 4 km from the river bank at the Janjira side. Within these boundaries consideration was given to potential changes in ecology, water use and management practices, dredge spoil disposal, agricultural and fishing practices which may occur due to the possible backwater effect, disrupted drainage, navigation and water transport. The Environmental Action Plan (EAP) was presented in seven reports. 198
The Padma Bridge design project received the 2010 award for “Best Safeguard System in Project Planning” by the South Asia Department, Asian Development Bank. 8 PROCUREMENT A key component of AECOM’s scope of work was assisting BBA with the procurement of the civil works. A proposed Procurement Strategy for the project was initially developed which included a review of contract packaging, methods of procurement, bid processes and prequalification that met project requirements and complied with both co-financier and GoB guidelines. In consultation with the GoB and co-financiers, it was determined that the project be packaged into a total of six works contracts comprising; (i) Main Bridge – river spans and viaduct spans; (ii) River Training Works; (iii) Janjira Approach Road, associated toll plaza and Service Area 3; (iv) Mawa Approach Road, associated toll plaza and Service Area 1; (v) Service Area 2; and (vi) construction yards on both sides of the river. The size, scale and complexity of most of the works contracts justified a customised approach to the procurement process. This customisation began with the prequalification of bidders, an essential step in procuring all six works contracts. Key qualification criteria covering an applicant’s eligibility, historical contract performance, financial capability and capacity, general and specific construction experience, personnel and equipment were all rigorously assessed and revised to take into account the unique nature of each works contract and ensure that only those firms and consortia capable of constructing and completing the works were prequalified. The next step in the procurement process was bidding. In the case of the two largest value contracts, the Main Bridge and River Training Works, it was agreed that a simplified ‘Two-Stage’ bid process would be adopted. This process gave bidders the opportunity to present their technical proposals for the contract, including construction methodologies, plant, equipment and personnel and receive feedback, before pricing their bids; and propose limited design alternatives and alternative completion timetables. All other contract packages followed a ‘Single-Stage One-Envelope’ bidding process following prequalification as the size and complexity of these contracts did not justify the additional time and cost of the ‘Two-Stage’ bid process. Tolley (2013) provides further details of the procurement process. 9 MANAGING THE PROJECT TEAM Managing a large team working in multiple locations and time zones; ensuring the safety, security and well-being of a team of national staff and short and long-term expatriate personnel; and operating effectively in a country with significant cultural differences, ongoing political instability and widespread corruption were all key challenges for the project. AECOM have been working in Bangladesh on highway and power projects since the mid-1980s, most notably the construction supervision for the Paksey Bridge (1998–2005) and the National Load Dispatch Centre, Dhaka (2005–2010). This ongoing presence and experience in country, albeit on a project basis, greatly assisted our mobilization efforts. Initial tasks included finding an appropriate project office in a suitable location to accommodate the large number of full-time and part-time staff members and hold meetings with clients and co-financiers; sourcing office furniture and equipment including computers, photocopiers and printers; installing a local area computer network with a secure server and reliable internet access; provision of a generator to power essential services during frequent daily power outages; provision of local project transport for travel within Dhaka and to site; establishing a robust project accounting system and facilitating visas and arranging suitable accommodation for both long and short-term international staff. The safety, security and health of project staff was paramount and a Project Safety Plan was prepared and issued to every staff member prior to mobilisation which detailed airport arrival procedures; accommodation arrangements; office location and contact details; guidance on personal safety, money and appropriate behaviour; medical contacts; embassy and consulate information 199
and emergency evacuation procedures. International staff were required to have appropriate vaccinations prior to arrival (Hepatitis A, typhoid, tetanus). Bouts of diarrhea and mild stomach upsets were not uncommon amongst the project team. More seriously we had two cases of typhoid and one case of appendicitis requiring evacuation. Dengue fever was also a significant risk. Road safety was a significant issue on highways and the use of competent drivers and roadworthy vehicles proved essential. Site visits introduced additional risks – isolation when away from the office, safety on water, access to suitable local food, poor communication and chance meetings with local village people aggrieved by land acquisition or other issues related to the project. Transparency International’s Corruption Perceptions Index has ranked Bangladesh one of the most corrupt countries for the last 5 years running. It is generally accepted that corruption in Bangladesh is widespread affecting every aspect of life. AECOM introduced measures to successfully mitigate corruption risk in all its dealings with the client, government, sub-consultants, contractors and other interested parties. 10 CONCLUSIONS The detailed design and independent proof checking of the main bridge and river training works was completed at the end of 2011. The prequalification process and preparation of bidding documents for the five main contracts were completed shortly thereafter, with funding for the estimated US $2900M construction costs agreed with the World Bank, Asian Development Bank (ADB), Japan International Cooperation Agency (JICA) and Islamic Development Bank (IDB). In September 2011, the World Bank suspended release of its $1200M share pending investigation of alleged corruption in the construction supervision consultancy bid process with the ADB, JICA and IDB following suit. At the end of June 2012, the World Bank cancelled its $1200M IDA credit because of inadequate response by GoB to the World Bank’s request to investigate corruption claims. In February 2013, the GoB withdrew its request to the World Bank to finance the Padma Bridge Project and announced its intention in April 2013 to construct the project using their own resources. In June 2013, AECOM was re-engaged to assist GoB in completing the procurement process under an Addendum to the original contract. The procurement of the three approach road and service area contracts were completed first with all three contracts in progress by the end of 2013. The evaluation of the first stage technical proposals and second stage financial bids for both the Main Bridge and River Training Works Contracts proceeded were completed by October 2014. The Main Bridge Contract was awarded to China Major Bridge Engineering Co. Ltd. on 2 June 2014 for approximately US $1547M and the contract for the construction of the River Training Works was awarded to Sinohydro Corporation Limited on 26 October 2014 for approximately US $1110M. The new Padma Multipurpose Bridge will provide a vital missing link in the transport network of Bangladesh. AECOM has effectively addressed a number of significant challenges to successfully deliver the detailed design and procurement stages of the project. REFERENCES Deery, M.J. 2011. Design of the viaduct spans on the Padma Multipurpose bridge project, Bangladesh. Proc. 8th AUSTROADS Bridge Conf., Sydney, November 2011. Neill, C.R., Oberhageman, K., McLean, D. & Ferdous, Q. M. 2010. River training works for the Padma multipurpose bridge, Bangladesh. Proc. IABSE-JSCE joint conf., Dhaka, 8–10 August 2010. Sham, S.H.R. & Tapley, M.J. 2010a. The design of the Padma Multipurpose bridge – challenges and solutions in design of the river spans. Proc. IABSE-JSCE joint conf., Dhaka, 8–10 August 2010. Sham, S.H.R., Yu, G.X. & De Silva, S. 2010b. Foundation design methodology for the Padma main bridge. Proc. IABSE-JSCE joint conf., Dhaka, 8–10 August 2010. Tolley, C., & Aves, R. 2013; Procurement of the Padma Multipurpose Bridge, Bangladesh. Proc. 14th REAAA Conference 2013, Kuala Lumpur, 26–28 March 2013 Wheeler, K., Sham, S.H.R., Aves, R., Tolley, C., & Islam, Md R. 2011. Design of the Padma multipurpose Bridge, Bangladesh. Proc. 8th AUSTROADS Bridge Conf., Sydney, November 2011.
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Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
The tied arch bridge of the Saale-Elster-Viaduct W. Eilzer, R. Jung, T. Mansperger & K. Humpf Leonhardt, Andrä und Partner Beratende Ingenieure VBI AG, Dresden, Germany
ABSTRACT: The Saale-Elster Viaduct is part of the future high-speed railway, connecting the cities of Munich and Berlin. It crosses the floodplain of the rivers Saale and Elster south of the city of Halle. With a total length of 8.6 kilometres it is Germany’s longest Railroad Bridge. Due to the branching off of the junction to Halle, the bridge has three abutments. The tied arch bridge is located in the heart of the viaduct, where the main route surpasses the junction to Halle. With its exposed position it is the optical and technical highlight of the entire building. Furthermore it is the first tied arch bridge of the German railroad network with a design speed of 300 km/h, which led to increased approval efforts during the design process. The dynamic assessment of the high-speed railway traffic, and fatigue resistant detailing of all connections governed the design.
1 THE SAALE-ELSTER-VIADUCT The Saale-Elster-Viaduct is part of the overall project VDE8 that stands for “Verkehrsprojekt Deutsche Einheit” (German unity transport project), which shows the origin of the Project in the early 1990ies. VDE8 will be part of the trans-European transport network, connecting the cities of Munich and Berlin with a travel time of about 4 hours. The project is divided in 3 major subprojects: VDE 8.1 is partly upgrade line between the cities of Nuremberg and Ebensfeld and partly a new line between Ebensfeld and Erfurt. VDE 8.2 is also a new line proceeding from Erfurt in northeastern direction to Halle and Leipzig. To reach both cities the line has a junction with the main route to Leipzig and the secondary route to Halle. The most northern section is VDE8.3, which is an upgrade line from Leipzig and Halle to Berlin.
Figure 1.
Overview VDE8 (picture www.vde8.de).
201
Figure 2. The entire viaduct under construction in 2012 (photo DB AG).
The Saale-Elster-Viaduct is on the new line VDE 8.2 between Erfurt and Leipzig, located in south of the city of Halle right at the junction of the secondary line. It crosses the floodplain of the rivers Saale and Elster with their very sensitive habitat for many species, especially birds. This circumstance led to several requirements regarding both, the construction period and the construction method. In some parts on site works were only allowed from October until March and it was not allowed to touch the ground besides the area of the foundation of the piers. Therefore the bridge had to be built with special formwork carriages that allowed building the next piers just ahead the previously finished superstructure. The viaduct is designed as a double tracked Railway Bridge with an intersection free junction. The main bridge has a total length of about 6.5 km and the junction bridge is about 2.1 km long. This makes a total length of the viaduct of 8.6 km, which is Germany’s longest railway bridge. The maximum height of the tracks above ground is 21 m. The main bridge consists of 143 spans, the junction bridge of 48 spans. The design speed for the main route is 300 km/h and 160 km/h for the secondary line to Halle. The Bridge is predominantly built out of two-span prestressed concrete beams, complemented by single-span beams and three-span beams as overpasses. In the heart of the viaduct, where the main route surpasses the junction, the tied arch bridge, the only steel construction of the entire building, is located. 2 GENERAL LAYOUT OF THE TIED ARCH BRIDGE The tied arch bridge with its exposed position is the optical and technical highlight of the viaduct. The span length of the bridge is 110 m while the rise of the arch is 19.60 m. This makes a slenderness of 5.6, which is a rather low value due to the severe requirements of high-speed traffic regarding the stiffness of the structure. The distance of the axis of the arches and the main girders is 11.2 m, while the clearance of the roadway slab is 10 m wide. A lateral bracing bound by portal beams stiffens the arches. The roadway slab is an orthotropic steel deck with a 14 mm thick steel plate, supported by flat bars with a spacing of 400 mm. It spans across about 2 m high cross beams with spacing of 2.5 m. The cross 202
Figure 3. Approaching the tied arch bridge from West.
Figure 4.
Longitudinal section.
sections of the main girders and the arches are box-shaped. Each of them is 1.2 m wide. The height of the arch is 1.6 m and the main girder rises up to 3.15 m. Most of the dimensions have been chosen to increase the rigidity of the bridge and to improve its dynamic behavior, which will be explained more deeply later on. Underneath each springing of the arches, a vertical elastomeric bearing is placed. For horizontal loads, transverse bearings are located in the axis of the bridge on both piers. A steering rod system has been installed for longitudinal loads. The steering rod distributes horizontal loads from breaking, acceleration and wind almost 50-50 on the two piers, while dilatation due to temperature is unblocked. To make sure, that the system doesn’t get any strain, a pendulum has been installed. The pendulum consist of a straight bar, with radial hinges on both ends. As these radial hinges are not covered by German codes, a single case approval for their usage is required. 3 SPECIAL REQUIREMENTS DUE TO HIGH SPEED TRAFFIC With a design speed of 300 km/h, the tied arch bridge will be the first of its kind in the network of the German Rail. As this construction type is only ruled for a design speed of up to 160 km/h, 203
Figure 5. Time history of main girder bending moment.
a so-called “single case approval” (Z.i.E.) was mandatory for the entire bridge. This procedure was very time consuming but necessary and led to various changes in the design. During the single case approval procedure, the tied arch bridge Wustermark on the high-speed line Berlin – Hamburg was found for an extensive monitoring program. It is currently in service for both, high-speed and freight trains and has a design speed of 250 km/h, which made it prefect as a reference building for the tied arch bridge of the Saale-Elster-Viaduct. Over a period of more than three months, deflections, accelerations and strain in various construction parts were monitored. The Wustermark arch also has a box shaped main girder, which increases its torsional stiffness. This torsional stiffness is very important to keep the deflections of the crossbeam to a minimum and to prevent resonant behavior of the hangers in transverse direction. This was one of the main findings to improve the dynamic behavior of the bridge in transverse direction and led to the requirement of widening the box cross-sections of the main girders. Dynamic assessment is a major topic on all high-speed bridges. As the tied arch bridge is the first of its kind with a design speed of 300 km/h, the procedure and the requirements have been tightened. Rolling stock of all High-Speed-Load-Models (HSLM) from the Eurocode and all existing and known future ICE trains were computed with travelling speeds of 160 km/h to 360 km/h in small intermediate steps. Additionally two freight trains with a maximum speed of 120 km/h were investigated. This led to about 400 train passages that had to be computed and evaluated. The results were compared to the static ULS section forces of all parts of the structure. It was found out, that for the main girders no real resonant behavior occurs. Due to the shape of the main eigenmode as a sinusoidal wave, only those positions of loads where the train is located on one half of the superstructure create a significant punch. This occurs when the train rolls on or off only, so there are two critical positions in each train crossing – to few to cause resonant response. For the hangers, dynamic results were governing for the design. The crossing of each wheel axle creates a significant punch on the crossbeams, causing torsional rotation of the main girders and finally exciting the hangers at their staying in the main girders. For some trains, especially one of the relatively slow travelling freight trains, resonant behavior of the hangers in transverse direction was detected. For the operational phase, an extensive monitoring program has been set up and a huge amount of acceleration sensors and strain gauges have been installed. It is planned to perform a long term monitoring on the bridge over the next years, starting with the operational test phase by the end of this year. Although numeric investigations don’t indicate it, the installation of damping measures 204
Figure 6.
Connection of a hanger to a diaphragm.
is provided, in case operation shows that incapable resonant behavior occurs. Tuned mass dampers (TMD) can be installed on the main girders subsequently and crossties can be added to damp the hangers. 4 FATIGUE ASSESSMENT Fatigue resistance was the other major topic in the design process besides the investigation of rolling stock. In the beginning, the standard calculations for steel bridges using LM71 and the λ-factor according DIN Fachbericht 103 were performed. Because of the lack of experiences with tied arch bridges for high-speed trains, the requirements were tightened. Unfortunately no load collectives of the future operation could be accessed or be predicted by the client, so an appropriate advanced fatigue assessment using the rain flow counting method and the Palmgren-Miner rule was not possible. Due to the fact, that high-speed train traffic causes vibrations and resonant behavior of some of the parts of the construction with many amplitudes, it was demanded, that the extreme stress variation ranges from the dynamic calculations should not exceed the endurance limit of the steel. In the end this requirement was governing for the design for the fatigue sensitive details. One of the most crucial details was definitely the connection of the hangers to the diaphragms of the main girder and the penetration point of the hanger through the upper chord of the main girder. Usually the main girder is an open double-T cross-section and for the connection to the hanger, the web is being extended above the upper chord and smoothly tied into the hanger. In case of a closed box, that kind of detail is not possible and because of corrosion protection, the penetration point has to be permanently airtight. It was decided to create a very smooth shape for the connection of the hangers to the diaphragms, to keep the connection flexible, without causing to high stress levels in case of resonant hanger movement. Therefore the cutouts of the diaphragms were enlarged and their shape was adjusted to the well-known detail of open cross sections. Extensive finite element 205
Figure 7.
Penetration point of an hanger through the main girder.
Figure 8.
Erection scaffolding (photo: DB AG).
calculations showed that all the dynamic defections and rotations caused by the high-speed trains are capable on a low-stress level. Additionally the construction parts consisting of the diaphragm and the beginning stump of the hanger were stress-relived at about 650◦ C before they were installed in the main girder. A collar out of triangular rubber elements to keep it waterproof and flexible at the same time closed the penetration point of the hanger and the upper chord.
5 CONSTRUCTION The bridge was manufactured in the shop in 24 major segments. Therefore the arches were divided in 5 parts each side. The main girders were each prefabricated in 3 segments and the roadway slab in 6 pieces. The heaviest parts were the end-cross-beams with a weight of about 100 t. On sight, the erection scaffolding, built out of three single span beams was set up. The scaffolding had a total steel weight of 500 t – enough to build a standard road bridge in the same length. A special demand for the scaffolding was to overpass the already built junction bridge underneath the tied arch. 206
Figure 9.
Installation of the arch segments (photo: DB AG).
Figure 10. The finished tied arch bridge (photo DB AG).
As all concrete bridges, main bridge and the two junction bridges were already built, the space for a sufficient crane was limited. For that reason, the prefabricated parts of the main girders, the end-cross-beams and the roadway slab could just be placed on one side of the scaffolding. When one third of the superstructure was finished, it was rolled along the scaffolding. Then the next parts were added and the rolling-procedure was repeated. After that, the last third was added and the superstructure was completed. For the construction of the arches and the lateral bracing, four auxiliary columns were placed on the main girders. After the parts of the arches and the lateral bracing were put in place and welded, the support has been removed, so the arches were under compression. Finally the hangers were added, beginning with the shortest, proceeding towards the middle. This procedure was very sensitive to make sure that all hangers are tensioned equally. To ensure this, the welding works were done at a certain temperature range and in the early morning hours, before the warming of the sun causes unsteady deflections. By measuring the eigenfrequencies, the tensional force in the hangers could be determined. The results showed very good accordance to predicted values from the numerical model. Finally the scaffolding was removed and the bridge was finished. 207
The structure of the bridge was finished in summer 2013 and currently the first operational tests are carried out. After a very elaborate test phase including an extensive monitoring program, the operational phase will start by the end of 2015. ACKNOWLEDGEMENTS Client: Project Management: Contractor:
Sub-Contractor Tied Arch Bridge: Basic Design: Detailed Design: Checking Engineers:
Monitoring and single case approval expertise tied arch bridge: Single case approval expertise steering rod system:
DB Netz AG, Germany DB Projektbau, Leipzig, Germany Joint Venture Saale-Elster-Talbrücke Hochtief AG, Berlin, Germany Adam Hörnig, Aschaffenburg, Germany Gerdum und Breuer, Fuldabrück, Germany ZSB Zwickauer Sonderstahlbau, Zwickau, Germany Krebs und Kiefer, Darmstadt, Germany Design Joint Venture Saale-Elster-Talbrücke Leonhardt, Andrä und Partner, Dresden, Germany (superstructures) Kinkel und Partner, Neuisenburg, Germany (piers and foundations) Dr.-Ing. Walter Streit, Munich, Germany (superstructures concrete) Dr.-Ing. Klaus Baumann, Greifswald, Germany (piers and foundations) Prof. Dr.-Ing. Gerd Albrecht, Munich, Germany (tied arch bridge; until 2010) Dipl.-Ing. Jörg Lutzens, Munich, Germany (tied arch bridge; from 2010) Prof. Dr.-Ing. Karsten Geißler, Dresden, Germany Prof. Dr.-Ing. Martin Mensinger, Munich, Germany
REFERENCES DB Projektbau GmbH 2013, German Unity Transport Project. [online] available: http://www.vde8.de [accessed: 27 November 2013]. Mansperger, T., 2014, The Tied Arch Bridge of the Saale-Elster-Viaduct, IABSE Young Engineers Colloquium, Dresden, Germany, 2014 (in German). Schenkel, M., Felgner, M. 2013, The Tied Arch Bridge of the Saale-Elster-Viaduct, EI – Der Eisenbahningenieur, Nr. 10/2013, pp. 63–68. (in German).
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Construction and design features of the bridge over the Danube River, Bulgaria J. Manterola, A. Martínez & J.A. Navarro Carlos Fernández Casado S.L., Spain
J.L. Alvárez & J.I. Díaz de Argote FCC., Spain
ABSTRACT: The main bridge over the Danube River between Vidin (Bulgaria) and Calafat (Calafat) has a total length of 1391 m. Expansion joints are included only in the three abutments, thus configuring an integral bridge of 1791 m. The same deck depth has been used for different span lengths using additional structural schemes such an extradosed cable stayed system and struts. All the longitudinal forces are held in the four 180 m bridge main piers. It has three different decks depending on the span lengths. The road bridge has an 80 m deck for the non-navigable part of the river and a 180 m deck for the navigable one, thus allowing two navigable channels of 150 m. The railway has 40 m span lengths with intermediate supports at one fourth over the road bridge on the 80 m spans. The span distribution is: 52.0 + 7 × 80.0 + 124.0 + 3 × 180.0 + 115.0 m. The deck is a precast prestressed concrete single box girder with transversal struts, with a total width of 31.35 m and a depth of 4.50 m. The bridge deck was built by free cantilever system with matched precast segments in the central core of the section.
1 INTRODUCTION For the design and construction of the bridge over the Danube River in Vidin some innovative aspects have been taken into account. Both the design of the bridge and the construction procedure have been considered in a unified way. The design and build contract allows this relationship in a proper way. The coordinated alignment design for the railway and railroad allows the construction of a unique bridge deck. The same deck depth has been used for different span lengths across the Danube River using additional structural schemes like an extradosed cable stayed system. The intermediate struts reduce the stress variation on the stays due to the important live loads from the railway. The structural alternative to these struts would have been an increase of the deck depth at supports. The deck is completely built in on these struts resulting an integral configuration of the bridge which allows resisting the horizontal forces due to braking and seismic loads on the main pier foundations. This also minimizes the number of joints in the bridge, designing a bridge with 1791 m without expansion joints; only three are included in each abutment. The bridge deck has had an evolutional construction both longitudinally by balanced cantilever construction and transversely. The main core of the box girder has been completely precast. The lateral cantilevers have been cast in place for the 80 m span deck and partially on the construction yard for the 180 m span deck. 209
Figure 1. View of the bridge over the Danube river between Vidin and Calafat.
Figure 2.
Main river bridge definition and final view.
2 MAIN RIVER DANUBE BRIDGE The main bridge over the Danube River has a total length of 1391 m. It has three different decks depending on the span lengths. The road bridge has 80 and 180 m span lengths, the railway has 40 m span lengths with intermediate supports at one fourth over the road bridge on the 80 m spans. The span distribution is: 52.0 + 7 × 80.0 + 124.0 + 3 × 180.0 + 115.0 m. The deck is a single box girder with transversal struts made of prestressed concrete. The box girder has a depth of 4.50 m all along the bridge, and a total width of 31.35 m. 210
Figure 3.
Construction of the main river bridge, both 80 m and 180 m deck.
The girder lower slab thickness varies from 0.45 m to 0.75 m. The webs have an initial thickness of 0.50 m which increases to 1.30 m over the supports. Considering this significant web thickness, no diaphragm is included over the piers in a second stage, greatly simplifying the construction of the deck. In order to achieve that, the bearing axes have to be as close as possible to the web axes. Thus, small spherical bearings have been designed, locating them close to the outer face of the pier, and separating the transversal guide from the axial bearings. To complete the cross section, there are two families of slanted struts which create two intermediate supports to the upper slab. The transversal struts are located every 4.30 m in the 80 m spans and every 4.186 m in the 180 m spans. The construction of the decks has been done through a balanced cantilever system with precast segments. The matched precast segments have been the main core of the box girder. The rest of the 211
Figure 4.
180 m main river bridge foundations and ship impact protections.
cross section has been cast in situ with a special form traveller in the 80 m deck, and partially in the yard for the 180 m deck. The 80 m span deck has been built by free cantilever system with a gantry girder. The precast segments are 2.15 m long, with a maximum weight of 1000 kN. The 180 m span decks has been built by free cantilever system with the extradosed stays and lifting the segments from the river. The precast segments are 4.186 m long, with a maximum weight of 2500 kN.
3 180 m BRIDGE PIERS AND FOUNDATIONS The piers PB9 to PB12 are for the 180 m span deck. They support the deck and the pylons for the extradosed cable system. The pier under the deck is a single box girder 9.70 m wide and a variable dimension longitudinally, from 4.90 to 5.40 m, with walls 1.00 to 1.20 m thick. The pylons above the deck level are two rectangular sections 1.20 m wide and variable dimension longitudinally, from 4.50 to 4.90 m. The total height varies from 38.7 to 44.8 m. From the foundation a pair of concrete struts connects the deck to the foundation 14.70 m apart the support axis. Piers PB9 to PB12 are founded over 24 concrete bored piles of diameter 2.0 m, with a pile cap 15.0 m long, 40.0 wide and 6.0 m deep. The four combined Piers PB9 to PB12 have a very important strength against longitudinal horizontal forces due to friction on the pot bearings under service conditions or seismic forces and ship impact forces under accidental conditions. Thus, the main river bridge abutments do not support any longitudinal forces from the bridge but the bearing friction ones, so they do not present any special feature. In order to provide protection against ship impacts, a new type of prestressed grillage has been designed, combining precast vertical pieces which have holes to join with longitudinal horizontal beams which give continuity between the precast pieces. These latter ones are cast in situ. The protection of the pile caps is located above the maximum water river level under emergency situation, 36.60 m, and the mimimum level, 25.50 m. The foundations have been verified to resist the vessel and barges impact forces according the design basis and the Eurocode-1 Part 7.
4 APPROACH STRUCTURES ON THE BULGARIAN SIDE The approach structure to the main bridge over the Danube has a total length of 400 m. The span distribution is: 32.0 + 9 × 40.0 + 8 m. It is a railway bridge which crosses above the road carriageways on embankment. The deck is a twin girder with transversal slab made of prestressed concrete. The girders are rectangular solid sections which have a depth of 1.90 m and a width of 1.0 m. The slab is 0.25 m thick. 212
Figure 5.
Elevation and final view of the 40 m railway approach bridge.
All the bearings from the railway bridge are elastomeric ones, free in the longitudinal direction and guided in the transversal one. Its dimensions rank from 600 × 700 × 160 mm in pier pA1 to 1550 × 1000 × 400 in pier pA10. Such a bearings are used to highly reduce the seismic forces, using both legs of the piers to resist lateral forces. 5 CONCLUSIONS The second bridge over the 600 km Danube border between Bulgaria and Romania has been finished including the following features: – One unique bridge for railway and road, improving its aesthetical view and minimizing its costs. – Integral bridge, with a 1791 m continuous deck and only three expansion joints in the abutments. – Concrete precast bridge, trying to minimize construction time and using the significant experience of the contractor with this kind of bridges, studying thoroughly all the connections between precast and cast in situ parts to ease its construction and reassure its performance.
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TUNeIT – Towards a global World E. Siviero IUAV University, Venetia, Italia
A. Ben Amara National Engineering School, Monastir, Tunisia
M. Guarascio Sapienza University, Roma, Italia
G. Bella & M. Zucconi Niccolò Cusano University, Roma, Italia
A. Adão da Fonseca University of Porto, Portugal
K. Slimi Sousse University, Tunisia
ABSTRACT: Developments in engineering of the last century open up the hypothesis of creation of a link between Sicily and Tunisia, which would mean a transcontinental territorial continuity between Europe and Africa, in line with other physical connections between Europe and Asia, between Asia and Africa and the hypothesized stable connection between Europe and Africa across the Strait of Gibraltar. This future project, called TUNeIT, is sustained by RMEI (Méditerranéen Réseau des Ecoles d’Ingénieurs), EAMC (Engineering Associations of Mediterranean Countries), PAM (Parliamentary Assembly of the Mediterranean) and UNIV. The project recommends repeating the design of the Messina Strait Bridge several times, although this time restricted to railways, incorporating new technologies of “intelligent” structures. To cover the 140 km extent of the crossing, four artificial islands are built in correspondence with shallows where the seabed depth is reduced to between 40 and 60 m, which divides the total distance into five subdivisions varying in length from 20 to 30 km.
1 BENEFITS AND ADVANTAGES OF THE LINK BETWEEN TUNISIA AND ITALY Planet Earth is becoming one single place. Globalization is mostly the result of easy communication, with information technology at the forefront, already providing wide “motorways” for digitalised data, and thus for financial products and knowledge. Movement of people is a corollary, and their concentration in mega polis is an uncontrolled driving force, no matter obstacles individuals have to overtake. Improving transport facilities is then a fundamental requirement for the future. Africa and Europe are two Continents with a common history, in recent centuries unfortunately with unequal development. If Africa and Europe are to walk together, side by side, into a better future for both, permanent and safe connections between the two Continents are to exist, and that must be a common objective of Europe and Africa. At the European extremes, connecting the two Continents implies a long stride through the Middle East or requires the Gibraltar Bridge. Well, connecting the centres is more efficient and the smaller distance between the two Continents is precisely from Tunisia to Italy (Sicily). 215
2 POSSIBLE ALTERNATIVES FOR THE REALIZATION OF THE CONNECTION Among the possible alternative projects for the realization of the connection, the most important are the ENEA project (ENEA, 2003) and the AUFO project. The first one was born from an ENEA survey analysing the consequences of the completion and integration of the Great Infrastructure Works currently in service, under construction and/or in the design phase. The principal aim is to create a balance between the use of road and rail transports and to mitigate their European environmental impact in the North-South direction to and from countries of North Africa. Feasibility studies, economic compatibility, environmental sustainability and geopolitical impact of the combined transport lead to the hypothesis of construction of a rail link through an undersea tunnel between Sicily and Tunisia, strictly correlated to the construction of the bridge over the Strait of Messina. The ENEA solution puts forward a rail tunnel Africa-Sicily on a total path length of 130–140 km and aims to optimize the European transport system (ENEA, 2003). AUFO Project (Architectural & Urban Forum), Infrastructurban, city-bridge between Italy and Tunisia was created by a non-profit research group based in Milan, involving European and African universities in the design of inhabited infrastructure. Between Marsala (Italy) and Kelibia (Tunisia) ten new cities with 1 million inhabitants each were designed, functioning both as line of communication and linear conurbation. Ports and international airports that connect directly the European and African continents can serve these cities. The TUNeIT project proposes to repeat the design of the Messina Strait Bridge several times, and, as with the ENEA project, covers the just over 140 km long crossing with four artificial islands. These islands are in correspondence with shallows where the seabed depth reduces to between 40 and 60 m, thus dividing the path into five sections varying in length from 20 to 30 km (Siviero et al., 2014).
3 POSITIVE AND NEGATIVE FEATURES OF THE POSSIBLE ALTERNATIVES The solution proposed in the ENEA project involves the construction of an undersea rail tunnel linking the territory of Bon (Tunisia) and Pizzolato (Sicily), north of Mazara del Vallo, for a total length of about 150 km. The hypothesis is to create one large tunnel formed by three adjacent smaller tunnels: the two external ones are dedicated to trains and shuttles for cars, while the middle tunnel is for maintenance and services (Siviero, 2014). The total length of the route divides into five sections of reduced length, through the formation of artificial islands built with materials from the excavation. All tunnel services are located in the artificial islands. The TUNeIT project regards the bridge as a smart-bridge, with all services necessary for the bridge operation sited in the bridge. In this bridge, as well as in the artificial islands, advantage is taken from innovative energy sources, such as solar, wind or marine currents power, becoming energetically autonomous. The islands will be epicentres for a variety of activities and functions, mostly touristic. In the TUNeIT project, construction of two high-rise buildings adjacent to the pylons of the bridge was also considered. These high-rise buildings would be 380 meters high, each with 80 floors (Siviero, 2014). This Euro-African Bridge will assume an extraordinary symbolic value in the blend of cultures. This fact provides a decisive contribution to the creation of a unique “Metropolitan City”. Sicily would be not only an integral part of the Italian Peninsula but also an important unification link to Countries around the Mediterranean and Europe. The Tunisian side is yet quite free from urbanization. Therefore, to build, urbanize and design the arrival of the bridge is facilitated, as infrastructures and roads are yet to be created. Contrary, landscapes and historic environmental features in the Sicilian side require a careful evaluation whether the crossing will arrive as a bridge or as a tunnel. 216
Figure 1. A schematization of the TUNeIT project on Google Earth satellite image. The S points indicate the 4 artificial islands.
Figure 2. A longitudinal section of the TUNeIT project across the Sicily Channel.
Figure 3.
Schematization of the Messina Strait Bridge.
4 THE BRIDGE PROJECT The project intends to repeat the design of the Messina Strait Bridge several times, although this time restricted to railways, incorporating new technologies of “intelligent” structures. To cover the 140 km extent of the crossing, four artificial islands are built in correspondence with shallows where the seabed depth is reduced to between 40 and 60 m, which divides the total distance into five subdivisions varying in length from 20 to 30 km. 217
Table 1. Messina Bridge characteristics. Characteristics
Dimensions
Central span Suspended side spans Overall suspended length Distance between anchorages Deck width Driveway lanes Railway tracks Traffic capacity
3,300 m 183 m 3,666 m 5,070 m 60 m 2 × 2 lanes + emergency 2 6,000 vehicles/hr – 200 trains/day
Figure 4. A possible cross-section the Bridge.
The implementation of the project, in analogy with other works of major magnitude, comprises some crucial steps, including: – geological survey and detailed design (lasting about four years); – construction works, including at least three phases (for an estimated period of about 10 years): 1. positioning of the four platforms for the construction of artificial islands and beginning of the implementation of the terminals; 2. construction of the various sections of the bridge and of the artificial islands; in particular, bridge works can be divided between the four islands and two terminal sites; 3. implementation of all technical systems. In these islands, access for different purposes has to be guaranteed. 2 railways lines would provide the freight and passenger traffic. If motorized vehicles are to be considered, dual carriageways for each direction of travel and emergency and service lines have to be added. Freight and passenger trains shall go in the centre, for a total width not exceeding 60 meters. The islands, in addition to serving as separate terminals for the multi-span suspension bridges, will be epicentres for a variety of activities and functions, especially in the tourism trade. Moreover, the presence of 400 m high pylons at the end of each bridges allow for the creation of new metropolitan areas with multiple functions (Siviero, 2004). 4.1 Use of innovative materials Use of steel cables in marine environment raises very serious problems concerning durability and maintenance costs. Composite Fibre-Reinforced Pultrusion materials (FRP) for cables and hangers would then be considered as an alternative, as they offer the highest ratio of strength to weight and have very low maintenance costs. In addition, use of complementary “organic” (Pacheco et al., 1996 and 2000) cable stays for variable loads would allow suspension cables and hangers to carry 218
Figure 5. A possible schematization of the pylons of the Bridges.
Figure 6.
Fibre-Reinforced Pultrusion materials (FRP) for cables and hangers.
permanent loads only. The great reduction of weight would make the work much less demanding from the point of view of structural behaviour and construction phasing (Anania et al., 2014). These innovative materials have great advantages for cable-stayed and suspension bridges: high geometric efficiency of cables due to, at the designed stress levels, Dischinger’s modulus near the Young’s modulus of the material; and reduction in the height of pylons of suspension bridges with equal spa (Meier, 2012; Noistering, 2000; Carpinteri, 2008). Span limit in suspension bridges, with equal cross-section area, equal cable and equal deck, is directly proportional to the existence of the specific load-bearing cable. Therefore, regardless of pylons height, cable area and deck area gain in length, when using composites, by a factor greater than 3; allowing for spans longer than 3000 m (Liang et al., 2011; Wu et al., 2008). 219
Pylons of the suspension bridge should be made with offshore technology, especially when the sea is deep.
5 POSSIBLE ENERGETIC SCENARIOS FOR THE ARTIFICIAL ISLANDS The project ought to minimize any environmental impact, mainly in respect to the artificial islands. Energy autonomy of the artificial islands will be achieved through use of hybrid and integrated energy from renewable sources. The exploitation of solar energy, wind energy and marine currents, given the latitude of the islands and the offshore location, is particularly advantageous, making the islands eco-friendly and technologically advanced. The project provides African-European new energy links and multimedia telecommunications to different operators (Bella, 2014). The project has obvious environmental impacts, but they can be mitigated. First by minimizing the energy dependence of the artificial islands from external sources, using hybrid generation and integrated energy from renewable sources, as proposed by the widely scientific literature for specific localizations (Andaloro et al., 2012; Brito et al. 2014; Calise et al., 2014; Carapellucci et al., 2012; Chen et al., 2007; Chua et al., 2014; Neves et al., 2014; Riva Sanseverino et al., 2014). One of the main features of the artificial islands is their location at a latitude characterized by high values of average annual solar irradiation, approximately 5.4 kWh/m2 /day with reference to Sicily (compared to 3.6 and 4,7 kWh/m2 /day, respectively, in the north and centre-south Italy). The other special feature of the artificial islands is related to their distance from the coast (offshore), typically with high values of wind, both in terms of intensity (wind speed) and in terms of frequency (annual hours with wind). The reference parameter to classify the potential for exploitation of the wind resource is the capacity factor, which shows higher values for offshore plants (40–50%), compared to onshore plants (25–35%), resulting in better potential manufacturability electricity (3504–4380 kWh/kW/year offshore vs. 2190–3066 kWh/kW/year on-shore), in terms of daily average European values (IRENA, 2012). The location of the wind turbines on artificial islands allows, therefore, a high wind energy exploit, typical of offshore installations, but simplified in terms of construction, once the islands are built. The waste produced by communities in artificial islands will be collected for energy reproduction. In particular, the project shall study possible alternatives to minimize the transport of waste to the outside of the islands. Therefore, the hybrid system based on the different types of renewable energy sources must include the integration of solar, wind and waste. The heating demand will be covered by the integrated use of solar and heat energies co-generated by the combustion of biogas and syngas in engines. The islands could even be centres of electricity production for export by exploiting the high potential of photovoltaics and wind energy in these latitudes. Furthermore, these islands could become places of experimentation about innovative technologies for the exploitation of renewable energy, including wave energy (Fadaeenejad et al., 2014). Obviously, economic and environmental evaluation related to different possible scenarios will be done.
6 SAFETY AND SECURITY Safety and security will be central for both the construction phases and the overall management of the rail, road and maritime traffic. With respect to the latter, 3000 m spans allow navigation in complete safety and security, with distinct single maritime transit lanes for the east-west and west-east directions. 220
7 CONCLUSION Mediterranean Europe needs an intermodal mobility rotational system that will enhance, on the southern territory, the crossing of the main intermodal corridors of the Sea (Guarascio, 2014). For this reason, the Network of Schools of Engineering in the Mediterranean RMEI (Réseau des Ecoles des Ingénieurs Méditerranéen) has developed a research project called MedTracking. The aim of this project is to draw future scenarios of an Intermodal Mobility system in the Mediterranean, suggesting the idea of the continuity of transport between Africa and Europe in its central axis, which sees the Italian Peninsula, Sicily and Cape Bon in Tunisia as the natural alignment of the Euro-African corridor. TUNeIT would create a transcontinental territorial continuity between Europe and Africa, like others connecting Europe with Asia (tunnels and bridges on the Bosporus), as well as the stable connection between Europe and Africa across the Gibraltar Strait. The main objective of this important scenario is the reduction of the time for exchanging goods and the easier communication between Europe and Africa (from 20 to 2 days). CNI, RMEI, EAMC (Engineering Association of Mediterranean Countries), PAM (Parliament Assembly of the Mediterranean) and Terna SpA have shown considerable interest for this fascinating hypothesis. TUNeIT is much more of an idea than an engineering challenge. It involves a process of composition, definition, re-aggregation of historical, social, economic and cultural elements. On one hand, this fact implies the involvement of a variety of skills interacting with each other and with all subjects as actors. On the other hand, it requires attention towards the different relationships between existing and future systems. This is a complex operation that involves, in addition to all engineering aspects, a look into the poetic interpretation of reality, which goes beyond simple functionality. TUNeIT and the Messina Bridge go beyond simple connections (it is a question which opens a wide cultural debate). It is also questioned how they connect. They could create, with their formal, chromatic and material articulation, new worlds, new waterfronts and a new symbol, in order to recovering the centrality of the Mediterranean.
REFERENCES Anania, L., D’Agata, G., 2014. Materiali compositi innovativi per cavi di ponti strallati e sposesi. Supplemento a Galileo 216: 38. Rome, Italy. Andaloro, A.P.F., Salomone, R., Andaloro, L., Briguglio, N., Sparacia, S., 2012. Alternative energy scenarios for small islands: A case study from Salina Island (Aeolian Island, Southern Italy). Renewable Energy 47: 135–146. Bella, G., 2014. TUNeIT, possibili scenari energetici per le isole artificiali. Supplemento a Galileo 216: 36–37. Rome, Italy. Brito, Mc., Lobato, K., Nunes, P., Serra, F., 2014. Sustainable energy systems in an imaginary island. Renewable and Sustainable Energy Reviews 37: 229–Z42. Calise, F., Cipollina, A., Dentice d’Acadia, M., Piacentino, A., 2014, A novel renewable polygeneration system for a small Mediterranean volcanic islands for the combined production of energy and water. Dynamic simulation and assessment. Applied Energy, volume 135: 675–693. Carapellucci, R., Giordano, L., 2012. Modeling and optimization of an energy generation island based on renewable technologies and hydrogen storage system. International journal of hydrogen energy 37: 2081– 2093. Carpinteri, A., Pugno, N.M., 2008. Super-bridges suspended over carbon nanotube cables. Journal of Physic: Condensed Matter 20. Chen, F., Duic, N., Alves, L.M., da Gracia Carvalho, M., 2007. Renewislands – Renewable energy solution for islands. Renewable and Sustainable Energy Reviews 11: 1888–1902. Chua, K.J., Yang, W.M., Er, S.S., Ho, C.A., 2014. Sustainable energy systems for a remote island community. Applied Energy 113: 1752–1763.
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ENEA Agenzia Nazionale per le Nuove Tecnologie, 2003. I Trasporti del XXI Secolo. Il Futuro dei Trasporti Europei e Fattibilità del Collegamento Europa–Africa con Tunnel Sicilia-Tunisia. http://www.tunnelsiciliatunisia.enea.it/documents.htm Guarascio, M, 2014. Mobilità e Sicurezza del Mediterraneo. Supplemento a Galileo 216: 9. Rome, Italy. Fadaeeneyad, M., Shamsipour, R., Rokni, S.D., Gomes, C., 2014. New approaches in harnessing wave energy: With special attention to small islands. Renewable and Sustainable Energy Reviews 29: 345–354. IRENA – International Renewable Energy Agency, 2012. Renewable energy technologies: cost analysis series. Wind power, Volume1: Power Sector Issue 5/5. Liang, P., Xiangnan, W., 2011. Non-linear properties of carbon fiber-reinforced plastic stay cables, International Conference on Electric Technology and Civil Engineering, 978-1-4577-0290-7/11/2011. Meier, U., 2012. Carbon fiber composites polymer cables: Why? Why not? What if? Arab J. Sci. Eng 37: 399–411. Neves, D., Silva, C.A., Connors, S., 2014. Design and implementation of hybrid renewable energy system on micro-communities: A review on case studies. Renewable and Sustainable Energy Reviews 31, 935–946. Noistering, J.F., 2000 Carbon fiber composites as stay cables for bridges, Applied Composite Materials 7: 139–150. Pacheco, P., Adão da Fonseca, A., 1996. Effector Systems in Structures. Conceptual design of structures – Proc. IASS Symposium, Stuttgart: 339–346. Pacheco, P., Adão da Fonseca, A., 2002. Organic Prestressing. Journal of Structural Engineering, ASCE: 400–405. Riva Sanseverino, E., Riva Sanseverino R., Favuzza, S., Vaccaro, V., 2014. Near zero energy islands in the Mediterranean: Supporting policies and local obstacles. Energy Policy 66: 592–602. Siviero, E., 2004. Tra Scilla e Cariddi un ponte abitato sullo Stretto di Messina. Le Strade, n◦ 10. Siviero, E., 2014. The Tunisia-Italy Bridge across the Mediterranean. EIR November. Siviero, E. & Stocco A., Viaro, N., 2014. TUNeIT, Ponte Mediterraneo tra Tunisia e Sicilia. Supplemento a Galileo 216: 32–35. Rome, Italy. Wu, Z.S., Wang, X., 2008. Investigation on a thousand-meter scale cable-stayed bridge with fiber composite cables, Fourth International Conference on FRP Composites in civil Engineering.
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The Russky Bridge: Pylons design approach optimization L.V. Miklashevich The Mostovik Company Ltd., Omsk, Russian Federation
V.E. Rusanov The Siberian State Automobile and Highway Academy (SibADI), Omsk, Russian Federation
ABSTRACT: A design of long-span bridges with reinforced concrete (RC) elements requires special approaches, because the most approaches used do not always result in an effective solution. This paper deals with The Russky Bridge experience which was constructed in 2012 in Vladivostok, Russia. The RC pylons were analyzed and designed with an optimized approach. Pylons were constructed with divisions of certain length and elevated in certain time period. Pylons also had a force function during the construction, so designer had to be as quick as possible with calculations. It was considered to use generic mathematical approach to improve the rate of calculations. The proposed approach allowed to increase calculation speed 3500–4000 times average and made real estimation of complex stress state of pylons during elevation, construction and operational process.
1 INTRODUCTION A design of complicated bridge structures is usually followed by complex stress state analysis of bridge structures elements, especially such structures as cable stayed bridges and suspended bridges. For the most of cases it is possible to use simplified approach (CII 52-101-2003, Eurocode 2), which can give appropriate solution and stable results. Even minor attempt to improve such kind of simplified approach can lead to very complicated algorithms which demand more time to find the solution even if using of FEM modelling. For some structures it is impossible to do so due to variable of conditions and complex stress state of its elements. This may lead to a lack of analysis data and additional using of construction resources or impossibility to implement some technical solutions. 2 THE RUSSKY BRIDGE PARAMETERS One of the recent bridges built in Russian Federation – The Russky Bridge – has three world records among such kind of bridges (Fig. 1). Firstly, it is cable stayed bridge with the central span of 1104 m length. The second record – reinforced concrete pylons has a height of 316 m from the top of foundation plate to the top of pylon observation desk (Fig. 2). And the last one – the bridge construction time lasted 48 month. The cross section of pylons is variable, from 13000 × 7800 mm with 2000 mm thickness at the bottom and to 7000 × 7000 mm with 700 mm thickness at the top of pylons. The separate inclined pylon’s legs are supported with each other by three struts at a height of 65.765 m, 187.58 m and 272.08 m. They also merged at the top of the pylons. Pylon legs have a steel core between the heights of 187 m and 320 m for cable anchorage. The total amount of cables for each pylon is 84. The construction of the pylon is achieved with self-climbing formwork in 72 sections of 4.5 m. Each section required 5 to 14 days to construct. 223
Figure 1.
General view of The Russky Bridge at the final stage of construction.
During the concrete casting of pylon legs, stage by stage, the cables were assembled and pylon was under certain loads. All construction stages should provide safe and sufficient reliability. Other than the strength of concrete was varied during pylon construction.
3 PROPOSED APPROACH In consideration of mentioned above the analysis required a quick calculations of the stress state for 72 sections with an extrapolation of the whole pylon stress state for 80 stages during construction and operational process. Every stage should be analyzed for 8 load cases. As a result data designers had to provide an optimal reinforcement of pylon. Huge amount of calculations should be automated to save time. The usual approach (CII 52101-2003) for design of a cross section (Fig. 3) was inefficient because the calculations for one load case of each section with finding of the angle of neutral axis α and boundary strains εb,max , εb,min took about 30 min. It was considered to use generic mathematical approach with hypothesizes taken as a basis in two dimensional coordinate system XOY : 1. Plane cross-sections remain plane (Bernoulli hypothesis). 2. Strain in concrete is the same as in reinforcing bars at the same point (bond between the steel and concrete is sufficient to keep them acting together). 3. Bending moments in planes OX and OY reduced to bending moment in plane with angle α. 224
Figure 2.
Parameters and form of pylons.
225
in and external bending moment M ex vectors are opposed to each 4. Internal bending moment M other that is satisfy to the statement (1) for polar coordinate system:
5. Internal normal force N in is equal to external normal force N ex that is:
This lead to finding of maximal k value:
Internal forces was defined according to:
where σb (ε) = stresses in concrete; σs (ε) = stresses in steel bars and prestressing strands; ε(r ) = strains in concrete; ε(ri ) = strains in steel bars and prestressing strands; r = radius-vectors of concrete elemental areas; ri = radius-vectors of steel bars and prestressing strands elemental si = cross sectional area of steel bars and prestress b = cross sectional area of concrete; A areas; A ing strands; N b = normal forces in concrete elemental areas; N si = normal forces in steel bars and prestressing strands elemental areas. The curved line of external perimeter and internal hollows of the reinforced concrete cross section was approximated by straight lines and defined by coordinates of nodes. Reinforcing bars were defined by coordinates and effective diameters. The optimized approach involved to variation by three independent parameters: angle of neutral axis α, minimal and maximal cross section strain εmin and εmax , that is ε (α, εmin , εmax ). Stress-strain relation for concrete and steel was defined as bi- or tri-linear functions σ b (ε) and σ s (ε) according to Russian Building Codes (CII 35.13330.2011, CII 52-101-2003) (Fig. 4). in (α, εmin , εmax ) are funcAs a result, vector-functions of internal forces N in (α, εmin , εmax ) and M tions of three continuous variables. This allowed to define integrals (4) and (5) by using of the Green’s Theorem in inverse order with further transformation to analytical formulas. It helped to reduce considerably the total amount of computing operations. Thereby, for finding of maximal k value (3) it should be found a set of α, εmin , εmax parameters, which satisfy statements (1) and (2). Taken external bending moment as Mex = const, it should be defined an internal bending moment Min → max. This was done with assumption that at least one elemental area of concrete or reinforcing bars reaches their strain limit. This condition allowed to use a variable separation method for solving the statement (6). The main principle of solving the algorithm is the defining of an angle α and next defining pair of strains εmin , εmax . in (α, εmin , εmax ) are non-linear by all theirs variables Since vector-functions N in (α, εmin , εmax ), M numerical bisection method was used for finding the internal forces. Each iteration had to give more precise value of angle α. 226
Figure 3.
Design cross-section of pylons.
Figure 4.
Stress-strain relations for concrete and steel.
For equation (2) at least one of εmin and εmax values should reach its bound. If so, the vectorfunction N in (α, εmin , εmax ) could be assumed as a function:
where N ct is a normal force, calculated for bound values of εmin & εmax . In general the flow-chart of the designing process for one cross-section provided on Fig. 5. 4 CONCLUSIONS During work and analysis of design process it was concluded that: – relatively limited amount of input/output data required for each pylon cross section design, such as geometric parameters of concrete and reinforcement, strengths, “stress-strain” relations, load cases; 227
Figure 5.
Flow-chart of cross-section analysis.
– using of input/output data has repetitive nature; – a huge amount of calculations should be done for all load cases of the same type for majority of cross sections. According to the mentioned above it was considered some general requirements for the developed software: – input/output data should be stored for a long period and should be structurized; – it should be used the most rapid algorithms for calculations. This all lead to next main concepts: – using the analytical methods instead of numerical methods where it is possible; – using the object-oriented programming (OOP) approach; 228
– developing the class library for simplifying of iterative using the data and processing algorithms (it was developed class library which involve more than 30 classes and parameterized classes using C++ programming language); – Standard Template Library (STL) was widely used due to optimized standard for structurized vector-like data storing and processing (STL container classes used for same-type data storing, data processing was done also with STL procedures); – virtualisation of functions was avoided due to performance; – code was optimized to reach highest speed rate; – multiple-processor with parallel arithmetic approach was used to increase speed of calculations. Proposed approach made it possible to increase speed of calculations drastically: 3500…4000 times average for uniprocessor computers. Parallel arithmetic procedures improved these achievements further more. As a summary – designers spent 30…40 minutes instead of supposed 3 years for full analysis of two pylons after each stage of construction (8-processor computer was used). This approach made it real to meet a deadline in regular, fail-safe and prompt construction of The Russky Bridge. REFERENCES
EN 1992-1-1:2004:E Eurocode 2: Design of concrete structures – Part 1-1: General rules and rules for buildings. EN 1992-2:2005:E Eurocode 2: Design of concrete structures – Concrete bridges – Design and detailing rules.
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Bridge across the Waschmühl Valley, Kaiserslautern, Germany: A harmonic symbiosis between a historic monument and a new innovative bridge K. Humpf, V. Angelmaier & W. Eilzer Leonhardt, Andrä und Partner Beratende Ingenieure VBI AG, Stuttgart, Germany
ABSTRACT: In the course of the extension of German federal highway A6 Mannheim – Saarbrücken from 4 to 6 lanes the existing Bridge across the Waschmühl Valley near Kaiserslautern has to be maintained and widened. For that reason a new bridge had to be built next to the historic sandstone arch bridge, built between 1935 and 1937 by architect Paul Bonatz. Due to this particular situation and the aesthetic importance of the historic bridge a design competition the client invited for a design competition. The winning design “… adapts the existing bridge in a brilliant way; on the other hand the new bridge represents an independent striking and modern structure.” (Jury’s comment). The design is based on the following principles of design and construction: semi-integral construction, extradosed beam, superstructure grillage, appropriate implementation of materials and perceptibility.
1 A HARMONIC SYMBIOSIS BETWEEN HISTORIC MONUMENT AND RECONSTRUCTION In the course of the extension of federal highway A6 Mannheim – Saarbrücken from 4 to 6 lanes the existing Bridge across the Waschmühl Valley near Kaiserslautern has to be maintained and widened. For that reason a new bridge had to be built next to the historic sandstone arch bridge. The historic bridge, built between 1935 and 1937 in collaboration with architect Paul Bonatz, is regarded as successful synthesis of engineering, architecture and landscape design and was included in the list of historic monuments in 1984.
Figure 1. Waschmühl Viaduct (1937) of Paul Bonatz, which stands under monument protection.
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Due to this particular situation and the aesthetic importance of the bridge designed by Paul Bonatz the Highway Authority Landesbetrieb Mobilität Kaiserslautern invited to a design competition. The jury gave the following explanation for its decision for the winning extradosed bridge: “This design clearly contrasts with the existing historic arch bridge in form and structure. But at the same time it adopts the structure of the existing bridge and emphasizes its rhythm and design. The small number of piers and the very slender extradosed superstructure offer an open view on the existing bridge. On the one hand the designer succeeded in adapting the existing bridge in a brilliant way; on the other hand the new bridge represents an independent striking and modern structure.” During the design process focus was laid on a respectful treatment with the bridge designed by Paul Bonatz regarding its quality and design. The primary objective was to create a structure with an own identity that fits perfectly in the landscape but at the same time forms a suitable counterpart to the historic monument. The design is based on the following principles of design and construction. 1.1 Semi-integral construction – Avoidance of bearings (except in the abutments) – Rigid connection between reinforced concrete piers and steel (composite) superstructure (Fig. 2) – Longitudinal and vertical rigid framework that allow extremely slender piers (reduction of buckling length)
Figure 2. Winning design: Elevation north and south and layout.
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1.2 Extradosed beam – Static system of a haunched girder with 4 spans (44.90 m – 68.10 m – 68.10 m – 45.55 m) – Disintegration of the haunch by steel posts and cable stays (Fig. 5)
Figure 3.
Rigid connection via anchor plate.
Figure 4.
Cross-section.
Figure 5.
Haunched birder.
233
Figure 6.
Bottom view.
Figure 7.
High-stressed tension members as high-tensile parallel-wire strands.
1.3 Superstructure grillage – Two stiffening girders as seal-welded box girders and cross beams in a distance of 3.25 m (Fig. 4) – 35 cm thick in-situ concrete deck – Composite action only in transversal direction 1.4 Appropriate implement of materials – High-stressed tension members as high-tensile parallel-wire strands (Fig. 7) – Girder grillage of welded construction steel 234
Figure 8.
Pylon with stiffeners placed outside – transparency.
Figure 9. angle.
Separation of material between the piers (concrete) and superstructure (steel) and cables with flat
– Realization of a rigid frame to the concrete piers via steel elements with shear studs and concrete dowels 1.5 Perceptability – Pylon with stiffeners placed outside – transparency (Fig. 8) – Separation of material between the piers (concrete) and superstructure (steel) – the system of the continuous beam become apparent (Fig. 9) – Cables with flat angle – differentiation between the extradosed girder and a cable-stayed bridge (Fig. 9). – Parallel wire strands with a fixed anchor at the top of the pylon and adjustable anchors at the bottom of the superstructure (visible) – obviousness of the suspension – The soffit of the bridge is characterized by the grillage – architecturally elegant appearance – Piers in view inclined downwards – analogous to the piers of the arch bridge 235
Figure 10.
Setting of the system via force controlled construction.
Figure 11.
Bigger anchor heads allow to add additional strands.
The construction has been carried out conventionally by crane. After construction of the deck slab the prestressing tendons have been threaded and tensioned to the required prestressing force (Setting of the system via force controlled construction, Fig. 10). In times of scarcity of raw materials demands concerning sustainability of structures become more and more important. As the roadway slab interconnects only in transversal direction it can be replaced without additional action and risks. An afterwards strengthening of the construction using plates is possible without any difficulty. The design of the suspension cables has been carried out using bigger anchor heads, so that additional strands can be added if necessary (Fig. 11). In the area of the rigid frame there is strengthening potential as well by adding concrete studs. The decisive factor for securing the long-life cycle of the bridge is the integral construction method. Integral bridges – concrete bridges without joints and bearings – distinguish themselves by a high degree of robustness (durability, maintenance, structural safety and redundancy) and moreover offer new opportunities regarding the design. With the new Bridge across the Waschmühl Valley a design has been realized that has faced the interaction with the existing historical monument at “eye level”. 236
The harmonic symbiosis was not achieved by subordination but rather in terms of a compulsive common attitude towards the building culture. With all the contrariness in design and structure the result is a high degree of complemented consonance. The new bridge forms a convincing counterpart to the existing bridge that becomes apparent in a variety of aspects: – The new bridge was reduced to a simple truss model and thus appears light and filigree. The bridge thus complies with the construction principle of the historic bridge of Paul Bonatz, who designed a light and filigree arch bridge. – A new cable-stayed structure with tension members above the superstructure is contrasted by a compression arch under the existing carriage way. – For the different load carrying system also different materials have been chosen. For the construction under compression concrete with sandstone was used while the new structure under tension uses high-tensile strands. – The existing bridge is a bridge that fits in the landscape, but also dominates it. The new bridge in contrast intentionally keeps the valley open.
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In 2014 the Washmühltalbrücke was shortlisted for the Structural Award 2014 in the category Highway or Railway Bridge Structures and received a commendation in the category Bridges of the Ingenieurpreis des Deutschen Stahlbaus (Engineer Award of German Steel Construction). ACKNOWLEDGEMENTS Owner
Federal Republic of Germany Landesbetrieb Mobilität Kaiserslautern, Germany Structural Design Leonhardt, Andrä und Partner Beratende Ingenieure VBI AG, Stuttgart, Germany Architect AV1 Architekten GmbH, Kaiserslautern, Germany Contractor JV Waschmühltalbrücke Kaiserslautern: Alpine Bau Deutschland AG, Leuna, Germany Plauen Stahltechnologie GmbH, Plauen, Germany REFERENCES Angelmaier, V. 2014. Innovation neben Tradition – Neubau der Waschmühltalbrücke in Kaiserslautern. Jahrbuch Ingenieurbaukunst 2014, p. 112–119. Dokumentation: Realisierungswettbewerb „Gestaltung der Waschmühltalbrücke“ erstellt: LSV Kaiserslautern; 2007; In: www.lbm.rlp.de Eilzer, W., Priebe, V., Stockmann, R., Lutz, R., Angelmaier, V. 2010. Neubau der Waschmühltalbrücke – Harmonische Symbiose, Baudenkmal – Neubau. Stahlbau 2010, Jg. 79, Nr. 5, 2010, Seite 370–388. Eilzer, W., Lutz, R., Priebe, V. & Angelmaier, R. 2013. Neubau der Waschmühltalbrücke Kaiserslautern – Harmonische Symbiose Baudenkmal – Neubau. Beratende Ingenieure 7/8 2013, p. 41–49. Lutz, R., Angelmaier, V. 2010. Symbiose aus Denkmal und Neubau, Ergänzung der Waschmühltalbrücke. 10. Symposium Brückenbau Leipzig, Februar 2010. Lutz, R., Winkler, B. 2013. Neubau der Waschmühltalbrücke im Zuge der A 6 bei Kaiserslautern, ein nicht alltägliches Bauwerk. In: 23. Dresdner Brückenbausymposium, März 2013. Winkler, B. 2012. Bau der Waschmühltalbrücke im Zuge der A 6 – überspannter Stahldurchlaufträger. Brücken- und Ingenieurbautagung 2012, München, BMVBS.
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Viaduct over river Deba in the “Y-Basque” high speed railway line in the north of Spain F. Millanes, M. Ortega, P. Solera, H. Figueiredo & J. Ugarte Ideam, S.A. Madrid, Spain
ABSTRACT: The bridge over the river Deba, on the Bergara-Bergara stretch of the ‘Y-Basque’ High-Speed Railway Line in the north of Spain, is located on a deep valley with a maximum elevation difference between the line’s profile and the ground of over 90 m. The height of the bridge, the length of the valley (900 m) and its slopes, together with the lower utility restrictions (with various roads and pathways), and the Deba River itself, make it necessary to adopt long spans. The solution finally constructed was a deck made of a prestressed concrete box girder with a slight haunch near the supports and the following span distribution: 50 + 80 + 70 + 60 + 3 × 65 + 70 + 65 + 70 + 3 × 65 + 45 m, with a main span of 80 m. The bridge has been built by using a movable scaffolding, and the 80 m span represents the span length record so far for a High-Speed Railway span built in Spain using this technique.
1 INTRODUCTION AND ANTECEDENTS The viaduct over river Deba, on the Bergara-Bergara stretch, allows the crossing of the railway over the valley along which flows the river Deba nearby Bergara. The valley is quite deep, reaching a 90 m difference between the trace and the ground below. It is 900 m wide at the viaduct’s height, with sheer slopes that form a considerably symmetric V shape that is slightly broken between the piers P1 and P2 by the Vitoria/Gasteiz-Eibar highway. The trace of the high speed railway has several restraints underneath the viaduct that are determining on the definition of its spans. The viaduct crosses with a great skew angle over the Vitoria/Gasteiz-Eibar highway over the GI-627 and the GI-632 roads, over the river Deba and over the new GI-632 road. Also, it crosses over several local roads, one of them very close to E1 Abutment 1. Due to the great height of the viaduct and its location on a very visible valley, the solution for the viaduct over the river Deba needed to combine landscape integration minimising the impact to the valley, all of it achieving an erection process independent form the ground. The original project for the Bergara-Bergara stretch, awarded in 2007 by ETS to the consortium formed by IDEAM S.A. and Euroestudios, showed a singular solution that solved all the restraints mentioned before. The large spans due to the great amount of restraints underneath the viaduct, around 80 m, leaded to the proposal of a steel-concrete launched composite truss for the tender. The proposed span distribution was 50 + 70 + 60 + 70 + 110 + 180 + 110 + 3 × 60 + 50 m, with two central V shape great piers, that reduced the number of the supports, overcoming the central zone with a great 180 m length span between supports (Fig. 1). This singular solution was adapted during the first typology studies, trying to reduce the visual impact of the V shape piers, simplifying and cheapening its execution with a steel-concrete composite truss with a 50 + 80 + 3 × 75 + 100 + 110 + 100 + 2 × 80 + 70 m span distribution and vertical piers (Fig. 2). After many modifications on the route that leaded to the delay of the basic project on almost 2 years, at the beginning of 2009, in the middle of the economic crisis, the restraints and the 239
Figure 1. tender.
Photomontage of the steel-concrete composite truss with two V shape central piers, winner of the
Figure 2.
Photomontage of the steel-concrete composite truss with vertical piers.
authority’s priorities were modified. This lead to the re-study of the typological alternatives to reduce the span lengths, taking into account the following concerns: – Mantain, as far as it was possible the location of the existing highway and roads, as well as taking care of the protected area of the river Deba. – Achieve a construction process that was independent from the ground. – Maintain the symbolic nature of the viaduct on the Basque High-Speed Railway. – Coordinate the previous restraints with the economy of the solution finally chosen. With this new restraints and having to maintain a minimum span of 80 m due to the skewed crossing of the Vitoria/Gasteiz-Eibar highway, the final solution was designed to be a steel-composite launched girder with two double T metallic beams connected by top and bottom concrete slabs that close the torsional flow of the girder. The bottom slab increased its depth above piers to achieve double composite action. Similar solutions were developed before by IDEAM for the Arroyo las Piedras viaduct, first steel-concrete composite viaduct of the high-speed Spanish railway, and the Archidona viaduct, the world longest viaduct without interior expansion joints and rail expansion devices. The total depth of the girder was 5,50 m, and the depth of the metallic beams 5,04 m, little below the standard transportation limits. It had a typical span of 80 m, with a very uniform distribution of 50 + 10 × 80 + 50 m with 11 vertical piers. Figures 3 and 4 show the elevation and the photomontage of the solution. The span distribution, very balanced and homogeneous with 10 typical spans 80 m length, achieved the crossing over the Vitoria/Gasteiz-Eibar highway, as well as the crossing over the roads GI-627 and GI-632 and the river Deba, only affecting to the local link road between Bergara and the Viotria/Gasteiz-Eibar highway (new GI-632 road). This affection, technically solvable without many difficulties, permitted to achieve an harmonic and balanced structure, with an absolutely homogeneous span distribution and an optimum structural design (Fig. 5). 240
Figure 3.
Elevation of the steel-concrete composite viaduct.
Figure 4.
Photomontage of the launched steel-concrete composite solution with double T metallic beams.
Figure 5.
Plan of the viaduct and its surroundings.
2 DESCRIPTION OF THE SOLUTION FINALLY CONSTRUCTED 2.1 Description of the superstructure After the awarding of the stretch Bergara-Bergara to the Abergara consortium, during the late 2011 it was proposed the possibility of the modification of the span distribution of the viaduct over the river Deba, oriented to avoid the affection to the link road between Bergara and the Vitoria/Gasteiz-Eibar highway, and trying to reduce slightly the spans on those areas that permit it. The span distribution proposed by the contractor, with typical spans of 70 m and one span 80 m length over the highway, has permitted the modification of the structure to a slightly cheaper solution, based on a prestressed box concrete deck. This new alternative has only been possible during the final execution project due to the recent progress of the technology used for the auxiliary resources needed to construct concrete decks with spans of this length by means of a movable scaffolding, developed by the Group Puentes y Calzadas. This group was the subcontractor of the Abergara Consortium for the construction of the deck, and in the last few years has developed a movable scaffolding capable of making high-speed concrete decks with spans up to 70–80 m length. Until the recent apparition of this movable scaffolding, on the Spanish High Speed Railway there was only experience on concrete prestressed box girder executed by means of a movable scaffolding with spans up to 65–66 m length, reason way during the definition of the original project between 2007 and 2009, this typology was dismissed for the viaduct over the river Deba. 241
Figure 6.
Elevation and plan of the constructed solution of prestressed concrete.
Figure 7.
Elevation view of the viaduct over river Deba after its completion.
The solution finally projected by IDEAM for the Abergara consortium, and that was recently satisfactorily tested on the static and dynamic load tests, has the following span distribution: 50 + 80 + 70 + 60 + 3 × 65 + 70 + 65 + 70 + 3 × 65 + 45 m, with a total length of 900 m and the largest span constructed by means of movable scaffolding so far on the Spanish high-speed railway. As it was mentioned before, the viaduct has been executed by means of a movable scaffolding from the Abutment 2 towards the Aabutment 1, having a single fixation point against horizontal forces on the Abutment 2 and a rail expansion device coinciding with the viaduct’s expansion joint on Abutment 1. The viaduct has 13 piers on the valley, with heights between 23 and 86 m. The cross section is a prestressed concrete box girder of variable depth nearby supports, 3,94 m depth on the mid-span area and 5,94 m depth over the piers. Therefore, it achieves a depth to span ratio for the largest span of 80 m of 1/21,05 on mid-span, and 1/13,46 over the piers (Fig. 8). The variable depth is limited to the first 15 m next to the piers, to achieve uniformity for the formwork and the possibility of modifying the length of the span only by the addition of new casts, without the modification of their form. This way, a large length of the mid-span area has constant depth (53% on the 65 m length spans and 62% on the largest 80 m span), conferring a slenderness sensation to the deck. The base width of the box section is variable nearby the piers to maintain constant the web’s inclination, while remains constant with 6,6 m on the mid-span area. The webs have a constant inclination that achieves an horizontal projection of 0,955 m in the constant depth area of midspan. The top width of the box section without the cantilevers of the top slab is approximately 8,521 m. This lateral cantilevers have a length of 2,739 m and its thickness is variable from 0,41 m near the webs to 0,20 m on the free ends. The interior trapezoidal void of the cross section is chamfered on the intersection between the webs and the top and bottom slabs as it is usual on this kind of structures, to improve the behaviour against transverse flexure and shear transfer, as well as to permit the disposition of the prestressing anchorages. 242
Figure 8. Typical mid-span cross section (left) and cross section over piers (right).
Figures 9a, 9b & 9c. View of one bearing, measurement system of the longitudinal movements and protection system against dirt and dust.
Over the piers, the concrete solid diaphragm has voids to permit the transit along the interior of the deck, achieving the deck to be inspectionable. The same way, there are vertical accesses to the top of the piers that permit the inspection and substitution of the bearings of the viaduct (Fig. 8). 2.2 Description of the substructure 2.2.1 Bearings The bearings of the viaduct are spherical, one fixed and the other free on the transverse direction on each pier/abutment, while on the longitudinal direction all the bearings are free except those of the 4 highest central piers (piers P6 to P9). In these piers the bearings are longitudinally fixed, achieving an elastic support of the head of the 4 highest piers that permits their deformation control. This way, the maximum longitudinal movements of the head of these 4 piers are limited to those due to the imposed movements of the deck: expansions (thermal) or contractions (thermal + shrinkage + creep) accumulated from the fixation point on the E2 abutment. This longitudinal support permits the control of the second order deformations of the highest piers due to buckling instability, reducing the longitudinal bending moments and improving their structural behavior. All the bearings follow the classical disposition that locates the stainless steel tray on the deck supported on the sliding plate that is fixed to the pier. Both, the stainless steel tray and the sliding plate are located horizontally, except on the E1 abutment, where they follow the deck’s inclination in order to avoid vertical movements of the expansion joint and thus of the rail, what could lead to comfort losses for the passengers. To achieve that the stainless steel tray, longer than the sliding plate to permit the movement of the deck, remains free of dirt and dust over the years in order to avoid the wearing of the sliding material and the consequent increase of the friction between the deck and the bearing system, a flexible protection system has been installed on the spherical bearings. The protection system permits the longitudinal movements of the deck avoiding the presence of dirt or dust between the stainless steel tray and the sliding plate, being also removable and replaceable (Fig. 9c). The sliding plate of the spherical bearings is made of polyethylene of high molecular density, with improved characteristics comparing to the classical Teflon plate, with guaranteed maximum 243
Figures 10a, 10b & 10c.
Geometrical definition, lateral view and skewed view of the piers.
friction coefficients of 2% with very low temperatures, below the classical 3% of the PTFE Teflon plates. Also, these kind of sliding plates have better durability, what permits the reduction of the future maintenance of the bearings. To achieve the control of the relative longitudinal movements between the stainless steel tray and the sliding plate, the bearings have lateral measurement rulers. This permits the knowledge of the exact relative movement between the deck and the piers during routine inspections. All the bearings are removable and replaceable. The design of the viaduct has been adapted from the beginning to permit this replacement in case it becomes necessary. 2.2.2 Piers and abutments The conception and design process of the piers was laborious, searching during project stages the complex equilibrium between the aesthetics and the construction simplicity. Taking into account this last aspect, special attention has been taken on the design of the piers, trying to achieve an elegant solution that gets integrated with the landscape, with a variable geometry of smooth shapes that differentiates from the classical rectangular wall piers, too sober and non-suitable for a valley as high and visible as the one of the river Deba. On a front view (Figs. 10a and 10c), the piers of the viaduct increase their width smoothly, with a circular variation, with a minimum width of 6,50 m on the top. On the longitudinal direction of the viaduct, the depth of the piers increases linearly, from a minimum depth of 3,50 m on the top to more than 6 m on the highest piers (Fig. 10b). Looking to the cross section, the classical rectangular shape is chamfered to achieve an octagonal cross section with always parallel edges, what creates a series of oblique planes that follow the geometry of the depth variation with variable width, giving to the pier a less abrupt aesthetic, much more elegant than the classical rectangular piers (Fig. 10c). On the centre of the lateral faces a V shape void of variable dimensions but with parallel faces has been designed. This void opens gently from the top of the pier to the bottom, what gives to the pier a spatial nature, achieving an elegant and slender geometry. The interior of the piers is empty with walls of variable thickness: 0,30 m on the top 25 m, 0,40 m on the next 25 m and 0,50 m on the base of the highest piers. The highest pier reaches 86 m next to the river Deba. The Abutment 1 accommodates the rail expansion device and it has been designed as an empty closed box. The Abutment 2 is the fixation point of the deck and on it there is rail continuity. All the 244
Figures 11 and 12 . View of the front of one stage and hanging system of the movable scaffolding, and general view of the viaduct during its execution with the movable scaffolding.
foundations are superficial, except those of the piers next to the river Deba, piers P7 and P8, that are supported by pile caps with 15 piles of 1,80 m of diameter. 3 CONSTRUCTION PROCESS Taking into account the construction process of the viaduct many aspects have been considered. Between them, two main aspects have to be outlined: the load transmission system between the movable scaffolding’s support rings and the piers, and the construction sequence of the viaduct. 3.1 Construction sequence of the viaduct The construction process of the viaduct was divided in 14 stages. In each one of them one of the spans of the viaduct and a cantilevered stretch of the next span were executed, by means of the movable metallic scaffolding of 155 m of total length. During the stages of disposition of the concrete, the movable scaffolding was simply supported on the free end of the previously executed cantilever by a hanging system (Fig. 11), and on the metallic structure (ring) situated on the next pier. The typical span for the metallic beam that forms the movable scaffolding was 50 m for the spans of 65 and 70 m, and of 55 m for the span of 80 m. The cantilevers of each stage varied depending on the length of the next span. This way it was achieved to maintain constant the typical span of the movable scaffolding during the concrete casting phases with spans of different length. The stages with cantilevers 15 m long were executed following the classical sequence of two concrete casting stages: first the partial cross section formed by the webs and the bottom slab was executed, and second, once the first partial cross section achieved a minimum strength, the top slab was executed. On the other hand, on the stages with cantilevers 20 or 25 m long that correspond to the previous stages to spans 70 and 80 m length, and due to structural restraints of the movable scaffolding, it was proposed an execution process based on 4 concrete casting stages. First the bottom slab and the webs of the area nearby the pier were casted (corresponding to the cantilever and its symmetrical length towards mid-span). Then the top slab of this area was executed. When this element achieved a minimum strength, the first post-tensioning family was prestressed, that corresponded to the negative prestressing over the pier. Third, the bottom slab and the webs of the mid-span region were casted (between the cantilever of the previous stage and the pier area casted on the first place). Then a second family of post-tensioning was prestressed (corresponding to the positive prestressing along the mid-span region). Fourth, the top slab of the mid-span area was casted, followed by the presetressing of the rest of the post-tensioning families. At this moment, the hanging system on the previous cantilever was removed and the metallic beam was lowered and then launched to the position for the execution of the next stage. Figure 11 shows the front of one stage (the free end of the cantilever), the transverse hanging beam of the movable scaffolding, as well as the prestressing anchorages of the continuity tendons 245
Figure 13.
General view of the viaduct after its completion.
(on the webs), and the negative tendons (on the top slab). The continuity between different stages of parabolic tendons was achieved by means of crossings on the web and wedges with double anchorages instead of the classical couplers, due to construction preferences of the subcontractor of the deck. 3.2 Support of the movable scaffolding over the piers The support rings of the movable scaffolding over the piers were two triangular elements tied by a prestressed horizontal brace that linked both triangular elements. The triangular elements were formed by two inclined struts that carried the load to the pier, while another vertical strut equilibrated the triangle between the inclined struts and the horizontal brace. The load of the movable scaffolding and the concrete was transmitted to the pier by a top and a bottom support on both lateral faces of the pier. The top support only introduced horizontal forces, while the bottom support introduced both, vertical and horizontal forces. As it was mentioned before, the top support transmited the transverse reactions to the pile, as well as the possible longitudinal reactions due to the launching of the metallic beams. This support was maintained in compression by means of the prestressed brace. The bottom support transmited the inclined compression that was carried by the inclined struts by a transverse force normal to the pier and by a vertical force applied on a little hole situated on the pier with this objective. 4 INSTRUMENTATION OF THE VIADUCT DURING EXECUTION In order to control the real stresses caused during execution on the deck, the hanging of the movable scaffolding was instrumented as well as the reaction on the support ring. This way, during all construction stages the real loads and the reaction distribution between the hanging system and the support ring were known. This control permitted the verification of the hypothesis made for the structural analysis during project stages, and the observation of how an important part of the casting loads were supported by the previously casted partial cross sections, avoiding the movable scaffolding to carry the 100% of the loads due to the concrete casting. As an overall conclusion, the instrumentation verified and confirmed the adequate behavior of the viaduct, the movable scaffolding and the interaction between them. The same way, 4 of the highest piers were instrumented to measure the movements on the top of the piers and efforts on the base. 246
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Structural solutions and construction methods for the main crossing of the Mersey Gateway Bridge Project G.D. Moir, S.H. Jang & Jacob Seo Merseylink Civil Contractors (Samsung C&T), Widnes, UK
P. Sanders Flint & Neill (COWI) Ltd., London, UK
ABSTRACT: The Mersey Gateway Bridge Project includes a 2.25 km landmark crossing of the River Mersey between Runcorn and Widnes in Northwest England. It was tendered using a Competitive Dialogue process in-tended to provide best value whilst meeting client specified core requirements. Merseylink studied various structural solutions for the main river crossing and selected a mono pylon multi-span cable stayed bridge with a single trapezoidal box girder. Main Bridge construction employs balanced cantilever travellers to construct an in-situ concrete deck whilst approach viaducts are to be constructed using a Movable Scaf-fold System (MSS). This paper discusses the process of the development and evaluation of the various potential alternative structural solutions and construction methodologies to meet the client’s requirements and provide best value for money. With construction now underway an evaluation of key construction challenges and solutions is made.
1 INTRODUCTION The Mersey Gateway Bridge Project comprises a central signature cable stayed bridge of 1000 m length along with approach viaducts of 706 m and 544 m to the north and south of the River Mersey respectively. This Main Crossing is linked to the wider highway network with an upgrading of approximately 8 km of the A-road network in Runcorn and Widnes. In addition to the main crossing structures eight single or multi-span highway bridges and viaducts and a large number minor highway structures are also to be provided. The six lane main bridge provides a significant capacity upgrade, when compared to the existing Silver Jubilee Bridge arrangement, and provides resilience to the wider highway network in the area. Additional key objectives of the project included improving air quality, enhancing the urban environment, minimising toll levels, maximising local and regional economic growth opportunities, improving transport links across the estuary and encouraging cycling and walking facilities.
2 PROCUREMENT AND BIDDING PROCESS The scheme was procured by Halton Borough Council (HBC) on a design, build, finance, operate basis with a 30 year concession period. A Competitive Dialogue tender procedure was followed by further negotiation with the preferred bidder which led to financial close and contract award in March 2014. The Merseylink Consortium includes Merseylink Civil Contractors (FCC, Samsung and Kier) and their design team (Flint & Neill and Fhecor, responsible for the Main Crossing, and URS and Eptisa, responsible for the Landside Works). 247
Figure 1.
In advance of the tender process a Reference Design had been produced and planning consent obtained. This Reference Design was then further refined into the Illustrative Design which provided flexibility to bidders and allowed value engineering solutions to be pursued. HBC had set a goal for the design and construction element at tender stage to ‘achieve best value using a performance based specification to provide maximum flexibility’. This was to be achieved through closely monitored requirements and regular dialogue sessions with the bidders. 2.1 Employer’s requirements Design quality was important to HBC as the promoter of the Project. The crossing, and in particular the new bridge, is be emblematic of the Borough. HBC intended that the quality of design, including the appearance of the new bridge, its approaches and landscape, would have a significant positive influence on the current environment and its future development. Significant Employer’s Requirements were established which were non-negotiable, these were relating to fixed planning conditions which had been imposed as a result of the planning approval process. They included specific land use, architectural principles and environmental restrictions which neither HBC nor Merseylink could change. These core requirements had significant potential to drive structural solutions. 2.2 Environmental Environmental studies identified the highly dynamic nature of the Mersey estuary above the Runcorn narrows with substantial movement of sediment and evidence of the deeper channels 248
migrating across the wider estuary. One of the primary objectives of the project was to ensure that the natural behaviour of the estuary would not be inhibited by the presence of the bridge which was achieved by limiting possible locations of foundations to “grey zones” within the estuary and ensuring minimum span lengths for approach viaducts. The Mersey estuary at this location is bounded by low lying salt marshes to the north and south. These marshes include rare specimens of flora and provide an important habitat for migratory birds. In addition, the wider estuary includes a European designated Special Protection Area, an internationally designated RAMSAR site and a Site of Special Scientific Interest (SSSI). It was stipulated within the Environmental Statement that as little as possible impact should be imposed upon these areas due to the new crossing, with the ethos of ‘treading lightly across the salt marshes’ as one of the project goals. Figure 1 shows the situation of the bridge within the upper Mersey estuary. 3 OPTIONS EVALUATION Through multiple technical and commercial submissions throughout the Competitive Dialogue process, Merseylink set out to investigate all possible structural solutions to the main crossing elements, including some which deliberately challenged the Employer’s Requirements and existing planning consents. By taking this approach, the team was able to test the flexibility of the Employer’s Requirements in relation to each other and to potential cost advantages of the various solutions. 3.1 Main crossing configuration Initially main bridge and approach viaducts were reviewed individually. The most significant main bridge options were: Multiple short spans: This configuration comprised 40 m spans across the full estuary. Whilst economically attractive there was a significant disadvantage in terms of environmental impact and risk. This configuration failed to remain within the designated ‘grey zones’ for river foundations and the environmental impact of multiple piers in the river led to an increase in scour and reduction in mobility of the river channels therefore leading to an unacceptable planning risk. Five span four pylon bridge adopting an extradosed configuration combining haunched deck elements and cable stayed spans. Adopting this configuration it was possible to remain within the designated ‘grey zones’ in the river. However, the requirement for a maximum three piers within the river was not met. A four span, three pylon bridge using cable stayed spans followed the reference and illustrative designs closely and was the most obviously compliant configuration. Nevertheless, multiple variations on tower and deck type were available as discussed below. Three span, two pylon cable stayed configurations for the main bridge were reviewed and were subject to rigorous evaluation. However, due to the increase in main span length this solution proved to be of little economic benefit. Additionally the necessary increase in pylon height led to an infringement of the allowable vertical envelope, determined by the close proximity of Liverpool Airport at Speke. 3.2 Pylon and deck Through dialogue it became clear that the compliant configuration adopting three unequal height pylons would be the only practical solution which would fulfil HBC’s requirements. Varying construction methodologies were investigated including: Balanced Cantilevers – Both steel composite and full concrete options were considered. Full width steel or precast concrete deck elements lifted directly into position would have been an 249
Figure 2.
3 pylon solution.
economical solution, were it not for the shallow river levels, severely limiting vessel access in the estuary. Incremental Launching – over a modified consistent radius alignment was unfortunately found not to be possible as the land availability was insufficient. Movable Scaffold Systems for the main crossing would require the adoption of temporary piers within the river. This was found to be an uneconomical solution. The main options investigated are illustrated in Figure 3. Through detailed evaluation and dialogue with HBC, the single box in-situ balanced cantilever option was finally adopted. Other options failed to comply with planning restrictions or proved economically unattractive.
3.3 Approach viaduct configuration With a length of approximately 706 m and 544 m for the north and south respectively the approach viaduct options were linked closely to the deck design options for the estuary bridge. Multiple precast concrete beams of spans up to around 40 m have proven to be economical in viaduct design both in the UK and internationally. Standard typical I, U and W beams are used extensively and each was considered with an in-situ reinforced concrete deck slab. Longer span composite precast I beams with in-situ stiches and deck were considered. Span lengths of up to 55 m were considered economically viable using this method. Composite steel I beams with an in-situ concrete deck was investigated. Span lengths of around 50 m were evaluated as providing an economical configuration. Steel and concrete box girders were evaluated, both single and twin boxes being considered. Both in-situ and precast segmental concrete were considered, there being many erection variations 250
Figure 3.
Bridge deck options.
available for this type of viaduct. In-situ concrete on a Movable Scaffold System (MSS) and precast segmental using a full span erection girder showed an economical span length of around 55 m would be achievable. Nevertheless, through the dialogue process it became clear that the environmental aspect of treading lightly over the salt marsh, combined with a strict requirement to adhere to a minimum 60 m span length, dictated that shorter span configurations would not be considered a conforming solution and were thus discounted. In the reference and illustrative designs the Manchester Ship Canal (MSC) was bridged with a single span of around 100 m. The limiting factor for this span being minimum horizontal clearance of 20 m to the canal and a vertical clearance of 28 m for shipping. The vertical clearance here dictating that the maximum height of the alignment would be at this point. HBC had indicated that horizontal clearance was negotiable dependent upon the MSC operator. It was found that a span length of 70 m would be acceptable by adopting either a steel composite or concrete box girder construction.
3.4 Option evaluation A combined evaluation was carried out considering both the estuary crossing and approach viaducts and is summarised in Figure 4. 251
Figure 4.
Option evaluation matrix.
Figure 5. Typical deck cross section.
When considered together, in order to meet the architectural stipulations of the Design and Access Statement imposed under the planning conditions, a main bridge configuration of monopylon with a single cell box girder concrete deck along with a similar deck configuration cast in-situ on an MSS for the approach viaducts was the preferred option (Option 4 in Figure 4). This allowed simple transition geometry between approach and main bridge and with approach spans of 70 m met both environmental requirements and the imposed limitations on the crossing of the Manchester Ship Canal. 4 CONSTRUCTION METHODOLOGIES 4.1 Main bridge construction The in-situ concrete cable stayed bridge deck is to be constructed using a conventional balanced cantilever method, adopting formwork travellers. Segments are constructed in a balanced configuration each side of the pylons. 252
Figure 6.
Formwork traveller.
Due to the 33 m width of the deck a segment length of 6 m was chosen limiting the segment weight to typically around 320 tonnes. As the length of cantilever increases, the unbalanced moment due to the construction sequence and wind loading will proportionally increase, for this reason temporary vertical piers are provided when the cantilever length exceeds a certain value. 4.1.1 Form traveller Whilst it is common for a derrick crane arrangement to support the movable formwork at the cantilever ends, the limited access below the deck to allow dismantling and removal with this arrangement led to the decision to adopt C-frame type travellers in this instance. There are advantages to the use of the C-frame type traveller for the project, it provides a more robust supporting system which allows free access to the formwork area from the top, it also permits use of the same formwork system for deck construction at the pylons. Three sets of travellers are to be used in order to meet the construction programme. 4.1.2 Construction cycle There is a target cycle time of 5 working days for the deck construction which is a challenging cycle given the complexity of rebar, PT and stay cable installation along with the sequence of form traveller operation. In order to fit into tight cycle time and to reduce at-height work, prefabricated rebar cages have been considered for bottom slab, web and top cantilevers. A full prefabricated cage was considered but was found to be impractical due to the number of variants in the internal geometry and post tensioning layouts. 4.2 Approach viaduct The approach viaducts are to be constructed over the salt marsh areas where ground conditions are poor. The MSS method allows construction, span by span, using a formwork system attached to special self-launching girder. Since this method does not require traditional false work for support from ground level it is widely applied for river crossings, flyovers located in congested areas or on poor ground conditions where substantial temporary works would otherwise be required. Where a normal MSS would construct a typical 40–50 m long span with a 2 lane cross section the approach viaducts demand a span configuration of 70 m length and a 6 lane cross section, thus 253
Figure 7. Traveller construction cycle.
Figure 8.
MSS.
the MSS girder in this instance is at the upper end of the technique. In order to reduce cycle times, deck construction will be phased such that a central trough is cast in advance of the deck slab and transverse beams, thus releasing the MSS earlier. The cantilever arms of the deck are to be constructed using separate ‘wing traveller’ equipment following behind the MSS. By adopting the phased casting sequence it is also possible to optimise the size of the MSS. 4.3 Construction sequence Due to the scale and complexity of the main crossing this element defines the critical path of the overall project programme. The main cable stayed bridge and approach viaducts are to be constructed simultaneously. Main bridge foundations commence with the south pylon foundations, spread foundations seated on a relatively shallow sandstone bearing strata, constructed within sheet pile cofferdams. North and central pylon foundations follow closely behind. Pylon and deck construction is to be carried out at all three locations simultaneously necessitating three sets of form travellers. Approach viaduct construction commences from the north approach viaduct, where the MSS is assembled immediately behind the north abutment. In order to avoid interference with the MSS 254
Figure 9.
Phased casting sequence of approach viaduct decks.
Figure 10. Wing traveller.
the abutment is constructed sequentially allowing advance from the lowest point in the alignment towards the tie down pier on the river bank. Upon completion of the north approach the MSS is withdrawn to a location on the saltmarshes where lowering and dismantling can take place. The south approach viaduct is similarly constructed from the south towards the main bridge tie down pier. 255
Figure 11.
Main crossing time chainage.
A summary of the programme can be seen in figure 11 in time chainage format. ACKNOWLEDGEMENTS The structural solution adopted for the Mersey Gateway Bridge has been selected through close collaboration between Client, Contractor and Designer. The selected design meets the Employer’s Requirements, Clients design, construction, planning and environmental requirements. The construction methods adopted provide programme security through tried and tested techniques and safe working methods. The authors wish to thank HBC and the Mersey Gateway Crossings Board team, for their permission to publish this paper and recognise the assistance provided by the project designers and construction team at the Merseylink Consortium. REFERENCES Marginson A. 2011. Preliminary Design of the Mersey Gateway Structures. IABSE Symposium, London. Sanders P. 2014. Design and Construction of the Mersey Gateway Bridge. IABSE Symposium, Madrid.
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Design of the long-span footbridge over the Bug River in Niemirów J. Biliszczuk & J. Onysyk Wrocław University of Technology/Research & Design Office Mosty-Wrocław, Wrocław, Poland
W. Barcik, P. Prabucki, K. Ste˛pie´n, J. Szczepa´nski, R. Toczkiewicz, A. Tukendorf, K. Tukendorf & P. Wo´zny Research & Design Office Mosty-Wrocław, Wrocław, Poland
ABSTRACT: This paper presents design details of the footbridge over the Bug River in Niemirów, near Polish-Belorussian border. Final design of the footbridge was based on the architectural concept selected in a competition organized in 2013 by the Municipality of Mielnik. The footbridge is a three-span stress-ribbon suspension structure with spans of 91.23 + 135.00 + 91.23 m and a total length of 336.5 m. The deck of variable width consists of reinforced concrete precast panels posttensioned by internal cables, placed on main bearing tendons and is additionally supported by suspension cables. Slender concrete piers have a “Y” shape. The paper also briefly describes numerical analyses and proposed construction technology of the footbridge.
1 INTRODUCTION Architectural concept of the footbridge over the Bug River in Niemirów (Poland) was selected in a competition organized in 2013 by the Municipality of Mielnik. The winning proposal, referring to the shape of the London Millennium Footbridge (Dallard et al. 2001), was a suspension hybrid structure of untypical configuration. The distinctness of the concept consisted in the combination of a typical stress-ribbon structure with a suspension structure, giving the footbridge the shape of a waving ribbon (Figure 1). Therefore, the authors of the winning concept gave it the name “White Ribbon of the Bug River” (Banaszewski et al. 2013). This paper presents the final design of the footbridge in Niemirów developed by the Research & Design Office Mosty-Wrocław.
Figure 1. Winning concept of footbridge in Niemirów – “White Ribbon of the Bug River” (Banaszewski et al. 2013).
257
Figure 2.
Final configuration of footbridge over the Bug River.
2 STRUCTURAL CONFIGURATION 2.1 General characteristics At the stage of developing the final design of the footbridge some changes were proposed in comparison to the initial concept. As a result of detailed static analysis, the following modifications were introduced: – the stress-ribbon concrete deck placed on six bearing cables running continuously along the entire length of the structure under the deck and anchored in abutments was considered the main load bearing element of the footbridge; – additional slender suspension cables carrying live loads, anchored in the deck and on top of the pylons were added; – thickness of the concrete deck was considerably increased near the supports, due to large bending moments acting in these sections of the spans; – cross-section of concrete piers (pylons) was optimized – stiffness of the bottom part of these supports was reduced. As a result of the above described changes the footbridge is a stress-ribbon suspension structure, where the main load bearing element is a concrete stress-ribbon deck, supported by additional cables anchored in the pylons, connected with the deck by hangers. Eventually a three-span structure with spans of 91.23 + 135.00 + 91.23 m and a total length of 336.5 m, shown in Figure 2, was designed. 258
Figure 3.
Cross-sections of precast section of span.
2.2 Structural details Deck of the footbridge is made of concrete class C50/60. Most of it consists of precast reinforced concrete panels with a thickness of 0.30 m, spaced every 3.00 m (Figure 3). Basic width of the deck equal to 4.30 m increases near the pylons to 5.90 m. Precast panels are placed directly on the bearing cables. Segments of the deck in each span are longitudinally prestressed by six internal cables of 13L15.7 and 7L15.7 type, tensioned after filling the segment joints with a non-shrink mortar. Sections of the deck situated near the supports with a length of about 7 m and height increasing from 0.30 m (in the joint with the prefabricated part of the deck) to 0.90 m (at the supports) are cast in-situ (Figure 4). The deck is fixed in the supports. Each of six main bearing cables consists of a bundle of 55 parallel 7-wire strands with a diameter of 15.7 mm made of high strength steel (1860 MPa), contained in HDPE sheath tubes (Figure 3). The cables are continuous over the entire length of the deck and are anchored in the abutments. They are running through the pylons in steel curved tubular deviators. On the length of 33.0 m on both sides of the pylons the deck is additionally supported by inclined hanger bars with a diameter of 30 mm. The hangers are anchored to the external suspension cables and to the deck every 3.00 m (spacing of precast segments). Side supports of the footbridge are massive concrete blocks encased in the ground, founded on driven prefabricated piles 0.40 × 0.40 m of varying inclination (up to 25◦ ). Large horizontal forces are carried by micropiles anchored in the abutments (Figure 2). Intermediate supports are reinforced concreteY-shaped pylons with a height of 16.7 m (Figure 4). Arms of the pylons have a form of transversally inclined double-column frames. Angle of inclination 259
Figure 4.
Configuration of pylons.
is 44◦ and height of the arms over the deck is 4.55 m. In the upper part of each arm there is encased a steel anchoring block providing fork connection for the suspension cables. The pylons are founded on driven reinforced concrete prefabricated piles 0.40 × 0.40 m inclined up to 15◦ . Upper surface of the deck is covered with a thin abrasion-resistant epoxy layer, which also acts as an insulation. The deck crossfall equal to 2.5% is directed to the longitudinal axis of the span (drainage axis), as in Figure 3. Water flowing along the deck axis is drained through inlets underneath the spans. Along the entire length of the deck balustrades are equipped with lighting installed under the handrails. Illumination and video monitoring system of the footbridge will also be installed. 3 ASSUMED CONSTRUCTION TECHNOLOGY Designed configuration of the footbridge quite directly implies the construction technology, which is typical for stress-ribbon structures (Stráský 2005). The main difference is the additional stage assuming installation of suspension cables and hangers. Main stages of footbridge construction, proposed in the design, are shown in Figure 5. 4 NUMERICAL ANALYSES The footbridge was designed to carry pedestrian live load taken as 4 kN/m2 , according to the Polish code. A possibility of passing a single 15-ton service vehicle along the footbridge was also 260
Figure 5.
Construction stages assumed in design.
assumed. Calculations were carried out taking into account the following construction phases: stage after placing precast segments on bearing cables, stage after casting the joints (continuous deck) and stage after installation of suspension cables (final static scheme). Numerical analysis was performed according to the 2nd order theory of deflections (geometric nonlinearity), using the linear elastic model of materials. The footbridge was discretized as a three-span stress-ribbon structure, suspended to the pylons by additional cables anchored in the deck. Geometry of the bridge is represented by a 3D bar 261
Figure 6.
FEM models of footbridge – stage before and after installation of hangers and suspension cables.
model. Construction stages assumed in the design of the footbridge were taken into account in the calculations by the use of a phased geometry model, i.e. by adapting it to the actual static scheme changing in subsequent stages of construction (Figure 6). The model was used to determine characteristic and design values of internal forces and reactions, as well as displacements resulting from dead and live loads. 5 CONCLUSIONS Development of detailed design of described footbridge, based on the winning competition concept, was difficult and time consuming. Cost of the construction determined on the basis of the design considerably exceeded originally planned budget, established by the investor. The tender for the construction of the footbridge was cancelled due to the lack of funding. Will the footbridge be build in the form presented in this paper? Time will tell. REFERENCES Banaszewski, K., Dec, O., Markocki, B., Sobala, D., Strasenburg, D. 2013. Konkurs na kładke˛ przez Bug w Niemirowie wygrały konstrukcje wste˛gowe (in Polish). Proc. of symposium Wrocławskie Dni Mostowe „Obiekty w infrastrukturze miejskiej”. Wrocław, 21–22 November 2013: 423–432. Dallard, P. et al. 2001. The London Millennium Footbridge. Structural Engineer 79(22): 17–35. Stráský, J. 2005. Stress ribbon and cable-supported pedestrian bridges. London: Thomas Telford Publishing.
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Design and proof checking of foundation, substructure and superstructure of Rail cum Road Bridge at Munger, Bihar, India H.M. Farook & G.S. Babu L&T Infrastructure Engineering Limited, India
ABSTRACT: The total length rail cum road bridge across river Ganga near Munger, in the state of Bihar, India, is about 3690.20 m. The bridge consists of 29 nos of 125 m spans and 2 nos of 32.60 m. The superstructure is of truss type to carry a broad gauge railway line at the bottom chord level and a three lane road carriage way at middle level of the superstructure. The truss planes are connected together through cross girders, portal bracings, and sway bracings and lateral bracings at top, bottom and middle level. The substructure comprises of RCC twin circular pier with RCC pier cap resting on well foundations. All the components of this Rail cum Road Bridge viz well foundations, substructure and steel superstructure have been designed for MBG loading-2008, IRS (1964) and IRC: 6 (2000). The superstructure at rail and road level rests on POT or POT cum PTFE bearing. All well foundations, substructure and superstructure of 125 m spans have already been constructed and construction of shore spans (32.6 m) is in progress. This bridge will be second longest rail road bridge of state bihar and third longest of the country. It is expected to be completed in 2015.
1 INTRODUCTION India has amazing network of railway lines across the country, with about 1,33,000 bridges. When the bridge construction is a challenging task as in the case of bridges across perennial rivers, it is prudent to go in for Rail cum road bridge. The mighty Ganga (or Ganges) River, makes its way from the western Himalayas to the Bay of Bengal, a journey of nearly 2,500 km. Its river basin is one of the most fertile and densely populated regions in the world and covers an area of about 1,000,000 square kilometres. The river Ganga traverses through the State of Bihar for about 460 km from Buxar to Rajmahal, dividing the into two parts – South and North Bihar. While the former is mineral rich area, the latter is the densely populated and agriculturally rich. The Railway plays a vital role as means of transportation. There are only two rail bridges across the river Ganga one near Mughalsarai, which is 410 km from Munger and other Rail Bridge is at Mokamah which is 55 km in the upstream. There are Road bridges at Buxer 306 km upstream of the Munger, the famous Mahatma Gandhi Setu in Patna 151 km upstream of the Munger and One near Bhagalpur which is 64 km at the downstream of the Munger. Construction of Rail cum Road Bridge at Munger will carry road and rail traffic on different levels. The contractor of the project is The Braithwaite Burn and Jessop Construction Company Limited, Kolkata and L&T Infrastructure Engineering Limited are the consultants for proof checking and re-design. 263
1.1 About the location The bridge was proposed to be located at Munger – Malhipur site as out of the two alternative sites i.e. Munger – Malhipur and Samastipur – Munger, Munger – Malhipur site is distinctly superior to other site due to the following reasons. i. The main bridge length will be shorter by 2.4 km. ii. Meandering of river and obliquity of flow will be lesser.
2 FINALISATION OF THE CROSS SECTIONS AND GENERAL ARRANGMENT The proposed bridge is to carry a single broad gauge rail track and a three lane highway flanked by footpaths on either side. Road deck on the bridge is given a cross slope of 2.5% from centre of bridge towards end for drainage. At the selected site, the bridge length is 3690.20 m, with 29 spans of 125 m c/c and end spans of 32.6 m. The entire alignment is in straight and has no longitudinal slope. 2.1 Superstructure The superstructure is primarily a truss type. The width of the superstructure is fixed based on minimum requirements for the road carriage way carrying three lane traffic and minimum clearances required for the passage of railway broad gauge trains. Sizes of the box sections of the superstructure are fixed from fabrication and maintenance considerations, besides structural requirements. Preliminary design for the following structural components of the superstructure was done to finalize the structural arrangement of the overall structure. – – – –
Design of rail supporting floor system. Design of road supporting floor system. Arrangement of bearings and design. Arrangement of seismic arresters.
2.1.1 Arrangement of truss The two planes of the truss are placed 12.25 m apart to accommodate Broad gauge railway line at the bottom level and a three lane road carriage way of 10.5 m width at middle level of the superstructure. The height of the truss is about 18.5 m. The cross sectional shapes of the truss members are: – The top chords, bottom chords and diagonals are hollow box sections. Middle chords are of twin channel sections placed face to face. – The diagonals, verticals, stringers (rail bearers), portal bracings, top cross girders, middle cross girders and bottom cross girders are of I-sections. – Wind and top lateral bracings at the bottom level are of star angle configuration with equal angle section. – Sway bracings are of twin channel sections placed back to back. 2.1.2 Arrangement of railway floor The rail tracks run on steel sleepers (as per Indian Railways Standards) supported on stringers (rail bearers). There are two stringers placed at spacing of 1.9 m. The stringers are connected to bottom cross girders. The centre to centre distance between bottom cross girders are chosen to suit the panel lengths. The stringers are designed to carry the train load, transverse forces due to wind on train and raking forces, and transferring them to the bottom chord members. Gangway is provided inside the truss planes on one side. 264
Figure 1. Typical Cross Section for 125 m span.
2.1.3 Arrangement of roadway floor The deck rests on six longitudinal stringers. The stringers in turn transfer the loads to the cross girders that are supported by vertical members. Footpath is also provided for the convenience of pedestrians. 2.2 Substructure The superstructure is supported by RCC twin circular piers with RCC pier cap. 2.3 Foundation Based on the characteristics of the strata at the bridge location, it was proposed to Double D shaped well foundations which are 11m wide along the direction of traffic and 18 m across. The wells were taken to a depth of 57 m below bed. Typical cross section of 125 m and 32.6 m spans are shown in the Figures 1 and 2 below. 3 ANALYSIS AND DESIGN APPROACH 3.1 Analysis Global and local models of the superstructure are analyzed for different loads and load combinations. The superstructure of the bridge has top panels provided with K bracings and bottom panels with X Bracings. 265
Figure 2. Typical Cross Section for 32.6 m span.
3.1.1 Load transfer The railway load is transferred from rails to the bottom chords through sleepers resting on longitudinal stringers. The stringers in-turn transfer the loads onto the cross girders that are connected to the bottom chords of the trusses at nodes. The loads from road deck are at deck at middle level of the truss. The deck rests on longitudinal stringers. The stringers in turn transfer the loads to the cross girders that are supported by vertical members. Wind and seismic forces acting in the transverse direction are transferred to the bottom chords and then on to the bearings. As envisaged in design basis note the models are analysed for various load cases. Based on the analysis of the model sections have been chosen and validated as per various clauses of design basis note and the relevant fs mentioned therein. 3.1.2 Structural model Three dimensional beam model of the superstructure is created using STAAD software and analyzed for various load combinations. The basic features of the model are: 1. 2. 3. 4. 5. 6. 7. 8. 9.
Linear elastic elements. All elements are beam elements. The joints in the plane of truss are idealised as moment resisting joints. The railway stringers are discontinuous over the cross girders. This is achieved by releasing the end moments at both the ends. The bottom and top cross girders are modelled as moment resisting members at the joints with truss plane. The cross girders of road deck have are made moment released at the joints with the truss. Lateral bracings at top and bottom have moment released at the joints with the truss. Bracings at the rail stringer level have moment released at the joints with the truss and the rail stringers. All loads are applied as static loads – either as member loads or nodal loads. Dynamic effect of moving load is accounted for by suitably modifying the loads by coefficient of dynamic augment as per codal provisions. 266
3.1.3 Joints a) In the plane of the truss all joints of the main truss members are modelled as moment resistant. b) The joints of the Rail cross girders to bottom chords are moment resistant about major axis. c) The joints of the Road cross girders to the verticals are not moment resistant about both minor and major axis. d) The joints of the Top cross girders to the top chords are moment resistant about both minor and major axis. 3.2 Provisions for fatigue Railway steel bridges are subjected to fatigue loading and the check of the members for failure against the fatigue is critical. Structural elements where live load is a large percentage of the total load are potentially susceptible to stress reversal and fatigue. Fatigue is a serviceability problem. Fatigue check is carried out for loads, which are normally less than the maximum design loads. Fatigue is critical at the joints. Residual stresses and stress concentrations control the stress levels to be adopted in the design. Detailing rules are perhaps the most important part of fatigue and fracture design. 3.3 Design approach The design is based on working stress approach as per Indian Railway Steel Bridge Code, IRS (1962) specifications. The design is further checked for serviceability limits such as: i. Deformation including vertical deflection, lateral deflection, rotation and twist. ii. Fatigue loading for members and connections. 4 FABRICATION AND ASSEMBLY OF STEEL TRUSSES 4.1 General The steel truss is fabricated using plates of strength up to E 350. All steel plates with thickness greater than 20 mm are tested for de-lamination. All steel plates subjected to tension perpendicular to the surface are also tested for de-lamination Full assembly of the end span truss was done in the open yard located behind one of the abutments so that it can be readily launched. After fabrication the truss was checked for pre-camber. Welding procedure and specifications shall be as per IRS Welded Bridge Code, IRS (2001). Welding quality and the geometric precision of all the members are the two key points for controlling quality of the truss. Automated welding process was used for the main structure’s seam and fillet welds. Other elements can be welded with manual welding equipment. Before final welding, various criteria regarding welding process shall be assured so that the mechanical behavior of the welded joint equals that of the parent steel. All welds are inspected with ultrasonic or radiographic equipment 24 hours after welding. 4.2 Connections i. Welded connections Based on structural configuration of the members fillet welded connections are used for the member sections. ii. Connections using rivets The number of members meeting at the joints, the plane at which they are meeting and the structural configuration of the members are different and hence the type of joints shall be 267
different. All connections other than welded connections are made using rivets and gusset plates. Bolts are used for connections related to bearings. 5 BRIDGE ELEMENTS AND EQUIPMENTS 5.1 Bearings As against the conventional rock cum roller bearings types, POT or POT cum PTFE bearings are recommended. These bearings are compact and easily available, of reliable quality, maintenance free and can be completely tested before erection. Anti seismic devices in the structure shall be provided to prevent unseating of the superstructure during an earthquake based on safety criteria. 5.2 Expansion joints Structure expansion joints are capable of accommodating translations and rotations between the superstructure and approaches without any damage. And they are capable of easy removal and replacement during the lifetime of the structure. Modular strip seal type expansion joints are used based on the design approach. 6 CONSTRUCTION SCHEME The steel truss is first erected/launched over the span. The concrete deck is cast next. The formwork is supported from the steel truss and the entire load of concrete deck is carried by the steel truss. Once the concrete attains adequate strength and behaves composite with the road stringers, the other imposed dead loads consisting of road and railway fixtures are erected. The construction sequence and method has been considered in the design of individual members. 7 DURABILITY, INSPECTABILITY AND MAINTAINABILITY The structural design and composition of construction materials have been done with adequate care to ensure durability, considering the structural details of which they form part as well as the effects of the environment to which they are exposed. Bridge components such as bearings, expansion joints and items of equipment are designed so that they can be easily replaced. Provision has been made to ensure easy access and adequate level of safety during inspection and maintenance operations. Gangways are to be provided for the full length of bridge at rail track level for inspection of track etc. Inspection platforms need to be provided around the piers for inspection of bearings. Suitable accesses are provided to the top of pier caps at selected locations. REFERENCES Indian Railways Bridge Manual 1998. Indian Railway Bridge Rules, Revised. 1964 (Including Correction slips up to no. 38). Indian Railway Standard (IRS), Code of Practice for the Design of the Substructure and Foundation of Bridges (Bridge Substructure and Foundations Code- adopted 1936, Incorporating up to Slip No. 22). Indian Railway Standard (IRS), 1985. Manual on the Design and Construction of Well and Pile Foundations, (Well and Pile Foundation Code), Adopted – 1941, Revised Edition – 1985. Indian Railway Standard (IRS), 1962. Specification for Steel Bridge Code Revised (Including Correction Slips up to No. 17). Indian Railway Standard (IRS), Welded Bridge Code for Steel Bridge Girders (Revised 2001).
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IRC: 6-2000. Standard specifications and Code of Practice for Road Bridges, Section-II – Loads and Stresses (Fourth Revision). IRC: 21-2000. Standard Specifications and Code of Practice for Road Bridges, Section-III – Cement concrete (Plain and Reinforced) (Third Revision). IRC: 83-2002 (Part III) – Standard Specifications and Code of Practice for Road Brides. Section IX, POTCUM-PTFE, PIN and Metallic Guide Bearings.
UNDER CONSTRUCTION AND COMPLETED PHOTOS OF THE PROJECT
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Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
Multi-span bridge bypass over the Dziwna Strait J. Hołowaty West Pomeranian University of Technology, Szczecin, Poland
ABSTRACT: The bridge structures for the Wolin bypass over the Dziwna Strait are presented in the paper. The by-pass carries the national road No 3 with 2-lane carriageway and hard shoulders. Wolin is a small historic town which could no longer bear the high volume of traffic on the national road going directly through the town. The bypass route followed the existing trunk railway line located on embankments along the northern historical outskirts of the town. Therefore, no other embankments for the road bypass were permitted and the flyovers were used along the town area in the accepted feasibility design. Alongside the town, the road was located on multi-span viaducts with four girders decks in prestressed concrete or steel-concrete composite. For the largest navigable span, a tied arch of the Nielsen type was developed with span 165 m long. Total length of the bridge structures in the by-pass is 1.1 km. 1 INTRODUCTION 1.1 General information The national road No 3 is the most important route in West Pomerania region, northwestern Poland. ´ It connects three major cities in the region, from north to south they are: Swinouj´ scie, Szczecin and ´ Gorzów Wielkopolski. Swinouj´scie and Szczecin are sea ports located in the Oder River estuary. The expansion of the road No 3 had been started in the seventies last century due to increasing traffic and congestions especially in summer months. One segment of the road near Szczecin had got the second carriageway, another one had got hard shoulders. The planned road extension on the island of Wolin had never been realistic due to disagreement of the Wolin National Park Authority and the lack of funding. The rapid increase of road traffic in Poland in the early 1990s accelerated the highway network modernization program also on the road No 3. Near the small town Wolin of 5000 inhabitants, road No 3 (DK3 see Figure 1) is crossing the Dziwna Straits, one of the three straits (the east one) connecting Szczecin Lagoon and the Pomeranian Bay of the Baltic Sea. The variation of the sea level makes sometimes a return flow of water in the Dziwna river (Strange river). The existing road crossing utilizes a low-rise bridge with a swing span. The road is going direct through the old center of the town. During summer month when the Baltic seaside is full of tourists there was a continuous flow of cars in both directions. The town itself is occupied by tourists during the annual Slav and Viking festival in July/August. Wolin is a very old town which origin are dated back to the 5th and 6th century. The old name of the town is Jomsborg, in legends the town is called Vinieta. The area of the town is statuary protected due to archeological findings from the 9th and 10th century when the town had flourished. Some archeological findings were located on the planned by-pass route so all the excavations for support foundations should be explored by archeologist at first. The timber pile structures and fortification had been discovered during flood protection works and not explored yet. 1.2 The Wolin National Park and road No 3 The main restraint in the road No 3 (DK3) extension on the Wolin island is the location of the Wolin National Park which takes nearly a half of the island area. The park area starts just a few 271
Figure 1.
Plan of main roads near Wolin.
kilometers from Wolin town due west. The road, in existing route through the park, had not been modernized for many years and had become very deficient and unsafe for drivers. It took nearly 20 years to negotiate the extension of the road No 3 segment in its route through the national park. However a second carriageway was not accepted. That influenced the cross section design for the Wolin bypass and one carriageway with hard shoulders was chosen despite that for all modernized segments of the road No 3 dual carriageway is adopted. The road segment in the Wolin National Park is going through forest and rolling terrain. It was rebuilt with correction of horizontal curves and construction of underbridge for the passage of wild animals. At the end of the segment a new junction giving access to Mie˛dzyzdroje, a sea resort, was constructed. It included a 280-m long trestle bridge structure over an environmentally protected swampy terrain. A little later a junction with two regional roads (DW) located due east from the Wolin town was constructed (Figure 1) and another jam-point on the national road No 3 was liquidated. The construction of the Wolin by-pass was one of the first projects which started the modernization of the national road No 3 in West Pomerania during the program for rebuilding the trunk road system in Poland launched at the end of 1990’s. The road is of international importance giving ´ access to the sea ports in Szczecin and Swinouj´ scie but it is also of national and regional importance giving access to the Polish seaside at the Baltic Sea. 2 BRIDGE STRUCTURES FOR WOLIN BY-PASS Environmental issues and protection of the national park enclaves had made great problems to the designers of the new route for the road No 3 that should omits the Wolin town. The large area of the national park on the Wolin island gave no possibility to radically change the route. Only the concept using north outskirts of Wolin town was acceptable. The new accepted route is parallel to the trunk railway line with truss railway bridge over the Dziwna Straits. The high level railway bridge gives the vertical clearance over the navigable waters. Location of the new route on the outskirts of the town required adoption of long multi-span trestle bridge structures on the west approach which give access to the existing terrain, the railway station and the statutory protected historic site. It gives the total length of the Wolin by-pass bridge structures equals to nearly 1.1 km. The total length of the by-pass is 2.6 km. An elevation of the bridge structures is shown in Figure 2. The Wolin by-pass carries 7.0 m carriageway with two traffic lanes 2 × 3.5 m and two 2.0 m hard shoulders as accepted for the extension of the National Park segment of the road. 272
Figure 2. Wolin by-pass elevation.
The bridge structures were divided into six sections with a major arch span of 165 m in length over the Dziwna Straits navigable waters. The Dziwna by-pass utilizes different types of bridge structures. Prestressed concrete slab and beam superstructures were cast in situ on formwork and falsework for section I and II. Steel-concrete composite superstructure for section III was used where falsework is inaccessible due to weak soils. The major bridge utilizes composite decks for the arch (section IV) and for the two span structure (section V). The division of the by-pass bridge structures into several sections allowed their parallel construction. The supports were constructed in reinforced concrete on pile foundation. All piers are of the same two-columns structure except for the arch span where frame structures were developed. The oval columns in piers have got circular cross heads to support two bearings each. The trapezoidal columns are developed in the arch span piers. For piles, caps and supports concrete grade C25/30 was used. The project was divided into two contracts due to the short construction schedule. The first contract included the main span (section IV) and composite steel concrete bridge structures (section III and V). The second contract included approach prestressed concrete (section I and II) bridge structures, approach roads and a viaduct over a county road on the east bank of the river. The contracts were planned to be executed simultaneously. It happened that for the two contracts, one contractor was selected.
2.1 Ground condition The location of the by-pass is characterised by complex postglacial geotechnical conditions, including clays, sands and organic soils like peats in different configuration. For all bridge structures the pile foundations were adapted. More complex soils were on the west bank where there is a top layer of peat 1.5–2 m thick (section III). This is also archaeological site with medieval remains so the commencement of foundation works was preceded by archaeological assessment of all bridge support areas. The contractor also assessed the soil condition and accepted the design pile foundation Reinforced concrete piles of diameter 1.5 m were developed with different lengths (up to 22 m) and in different number. For the section I and II which decks were cast in situ the soil condition were assessed to be good for direct supporting of falsework foundation except for the last two span of the section II where top layer of peat is present. As it was no possibility for removing the weak soil layer it was decided to consolidate the soils by extra ballast. That allowed to avoid a time-consuming archaeological excavation.
2.2 Prestressed concrete sections From the west side the by-pass has got two sections I and II with decks constructed in prestressed concrete. The number of spans is: 7 for section I and 6 for section II, the span lengths are 27 + 5 × 36 + 27 m and 27 + 4 × 36 + 27 m, respectively. The adopted cross section is a typical PC slab and beam deck with four ribs (Figure 3). The PC girder centering is 3 m. The concrete grade is C32/40. The decks were designed to be built twospans by two-spans and it was followed in construction to speed up the rate of construction. As the access was easy the formwork was supported by a full height falsework from the ground level. 273
Figure 3.
Cross section of the deck for section I and II.
Figure 4. Formwork and falsework for the first deck segment (two spans) of section I.
Figure 5. Stage construction of the decks for section I and II on the full height falsework system.
The total length of PC decks is 434 m. The deck construction started at the west abutment and progressed two spans at a time, once a one span stage was used (Figures 4 and 5). The segments of the viaduct were cast in one pour and construction joints were introduced in the decks. The prestressing tendons have simple profiles with couplers at construction joints. Additional short cables are set over pier Staged construction allowed for avoiding hydration effects and shrinkage cracks, as well as stressing the prestressing tendons effectively. The completed decks were used for access. For the last segment of section II the consolidation of the top peat layer was executed to avoid excessive settlement of the falsework (Figure 5). A 1-m fill and stockpiles of precast slabs were used as a ballast for a month. The casting of the viaduct decks were to be executed a in cold winter season which happened that year so special measures for winter concreting were necessary. The temporally concrete-mixing plant was located close enough to the construction site. To transport concrete mixtures a working road was constructed along the construction site. 2.3 Composite steel-concrete sections The next section has got a composite steel-concrete deck of total width 13.3 m. The plate girders centering is 3.2 m. The section III is a continuous six span structure, the span lengths are 54 + 4 × 66 + 54 m. Total length of the deck is 373 m. The adopted cross section is a typical composite deck with four steel girders (Figure 6). The same deck is developed for the section V where the two span deck is used with span lengths 2 × 60 m. The section III is built in a horizontal curve with arc radius R = 1200 m and two transition curves. The depth of the deck slab is 240 mm, the concrete grade is C25/30. The plate steel girders were assembled in two-girder segments on temporary supports by mobile cranes (Figure 7). The steel girder segments were delivered by road transport from Puławy (the midst of Poland). All the site connections were welded. As the steelwork was ready the temporary supports were dismantled and slab formwork was supported on girders. The concrete deck slab was cast in stages using several 274
Figure 6.
Cross section of the deck at midspan for section III and V.
Figure 7.
Plate girder segments assembly for section III on temporary supports.
fronts for concrete placement. At first the mid-span parts of the deck slab were cast. For concrete placement a temporally horizontal bracing was developed with bolted connections. The two steel segments for the section V of the total length 121 m were assembly one by one from supporting towers on a barge. The segments were transported by water on two barges from Szczecin. The deck slab was also cast in stages to reduce tensile stresses in the deck slab. 2.4 Arch section The main navigable section of the Dziwna by-pass is the 165-m long arch span. The arch rise is 24 m, it gives the ratio of span to rise nearly 7:1. The cross section of the span at midspan is presented in Figure 8. The navigable waters are located under the east part of the arch span. The clearance is 50 m horizontal and 12.6 m vertical. The vertical alignment of the arch deck was design in a vertical curve with a radius 20000 m. Two arch ribs of rectangular section 1 × 1.8 m are inclined. The inclination ratio is 24:7. At supports the steel ribs are made composite with reinforced concrete skewbacks. The pot bearings are located under the skewbacks, bearing capacity of the bearings is 12 MN. The fixed bearings are located on the west pier. The composite steel-concrete deck is suspended by inclined hangers of 41 mm diameter. The deck steel grid consists of the three longitudinal stingers and 26 transverse girders. The spacing of stingers is 6.4 m. The transverse girders are spaced at 6 m. The inclined hangers from the arch ribs are attached to transverse girder cantilevers. The concrete slab of the deck is supported on stringers and transverse girders and made composite with them. Depth of the deck slab is min 240 mm. The concrete grade for the deck is C32/40. 275
Figure 8.
Cross section of the arch span.
Figure 9.
Stages in construction of the arch ribs.
The Nielsen type of the arch which consists of two symmetrical planes of inclined hangers gives sufficient stiffness to the composite deck. To balance longitudinal forces in the arch span six longitudinal cables were developed in the deck, three cables 37 × 15.5 mm for each rib. The cables are anchored at the RC skewbacks. The tie cables are necessary to resist the horizontal thrusts. The design camber of the deck was a little over 200 mm, and the cables were stressed in staged according to the technology of concrete slab placement. The site construction of the arch span started with construction of two auxiliary towers in the river bed which were used for assembly of steel arch ribs (Figure 9). At the same time casting of RC skewbacks was finished. The paired ribs segments were transported on barges from Szczecin were they were fabricated. Using the temporary towers they were installed, lifted to the final position in a sequential erection process and finally welded. The steel rope hangers were installed to allow the partial removal of the temporary towers and erection of deck steel grid segments and longitudinal prestressing cables. The concrete for the deck slab was pumped from a barge in the design sequence with partial stressing of the tie cables. 3 CONCLUSIONS The Wolin by-pass was completed in October 2003 and opened to traffic in December 2003. Designed to take traffic away from town areas, and improve access to see port in Szczecin and ´ Swinouj´ scie the project was executed on time and to budget. The design team was recognized by several rewords as well as the contractor. The Wolin town received not only a new long awaited bridge crossing but also an iconic bridge with an elegant arch steel structure. The arch span was to be the longest in Poland but a little earlier a 3-m longer arch bridge was completed. The bridge structures and the method of construction were adjusted to the existing ground conditions and the required navigational clearance. The western approach was divided into prestressed concrete and composite structures. Composite structures were used where the ground conditions did not permit falsework foundations. Dividing the bridge bypass into sections allowed the parallel construction of the bypass structures within the very limited time schedule.
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Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
Large multi-span bridges built in recent years in Poland J. Biliszczuk & J. Onysyk Wrocław University of Technology/Research & Design Office Mosty-Wrocław, Wrocław, Poland
P. Prabucki & R. Toczkiewicz Research & Design Office Mosty-Wrocław, Wrocław, Poland
ABSTRACT: The road and railway infrastructure in Poland has been intensively developed and modernized over the last years. During the last decade more than 2500 new bridges were built along the main roads and motorways. Among them several large multi-span bridges with a length of more than 500 m were constructed. This paper presents the following most interesting and complex structures: – Multi-span post-tensioned concrete flyovers along the A8 motorway; – The North Bridge in Warsaw – one of the largest steel beam bridges in Poland; – The new Warszawski Bridge in Wrocław – road-tram urban bridge, a six-span continuous steel beam structure with a Langer arch over the water channel; – The bridge over the Vistula River in Puławy – continuous multi-span steel-concrete beam, the main span is a steel tied through arch with a length of 212 m; – The Re˛dzi´nski Bridge over the Odra River in Wrocław – concrete structure consisting of two multi-span access flyovers and the main cable-stayed bridge with two spans of 256 m suspended to a single pylon; – The Solidarity Bridge over the Vistula River in Płock – steel cable-stayed structure with a span of 375 m (the longest span in Poland) suspended to two I-shaped pylons fixed in deck. The paper also describes construction technology of the aforementioned bridges. 1 INTRODUCTION During the last years several large bridges of different types, with a length of more than 500 m, were built in Poland. Most of the new long-span bridges are situated along main roads crossing major rivers (the Vistula River and the Odra River). They are also often built in cities, where they have a chance to become their distinctive landmark. Multi-span structures with shorter spans (usually not exceeding 60 m) are built in the case of necessity of crossing wide obstacles, when it is convenient to build a large number of piers. They are mostly erected as beam flyovers over urbanized areas or floodplains. Multi-span beams are also constructed as access viaducts leading to the main bridge crossing river bed. The paper presents selected examples of recently built Polish long-span and multi-span bridges: beam, arch and cable-stayed structures. 2 BEAM STRUCTURES 2.1 Flyovers along the A8 motorway (2011) Multi-span flyovers denoted as WA-19 and WA-17 were built along the A8 motorway ring road of Wrocław (Biliszczuk et al. 2012). WA-17 viaduct is a seven-span horizontally curved structure, crossing several streets and a railway line. Its total length is 300.0 m, the length of spans varies from 30.0 to 60.0 m. The 277
Figure 1.
Completed WA-17 viaduct (left photo) and view of WA19 viaduct during construction.
Figure 2. WA-17 viaduct – scheme of the fabrication area and cross-section of the superstructure.
superstructure is a single-cell box girder. Due to the significant length of deck plate overhangs, they are supported by steel tube struts forming a “V” shape in side view (Figure 1). The struts carry loads in transverse and longitudinal direction, and dominate the architectural reception of the flyover. Superstructures for both carriageways were longitudinally launched. After one of them had been completed, the temporary supports were moved transversally to launch the parallel structure. The longest span during the launching was 40 m. The scheme of the fabrication area and the sequence of works (concrete casting, struts assembly) are shown in Figure 2. WA-19 flyover has a similar structural configuration. It crosses a city road interchange and a railway line in the vicinity of EURO 2012 football stadium. For that reason it was proposed to be constructed using the longitudinal launching method. Commencing the construction works, the contractor proposed to use traditional scaffolding for the construction divided into five multispan sections (Biliszczuk et al. 2012). According to the change of construction technology, new arrangement of external and internal prestressing cables was designed. WA-19 flyover consists of two structures curved in plan, separate for each of the carriageways. 16-span continuous beam superstructures are made of prestressed concrete class C50/60. Their cross-section is similar to the one shown in Figure 2. Span lengths vary from 35.0 to 58.0 m and the total length of the flyover is 750 m. 2.2 North Bridge over the Vistula River in Warsaw (2012) North Bridge is an important part of communication network in Warsaw. It consists of three parallel structures: a tram bridge and two road bridges (Ba˛k et al. 2011). Each of them is a continuous 278
Figure 3.
North bridge – side view and cross-section in spans (Ba˛k et al. 2011).
Figure 4. Assembly of main span (Ba˛k et al. 2011) and view of completed bridge (www.pomost.com.pl).
10-span beam with a total length of 795 m (Figure 3). Length of the main span crossing the Vistula River is 160.0 m. The superstructure is a steel, single-cell box girder with inclined webs, interacting with a reinforced concrete deck slab (Figure 3). Height of the girder is constant in side spans situated over inundation areas. In three main spans height of the box girder varies from 3.75 m in the spans to 9.00 m over the supports. In the spans of variable height width of the bottom flange changes from 6.05 m to 4.70 m. The webs and bottom flange are stiffened with closed ribs. During construction the superstructure was divided into five parts erected using various methods (Ba˛k et al. 2011). Right embankment section with a length of 317 m was longitudinally launched, using truss launching nose and no temporary supports. Left embankment section with a length of 125 m was assembled using cranes and two temporary supports. Steel segments of the above described sections had a maximum length of 28.0 m and weighed up to 75 tons. Support sections of the main span were assembled using cranes. Due to the heavy weight those segments were transported to the construction site by river. Each segment was divided into two elements weighing 290 tons (lower sub-element) and 220 tons (upper sub-element). The segment was assembled using temporary support situated in the river. Central parts of spans adjacent to the main span, with a length of 65 m were assembled using cantilevering. Segments were joined on the embankment and lifted using hydraulic jacks installed on the previously erected part of the span. 279
Figure 5.
New Warszawski Bridge – cross-section in the main span.
Middle part of the main span with a length of 138 m was cantilevered from barges (Figure 4). This section consisted of five segments, of which the largest (middle) segment with a length of 64.0 m had the weight of approx. 510 tons. Concrete deck plate was cast in sections (first in spans, then over supports) using light movable scaffolding. 3 ARCH BRIDGES 3.1 New Warszawski Bridge in Wrocław (2008) The new Warszawski Bridge (Biliszczuk et al. 2013) was built parallel to the existing concrete bridge that has been used since 1916. It is a road-tram six-span steel-concrete beam structure. The longest span crossing the water channel is strengthened by an arch (Langer arch type) with a length of 61.9 m. The bridge has a variable skew angle of spans, aligned to the axes of existing supports of the old bridge, changing from 77.0◦ to 57.9◦ . The superstructure consists of two steel box girders composite with a concrete deck slab. Height of the superstructure is constant along its length. The girders with a transversal spacing of 14.5 m have a double-cell box section in the beam spans and a triple-cell box section in the arch span (Figure 5). Hangers are made of multi-strand tendons of 19L15 type, contained in white HDPE outer sheath pipes. Steel segments of the spans and the arch were assembled on the embankment and then longitudinally launched into the final position (Biliszczuk et al. 2013). The launching was conducted using a steel launching nose and a single temporary support situated in the water channel. The longest span during the launching was 38 m. The last stage of construction was casting the deck plate. The slab has a composite structure consisting of precast panels interacting with a layer of in-situ cast reinforced concrete. After placing the prefabricated panels, upper reinforcing bars were added and unbonded prestressing tendons over the supports were installed. Deck plate concreting was carried out in stages: first in the spans and then over the supports (Biliszczuk et al. 2013). Due to the location of the bridge in urban area, particular attention was paid to the architectural design. Night-time illumination lights were installed on the additional non-structural arches, below the main arch structure. View of completed bridge is shown in Figure 6. 3.2 Bridge over the Vistula River in Puławy (2008) John Paul II Bridge over the Vistula River is situated along the ring road of Puławy – a part of the S12 expressway, section Radom – Lublin (Grej & Biliszczuk 2012). The 280
Figure 6. Assembly of steel structure and view of completed bridge.
Figure 7.
Side view of the bridge over the Vistula River in Puławy.
Figure 8.
Bridge in Puławy – construction of the main span and view of completed structure.
total length of the bridge is 1038 m. It is a continuous 14-span structure with spans of 44.0 + 3 × 56.0 + 6 × 64.0 + 80.0 + 212.0 + 80.0 + 44.0 m (Figure 7). The deck of the main tied through arch span crossing the riverbed is suspended by 28 units of hanger bars. Rise of the arch over the roadway is 24.0 m. The deck is a steel-concrete composite structure. The tie beam consists of four plate girders of a constant height of 3.00 m grouped in two tandems (two plate girders spaced at 2.5 m in each tandem) with spacing of 12.5 m. The girders are braced by crossbeams with regular spacing of 4.0 m. The reinforced concrete deck slab has a thickness of 0.27 m. The arch girders are inclined to the bridge axis and have a rhomboidal box cross-section, varying from 2.50 × 3.00 m over the support to 2.50 × 2.00 m in the crown. The arches are braced by box struts. Each hanger consists of four tension bars with a diameter of 81 mm, anchored in the crossbeam and in the webs of the arch using fork connectors. Spacing of the hangers is 12.0 m, their length varies from 3.5 to 24.0 m. Steel deck of the main span was constructed using two auxiliary supports situated in the river (Figure 8). Segments of the arches (each of them was divided into three units) were assembled on the riverside. Each segment was then launched on trolleys along the tie beam into the final position, lifted, supported by assembly towers and joined by welding. The reinforced concrete deck slab was cast after installation of hangers. 281
Figure 9.
Figure 10.
Re˛dzi´nski Bridge over the Odra River – basic dimensions.
Re˛dzi´nski Bridge – construction and view of completed structure.
Due to the complexity of the structure a monitoring system has been installed on the bridge. It consists of three subsystems: structure monitoring, video monitoring system and a weather station. 4 CABLE-STAYED BRIDGES 4.1 Re˛dzi´nski Bridge over the Odra River in Wrocław (2011) The Re˛dzi´nski Bridge along the A8 motorway ring road of Wrocław consists of three structures (Biliszczuk et al. 2014): two multi-span prestressed concrete beam flyovers with spans up to 60 m and a cable-stayed main bridge – 612 m long, span lengths 50 + 2 × 256 + 50 m, with two separate superstructures suspended to a single 122 m high pylon (Figure 9). The superstructure of the main bridge consists of two post-tensioned concrete box girders suspended to an H-shaped concrete pylon. Foundation of the pylon is a concrete massive slab with base dimensions of 67.4 × 28.0 m and thickness variable from 2.5 to 6.5 m, placed on 160 bored piles. The pylon is a hybrid structure: its legs and the lower parts of the arms are made of reinforced concrete, the upper part of the arms (above the deck level) is a hollow composite structure. The lower crossbeam is a post-tensioned concrete element, the upper crossbeam is a post-tensioned steel-concrete box structure. Inside the pylon, a steel core was placed (Biliszczuk et al. 2014). It formed the inner formwork during erection of the pylon and it interacts with the reinforced 282
Figure 11.
Solidarity Bridge in Płock – basic dimensions.
concrete shell, transferring vertical forces and bending moments. In the cable anchorage zone the core carries horizontal forces transmitted to the walls of the pylon by cable stays. The steel core and the reinforced concrete shell act as a composite section due to shear studs welded to the core’s side plates. An important load carrying element of the pylon is the upper crossbeam, which was vertically and longitudinally post-tensioned. The superstructure of the main bridge was longitudinally launched on steel column temporary supports and large truss supporting structures over the channel locks (Figure 10). Fabrication of the segments was carried out in stages including placing side precast elements, casting the bottom slab and webs, casting the deck slab, tensioning the cables and launching the segment. The decks are suspended by 160 stay cables. 4.2 Solidarity Bridge over the Vistula River in Płock (2005) The Solidarity Bridge was built as an alternative to the only rail-road bridge, connecting districts of Płock situated on both sides of the river, built before the Second World War. The total length of the new bridge is 1200 m. It consists of two structures: – main bridge over the Vistula River with a length of 615 m – it is a cable-stayed structure suspended by a single plane of stays (along the axis of the bridge) to two pylons, with a main span of 375 m and four back spans (Figure 11); – access flyover with a length of 585 m crossing left riverside flood plains – it consists of two continuous five-span beams with spans of 58.50 m. The girder of the main bridge is a three-cell box section with a constant height of 3.56 m and a width of 27.25 m (Figure 11). Internal vertical webs form a central cell, where the steel pylons are fixed and the anchor blocks are located. The deck is an orthotropic plate with closed longitudinal ribs under the roadway and open ribs under the sidewalks. Transverse bracing bars and sidewalk cantilevers are spaced every 3.75 m. Diaphragm plates are located above supports and in sections where the stays are anchored. All connections are welded. Steel pylons with a height of 63.7 m above the deck level are fixed in the deck axis. They have a rectangular cross-section varying from 3.75 × 2.25 m at the base to 3.12 × 2.25 m at the top. Anchorages are spaced every 6.25 m in the pylon and 22.5 m in the deck. The main span is suspended to each pylon by seven pairs of harp-arranged cables. The number of backstays is the same. Length of the cables varies from 39 283
Figure 12.
Construction of bridge in Płock and view of completed bridge (photo: M. Hildebrand).
to 187 m. Each cable consists of a bundle of 47 to 84 parallel high strength steel strands with a diameter of 15.7 mm. Superstructure of the main span was cantilevered from both sides of the river at the same time (Figure 12). Preassembled segments with a length of 22.5 m and a weight of about 240 tons were transported on barges, lifted and welded to the previously erected structure. Joined segments were consecutively suspended to the pylons. Connection of both cantilevers was the last stage of construction. The Solidarity Bridge in Płock has so far the longest span in Poland. It is also the first structure that has been equipped with a structural monitoring system (Hildebrand 2012). 5 CONCLUSIONS Poland is generally a lowland country, with a large number of wide horizontal obstacles like rivers. Therefore, due to the intensive development of road and rail network, a relatively large number of multi-span and, in the necessity of crossing wide rivers, long-span bridges was built in recent years. Multi-span structures are also often erected along the routes in urban areas. This paper presents some examples of concrete and steel large multi-span bridges of various types – beams, ceilings and arched. Due to the continuous development of the communication network many new structures of this type are expected to be build in the next years. REFERENCES Ba˛k, J., Grej, K., Oleksiak, C., Sałach, W., Wyrzykowski, M., Reczko, R. 2011. Most Północny przez Wisłe˛ w Warszawie. Technologia monta˙zu (in Polish). Proc. of symposium Wrocławskie Dni Mostowe, Aktualne realizacje mostowe”. Wrocław, 24–25 November 2011: 87–96. Biliszczuk, J., Barcik, W., Onysyk, J., Prabucki, P., Sadowski, K. 2013. New Warsaw Bridge over the Odra River in Wrocław. Proc. of the 7th International Conference on Arch Bridges. Trogir – Split, 2–4 October 2013: 375–382. Biliszczuk, J., Barcik, W., Onysyk, J., Toczkiewicz, R., Tukendorf, A., Tukendorf, K. 2014. Re˛dzi´nski Bridge in Wrocław - the largest concrete concrete cable-stayed bridge in Poland. Structural Engineering International 24(2), May 2014: 285–292. Biliszczuk, J., Onysyk, J., Barcik, W., Prabucki, M., Sułkowski, M., Sadowski, K., Toczkiewicz, R. 2012. Construction of multi-span motorway viaducts using various technologies. Proc. of fib Symposium “Concrete Structures for Sustainable Community”. Stockholm, 11–14 June 2012: 593–596. Grej, K., Biliszczuk, J. 2012. Arch bridge over the Vistula in Puławy. In Zidek, S., Benko, V., Paulik, P., ˇ Rawicki, Z., Csaba, H. (eds), Engineering Structures of Visegrad four countries. Prague: IC CKAIT. Hildebrand, M. 2012. Seven years of structural monitoring of a large steel cable stayed bridge. Proc. of Workshop on Civil Structural Health Monitoring (CSHM-4) “SHM systems supporting extension of the structures’ service life”. Berlin, 6–8 November 2012.
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Kassuende Bridge over Zambezi River in Tete, Mozambique T. Mendonça, V. Brito & M. Monteiro BETAR Consultores Lda, Lisbon, Portugal
ABSTRACT: The new bridge over Zambezi River near the city of Tete in Mozambique is integrated in the national road N103 which is the main connection between Mozambique and Zimbabwe. This road is also fundamental for the transportation of goods from Malawi and Zambia to Beira’s Port in the Pacific coast of Mozambique. The recent growth of the city of Tete is related with coal extraction and has encouraged the construction of a new bridge. This article describes the main aspects of the bridge design and its construction.
1 SCOPE The new bridge over Zambezi River near the city of Tete in Mozambique is integrated in the national road N103 which is the main connection between Mozambique and Zimbabwe. This road is also fundamental for the transportation of goods from Malawi and Zambia to Beira’s Port in the Pacific coast of Mozambique. The recent economic growth in Tete region is mainly due to the coal extraction mines located in Moatize which led to the increase of traffic namely in heavy trucks. In the past, the construction of the now called Samora Machel Bridge (designed by a Portuguese engineer called Edgar Cardoso) was motivated by the construction of the Cahora Bassa dam. In the XX century the economic development was strongly dependent of those two activities: coal mining and Cahora Bassa energy production. At the beginning of the XXI century this region is again under economic growth promoting infrastructures construction and improvement. The old bridge did not fulfill the required service level which had encouraged the construction of a new bridge able to carry out the increasing of heavy traffic loads. Three different locations were studied for the Zambezi crossing near the ancient bridge but the difficulties to accommodate heavy traffic inside the urban areas of Tete lead to a fourth location about 5 km downstream, after the Revúbué river mouth in the Zambezi river (Fig. 2). The new
Figure 1.
a) Main Road Network in the Tete Province; b) Overview of the old Samora Machel Tete Bridge.
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Figure 2. Alternative bridge locations. Corridor #4 and the new road alignment downstream Tete city.
crossing includes beside the bridge itself a road with 15 km of extension to assure the proper connection to the existing roads. A Portuguese/Mozambican joint-venture called “Estradas do Zambeze” was created to build and operate the bridge and connection roads and to rehabilitate and operate the national roads represented in the previous Figure 1 a). All of them are integrated into a concession that will last 30 years.
2 MAIN CONSTRAINTS 2.1 Zambezi River, the fourth biggest river in Africa Zambezi River is the fourth biggest river in Africa after Congo, Nile and Niger rivers. Its basin is the biggest of Austral Africa with about 1 385 300 km2 from which more than 950 000 km2 are upstream from the new crossing bridge. The hydrologic study that was carried out to evaluate the design discharge values and maximum water levels has defined the 1000 years return period design values. It also allowed defining the design values to guarantee during the construction period. The local scour was prevented using counter measures placing riprap in the river bed around the pile caps designed in order to take into consideration the flow velocity. The bridge elevation allows the same vertical clearance that the existing bridge. The same navigation conditions are guaranteed. 2.2 Geologic and geotechnical conditions The bridge is placed into a plan area. Margins distance is about 1.8km measured over the river flood limits. From the geologic point of view very different scenarios were identified: sand deposits, fluvial terraces and outcrops from the Quaternary and from the inferior Karoo. The sand layers vary from 20 meters depth near Benga margin to about 67 meters in the opposite side. The load bearing layers are sandstone with some coal intercalations (Fig. 3). As a result of the local geologic conditions the foundations of the bridge are made using deep piles. Exception made for the right side margin abutment which was directly founded. 286
Figure 3.
Geologic profile at bridge alignment.
Figure 4. Typical cross section.
2.3 Tete, the warmest place in Mozambique The local average air temperature is about 33◦ C but during the raining season at 8 a.m. the air temperature can reach 38◦ C increasing up to 47◦ C. The temperature at sun exposure can reach 50◦ to 60◦ C. The weather conditions created some additional difficulties for construction namely during concrete cure. The high gradients in the concrete elements due to the sun/ shadow different exposure conditions have conditioned the cantilever construction namely the closure operations. 3 BRIDGE DESCRIPTION The total length of the bridge is 1586 meters. The structure is divided into an approach bridge with a length of 869.6 meters and a main bridge with 716.8 meters length. The bridge typical cross section accommodates a two lane road 10.20 meters wide, including 2 lanes of 3.60 meters and 2 right shoulders with 1.50 meters each (Fig. 4). 3.1 Main bridge The main bridge has a maximum span of 135 meters. The span distribution is the following (Fig. 5): 75.90 + 135.0 × 3 + 85.0 + 55.0 + 50.0 + 45.9 = 716.8 meters. The deck is a prestressed concrete box girder cross section with variable depth from 7.50 meters at piers alignment to 3.50 meters at mid span. The ratio depth/span is 1/18 (over supports) and 1/38 (at mid span) which are in the usual range of values to assure a good relation between a proper structural behavior and construction costs. The deck segments have a variable length due to the construction equipment load carry capacity limitation. For this reason near the supports the segment starts with a length of 3.85 meters increasing up to 5.0 meters. 287
Figure 5.
Longitudinal section of the main bridge.
Figure 6. a) Main bridge deck cross section at supports and mid-span; b) Picture of the deck during the cantilever construction.
Figure 7.
Longitudinal profile of the deck representing the segments and the prestress layout.
The prestress layout is formed by cantilever tendons and by span tendons. The cantilever tendons are composed by two pairs of 19 strands tendons up to segment #9 and by two pairs of 12 strands from that point forward. The span tendons longitudinal profile follows the parabolic configuration of the bottom slab of the box girder. It were considered empty ducts to allow a possible strengthening of the deck (Fig. 7). The central piers are made using a double wall with a rectangular cross. Each wall section is 7.0 × 1.20 meters. At the base, due to hydraulic constraints the walls are connected through lateral walls creating a hexagonal hollow section. The foundation are made through 6 concrete piles connected by a pile cap. The piles have 2.0 meters diameter and circular section. The remaining piers of the main bridge have a cross section that results from the extension of the piles disguised by a wall. Those piers are founded through 2 piles also with 2.0 meters diameter and a circular section (Fig. 9). The north abutment is formed by walls creating a U shape with the lateral walls. The abutment is founded by 16 piles with a 1.20 meters diameter circular section. Bridge’s two central piers are rigidly connected with the deck. In the remaining piers the deck is supported through bearings. 288
Figure 8.
a) Design drawing of the central piers; b) Image of a pier during construction.
Figure 9.
a) Design drawing of the South side piers; b) Image of the piers construction phase.
Figure 10. Approach bridge longitudinal section.
3.2 Approach viaduct The approach viaduct is formed by two independent viaducts with a length of 434.8 meters each. The typical span has a length of 55.0 meters (Fig. 10). The deck is a prestressed box girder with constant depth of 3.0 meters and with the same total width of the main bridge deck. The deck was built using a movable formwork. Each stage consisted in concreting a single span plus a cantilever of the next span with approximately 1/5 of the span length. The prestress is made span by span with a typical layout. The conception included deviation saddles and empty ducts to allow future installation of additional tendons. Each pier alignment is formed by two columns with circular shape with 1.80 m diameter. The columns are connected by a transversal beam. This beam was fundamental to assure the stiffness and resistance during the construction of the deck (Fig. 11). The south abutment is similar to the north except for the foundation which is direct. 289
Figure 11.
Images of the approach bridge during deck construction.
Figure 12.
General overview of the temporary road for foundation and piers construction.
4 MATERIALS The quality of the concrete was conditioned by the quality of the cement and aggregates that was available in the region. The maximum concrete class was C35/45 and was used in the decks. The mild steel has yield strength of 450 MPa. The prestress steel has an ultimate strength of 1860 MPa. 5 STRUCTURAL ANALYSIS The structure was analyzed using three computer models considering the whole structure and several partial models to analyze local behavior. It was used the computer software TDV RM 2006 in which it was simulated the construction stages. The interaction soil-structure as well the partial models was analyzed in SAP2000 software. The actions considered in the design followed the SATCC regulations (Southern Africa Transport and Communications Commission). 6 CONSTRUCTION 6.1 Preliminary works In the south side where the approach bridge is located due to the high ground water level and soft soils it was needed to build an embankment for a temporary road for machinery access (Fig. 12). The integrity of all piles was tested using sonic tests (cross-hole). To evaluate the behavior of the piles regarding to its load capacity 3 specimens were tested using load cells (Osterberg®). The results of the tests have allowed calibrating the design values for friction and for base resistance (Fig. 13). 290
Figure 13. a) Details of the load cells placement; b) monitoring during the test using gauge meters and reference beam.
Figure 14.
Piers construction – two different points of view.
6.2 Approach viaduct – Construction details The following images show some particular aspects of the construction sequence of the approach viaduct. 6.3 Main bridge – Construction details The construction of the piers in the river bed was one of the most challenging parts of the construction. To execute the pile caps in order to avoid temporary cofferdams it was used prefabricated elements. They were built above the water level and afterwards leveled into the final position. To help the operation was built a platform supported by the piles already concluded. This prefabricated element was used as shell that allowed the execution of all the works to build the pile caps out of the water. Due to the different constraints the bridge was built using the cantilever method. The piers were designed to resist any unbalanced loads during construction even for accidental loads resulting by the fall of the equipment. The distance between pier walls allowed a double support and the possibility to avoid additional stability systems. Near the north abutment the last 3 spans were built using traditional formwork founded into an additional platform created for that purpose (Fig. 16). 6.4 Main bridge – deck geometry control The geometry control of the deck during its construction was performed by a strict survey of the elevation obtained on each segment at each stage. All the relevant operations were simulated to evaluate onsite the real deviations in comparison with the theoretical behavior of the structure. 291
Figure 15. a) Driving the steel casing for pile construction, b) Work Platform and shell during the lowering stage, c) shell in its final position.
Figure 16.
a) View of the 3 spans near north abutment that was built using traditional formwork.
Figure 17.
General overview of the bridge after the last closure.
The geometry design has included a stage-by-stage detailed analysis of all relevant actions including the creep and shrinkage effects. Pre-camber values were so provided for each stage of the construction sequence. During the execution the Contractor decided to change the closure sequences which create some additional complexity leading to a reevaluation of the bridge geometry. Nevertheless, the deck final geometry was very close to the theoretical and the maximum deviation was less than 17 mm. 7 FINAL REMARKS The design and the technical assistance to the construction works was a very challenging work carried out by Betar. The dimension and complexity of the project have mobilized several specialists from different areas of expertise that have given their contributions to the success of the project.
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Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
Multi-span bridge crossings for improved road access to Szczecin sea port J. Hołowaty West Pomeranian University of Technology, Szczecin, Poland
ABSTRACT: The paper presents the major bridge crossings built to improve the road system in Szczecin, a city and sea port in northwestern Poland. Transportation problems had arisen following an increase in traffic and delays in the construction of a new bridge crossing on a major city route. Two crossings were constructed: the Regalica River Crossing on the major route and the Parnica River Crossing, which gives direct access to the sea port main gate. A wide range of spans and bridge types were used. For the largest spans, twin girder or box girder steel-concrete composite superstructures were built and for structures with limited depth, multi-beam composite superstructures were used. For the approach viaducts, reinforced concrete decks were utilized in the form of voided slabs, solid slabs or beam and plate.
1 INTRODUCTION Szczecin is a both a city and sea port located on the Oder River estuary in northwestern Poland. The Oder River estuary consists of three main branches, several canals, rivers and lakes. The waterways divide the city and the port; all the main waterways are highest class navigable inland waters of international importance. The oldest areas of the city are located due west and east of the River Oder estuary. This forms the main transportation corridor connecting the oldest parts of the city located on the banks of the waterways via the industrial and port midland. Major bridges are necessary in order to cross the three branches of the river, the Western Oder, the Regalica and the Parnica, along with many viaducts to cross over the railway tracks which access port areas on the river islands. Figure 1 is a map of southern Szczecin with transportation routes showing the crossing locations. The old road is still the major access route for Szczecin with its many port terminals and wharfs.
Figure 1.
Map of southern and eastern Szczecin with transportation routes.
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In the 1930s, two west-east routes were added in the south of the city. One is along the present city limits and the second is a motorway (now the A6) which forms the Szczecin southern bypass. Following the war, bridges providing a railway connection to the port were first to be rebuilt. Next, fixed crossings on the major road route were rebuilt one by one, replacing temporary bridges. The major city route received entirely new bridges including at least four traffic lanes and a tram line. A high level bridge over the Regalica River was constructed in prestressed concrete and opened to traffic in 1960 (Customs Bridge – Fig. 1). A little earlier the Czajkowski viaduct was constructed over railway tracks to a coal terminal. The major route with four traffic lanes was sufficient for existing traffic for many years with a low rate of increasing traffic. This route is currently national road No 10 (DK10). In the 80s and 90s the Castle Route including four steel box bridges over the West Oder River, the Parnica River and a multi-span trestle over the port area was constructed, connecting the city centre with the midland of the Oder River. The crossing utilises separate bridge structures for each carriageway with a minimum of three traffic lanes in each direction. Over the years, the growth in traffic caused some problems with the Customs Bridge over the Regalica River, in that it was becoming too narrow. In November 1996, daily traffic volume on the bridge was measured at nearly 60 000 nominal vehicles; this equated to 3400 vehicles per hour and was much higher than the design capacity. Feasibility studies were drawn up for a Regalica Crossing with a new road alignment. The first concept assumed the development of a double-deck steel truss bridge with six traffic lanes on the upper deck and a tram line on the lower deck. Foundation works were even started but were halted due to the country’s economic and political problems. In the early 90s, new requirements for the crossing were made and a new concept was designed, which included six traffic lanes, three in each direction and double-track tram line supported by three parallel separated decks. A little later, to solve the problem with road access to the port main area, the new Parnica Crossing design was drafted. Meanwhile the majority of the A6 motorway was reconstructed in stages (along the southern outskirts of Szczecin), and almost all the bridge structures on it were replaced (Fig. 1). The Oder River estuary is characterised by complex geotechnical conditions, including postglacial organic soils and peats. As a result, the two new bridge crossings (Regalica and Parnica) were designed with pile foundations in risky soil conditions. For both projects, steel-concrete composite structures were used for the major bridges and reinforced concrete viaducts on the approaches (Furtak 1999). They were designed to increase capacity of the main city route with improving access to Szczecin sea port. The Regalica River crossing was opened in stages to road traffic in 2001 and 2002, while tram traffic is expected to commence in 2015 when a new tram line linking the city centre with its eastern districts is completed. The Parnica River crossing was opened in 2002. The locations of the bridge crossings are shown in Figure 1. Bridge construction in urban and industrial areas creates particular difficulties. The presence of navigable waters, streets, tram lines or railways rules out construction technology. Equipment operation is limited and the structural solutions selected should follow the requirements for economical construction. The projects presented in the paper utilise different types of structures for defining the relevant parameters for the highest loading class of bridges, as determined by the Polish design codes. The bridges are exposed, as is all the transport and port infrastructure, to weak ground conditions. The thickness of the weak soil layers is 3–9 m, so a pile foundation was necessary. All the bridge structures are supported on pile foundations with bored and cast-in-situ piles. For the falsework foundation, different techniques were used (Hołowaty 2009, 2013). 2 THE REGALICA CROSSING The new crossing is located about 400 m south of the existing Customs Bridge and 4 km from Da˛bie Lake (Fig. 1). It was constructed to alleviate traffic on the Customs Bridge and to allow a new tram connection with the eastern districts which have a growing population. The major bridge structures of the crossing are three composite steel-concrete bridges separated for each carriageway and a 294
Figure 2. Plan of the Regalica Crossing.
Figure 3. Regalica Crossing elevation.
tram line, each with a length of 535 m. The total length of the bridge crossing is over 950 m and the length of the construction works is 2.8 km. A plan of the bridge structures is shown in Figure 2. The Regalica Crossing is a short cut with new wide carriageways on the existing main city route (Fig. 1). The Regalica Crossing consists of two auxiliary viaducts E-1 and E-2, western approach viaducts WD-1, WT-1, WD-2 and WD-3, WT-2, WD-4 and three major bridges marked as M-1, M-2, M-3 (Fig. 2). The eastern approach is constructed on embankments with strengthening of the subsoil. The navigable sections of the crossing are major bridges with two navigational clearances of 60 m horizontal and 7 m vertical. The bridge section of the project was divided into seven bridge contracts and started with the construction of two auxiliary viaducts, E-1 and E-2. The next two contracts were slated for the southern section of the crossing: the first for bridge M-1 and the second for approach structures WD-1 and WD-3. This allowed some relief for the Customs Bridge and the demolition of the onespan Czajkowski viaduct over the railway tracks to the port coal terminal, which was necessary in order for the new structures and roads to be built. The last four contracts were executed almost simultaneously, for the two major bridges M-3 and M-2, and two approach viaducts WD-2, WD-4 (road) and WT-1 and WT-2 (tram). For the two approach viaduct contracts, one contractor was selected. The construction schedule followed the contract plans to keep the existing heavy traffic and expected financing. Viaducts E-1 and E-2 had been constructed in the first phase to take over the traffic and gave space to construct the WT-1 and WD-2 viaducts. Now the E-2 viaduct is a part of the port area. The Regalica Crossing utilizes different types of continuous system bridge structures. Steelconcrete composite superstructures were used where falsework is inaccessible, in the major bridges M-1,-2,-3 and approach viaducts over the railway tracks E-1,-2, WD-1,-2 and WT-1. Reinforced concrete voided slab superstructures were cast in situ on formwork and falsework for WD-3,-4 and WT-2. Steel structure segments were assembled on temporary supports on land and by cantilever from barges on the water. For the major bridges, a gantry was used for the cast-in-situ deck slab segments. Different erection methods allowed the construction to be both economical and efficient. 295
Figure 4. Midspan cross sections of the major bridges for the Regalica Crossing.
Figure 5. General view of the major composite bridges for the Regalica Crossing.
Figure 3 shows the Regalica Crossing southern elevation, comprising the structures selected for construction. From east to west, the structures are as follows: – A steel-concrete composite bridge M-1 with continuous spans 60 + 90 + 116.25 + 116.25 + 90 + 63 m long. Haunched twin main girders with a middle stinger were used in the cross sections for all three bridges (Fig. 4). The road bridges carry three traffic lanes of 3 × 3.5 m and a sidewalk; the total width of each deck is 14.7 m. The total width of the tram bridge is 12.5 m. The main girder centring is 8 m for the road bridges and 7 m for the tram bridge. The reinforced concrete deck slabs are of 210 mm in depth. The bottom RC slabs are used at the pier sections to develop double composite action. The depth of the bottom slab is variable, from 200 mm to 700 mm. Normal concrete C45 is used for the slabs. The major bridges are founded on 1.5 m and 1.2 diameter RC piles, 20–23 m long. The north view of the three major bridge is shown in Figure 5. – An approach viaduct marked as WD-3, a multi-span reinforced concrete voided slab structure constructed in three sections (A, B, C) which are connected by hinges. The total number of spans is thirteen in each structure. A cross section of the RC approach viaducts is shown in Figure 6. The void formers are Spiro pipes. The width of the road viaducts is 13.72–20.69 m and the width of the tram viaduct is 11.82 m. All the viaducts are divided into three sections with expansion joints. In addition, there is a two-span viaduct to connect the crossing with the road to the Customs Bridge. The overall length of the viaducts is 320 m, with 328 m for the road viaducts (WD) and 280 m for the tram viaduct (WT). The span lengths are up to 27 m, while the deck depths are 1.3 m and 1.4 m respectively. Normal concrete C35 is used for the voided slabs. The south view of the approach viaducts (WD-3) is shown in Figure 7. – A steel-concrete composite viaduct WD-1 with continuous spans and a small cantilever to connect the voided slab span. Span length is 1.9 + 27.3 + 40 + 27.3 m. A five-girder deck with 210 mm concrete slab was built (Fig. 8). A similar structure was used for the WT-1 and WD-2 viaducts. The last one structure utilizes a seven-girder deck. The 40 m spans are over the rail tracks to the coal terminal of the port. – Steel-concrete composite viaducts E-1 and E-2 over the rail track are parallel to the previous viaducts. They are situated along auxiliary roads. Span length is 27.3 + 40 + 27.3 m. Four-girder decks with 210 mm concrete slabs were constructed. The width of the carriageways is 7 m. The structures were constructed in the first phase of the project in order to take the traffic during the next stages of the Regalica Crossing construction. The technical designs and construction materials are in accordance with Polish codes. The structural steel in the plate girders is 18G2ACu with yield strength fy = 355 N/mm2 . Reinforcement 296
Figure 6. Cross sections of the RC voided slab viaducts for the Regalica Crossing.
Figure 7. General view of the approach RC viaducts for the Regalica Crossing.
Figure 8. Cross sections of the composite viaducts for the Regalica Crossing.
bars of yield strengths fyk = 355 and 230 N/mm2 were used. Concrete grades for the decks were C45 for the major bridges and C35 for the viaducts. An ordinary Portland cement was used in the concrete mixtures. 3 THE PARNICA CROSSING The new crossing is located next to the existing railway bridge. The Parnica Crossing was constructed to improve road access to the main gate of Szczecin sea port and its customs buildings from the major dual urban thoroughfare. The crossing was constructed for one carriageway, but a second carriageway may be added in the future to also provide a connection to a planned new thoroughfare. A plan of the bridge structures is shown in Figure 9. The total length of the bridge crossing is 509 m. The major section of the crossing is a composite steel-concrete box bridge over the navigable waters of the Parnica River. The bridge is marked M-2 and is a two-span structure with a cantilever of total length 165.1 m. Two other spans were constructed to connect with the RC loop viaducts, which are marked as M2A and M0. They are simply-supported structures with spans of 51.2 m and 57 m respectively. In the future, the two spans are assigned for replacement when the extension of the crossing is constructed. On the approaches, multi-span reinforced concrete viaducts were constructed, marked as E2, E3 and E4. The E2 and M2 structures carry a 7 m carriageway and a 1.5 m sidewalk. The M2A and E4 structures carry a 6 m carriageway and a 1.5 m sidewalk. The sidewalk is on the west side. The bridge structures of the crossing are located on an interchange ramps and overpasses of the main road. The approach viaducts, both steel and concrete, are shaped in complex geometry of the crossing with allowance of a future extension. The bridge section of the project was divided into two bridge contracts. The first consisted of the main bridge sections marked as M2, M2A and M0, while the second consisted of the 297
Figure 9.
Plan of the Parnica Crossing.
Figure 10.
Parnica Crossing elevation.
approach viaducts marked as E2, E3 and E4. The contracts were executed simultaneously. The Parnica Crossing utilises bridge structures of continuous and simply-supported systems. Simple spans are used for future replacement during the crossing extension. Steel-concrete composite box superstructures were used for the major section of the crossing. The M0 span is a steel box structure with an orthotropic deck. On the concrete decks, asphalt pavement was laid. For the orthotropic deck a thin epoxy-asphalt pavement was used. Figure 10 is the Parnica River Crossing eastern elevation comprising the structure selected for construction. From north to south, the structures are as follows: – A approach viaduct marked as E-2, a multi-span reinforced concrete beam and slab structure. The total number of spans is five. A cross section of the viaduct is shown in Figure 11. The width of the viaduct is 11.6 m. The overall length of the viaduct is 119 m. The span lengths are up to 28 m, with the depth of the deck 1.35 m. The depth of the slab deck is 320 mm. Normal concrete C25 was used. – A steel-concrete composite bridge M-2 with two continuous spans and a cantilever 69.30 + 83.0 + 12.8 m (cantilever). A two-cell box girder with composite RC slab was used in the constant depth section (Fig. 12). The total width of the bridge is 11.6–20.5 m. The bridge cantilever supports two simply-supported spans, M2A and M0. To balance loads, a box counterweight is used under the cantilever. The web centring is 5.6 m. The reinforced concrete deck slab is 210 mm thick. In addition, normal concrete C25 is used for the deck slab. The major bridge is founded on 1.5 m diameter RC piles, 21–23 m long. – A connecting span marked as M2A. This is a simply-supported curved box girder with a total length of 62 m. The width of the deck is 8 m (Fig. 12). It connects the main bridge with the 298
Figure 11.
Cross sections of the approach viaducts for the Parnica Crossing.
Figure 12.
Cross sections of the box bridges for the Parnica Crossing.
western approach viaduct (E4). Span M0, which is parallel to span M2A, connects the main bridge and the eastern approach slab viaduct (E3) with a total length of 133 m including seven spans of up to 20 m. The width of the viaduct is 8–8.5 m. – A multi-span reinforced concrete beam and slab structure marked as E4. The viaduct connects the south carriageway of the main road with the bridge. The total number of spans is seven with span lengths up to 25.5 m. A cross section of the RC approach viaducts is shown in Figure 11. The depth of the cast-in-situ slab and beam deck is 1.3 m. The width of the viaduct is 9.6 −10 m. The overall length of the viaduct is 167 m. The structural and reinforcement steels are the same grades as for the Regalica Crossing. The concrete strength of the decks was C25 for all the structures. 299
4 MATERIALS In both projects, the designs and materials for construction are in accordance with Polish codes. The Eurocodes have not been fully implemented in Poland due to administrative procedures. For bridge structures, care is taken to obtain structural concretes with the required strength and durability. The minimal prescribed concrete grade was C25 for the reinforced concrete decks. Only ordinary Portland cement was permitted in the production of concrete mixtures with water cement ratio w/c ≤ 0.5. High technical specification requirements for absorbability, permeability and frost/thaw resistance of structural concrete were used for the bridge structures. A minimum depth for the concrete members was also established. For the concrete deck in the road bridges it was 210 mm. For the decks, concrete grades from C25 to C45 were used in both projects. The requirement of minimum thickness in the concrete members was also adhered to. The structural steel in the plate and box girders is 18G2ACu with yield strength fy = 355 N/mm2 and improved resistance to atmospheric corrosion. The structural steel meets supplementary requirements for toughness, with prescribed impact toughness of 29 J/cm2 at temperature −40◦ C for a Mesnager specimen. Reinforcement bars of yield strengths fyk = 355 and 230 N/mm2 are used in the reinforced concrete members. The structural and reinforcement steel grades which were used in both projects are of Polish origin. They are now being replaced by steel grades according to European standards. Slightly higher values in minimum concrete grades and minimum thickness of concrete members for road bridges are also being introduced this year. 5 CONCLUSIONS In the paper two bridge crossings in Szczecin opened to traffic at the beginning of the 21st century are presented. By improving road access to Szczecin sea port, the Regalica and Parnica Crossings have provided increased comfort and safety to road users. The Regalica Crossing in particular had been a long-awaited structure aimed at solving permanent traffic congestion on the road. Development of the city on both banks of the Oder River estuary required modernisation of the main road connecting the western and eastern districts and improved access to the sea port gates. In the bridge crossings presented, different type of bridge structures are used. Different erection methods allowed construction to be both economical and efficient. The bridge crossings utilise a wide range of spans (Collings 2013, Furtak 1999). For the spans over the water and railway tracks, steel concrete composite structures were designed. For the large spans, two girder or box girder superstructures were designed and for structures with limited depth, multi-beam superstructures were used. All the structures were designed to meet the highest highway class loading specifications in Poland. The existing Customs Bridge was repaired and its sidewalks were widened to accommodate cycle traffic (Hołowaty, 2014). This year a long-awaited tram line to the eastern districts is expected to be opened using the central section of the Regalica Crossing. REFERENCES Collings, D. 2013. Steel-concrete composite bridges. ICE Publishing: London. Furtak, K. 1999. Composite bridges (in Polish). PWN: Warszawa, Kraków. Hołowaty, J. 2009. Bridge falseworks in construction of approach viaduct for the Regalica Bridge Crossing in Szczecin (in Polish). In˙zynieria i Budownictwo No 8/2009: 441–445. Hołowaty, J. 2013. Falseworks for concrete bridge construction in subsidence of soft soils in Szczecin. In CCC 2013 – Central European Congress on Concrete Engineering – Concrete Structures in Urban Areas. Wrocław, September 4–6 2013. Wrocław: DWE. Hołowaty, J. 2014. Adapting a prestressed concrete bridge for cyclists and to provide nesting sites for house martins. In 37th IABSE Symposium Madrid 2014, Engineering for Progress, Nature and People. Madrid, September 3–5 2014. Zurich: IABSE.
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Armado Guebuza Bridge over Zambezi River in Caia, Mozambique T. Mendonça, V. Brito & M. Almeida BETAR Consultores Lda, Lisbon, Portugal
ABSTRACT: The present article concerns the project and the construction issues of the Bridge over the Zambezi River in Caia, named Armando Guebuza Bridge, in Moçambique. This bridge is located on the National Road 1, N1, and allows the connection between Sofala (in the south) and Zambézia (in the north) provinces. One can subdivide this structure in Approach viaducts and the main bridge. Both are executed in prestressed reinforced concrete. Approach viaducts are composed by three independent structures. A movable formwork – steel launching girder was considered for the Approach viaduct’s deck construction. Viaduct V1 has a total length of 546 m, viaducts V2 and V3 are both 560 m long. The total length of the Approach viaducts is, therefore, 1660 m. Excluding the first span of viaduct V1 that has 42 m, all the remaining spans have equal length of 56.0 m. The main bridge deck was built by the balanced cantilever method and has a total length of 710.0 m. Intermediate spans have 137.5 m of extension; External spans have 80.0 m long.
1 SCOPE The bridge over Zambezi River near Caia, named Armando Guebuza Bridge, in Moçambique is integrated in the national road N1and permits the connection between Sofala (in the south) and Zambézia (in the north) provinces. The conclusion of this bridge represents the end of a process initiated in 1976 when the construction was attributed to a Mozambican company called SOMOP, under a design made by
Figure 1.
Bridge location.
301
Figure 2.
Bridge structures.
Table 1. Bridge length. ID Approach Viaducts Main Bridge
Figure 3.
V1 V2 V3
Kmi
Kmf
Length (m)
Span distribution
1 + 445.00 1 + 991.00 2 + 551.00 3 + 111.00
1 + 991.00 2 + 551.00 3 + 111.00 3 + 821.00
546.00 560.00 560.00 710.00
42 + 9 × 56 10 × 56 10 × 56 80 + 4 × 137.5 + 80
Main bridge longitudinal section.
Prof. Edgar Cardoso. In 1981 construction works and the project development were totally suspended due to the Civil War. At the time only the abutments and the piles of three of the piers were already built. In 1996, Betar has developed a preliminary design for this bridge, together with Somague and Banco Mello, for construction and operation respectively. Following the Road National Administration (ANE) tender, in 2005 won by the consortium composed by WSP/Grid/Louis Berger Group has developed the bridge basic design and gatherer the construction supervision. One year later, in 2006, the consortium Mota-Engil/Soares da Costa has won the tender for the bridge construction. Betar Consultores was then chosen to prepare the detailed design. 2 BRIDGE DESCRIPTION The bridge is subdivided in 4 independent structures with the following characteristics: This project also includes the rehabilitation of the national road N1 within 5 km. Since different structural solutions and construction methods were adopted for the Approach viaducts and the bridge, their description is presented independently. 2.1 Main bridge The bridge has a total length of 710 m split in four intermediate spans of 137.5 m long and two external spans of 80.0 m. 302
Figure 4.
Main bridge – Deck cross-section over the piers.
Figure 5.
Main bridge – Deck cross-section at mid-span.
Figure 6. Typical cantilever longitudinal section.
Figure 7.
Main bridge – pre-stress cables layout.
2.1.1 Deck The box girder cross-section adopted for the bridge varies in height, with 7.50 m near the piers, and 3.50 m at mid-span. The deck’s webs have 0.70 m thick near the piers and 0.45 m at mid span. A linear variation was adopted for the web thickness transition. This occurs in 3 segments on each cantilever. The deck’s bottom flange thickness varies parabolically until the 10th segment between 1.10 m over the piers and 0.22 m. 303
Figure 8.
Main bridge piers cross-sections.
Figure 9.
Main bridge – Piers PB.2, PB.3, PB.5 and PB.6.
The bridge deck was constructed using the balanced cantilever method. The first segment have 9.30 m long. The next 14 segments have different lengths: segments #2 and #3 have 3.50 m; segments #4 to #6 have 4.20 m; segments #7 to #10 have 4.50 m; segments #11 to #15 have 5.00 m. The bridge deck is longitudinally pre-stressed using 3 different cable families: the balanced cantilever cables in the top, tensioned as each segment is executed; the continuity cables located in the bottom flanges, tensioned after the cantilever locking is done, and finally external pre-stressed cables. 2.1.2 Piers An hexagonal cross-section, with exterior dimensions of 7.00 × 4.70 m was adopted for the piers. Piers walls have a constant width of 0.70 m. To improve the hydraulic behaviour, in the bottom piers cross-section becomes elliptic with 8.80 × 5.40 m. Bridge piers are founded using cast-in-place piles with 2.0 m diameter. Piers PB2, PB3, PB5 and PB6 have 6 piles each. PB4 has 8 piles. 304
Figure 10.
Main bridge – Pier PB.4.
Figure 11. Approach viaduct V1 longitudinal section.
Figure 12. Approach viaduct V2 longitudinal section.
2.2 Approach viaduct Approach viaducts have a total length of 1666 m subdivided in three independent structures, V1, V2 and V3. V1 has 546 m long, V2 and V3 have 560 m long. The current span length is 56.0 m. There is only one exception that is the first span of viaduct V1 that has 42.0 m long. 2.2.1 Deck The deck is composed by a box girder cross-section with constant height equal to 3.50 m. 305
Figure 13. Approach viaduct V3 longitudinal section.
Figure 14. Approach viaduct deck cross-section at piers.
Figure 15. Approach viaducts piers cross-section.
Over the supports diaphragms were adopted. In South abutment and transition piers alignment diaphragms are 1.0 mm thick and at the other piers diaphragms are 0.80 m thick. For maintenance and inspections purposes, openings were left in the diagrams. Longitudinal pre-stress in the deck webs is composed by 16 strands 0.6 (1.4 cm2 ) cables with parabolic layout. 56.0 m end spans were strengthen with additional straight cables. 2.2.2 Piers A hexagonal box cross section was adopted for piers. Each pier is supported by 4 piles of 1.50 m diameter. Pile caps have the following dimensions 10.00 × 6.75 × 2.50 m.
3 CONSTRUCTION PROCESSES Two different constructions systems were adopted. For the bridge it was considered a balanced cantilever method using two pair of cars. The Approach viaducts deck was built using a steel launching girder. 306
Figure 16. Approach viaducts piers front and side views.
Figure 17.
Main bridge construction stages – example for stages 2 and 7.
Figure 18. Approach viaducts construction stages – representation of the first 5 stages of the deck assembly.
4 CONCLUSIONS Betar Consultores has used all its knowledge, expertise and commitment for designing this bridge. Considering the specificity and complexity of the task embraced. For several times the specialists involved in this ambitious project had to developed and solve complex and sensible issues during construction that required all their determination and effort. 307
Figure 19.
Bridge photograph after construction.
Developing the bridge detailed design has required the mobilization of numerous specialists for several years, allowing essential bridge for the Mozambican economical competitively. This communication is dedicated to everyone involved.
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Design and construction of a long-span continuous fin-back bridge Y. Lu, M. Fu, X. He & C. Zhou Shanghai Urban Construction Design and Research Institution, Shanghai, China
ABSTRACT: Fin-back Bridge is a novel bridge type in China, which consists of the girder and the fin as the bearing components. Dazhi River Bridge in Shanghai is the first long-span prestressed concrete continuous fin-back bridge with wide deck in China. It has the merits such as low building altitude, high stiffness, full prestressed concrete, etc. Besides, balanced cantilever construction method is applicable for the bridge. To popularize the bridge type and share design experience with other engineers, the general arrangement, design of the superstructure and infrastructure, seismic measures, landscape design, and the key construction technology of the bridge are introduced.
1 INTRODUCTION Lianggang Road located at Pudong New District in Shanghai is the main road connecting Pudong International Airport and the New Town of Lingang. Dazhi River Bridge on Lianggang Road is the first long-span prestressed concrete continuous fin-back bridge in China. To coordinate with the surrounding environment and meet the demand of the traffic condition at the bridge site, the proposed bridge should have large span, low building height, easy maintenance and beautiful landscape. A three-span prestressed concrete fin-back bridge is adopted after a scheme comparison, which perfectly meets the above requirements. 2 DESIGN 2.1 General arrangement Dazhi River Bridge is a prestressed concrete continuous fin-back bridge with three spans of 92 m, 158 m and 92 m. The deck width of the bridge is 35 m with eight-lane in both directions. Two fins over the deck are set continuously along the bridge. The outline of the fin along the bridge is parabola with the maximum height of 20 m. The general arrangement of the bridge is shown in Figure 1. In order to increase the stiffness of the structure, the fin containing prestressing tendons is set above the bridge deck rather than increase the effective depth of the section, so that the efficiency of the cross-section over the piers increases greatly (Gee, 1991). Meanwhile, most of the prestressing tendons are arranged in the fin, which increases the bending and shearing resistance of the cross section greatly, hence the problem of deflection and cracking of long-span concrete bridges is solved. Eventually, the engineering quantity decreases about thirty percent, which makes the construction cost more competitive. 2.2 Superstructure The superstructure consists of the fin and the box girder. The cross-sections are indicated in Figure 2. Four-cell box girder with large cantilever and inclined webs are used. The deck and the bottom width of the box are 34.7 m and 14 m, respectively, while the height of the box is 3.5 m. The depthspan ratio of the bridge is 1/45. Transverse beams are set in the box at the pier while diaphragms are 309
Figure 1.
General arrangement of the bridge (unit: m).
Figure 2. Arrangement of the cross section (unit: cm).
set elsewhere. The interval between two diaphragms is 6 m. The fin is set at the median separator of the bridge. Its longitudinal outline is parabola with the effective height of 15 m at the middle pier. The width is 1.6 m at the top of the fin and 1.9 m at the junction with the box girder. Concrete with hollow cross section is cast at the top of the fin, which does not bear load and only be used for decoration. 310
Figure 3. Arrangement of prestressing tendons (unit: cm).
Figure 4.
Layout of the foundations (unit: cm).
Stranded wires and fine rolled twisted bars are used as the prestressing tendons. They are protected by plastic corrugated pipe and iron sheet tube, respectively. The major tendons are arranged in the fin, and they are all stranded wires. Three-dimensional prestressing tendons are set in the box girder, which consist of stranded wires and fine rolled twisted bars. The arrangement of prestressing tendons is shown in Figure 3. 2.3 Infrastructure Solid piers are adopted. The supports are directly set at the top of the piers. In order to meet the requirements of installation of precast approach bridge, T-shape prestressed concrete cover beam is set at the side pier. Cast-in-place pile foundations are adopted. The diameter and length of the pile at the middle pier are 1500 mm and 70 to 71 m, respectively. Plane dimension of the corresponding cap is 24 m wide and 31 m long. The thickness of the cap is 5.5 m with two layers (2.0 m and 3.5 m). The diameter and length of the pile at the side pier are 1200 mm and 54 m, respectively. Plane dimension of the corresponding cap is 11.8 m wide and 31 m long. The layout of the foundations is shown is Figure 4. 2.4 Seismic design Because of the low altitude and large quality of the pier, seismic isolation and reduction method is needed. Response spectrum and nonlinear time history analysis (TU, 2011) based on the finite element method was conducted. After doing research on the seismic behavior of the bridge, longitudinal E-type damper is set on the fixed pier to decrease the earthquake response. Compared to that no damper is set, the shearing force and flexural moment at the bottom of the fixed pier 311
Figure 5.
Segment division of layered construction.
decrease thirty-three percent, while the shearing force and the flexural moment of the pile decrease above fifty percent. 2.5 Landscape design The size of bridge structure varies with the bearing load. Beauty of mechanics is the main characteristics of bridge aesthetics, so the bridge not only has proper bearing capacity, stiffness and stability, it also needs to meet people’s aesthetic demand and be harmony with the surrounding environment. Firstly, the configuration of the fin is introduced. The surrounding terrain is very flat and the building height of the bridge is very low, which makes the fin seem like “distant hill”. The fin itself is not only a complete landscape but also is harmony with the surrounding environment (Sheng, 2011). Through modelling, the eventual outline of the fin looks like a trapezia, which is 1.6 m wide at the top and 1.9 m wide at the bottom. The slope of the fin gives one a feel of going upward and relieves the sense of weight of the structure. Then it is the wall decoration of the fin. The simplest vertical lines are used for decoration to ensure the safety of driving. The vertical lines are groove shaped. The width and depth of the groove are 60 cm and 20 cm, respectively. Dark coating is used, which is in sharp contrast to the beam and makes the bridge be of rich levels. Lastly, nightscape lighting is presented. The nightscape lighting should meet the demands of the road lighting specification and traffic safety as well as energy conservation and equipment maintenance. Decorative lightings are installed on the fin and at the bottom of the highway guardrail. LED downlights are set on the piers. The light casting from bottom to top highlights the origin design concept and demonstrates the rhythm of irregularity. 3 CONSTRUCTIONS 3.1 Forming process of the fin The fin and the box girder are constructed synchronously, so the forming process of the fin has effect on the mechanical properties and construction difficulties of the bridge. Layered construction is mostly adopted for this bridge type built abroad (Fig. 5). However, many problems exist in layered construction as follows: (1) Superimposition of multilayers increases the complexity of the structure performance. (2) Pre-tensioned prestressing tendons do not provide pressure stress for concrete cast later, so more prestressing tendons is needed. (3) The construction method is complex, so the junction surface between two layers is easy to crack. 312
Figure 6.
Segment division of this bridge.
New construction methods by dividing the whole bridge into several parts and dividing the local region into several layers are firstly adopted for Dazhi River Bridge (Fig. 6). Firstly, a 60 m long segment of box at the middle pier was constructed, and then the fin above the box was cast layer by layer. Afterwards, cast the fin with two segments of the box girder each time. This method has the following advantages: (1) The stress state of the cross section is clear, so it is easy to analyze the actual stress state to ensure safety of the structure. (2) Not like the layered construction, no new equipment is needed. (3) Much less joint surfaces than the layered construction reduces the complexity of construction.
3.2 Cantilever construction method The balanced cantilever construction method covers the temporary anchorage between the girder and pier, cantilever casting and closure technique of the bridge. 3.2.1 Temporary anchorage Temporary anchorage should bear the tensile stress as well as the compressive stress caused by bending moment. To meet the requirement, the following measurements are adopted: (1) Three reinforced concrete square columns are set at each side of the box girder to bear compressive stress caused by bending moment. (2) Beside each column, four fine rolled twisted bars are set from the top of the box girder to 1.5 m depth of the pile cap to transfer tensile stress. (3) The permanent support is fixed to bear force with the columns together during construction, and no temporary support is set. 3.2.2 Cantilever casting Four hanging baskets are used during cantilever casting (Fig. 7). The hanging basket consists of load-bearing system, suspender system, running system and formwork system, which weighs about 95 tons in total. Six longitudinal beams and four transverse beams are set at the top of the box girder. During cantilever casting, the maximum load difference between both sides of the box girder is limited to 150 tons. The middle span of the bridge is divided into twenty four segments. Segment zero is cast in place on brackets, while segment one to segment twenty-two are constructed by cantilever casting. Segment twenty-five is the closure segment. The side span is also divided into twenty four segments. Segment twenty-four is the closure segment and segment twenty-three is cast in place on brackets. Other segments are constructed by cantilever casting. 313
Figure 7.
Elevation drawing of the hanging basket.
3.2.3 Closure of the bridge According to the design requirements, closure of the side span comes first. Then temporary anchorage is removed. Finally closure of the middle span is carried out. The following measures are adopted to avoid shrinkage crack of the closure segment: (1) Counterweight with the same weight of the closure segment is added to the cantilever end to guarantee stability of the structure during concreting. (2) The moment to remove the temporary connection between the closure segment and the adjacent segment is precisely chosen to make the temporary connection be in compressive state under temperature stress. (3) Longitudinal restraint of the support is removed immediately after the temporary connection between the closure segment and the adjacent segment is erected. (4) The time for wet curing is greater than 7 d to avoid shrinkage crack of the joint between new and old concrete. 3.3 Prestressing technology The height difference and tensioning force of prestressing tendons in normal continuous bridges are generally very small. However, most of the prestressing tendons in the fin are very long and have large height difference. The length and height difference of the longest tendon in the fin reach 156 m and 20 m, respectively. In this project, special prestressing technology are adopted. It includes the tendon passing through, stretching and grouting. 3.3.1 Tendon passing through the pipe Push-type tendon duct machine is used to push tendons into the pipe one by one. The construction process is as follows: (1) Push one tendon from one side to another. Ensure that the steel strand does not bend in the pipe. (2) Number every steel strand. Ensure that both ends of the tendons are stretched at the same time to decrease the frictional resistance. (3) Repeat above steps until all strands are pushed into the pipe. 3.3.2 Stretching of the prestressing tendons Stretch the prestressing tendons by jack. For long prestressing tendons, adjust initial stretching force and increase the load duration time to reduce the loss of prestress. Improve the stretching method by contrasting the values between tube friction test (Wang, 2014a) and the theoretical 314
calculation. After improvement, the prestress loss decreases a lot, and no additional equipment is needed. 3.3.3 Grouting Grouting test (Wang, 2014b) of the full size tube model is carried out. Based on the test results, the grouting method is improved by grouting from both ends of the pipe and ejecting mortar from multiple vent holes. The vent holes are set at the top of the girder to ensure compactness of the grouting. Compared to regular grouting method, the improved method increases the compactness of the grouting greatly. 4 CONCLUSIONS Dazhi river bridge is completed and comes into service in December 2012. Based on the experience of design and construction, the following conclusion is drawn: (1) Compared to regular continuous bridge, the building height of the fin-back bridge is much smaller, and the engineering quantity decreases about thirty percent. Hence the bridge is very economical. (2) Compared to regular continuous bridge, the stiffness and the prestressing efficiency of the bridge are much higher, so the problem of deflection and cracking of the bridges is solved. (3) The bridge is beautifully shaped and in harmony with the surrounding environment. (4) Based on the mature balanced cantilever construction method, new forming process of the fin and new prestressing technology are introduced, which guarantees the smooth construction of the bridge. REFERENCES Gee, A. 1991 Concrete fin-back bridge in USA. ICE Proceedings. London: Thomas Telford. Sheng, H. 2011 Aesthetics of bridge architecture, Beijing: China Communications Press. TU (Tongji University) 2011 Research report on the seismic performance of Dazhi River Bridge. Shanghai, China: Tongji University. Wang, S. 2014a Frictional loss test of plastic prestressed pipe. China Municipal Engineering 2014(1): 18–19. Wang, Z. 2014b Study on grouting compactness detection of prestressed duct. China Municipal Engineering 2014(1): 30–32.
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Pinhal Interior Motorway Concession – IC3 – Section Condeixa – Coimbra – Special engineering structures – Construction processes T. Nogueira, A. Hipólito & N. Amaro
Mota-Engil Engenharia e Construção SA, Linda-a-Velha, Portugal
ABSTRACT: The paper describes the construction process of the Special Engineering Structure bridges/viaducts, part of the 1st section of IC3 Condeixa/Coimbra, Pinhal Interior Motorway Concession.
1 INTRODUCTION The 6 bridges/viaducts have an overall length of 2600.5 m, with two independent decks and current spans that vary between 37 m and 52 m (Fig. 1). The maximum height of the viaducts/bridges piers varies between 50 m and 80 m. Five bridges/viaducts were constructed span by span, cast in situ with Movable Scaffolding Systems (three overhead and one under slung) and one with prefabricated beams. Structural Designers of each viaduct: – – – – –
Vale do Inferno: Braçais: Corvo: Sobral: Ceira:
TRIEDE ARMANDO RITO Engenharia NSE Engineering QUADRANTE Engenharia e Consultoria LCW Consult
The viaducts substructures generally comprise reinforced concrete piers, with flared pier heads to support the deck, and reinforced abutments. Each viaducts were constructed span-by-span with vertical joint along the spans, 20% of the typical span. On each viaduct the MSS rear end was supported on the front cantilever in the overhead MSS case, and suspended from the cantilever in the underslung MSS case. In both underslung and overhead, the front support was mounted on the piers.
Figure 1. Viaducts along the route.
317
Figure 2.
Elevation view and cross-section of the deck.
Figure 3.
Cross section of the MSS; Kinematic.
2 VIADUCTS 2.1 Vale do Inferno (Fig. 4) The viaduct has a total length of 390.0 m distributed over 8 spans of 39.0 m + 6 × 52.0 m + 39.0 m, with a maximum height of 65.0 m above ground level. The superstructure is fully continuous in its total length and it is made of two parallel decks with a total width of 29.6 m. Each deck has two longitudinal beams 3.35 m high, connected by a reinforced concrete deck slab (Fig. 2). The average deck weight per meter is approximately 24 t. The MSS operated downhill and uphill with 2% longitudinal slope and a 6.0%–2.5% transverse slope. The underslung MSS used was of the double box girder type, with 116 m in length (front nose 30 m + main girder 56 m + rear nose 30 m), and supports on brackets anchored to the sides of the piers by stressed bars. The approximate total weight of the MSS was 650 t (Fig. 3). 2.2 Braçais (Fig. 7) The viaduct has a total length of 392.5 m distributed over 11 spans of 31.0 m + 9 × 37.0 m + 28.5 m, with a maximum height of 60.0 m above ground level. The superstructure is fully continuous in its total length and consists of two parallel decks with a total width of 29.6 m. Each deck has two longitudinal beams 2.6 m high, connected by a reinforced concrete deck slab (Fig. 5). 318
Figure 4.
Overview.
Figure 5.
Elevation view and cross-section of the deck.
The average deck weight per meter is approximately 19 t. The MSS operated uphill with 4.8% longitudinal slope and a 7.0% transverse slope. The overhead MSS used was of the single truss type, with 100 m in length (front nose 27 m + main girder 45 m + rear nose 27 m), and supports on towers anchored to the pier caps that resist uplift forces. The approximate total weight of the MSS was 530 t (Fig. 6). 2.3 Corvo (Fig. 9) The viaduct has a total length of 378.5 m distributed over 8 spans of 45.5 m + 5 × 52.0 m + 41.0 + 32.0 m, with a maximum height of 80.0 m above ground level. The superstructure is fully continuous in its total length and it is formed by two parallel decks with a total width of 29.6 m. Each deck has two longitudinal beams 3.35 m high, connected by a reinforced concrete deck slab (Fig. 8). The average deck weight per meter is approximately 24 t. The MSS operated uphill with 4% longitudinal slope and 7.0% transverse slope. 319
Figure 6.
Cross section of the MSS; Kinematic.
Figure 7.
Overview.
Figure 8.
Elevation view and cross-section of the deck.
320
Figure 9.
Figure 10.
Overview.
Elevation view.
The underslung MSS used was the same as for Inferno. 2.4 Sobral (Fig. 13) The viaduct has a total length of 249.0 m distributed over 7 spans of 23.0 m + 5 × 40.0 m + 26.0 m, with a maximum height of 60.0 m above ground level (Fig. 10). The superstructure is fully continuous in its total length and it is formed by two parallel decks with a total width of 29.6 m. Each deck has two longitudinal beams 2.63 m high, connected by a reinforced concrete deck slab (Fig. 11). The average deck weight per meter is approximately 19 t. The MSS operated uphill with 4% longitudinal slope and 7.0% transverse slope. The overhead MSS used was of the single truss type, with 75 m in length (front nose 30 m + main girder 45 m), and supports on towers anchored to the pier caps that resist uplift forces (Fig. 12). The approximate total weight of the MSS was 580 t. 321
Figure 11.
Cross-section of the deck.
Figure 12.
Cross section of the MSS; Kinematic.
Figure 13.
Overview.
2.5 Ceira (Fig. 17) The engineering structure is made of two distinct structures: the crossing over the River Ceira and the access viaduct. Geometrically, the work is S-shaped and each of the structures shows always a same-side slope, meaning that the transition pile between both structures corresponds to an inflexion area. Crossing over the River Ceira is made by a multiple frame bridge with continuous pre-stressed concrete deck between abutment E1 and pile P4 extending over 582 m and spans of 65, 140, 250 and 128 m (Fig. 14). The deck is a single-cell box-girder deck with inclined webs varying between 322
Figure 14.
Elevation and cross-section of the deck; Bridge (left) and Viaduct (right).
14.5 m of height on piles P2 and P3 and 5.5 m of height on the central closure joint area of the span between those pies and on sections between abutment E1 and pile P1 and next to pile P4. The access viaduct is made of a continuous pre-stressed concrete beam slab with 4 beams extending over 345.75 m distributed over 7 spans of 36.7 m + 7 × 40.0 m + 30.0 m. 2.5.1 Viaduct (span by span) The viaduct has a total length of 346.75 m distributed over 9 spans of 30.0 m + 7 × 40.0 m + 36.75 m, with a maximum height of 50.0 m above ground level. The superstructure is fully continuous in its total length and it is formed by one deck with a total width of 26.4 m. The deck is constituted by four longitudinal beams 3.0 m high, connected by a reinforced concrete deck slab. The average deck weight per meter is approximately 22 t. The MSS operated uphill with 2.6% longitudinal slope and 2.5–7.0% transverse slope. The overhead MSS used was of the single box-girder type, with 92 m in length (front nose 23 m + main girder 46 m + rear nose 23 m), and supports on towers anchored to the pier caps that resist uplift forces. The approximate total weight of the MSS was 400 t (Fig. 15). 2.5.2 Bridge (balanced cantilever) Each 125 m long cantilever is divided into 21 dowels, plus the pier head and closure stitches. Dowels vary between 3.5 m and 5.0 m in length (Fig. 16). The closure stitch is 4.0 m long. Maximum dowel weight is 490 t. The construction of the deck was carried out by form travellers, on each cantilever – two pairs in total. The capacity of each form traveller was 500 t (concrete) and 100 t self-weight. 3 CONCLUSION On a final note, one should mention the technical follow-up of the project during design stage of the engineering structure, covering features below: – Layout development; – Cross sections of the shafts of columns; 323
Figure 15.
Cross section of the MSS; Kinematic.
Figure 16.
Distribution of dowels for 125 m long cantilever.
Figure 17.
Overview.
– Length of viaduct spans and cross sections; – Modeling of length of the bridge dowels and maximum self-weight. Among other features, those mentioned had an impact on the development of the construction job and they were adjusted to available means during design/project stage wherever possible. Action taken in conjunction with all parties involved (client, design engineer, technical management, site staff) enabled an efficient planning of construction process as well as resorting to means and equipment available at that time to cope with a highly complex and technically demanding project.
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Viaduct Araranguá – The alternative design of viaduct of 1661.59 meters in the BR-101/SC Brazil I.C. Santos Universidade de Brasília (UNB), Faculdade de Engenharia da Universidade do Porto (FEUP) and Departamento Nacional de Infraestrutura de Transportes (DNIT), Brasília, Brasil
F.P.S. Nunes Departamento Nacional de Infraestrutura de Transportes (DNIT), Brasília, Brasil
ABSTRACT: The technical solution initially planned for the project for Contour of City Araranguá in BR-101/SC was balance berms. Thus, the embankment with balance berms would begin on the boundary of the crossing to the riverbank Araranguá. With a 2.160 m length extension, and that because of the low support capacity of the ground, with a width of up to 250 m and height of 5 m. This embankment would develop in a region over rice crops with high productivity, and land worked by farmers for decades. Another problem caused by berms would be the effect of a dam caused by the embankment, since the planned bridge over the river Araranguá would be the only way out of the water. However, National Department of Transportation Infrastructure – DNIT proposed an alternative project, replacing the balance berms for a viaduct with metal columns and pillars, crossbars, beams and deck in reinforced concrete. The viaduct was designed as an alternative solution for the replacement of balance berms was designed in reinforced concrete, Class 45 as defined in Brazilian Standard ABNT NBR-7188 to load 45 tons, and double track. The viaduct consists of two segments, connected by the existing viaduct in the middle the stretch, the first section is 587.37 m long, and the second section is 1072.22 m long, with a total length of 1661.59 m. The isotactic structural system consists of multiple trusses of five longitudinal beams, supported transversely for two crossbeams and a slab of reinforced concrete, molded in situ. The deck is 24.10 m wide; with two lanes of 7.20 m each, two road shoulders of 1.10 m, and the extremities is barriers of 0.425 m and double barrier of 0.65 m in the center. The mesostructure is composed from transverse blocks constant section, and support for the main beams. The foundations are metal columns in laminated steel profile. The solution viaduct certainly was the most suitable for the situation described. Technically, it is not subject to any risk of consolidation, deformations and settlements of land, which would require continuous maintenance. The viaduct is rigid structure, easily defined and quantified, unsurprisingly during its execution, and maintenance costs will be lower than other solutions. This solution reduced practically zero the socio-economic and environmental impacts in the region. The change of the solution still is by public interest having the support of the local population, and is based by the principle of efficiency that should govern public administration.
1 INTRODUCTION 1.1 History In May 1998, the DNIT started the engineering design for the works to expand the capacity of the road BR-101/SC in the stretch by Km 411 and Km 437, in November 1999 the project was completed. The first contract for the works of the stretch known as the Contour of City Araranguá, was signed on 30/11/2004, the service order issued on 02/12/2004. The work had beginning in 325
March 2005, however, in June 16, 2006 the contractor requested the termination of the contract. Completed the process of termination on 03/05/2007 was signed new contract for the works, with the second place in the bidding, and the services was started under the new contract on 05/05/2007, this contract was also terminated when a small portion the work had been performed in October 2010. After the second contract termination, the supervisor of the project produced new project, compiled with the remaining quantity and by decision of the DNIT, the segment 409 km to 411 km, was changing the original solution in embankment with balance berms for solution viaducts, the location of works is appointed in Figure 1.
Figure 1.
Situation map.
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1.2 Overview The project was compatible with the Functional Plan and developed by Traffic Studies preliminary, as well as with the functional class was established according to criteria and concepts of DNIT Rules. In development of the project considered the functional aspect of the highway, the occupation of the existing size the current track, reducing the environmental impact and minimizing deployment costs. The highway, because its functional class and traffic volumes was framed in technical characteristics of Class IA, according to the classification of DNIT Rules. The project was developed in two tracks, provides geometric conditions for the development of speed of 100 km/h. 2 STUDIES AND SOLUTIONS 2.1 The embankment solution (original) The proposed solution of original design, were the implementation of two embankments, this would develop in a region over rice crops with high productivity, and low ground resistance, in accordance Figure 2. For the implementation of the embankment was considered that when ready it will not have settlement, caused for low bearing capacity of the ground of the region. Thus, for the implementation of the original design would need to expropriate an area of 56 ha, divided in 12 properties, where the estimated cost for expropriation could reach 10 million dollars. The construction would affect other areas of the segment along the highway, since the embankment would cause a discontinuity of the existing irrigation channels that control the water level and ensure the irrigation of crops. The large proportions and weight of the berm would create an excess load on the ground, which would cause grounding of the region in neighboring and therefore, the rice planting activity in the region would be impossibility. The original solution foresaw the densification of the berm in 12 months, but the experience showed a high risk that the end of this period the embankment would not have the support needed to get the paving, which would increase the time and costs provided for the work.
Figure 2.
Projection area of embankment with balance berms.
327
Figure 3.
Structure cross section.
Figure 4.
Construction of the viaduct – Top view.
2.2 Solution viaduct The DNIT proposed an alternative project, replacing the balance berms for a viaduct with metal columns and pillars, crossbars, beams and deck in reinforced concrete. The viaduct was designed as an alternative solution for the replacement of balance berms was designed in reinforced concrete, Class 45 as defined in Brazilian Standard ABNT NBR-7188 to load 45 tons, and double track. The reinforced concrete viaduct adopted have the superstructure compost of 33 spans, with a width of 11.50 m and length of 33.0 m. The cross section consists of 336 precast beams and slab 15.0 cm thick, consisting of mesostructure. The infrastructure comprising of reinforced concrete beams and pile in metal profile W 360 × 110, as shown Figure 3, 4 and 5. 2.3 Modeling and structure calculation The structural design was developed by Enescil Engineering. The cross section was patterned (3D) considering the stiffness by applying a uniform force on the top of the beam. Thus, with the stiffness were determined portion of the loading force corresponding 328
Figure 5.
Construction of the viaduct – Bottom view.
Figure 6.
3D modeling and porch Bars.
to each horizontal support, added to these horizontal and vertical forces adopted in superstructure calculation memory, done that were applied every effort advocated in Norma. In the top of the track, acting simultaneously with the respective weighting the combinations commonly used for this type of structure where longitudinal and transverse efforts occur. The beams designed considering the live load acting on the most unfavorable positions, maximizing positive and negative moments. The same were done with the blocks. The software used for the structural analysis was STRAP, as shown in Figure 6. 2.4 Advantages of proposed viaduct Initially the term of the project was set at three years using the solution in embankment with balance berms, the alternative project reduced the time to 18 months. 329
Regarding the environmental impact of the project in the solution berms would need a volume of 2,600,000 m3 of sandy material originating from fields, already the solution in viaduct would have less impact, only of concrete blocks, pillars, beams, crossbeams and board beam. The solution berms would be necessary to expropriate large areas of rice planting that would affect the economy of the region where the project would be built already viaduct solution would bring minimal impacts on the regional economy. The great movement of trucks to transport 2.6 million m3 would cause logistical problems for loading and unloading of the material may result in accidents and delay in the works. 3 CONCLUSION After analyzing the schedules and financial impact DNIT decided by the viaduct in solution whose impact financial was approximately 16% over the solution in embankment with balance berms, however, the reduction in run time was crucial for choosing the solution viaducts. The solution viaduct certainly was the most suitable for the situation described. Technically, it is not subject to any risk of consolidation, deformations and settlements of land, which would require continuous maintenance. The viaduct is rigid structure, easily defined and quantified, unsurprisingly during its execution, and maintenance costs will be lower than other solutions. This solution reduced practically zero the socio-economic and environmental impacts in the region. The change of the solution still is by public interest having the support of the local population, and is based by the principle of efficiency that should govern public administration. Currently DNIT is opting for solutions viaduct instead of adopting more conservative geotechnical options this is due in part to the success in the implementation of project the Contour of City Araranguá. REFERENCES ABNT:NBR-6118. 2014. Projeto de estruturas de concreto — Procedimento. ABNT – Associação Brasileira de Normas Técnicas. São Paulo: ABNT – Associação Brasileira de Normas Técnicas. ABNT:NBR-7187. 2003. Projeto e Execução de Pontes em Concreto Armado e Protendido – Procedimento. ABNT – Associação Brasileira de Normas Técnicas. São Paulo: Associação Brasileira de Normas Técnicas. ABNT:NBR-7188. 1984. Carga móvel rodoviária e de pedestres em pontes, viadutos, passarelas e outras estruturas. ABNT – Associação Brasileira de Normas Técnicas. São Paulo: ABNT – Associação Brasileira de Normas Técnicas.
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Design and construction of flyovers in Outer Ring Road, Delhi K. Ganesh & V. Shanmugham L&T Infrastructure Engineering Limited
ABSTRACT: The four elevated flyovers in Delhi have a cumulative length of about 15 km which are constructed to ease the traffic in outer ring road. The width of 6 lane deck with edge barrier is maximum 24.20 m, with standard spans of 35 m. There are special spans such as varying width and curved spans to cater for mergers and geometrical requirements. The deck is of spine beam segmental construction, which is longitudinally and transversely prestressed. Deck is supported on POT cum PTFE bearings. Expansion joints are provided at every third span, i.e at every 105 m, to ensure smooth riding surface. The substructure consists of single pier, supported on pile foundation.
1 INTRODUCTION As a part of improvement of urban infrastructure, and to ease the traffic congestion on the ring road, Government of New Delhi had decided to build four flyovers to cater for all the at-grade junctions, and to make the flow of traffic, signal free. With the construction of this elevated corridor, the travel time from Mukarba Chowk to Vikaspuri would be shortened and signal free passage for approximately 15 km would be possible. The project is implemented by Delhi Public Works Department, with L&T Infrastructure Engineering Limited as its detailed design consultant, and construction contract was item-rate. 2 GENERAL ARRANGEMENT 2.1 Development of concepts The following criteria were governing the selection of structure type.
Figure 1.
Mukarba Chowk to Mangolpuri.
331
Figure 2.
Madhuban Chowk to Mangolpuri.
Figure 3.
Meerabagh to Vikaspuri.
a) Construction friendly, including faster output. The proposed elevated corridor passed through one of the main arterial roads of Delhi. The complete stretch had developments on either side and hence the traffic could not be blocked for construction. The construction needed to be taken up with traffic plying below. The construction period of this elevated highway was initially set as 3 years. Transportation of wide segments was not possible in the project road. So the efforts were channelized for making the standard system as far as possible, so that mass production is achieved without any significant variation either in dimension, or in reinforcement detailing as far as possible. The in-situ works were minimised. The production was continuous through all the seasons. More over since the width of 24.6 m was beyond the limit of casting, handling and transporting, the system should be segmental in the transverse direction also. Thus a segmental spine and wings precast superstructure was conceptualized. This would help in minimising the site works and would help in handling the segments safely with the traffic plying below. In order to keep the at-grade roads operational, it was decided to adopt a system of single pier with widened pier cap. 332
Table 1. Details of flyovers. Stretch
Nomenclature
Length of corridor (km)
Stretch No. 1 Stretch No. 2 Stretch No. 3 Stretch No. 4
Mukarba Chowk to Madhuban Chowk Mangolpuri to Madhuban Chowk Vikaspuri to Meerabagh Azadpur to Prembaripul
3.84 4.19 4.41 2.46
Figure 4. Typical cross section.
b) Aesthetics As an urban structure, which is going to act as a gateway to the one of world’s best airports, developers were very keen to make the structure elegant. The structure need to be unique, iconic to the city and shall necessarily blend with the urban backdrop. Any structure which is sleek and has less concrete in it, would be more appreciated than a massive one. So an aesthetically treated central spine beam with sleek wings on both sides was chosen to satisfy all the above requirements and design provisions. 2.2 Layout and cross section The total stretch is divided into four stretches, from the point of execution control. They are as follows. The typical span is 36.0 m. This span was kept constant throughout all the stretches and minor adjustments were done for at-grade junctions. In order to ensure proper drainage, a gentle longitudinal slope of 0.3% is also provided. In order to reduce the number of expansion joints and to have a better driving comfort, tied decks are provided for three spans, and EJ at every third pier. The cross section of the flyover is as given in the Figure 4. The cross section consists of central spine of about 6.0 m, and two equal wings of 9.1 m each. The ribs are kept at a spacing of 3.0 m, which is found to be ideal both from casting and design point of view. The ribs were stitched to the central spine by transverse stressing. While match cast technique was adopted for spine segments, in-situ stitch was adopted for connection between the wings and between the wing and spine. Bidirectional camber of 2.5% is provided in the deck for the purpose of drainage. The overall width of the superstructure is 24.2 m, which includes two carriage ways of 11.20 m each central median of 1.0 m. The system of substructure consists of a pair of POT & POT-Cum-PTFE bearings for each superstructure, supported on a piercap, flared from a single pier. Foundation system is a group of 1.20 m dia pile, length depending on the type of subsoil at each location. There is one locations where the bridge crosses a stream, in which the existing bridge to be retained and additional at-grade bridge to be constructed. This was a challenge for planning the 333
Figure 5.
Cross Section at river crossing.
geometry, and including the construction methodology to be adopted. The cross section of the bridge at this location is given in Figure 5. Due to the various constraints of existing at-grade bridge, portal type of pier is adopted. Steel concrete composite superstructure is adopted at this location. 3 DESIGN ASPECTS 3.1 General The design is carried out as per Indian Roads Congress (IRC) codes of practice. The following loads are considered for the design as per IRC-6. 3.1.1 Live load Live load types of Class 70R, with a single train totaling to 100 MT occupying two lanes and Class A, totaling to 55.40 MT occupying single lane, individually or in combination to give the maximum effects on various elements are considered. 3.1.2 Materials Grade of concrete for superstructure and crash barrier/median adopted is M60 and grade of steel is Fe 500. Prestressing steel is of nominal diameter 15.24 mm, conforming to IS: 14268 class II, with yield strength 1860 Mpa. 3.2 Method of analysis and design The following procedure has been followed for the design of spine beam. Since the deck is simply supported for all the gravity based loads, a regular line model analysis is carried out. 1. Longitudinal bending moments, shear forces and torsional moments are estimated on single beam model from STAAD. 2. For the moments and forces due to restraints of warping torsion at the ends, the design shear forces and bending moments in the longitudinal direction due to live load are increased by 10% and the warping moment from analysis has been found to be within this assumed value. 3. The transverse analysis is carried independently for the spine beam, with loads moving on deck slab. The moments in the wings are converted to tensile forces on top and compressive forces on bottom of the spine beam. 334
Figure 6.
Bearing articulation.
3.3 Design philosophy 3.3.1 Superstructure It is observed that a superposition of the line model analysis coupled with a unit slice analysis in the transverse direction, gives, though conservative, a fairly accurate result for all practical purposes. The effect which is not obtained directly from the above analysis is the distortion, which will impart additional torsional and longitudinal warping effects, which is accounted by enhancing the corresponding unsymmetrical (in this case Live load) bending moment by about 10%, though it is seldom more than 5 to 7%, observed from rigorous analysis B.E.F analysis. In the transverse direction, the analysis is carried out with unit slice method, meaning a slice of unit length is modeled as a closed frame with support locations at web, reflecting the shear transfer through webs. In the transverse direction, the wings are designed as flanged beams, with prestressing on top, and the spine deck is designed as an RC slab with axial force in it. In the longitudinal direction, the top slab acts as a continuous member supported on ribs. 3.3.2 Substructure and foundation Substructure consists of system of bearings piercap & pier. The bearing articulation for a three span continuous unit is given Figure 6. The system of bearings is designed in such a way that all the horizontal loads in the longitudinal direction is taken by pin bearings, and lateral loads are shared by guided and pin bearings equally. All the vertical loads are taken by the combination of free and guided bearings. Hence the substructure and foundation design is differentiated as fixed, free and EJ depending upon the type of bearings in the particular location. The stability of the structure is checked for all the service condition. Also enough safety factors are accounted during all the stages. Foundation system consists of piled foundation of 1.20 m diameter anchored into the rock. 4 CONSTRUCTION ASPECTS The construction of this long flyover on an urban stretch requires lot of planning and monitoring. Construction of foundation & substructure is taken up in stages after locally diverting the traffic. Any utility services which are encountered on the median where piers are planned were shifted to designated locations. While this was being carried out, precasting of segments were initiated in the casting yard. The spine segments were match cast with shortline method, and stacked in two tiers. Wings were precast separately in the yard, for the required profile and stacked separately, supported on designated location. Since the deck has a constant bi-camber of 2.5%, geometry control was comparatively easier. The demoulding time, and lifting time were so optimized that there was a continuous supply of segments to the site. At site one over head launching girder, marching from one end, was put to service. The weight of each spine segment was 40MT. The launching girder capacity was 550 MT. Each segment was transported to the site by trucks, lifted and held in position by LG. After lifting of one complete span, all the segments are dry matched and checked for the geometric accuracy. Then the segments are separated and epoxy glue is applied on the face. Temporary prestress is applied in between each segment by means of tension bars. After all the segments are glued, then 335
permanent prestress is applied and released from LG to the bearing. LG is marched forward to the next span. After this the wings are erected by cantilever frames supported on spine. There are longitudinal stitch between adjoining wings and transverse stitch between wing and spine. After these stitches are done, cross prestressing is carried out. Other finishing works were taken up subsequently. The total cost of the project is 25000 million Indian Rupees. 5 ARTISTIC IMPRESSIONS OF THE FLYOVER
6 CONSTRUCTION IN PROGRESS
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Haramain high speed railway line J.M.G. Parejo, M.T. Serrano, M.M. Cañueto, M.B. García & F.J.M. López Acciona Ingeniería, Alcobendas, Madrid, Spain
ABSTRACT: Acciona Ingeniería developed the detailed design of six railway bridges with a total length of 3090 m. located along Al-Haramain High Speed Railway Line (HSR) in the Kingdom of Saudi Arabia, whose owner is the Saudi Railways Organization (SRO). This railway line is a worldwide landmark for HSR design due to the climate challenges and foreseen operation peaks. Special seismic parameters from the Saudi Geological Survey (SGS) were also taken into account in the design of these six bridges. Since there was no local regulation on HSR design, bridge design for Al-Haramain High Speed Rail (HHR) was carried out in accordance with international best practices and standards as Eurocodes and AASHTO. The scope of the line is improving the access to Makkah, the epicenter of Islam. It will be a connecting path among the three Holly Cities of Islam: Madinah, Jeddah and Makkah and the first milestone to develop a transnational HSR on the GCC region. Four of the six viaducts are designed as post-tensioned concrete box girder with a span by span erection process and the deck of the other two bridges consists of two simply supported post-tensioned precast U-beams. 1 INTRODUCTION The current article tries to show the design and construction of six bridges for the Haramain High Speed Railway Line among the cities of Madinah, Jeddah and Makkah. These three cities are located in the west of Saudi Arabia Kingdom. Madinah and Makkah are the most important holly cities of Islam, while Jeddah is considered the entrance gate to Makkah. This railway line is a worldwide landmark for HSR design due to the climate challenges and foreseen operation peaks. The scope of the line is improving the access to Makkah, the epicenter of Islam. Moreover, it will not be only the entrance gate to Makkah for several millions of pilgrims every year from all over the world, but also it will be a connecting path among the three Holly Cities of Islam (Madinah, Jeddah and Makkah) and the first milestone to develop a transnational HSR on the Gulf Cooperation Council (GCC) region. Haramain High Speed Railway Line is a project of extraordinary importance due to not only the amount of investment but also the fact that it is the first project of this nature which is opened to the international engineering market in the Kingdom of Saudi Arabia, to take advantage of the recent experience and innovation in other countries. The whole project involves over 400 km. of high speed railway line. The client of the contract is Saudi Railways Organization (SRO) which is the authority in charge of the line. Works developed by Acciona Ingeniería and the authors of this article can be organized into the two following phases: – The first stage when the Preliminary Design and the Detailed Design of six bridges were developed. This phase lasted almost two years. – A second stage, currently in course, in which a technical assistance to the site engineers is being carried out. This second phase has a huge importance since every change to the original project proposed on site needs the approval of the designers. In this way, the relation between the site engineers and the designers gets high fluent. 337
Table 1. Bridges designed by Acciona Ingeniería in Al-Haramain Railway Line. Bridge
Single or Double Track
Starting Sta.
Ending Sta.
Length (m)
Type
RB 5+078 ST RB 5+078 DT RB 8+674 ST RB 8+674 DT RB 175+302 RB 184+193
Single Double Single Double Double Double
4+835.53 4+837.4 8+168.25 8+171.25 175+131.85 184+017.5
5+325.03 5+326.9 8+885.75 8+888.75 175+469.75 184+355.4
489.5 489.5 717.5 717.5 337.9 337.9
Post-tensioned box girder Post-tensioned box girder Post-tensioned box girder Post-tensioned box girder Pre-casted U-beams Pre-casted U-beams
Calculations developed during the detailed design of these six bridges involved all aspects usually related with this type of structures, as the following ones: fatigue design, dynamic analysis, trackstructure interaction, passengers comfort, seismic analysis, stress analysis of deck, derailment situations, etc. Furthermore, the design of these bridges takes into account some features due to the special conditions of this project: – The detailed design of the bridges belonging to the Al-Haramain High Speed Rail (HSR) was developed taking into account the criteria stated in the report: Terms of Reference (TOR) for the Preliminary and Detailed Design of the Al-Haramain High Speed Rail (HHR) – Phase 1. – Special conditions of temperature in Saudi Arabia with the following values for the bridges: • Temperature Rise (concrete structures): 30◦ C • Temperature Fall (concrete structures): 30◦ C • Temperature Rise (rail): 40◦ C • Temperature Fall (rail): 45◦ C – A specific aspect of this project was the seismic analysis of the bridges. The region where the bridges are being built is dominated by the presence of the Red Sea Fault. The Arabian Plate is a single tectonic plate, which is surrounded by relatively high active tectonic zones. The tectonic activity of the region is ruled by the collision of the Arabian Plate and the Eurasian Plate along the Zagros and Bitlis thrust systems. Calculations were developed in accordance with the National Seismic Standard named Guideline for Earthquake Load Calculation and Structural Design Systems in Saudi Arabia and Eurocode 1998 Design of Structures for Earthquake Resistance. Concerning this topic, it was quite important the information available in the report Seismic Hazard Assessment along Haramain High Speed Rail Project (Makkah – Madinah) developed by the Saudi Geological Survey (SGS). All in all, soil acceleration is determined taking into account the distance from each structure to the Red Sea coast, and two seismic levels, type I (severe) and type II (moderate), shall be considered. A table with the structures designed by Acciona Ingeniería is exposed in Table1. Obviously, the station of Makkah is the beginning of Al-Haramain railway line, so that, the four bridges located at station 5+078 and 8+674 are located inside Makkah city. 2 BRIDGES DESIGNED BY ACCIONA INGENIERÍA 2.1 Rail Bridges 5+078 Single Track and Double Track Railway Bridge 5+078 is one of the bridges designed for the Segment A (Makkah – Jeddah) of Al-Haramain High Speed Rail Link between Makkah and Madinah. Bridge 5+078 consists of two parallel structures carrying double and single track respectively and spaced 1.8 m transversely. The span arrangement is the same for both structures: 15 spans 29 + 2 × 36.5 + 27 + 2 × 34 + 2 × 36.5 + 29 + 23 + 29 + 3 × 36.5 + 29 m long, with a total length of 489.5 m from station 4+835.530 to station 5+325.03 (single track axis). This bridge allows the railway link to cross over several existing roads together with Makkah Fourth Ring Road, which passes under fifth and sixth spans. 338
Figure 1.
Elevation of rail bridge 5+078 single track.
Due to its length, which is greater than 450 m, and in order to comply with the maximum expansion length specifications the deck is divided into three continuous sections separated by means of structural expansion joints. The first continuous section, 299 m long, goes from abutment A1 to pier P9. Between pier P9 and P10 a simply supported span is provided, and the final section, from pier P10 to abutment A2 is 167.5 m long. This arrangement allows to withstand horizontal actions providing fixed points for the longest deck sections at the abutments A1 and A2 respectively. The simply supported span is fixed at pier P9. Therefore, two rail expansion devices have to be located at pier P9 and P10, since the expansion length of both decks is longer than 90 m. This document covers single track bridge design. The deck for single track consists of a posttensioned box girder for the longest two continuous sections into which the bridge is divided and a reinforced box girder for the central single span section. This deck platform is only 7.50 m wide to carry the single ballasted track and two side walkways. The width of the deck soffit is 4.70 m and, therefore, each side cantilever slab is 1.40 m long. The deck has constant depth of 2.70 m along all the bridge length. At both sides wall concrete barriers 1.70 m high are provided, together with space for utilities under the walkways. The thickness of the upper slab is 0.375 m in the central part of the box girder and varies down to 0.30 m at the end of the cantilever slab. A transverse slope of 2% towards both sides is provided at the slab top surface under the ballast bed to ease drainage. In the other hand, deck for double track consists of a post-tensioned box girder for the longest two continuous sections which the bridge is divided into and a reinforced box girder for the central single span section. This deck platform is 12.50 m wide to carry a double ballasted track and two side walkways. The width of the deck soffit is 6.10 m and, therefore, each side cantilever slab is 3.20 m long. The deck has constant depth of 2.70 m along all the bridge length. At both sides wall concrete barriers 1.70 m high are provided, together with space for utilities under the walkways. The thickness of the upper slab is 0.325 m in the central part of the box girder and varies down to 0.30 m at the end of the cantilever slab as shown in the following figure. A transverse slope of 2% towards both sides is provided at the slab top surface under the ballast bed to ease drainage. 339
Figure 2. Typical cross section of rail bridge 5+078 double track.
Figure 3.
Couplers section of rail bridges 5+078 single and double track.
Both decks are designed with C50/60 concrete. Reinforcing steel has been M31 grade 60 (AASHTO), with yield stress of 415 MPa. The prestressing steel was A416/80 (ASTM) – low relaxation with cable type T15 (section = 140 mm2 ). The definition of prestressing is as follows: Rail Bridge 5+078 Single Track: Prestressing of deck consists of 12 cables (six cables per web) with 31 strands each cable. These 12 cables are distributed into two families with 6 cables each family. At every couplers section, one family of cables has continuity while the other family ends at this section and it is connected with the family of the next phase by means of couplers. In this way, only half the cables are coupled at each coupler section, following the criteria of Eurocodes. In addition to this prestressing, 4 ducts are left at both, top and bottom slab, with enough section for 14 strands each duct, in order to have available space for a future additional prestressing. Rail Bridge 5+078 Double Track: Prestressing of deck consists of 16 cables (eight cables per web) with 31 strands each cable. These 16 cables are distributed into two families with 8 cables each family. At every couplers section, one family of cables has continuity while the other family ends at this section and it is connected with the family of the next phase by means of couplers. In the same way of single track bridge, only half the cables are coupled at each coupler section, following the criteria of Eurocodes. Moreover, there is an extra prestressing of two cables with 12 strands each cable at the top slab at the bearings surroundings and at the bottom slab at the middle of span surroundings. In addition to this prestressing, 7 ducts are left at both, top and bottom slab, with enough section for 12 strands each duct, in order to have available space for a future additional prestressing. Due to the high number of passengers that will use this new railway line, the Saudi Administration stated the necessity of avoiding any tension stress at any fiber of the deck under the characteristic 340
Figure 4.
Elevation and side view of piers belonging to rail bridges 5+078 single and double track.
Figure 5.
Detail of pier caps of rail bridges 5+078 single and double track.
combination of actions. This circumstance implied the design of a rigid deck with an important ratio Depth/Span = 2.7/36.5 = 1/13.5 and a high quantity of prestressing steel. The intermediate concrete piers have maximum heights of about 21.0 m from ground level. Their section has rectangular shape with rounded corners, being 3.5 m long (bridge transverse direction) for single track and 5.0 m long for double track and 2.5 m wide (bridge longitudinal direction) for both structures. At the top of the pier a longer monolithic pier cap is provided in order to place the bearings. At all the piers, between both bearings, a depressed area is left to place the transverse shear key from the deck. Both abutments A1 and A2 are similar. They consist of a front wall with a height of about 11 m, for A1, and 7.5 m, for A2, from top of foundation level to bearings level. The abutment is completed with two wing walls on both sides to contain the embankment fills. An inspection gallery is provided 341
Figure 6.
Detail of fixed point at abutment 1 of rail bridge 5+078 double track.
Figure 7.
Construction of piers at both rail bridge 8+674, single and double track.
at the top of the backfill and monolithic with the abutment wall. From this gallery and trough an opening at the end of the deck it is possible to access to the interior of the deck. 2.2 Rail bridges 8+674 single track and double track Both Rail Bridges 8+674 Single and Double track have the same structural scheme as the two bridges of 5 + 078. In this case the span arrangement is as follows: 21 spans 27 + 33 + 7 × 36.5 + 29 + 23 + 29 + 8 × 36.5 + 29 m long, with a total length of 717.5 m. 2.3 Rail bridges 175+302 and 184+193 Both Railway Bridges 175+302 and 184+193 have been designed for the Segment C (Makkah – Jeddah) of Al-Haramain High Speed Rail Link. Both bridges are designed with the following arrangement: 13 spans 25.40 + 11 × 26.10 + 25.40 m long, with a total length of 337.90 m. 342
Figure 8.
Falsework for the construction of rail bridge 8+674 double track.
Figure 9. Typical cross section of rail bridges 175+302 and 184+193.
The deck of both double track bridges consists of two simply supported precast U-beams, designed with a distance of 5.0 m between axis, and an upper slab with a thickness of 0.425 m in the central part of the section that varies down to 0.30 m at the end of the cantilever slab. Each cantilever slab is 2.11 m long, and the beams have a constant depth of 2.2 m along the span. The deck platform is 12.50 m wide to carry a double ballasted track and two side walkways, and at both sides wall concrete barriers 1.70 m high are provided, together with space for utilities under the walkways. A transverse slope of 2% towards both sides is provided at the slab top surface under the ballast bed to ease drainage. A special feature of these bridges consists of the design of several piers as portal frames. These frame structures are formed by a top concrete transverse girder of 26.0 m long between the axis of the columns (bridge transverse direction), 2.7 m wide and 2.8 m thick. This girder rests on two columns designed with a rectangular shape of 2.0 m long (bridge transverse direction) and 3.20 m wide (bridge longitudinal direction), and with rounded corners of 0.75 m of radius. The columns have maximum heights of about 8.90 m from upper face of foundation to bottom face of transverse girder. 343
Figure 10. View of rail bridge 175+302 in construction.
Figure 11. View of rail bridge 184+193 already built.
At the top of the transverse girder, the same monolithic concrete area designed in the rest of the piers is placed between the bearings, in order to set the lateral bearings of the beams. This specific area is 2.40 m wide, 2.70 m long and 0.85 m thick. ACKNOWLEDGEMENTS Special acknowledgements to engineers belonging to Saudi Railway Organization and to Al-RajhiAlliance Contractor and very special acknowledgement to Technical Director of the Project Management Office Fadi Daher and to the geotechnical engineer Abdulwahab Al Shawaf. 344
Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
Design and construction of viaduct to Mumbai International Airport P.G. Venkatram & K. Ganesh L&T Infrastructure Engineering Limited, India
ABSTRACT: The 4.5 km long elevated corridor, connecting the Western Express Highway and the Mumbai International Airport, India, caters for the main connectivity and also for different movements from and to the various main roads and multi level car parking. The width of the 6 lane deck with planter box is maximum 27.60 m, with standard spans of 35 m. There are special spans such as varying width and curved spans to cater for mergers and geometrical requirements. The deck is spine beam segmental construction, which is longitudinally and transversely prestressed. Deck is supported on POT cum PTFE bearings. Expansion joints are provided at every third span, i.e at every 105 m, to ensure smooth riding surface. The substructure is a single pier, supported by pile foundation.
1 INTRODUCTION As a part of modernization of Mumbai International Airport and to upgrade the capacity to 50 million passengers per year, Government of India had decided to build a new terminal T2. As a part of this modernization, the proposal of a forecourt, along with the connectivity to Western Express Highway and other main roads and to multi level car parking was conceived. The network of at-grade roads, ramps, underpasses and elevated corridor took two years of planning and took five years to build. With the construction of this elevated corridor, the travel time from the Western Expressway Highway would be reduced and facilitated a congestion free movement between various facilities such as departure, arrival, multi level car parking etc. The project was implemented in Public-Private Partnership (PPP) and the developer is M/s MIAL (Mumbai International Airports Limited), which has M/s GVK as the majority share holder.
Figure 1.
Layout of the project.
345
The EPC contractor of the elevated corridor is L&T Construction, and the detailed designers are L&T Infrastructure Engineering Limited. 2 GENERAL ARRANGEMENT 2.1 Development of concepts The following criteria were governing the selection of structure type. a) Construction friendly, including faster output. The proposed elevated corridor passed through one of the two main approaches to the airport. The 1.20 km stretch had multi storied hotels and food supply chains to aircrafts, on both sides of the road, for which the access could not be blocked. Since there were no service roads and alternate diversion road was not permitted, the wide deck construction needed to be taken up with traffic plying below. The construction period of the part of this elevated highway was initially set as 3 years in line with the main terminal construction, including time to plan the development works, movement of traffic near the terminal buildings etc. Transportation of wide segments was not possible in the project road. So the efforts were channelized for making the standard system as far as possible, so that mass production is achieved without any significant variation either in dimension, or in reinforcement detailing as far as possible. The in-situ works were minimised. The production was continuous through all the seasons. More over since the width of 27.60 m was beyond the limit of casting, handling and transporting, the system should be segmental in the transverse direction also. Thus a segmental spine and wings precast superstructure was conceptualized. This would help in minimising the site works and would help in handling the segments safely with the traffic plying below. In order to keep the at-grade roads operational, it was decided to adopt a system of single pier with widened pier cap. b) Aesthetics As an urban structure, which is going to act as a gateway to the one of world’s best airports, developers were very keen to make the structure elegant. The structure need to be unique, iconic to the city and shall necessarily blend with the urban backdrop. Any structure which is sleek and has less concrete in it, would be more appreciated than a massive one. So an aesthetically treated central spine beam with sleek wings on both sides was chosen to satisfy all the above requirements and design provisions. 2.2 Layout and cross section The total layout is broadly divided into three zones as given in the Figure 1 above, from the point of construction methodology adopted. First zone consisted of the entry from the Western Express Highway which had an underpass taking off from the median of the highway, and the approach upto the elevated corrdior. The second zone is the six lane straight elevated corridor for a length of about 1.80 km, upto the point of bifurcation to third zone. Here the typical span is 35.0 m. This span was kept constant throughout the initial straight length (zone-1), and minor adjustments were done for at-grade junctions. The third zone of elevated corridor is the part of network leading towards to the forecourt including one ramp leading to an at-grade junction, one up ramp from at-grade junction, one down ramp to arrival area, connecting arms to multi level car parking, one up-ramp from arrival area etc. The cumulative length of this zone is about 3.0 km which has unique spans, varying widths, challenging geometry in the forms of acute curves, mergers and demergers and differential levels. In order to ensure proper drainage, a gentle longitudinal slope of 0.3% is also provided. In order to reduce the number of expansion joints and to have a better driving comfort, tied decks are provided for two spans, and EJ at every third pier. The cross section of the flyover is as given in the Figure 2. 346
Figure 2. Typical cross section.
The cross section of second zone is three separate units consisting of central spine of about 9.0 m, and two equal wings of 9.3 m each. The ribs are at a spacing of 3.0 m, which was found to be ideal both from casting and design point of view. The ribs were stitched to the central spine by transverse stressing. While match cast technique was adopted for spine segments, in-situ stitch was adopted for connection between the wings and between the wing and spine. In the third zone, since the geometry varied completely, and standardization was impossible, the spine was cast-insitu and connecting wings were precast. However the form and shape of the superstructure and substructure of the whole bridge was kept same. With this general over view of the whole project, further discussions are limited to the second zone only, since the design and construction of this zone was both innovative and challenging. Bidirectional camber of 2.0% is provided in the deck for the purpose of drainage. The overall width of the superstructure is 27.60 m, which included six lanes of 22.0 m, central median of 1.0 m, planter box along with service walkway of 1.50 m on each side. The system of substructure consists of a pair of POT & POT-Cum-PTFE bearings for each superstructure, supported on a piercap, flared from a single pier. Foundation system is a group of 1.20 m dia pile, length depending on the type of subsoil at each location. 3 DESIGN ASPECTS 3.1 General The design is carried out as per Indian Roads Congress (IRC) codes of practice. The following loads are considered for the design as per IRC-6. 3.1.1 Live load Live load types of Class 70R, with a single train totaling to 100 MT occupying two lanes and Class A, totaling to 55.40 MT occupying single lane, individually or in combination to give the maximum effects on various elements are considered. 347
3.1.2 Materials Grade of concrete for superstructure and crash barrier/median adopted is M60 and grade of steel is Fe 500. Prestressing steel is of nominal diameter 15.24 mm, conforming to IS: 14268 class II, with yield strength 1860 MPa 3.2 Method of analysis and design The following procedure has been followed for the design of spine beam. Since the deck is simply supported for all the gravity based loads, a regular line model analysis is carried out. 1. Longitudinal bending moments, shear forces and torsional moments are estimated on single beam model from STAAD. 2. For the moments and forces due to restraints of warping torsion at the ends, the design shear forces and bending moments in the longitudinal direction due to live load are increased by 10% and the warping moment from analysis has been found to be within this assumed value. 3. The transverse analysis is carried independently for the spine beam, with loads moving on deck slab. The moments in the wings are converted to tensile forces on top and compressive forces on bottom of the spine beam. 3.3 Design philosophy 3.3.1 Superstructure It is observed that a superposition of the line model analysis coupled with a unit slice analysis in the transverse direction, gives, though conservative, a fairly accurate result for all practical purposes. The effect which is not obtained directly from the above analysis is the distortion, which will impart additional torsional and longitudinal warping effects, which is accounted by enhancing the corresponding unsymmetrical (in this case Live load) bending moment by about 10%, though it is seldom more than 5 to 7 %, observed from rigorous analysis B.E.F analysis. In the transverse direction, the analysis is carried out with unit slice method, meaning a slice of unit length is modeled as a closed frame with support locations at web, reflecting the shear transfer through webs. In the transverse direction, the wings are designed as flanged beams, with prestressing on top, and the spine deck is designed as an RC slab with axial force in it. In the longitudinal direction, the top slab acts as a continuous member supported on ribs. 3.3.2 Substructure and foundation Substructure consists of system of bearings piercap & pier. The bearing articulation for a three span continuous unit is Figure 3. The system of bearings is designed in such a way that all the horizontal loads in the longitudinal direction is taken by pin bearings, and lateral loads are shared by guided and pin bearings equally. All the vertical loads are taken by the combination of free and guided bearings. Hence the substructure and foundation design is differentiated as fixed, free and EJ depending upon the type of bearings in the particular location. The stability of the structure is checked for all the service condition. Also enough safety factors are accounted during all the stages. Foundation system consists of piled foundation of 1.20 m diameter anchored into the rock. The deck continuity is achieved by tied deck arrangement, which is given in Figure 4. 4 CONSTRUCTION ASPECTS The construction of this long flyover on an urban stretch requires lot of planning and monitoring. Construction of foundation & substructure is taken up in stages after locally diverting the traffic. Any utility services which are encountered on the median where piers are planned were shifted to 348
Figure 3.
Bearing layout.
Figure 4. Tied deck arrangement.
designated locations. While this was being carried out, precasting of segments were initiated in the casting yard. The spine segments were match cast with shortline method, and stacked in two tiers. Wings were precast separately in the yard, for the required profile and stacked separately, supported on designated location. Since the deck has a constant bi-camber of 2.5%, geometry control was comparatively easier. The demoulding time, and lifting time were so optimized that there was a continuous supply of segments to the site. At site one over head launching girder, marching from one end, was put to service. The weight of each spine segment was 40MT. The launching girder capacity was 550 MT. Each segment was transported to the site by trucks, lifted and held in position by LG. After lifting of one complete span, all the segments are dry matched and checked for the geometric accuracy. Then the segments are separated and epoxy glue is applied on the face. Temporary prestress is applied in between each segment by means of tension bars. After all the segments are glued, then permanent prestress is applied and released from LG to the bearing. LG is marched forward to the next span. After this the wings are erected by cantilever frames supported on spine. There are longitudinal stitch between adjoining wings and transverse stitch between wing and spine. After these stitches are done, cross prestressing is carried out. Other finishing works were taken up subsequently. The total cost of the project is 5000 million Indian Rupees 349
5 SOME COMPLETED PHOTOS
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Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
Meriç Bridge: Construction and quality control S. Uluöz, S. Düzbasan & T. Uluöz ˙ saat Tic. Ltd. S¸ ti (Ilgaz Construction and Trade Co.), Istanbul, Turkey Ilgaz In¸
E. Yakıt ˙ Ihr. ˙ Ltd. S¸ ti., Ankara, Turkey Railone Ilgaz Demiryolu Sis. Ürt. Ith.
U. Akyazı General Directorate of Highways, Edirne, Turkey
ABSTRACT: The construction of TEM Edirne-Karaa˘gaç Connection Road Project has been commenced in 14 July 2013 in order to activate Pazarkule Border Gate that opens Turkey to Europe and protect historical bridges. Meriç Bridge is the longest bridge of the 10 bridges within the project with 53 spans and 1943 m. length. For the construction of the bridge; the production of 26.142 m. long 1381 ea bored piles, 106 ea foundations, 216 ea elevations, 3.996 m. slab concrete, different type and length of 1232 ea pre-stressed beams have been planned to construct. Within the scope of the project; C45/55 and C25/30 class 115.380 m3 concrete, 17.336 tons of structural steel and 1.173 tons of pre-stressing wire will be used. In order to be able to complete the project within the scheduled date, concrete production and pre-stressed beam installations in the construction site continue perpetually 24 hours a day. Ilgaz Construction R&D technical staff has been conducting necessary controls regarding concrete quality and research and development studies in order to solve quality issues in concrete production as well. To illustrate, it has been observed that the compressive strengths of concretes produced in 20-28.08.2014 decreased approximately 20%. After conducting the R&D studies, it is confirmed that the bacteria is produced in lignin sulphonate and poly carboxyl ether admixtures which are used in concrete production, thus the admixtures could not affect on cement particles effectively and consequently caused increase in water/cement ratio in concrete. To solve the problem, the admixtures in the tank have been sent back to the producer company. It is confirmed that the admixture tank inner wall is rough due corrosion, thus the bacteria which could remain there would transmit to the cement admixtures, by consideration of these admixture tank is changed.
1 INTRODUCTION Meriç, Tunca and Arda rivers which passes through Edirne, the city of water and bridges, overflow in almost every winter. For such occasions, it is not possible to use the historical bridges and because of this approaching difficulties to some residential areas, loss of lives and properties occurs. The construction of TEM-Edirne Karaa˘gaç Connection Road Project has been initialized by 14 July 2013 in order to prevent the problems occurred due overflow of Meriç, Tunca and Arda rivers, make Pazarkule Border Gate more active which is in between Turkey and Greece, protect the historical bridges on rivers, increase the commercial and touristic activities in Edirne and surroundings, affect positively the border trade with Greece. When the project will be completed, Pazarkule Border Gate’s Greece side E85 main road and Edirne Beltway which comprises some part of TEM Highway will be connected. 351
˘ 2 TEM EDIRNE – KARAAGAÇ CONNECTION ROAD PROJECT Project is realized in order to make connection with Pazarkule Border’s Gate Greece side E85 main road and Edirne Beltway which comprises some part of TEM Highway. TEM-Edirne Karaa˘gaç Connection Road Project, which will connect Pazarkule Border Gate to Edirne Beltway, is a 5.6 km long high standard road project. Within scope of the project; 123.534,53 m3 concrete, 18.324 tons of structural steel, 1.173 tons of pre-stressing wire and 841 tons of steel profiles will be used. 2.1 Meriç bridge In the project of TEM Edirne-Karaa˘gaç Connection Road, there will present 10 bridges in which the Meriç Bridge with its 1943 m. length and 2*2 highway standards construction and 53 spans will be the longest amongst. The construction of the Meriç Bridge is ongoing which can be seen in Figures 1 and 2. 2.2 Benefits of the project Meriç Bridge Construction is the most important part of the project when it will be completed, will provide below mentioned benefits. a. Transportation problems due overflows in Meriç, Tunca and Arda rivers in settlements around Edirne city will be solved. b. Pazarkule Border Gate will be more active. c. Heavy vehicle traffic, which is not permitted in between Karaa˘gaç-Edirne to protect historical bridges, will be provided. d. Border trade with Greece will be utilized. e. City traffic will be relieved due the vehicles which enter from the border gate will go through Edirne Beltway Highway. f. Export and import capabilities of the industrial plants in Edirne and surroundings will get easier. g. Commercial and touristic activities in Edirne center will increase. h. The transportation in between the countries as Greece and Russia, Ukraine which will be passing through Turkey-Istanbul will provide important advantages to our country. In order to finish the construction of TEM-Edirne – Karaa˘gaç Connection Road within the scheduled period, Ilgaz Construction Co.’s technical staff, foreman and qualified workers work on shifts with 24 hours a day continuously. 2.3 The production of the concrete used in the project With the purpose of producing 123.534, 53 m3 concrete required within the scope of TEM-Edirne – Karaa˘gaç Connection Road Project, 2 concrete plants with 100 and 120 m3 /hour capacities have been established in the Construction Site by Ilgaz Construction.
Figure 1, 2.
Meriç Bridge.
352
2.4 Concrete components Aggregate; in the concrete production, aggregate at 4 different grades in conformity with TS 706 EN 12620 standard is used. Cement; in the concrete production, CEM I PC 42,5 R class cement in conformity with TS EN 197-1 standard is used. Since active silica is detected in the river sand used in the concrete production, it is used at most 15% ratio in the concrete production and the amounts of Na2 O, K2 O and SO3 in the cement are tested in laboratory tests. Concrete additives; in the production of pre-stressed beams within the scope of the project, poly carboxyl ether concrete admixture is used in C 45/55 class concrete and in the other RC structures, concrete admixture including lignin sulphonate is used in the C 25/30 class concrete production. Concrete mix and curing water; the water available in the plant in conformity with the TS EN 1008 standard criteria is used as concrete mix and curing water. 2.5 Designation of concrete designs Before starting the designation of concrete designs, the meteorological reports of Edirne and nearby for the last 5 years have been examined and the precautions to be taken against the unfavorable weather conditions that may be faced during the concrete productions have been determined. The samples taken from the concrete produced in the laboratory environment are kept in different ambient temperatures and their compressive strengths have been determined and the impacts of unfavorable weather conditions on the concrete quality are identified. The concrete designs used within the scope of the project are given in Table 1. 2.6 RC structures in Meriç Bridge In the construction of Meriç Bridge, of 112.313,32 m3 concrete; 24.924,32 m3 has been used in the pre-stressed beam construction and the rest 88.389,4 m3 has been used in the structures specified below. Bored pile; in the construction of Meriç Bridge, in total 30.012 m. length 1.3981 ea bored piles will be constructed by using 29.650 m3 C 25/30 class concrete and 4.100 tons of structural steel. In order to be able to complete the project in the scheduled period, 4 bored pile machines have been operated continuously 24 hours a day. Foundation; in the manufacture of 106 ea foundations in the construction of Meriç Bridge, 19.604 m3 C 25/30 class concrete and 2.165 tons of structural steel will be used. Elevation; in the manufacture of 216 ea elevations in the construction of Meriç Bridge, 9.963 m3 C25/30 class concrete and 1.371 tons of structural steel will be used. Pier cap; in the manufacture of 104 ea pier caps in the construction of Meriç Bridge, 11.225 m3 C 25/30 class concrete and 1.636 tons of structural steel will be used. Slab Concrete; In the manufacture of 106 ea slabs for 53 spans in the construction of Meriç Bridge, 13.430 m3 C 25/30 class concrete and 1.782 tons of structural steel will be used. Table 1. Concrete designs used within the scope of the project. 1 m3 concrete production Concrete components
C 25/30
C 45/55
River sand (0–3) Crushed limestone (0–5) Crushed limestone (5–12) Crushed limestone (12–22) Cement (CEM I PC 42.5 R ) Additive (Lignin sulphonate) Additive (Poly carboxyl ether) Mixing water
296 658 791 263 290 %1.3 – 165
285 752 752 272 450 – %1.0 141
353
Table 2. The current situation of Meriç Bridge construction as of 13.02.2015. Completed section
Figure 3, 4, 5.
Name of the structure
To be constructed (ea)
(ea)
(%)
Bored pile Foundation Elevation Pier cap Slab concrete
1381 106 216 104 106
1344 69 131 60 17
97,3 65,1 60,7 57,7 16,0
Studies for the installation of pre-stressed beams in Meriç Bridge.
The information about the completed section of Meriç Bridge construction as of 13.02.2015 is given in Table 2. 2.7 Production of pre-stressed beams In the manufacture of 1232 ea I 150 and I 90 types pre-stressed beams to be used in the construction of Meriç Bridge, 24.924,32 m3 C 45/55 class concrete, 5.898,91 tons of structural steel and 1.173 tons of pre-stressed rope will be used. The pre-stressed beams to be used within the scope of the project are manufactured in special molds which may be opened and closed by a hydraulic system. As of 13.02.2015 for the construction of Meriç Bridge, the production of 346 ea (28,1 %) pre-stressed beams have been completed. The production of pre-stressed beam is performed in 3 stages as follows: 1st Stage; Stress operation is applied on the pre-stressing ropes which are put in between the reinforcement located in the beam mold in 3 stages (110, 210 and 278 bars). 2nd Stage; Concrete is filled in the beam mold gradually and then vibration is applied so that the production is completed. 3rd Stage; After 2 hours of resting the concrete in the mold, 12 hours of steam curing is applied and when the compressive strengths of 15 cm cube samples which are treated with the same conditions of steam curing with beam concrete reaches to 41,3 N/mm2 pre-stressing wires are cut and the beams which are removed from the mold are carried to the stockyard (Uluöz et al., 2014). 2.8 Installation of pre-stressed beams Even though launching girders used in the beam installation are available in the equipment pool of Ilgaz Construction, since the location of the place that the beam installation will be performed is taken into consideration, telescoping-boom mobile crane with 200 tons capacity is used in the prestressed beam installation. Images of transportation and installation of pre-stressed beam are given in Figures 3, 4 and 5. 354
Table 3. Compressive strengths of the RC structures in Meriç Bridge.
Name of the structure Bored pile Foundation Elevation Pier cap Slab concrete
Structure no.
Date of production
Concrete class
A2-15-L. A2-16-L A2-4-L. PY-47-R. PY-49-L. PY-49-R. PY-47-L. PY-48-R. A1 Abutment PY-11-L. PY-12-R. PY-12-L. PY-13–15-R. PY-7–10-L PY-13–15-L.
17.11.14 18.11.14 19.11.14 27.11.14 12.11.14 15.11.14 02.12.14 06.12.14 27.12.14 24.11.14 30.11.14 29.11.14 28.11.14 23.12.14 28.12.14
C 25/30 C 25/30 C 25/30 C 25/30 C 25/30
Compressive strength (N/mm2 ) 7 days
28 days
28,9 29,9 29,5 30,9 31,9 30,7 29,1 33,1 31,4 30,3 31,4 29,1 32,1 33,4 32,3
41,9 40,5 41,6 44,6 41,5 40,1 40,9 40,8 40,1 38,2 39,9 43,2 40,6 39,4 39,6
L = Left, R = Right
3 CONCRETE QUALITY AND RESEARCH & DEVELOPMENT STUDIES A laboratory has been established in the construction site in order to determine the quality of concrete production to be performed within the scope of the project and carry out research and development studies regarding the quality problems to be faced. The technical staff and research technicians assigned in the laboratory are working continuously 24 hours a day in shifts in parallel to the construction site activities. 3.1 Concrete quality in the construction of Meriç Bridge 3.1.1 Concrete quality of the reinforced structures in Meriç Bridge In order to determine the concrete quality in the RC structures manufactured within the scope of the project such as bored pile, foundation, elevation, pier cap etc., 6 ea 15 cm cube samples are taken for each. After the concrete samples are kept in the same environment with the concrete production, they are preserved in the curing pool in the laboratory and their compressive strengths are determined in 7th and 28th days. In the period from the beginning of the project in 14 July 2013 to 13.02.2015, 12.800 ea concrete samples have been taken and their compressive strengths have been determined. To illustrate, the compressive strength results of some of these concrete samples taken within this scope are given in Table 3. 3.1.2 Concrete quality in the production of pre-stressed beam Together with the beams, steam curing is applied on 4 ea of 10 ea 15 cm cube samples taken from the fresh concrete used in the production of pre-stressed beam and the compressive strengths of 3 ea samples have been determined in both 7th and 28th days. To illustrate, the compressive strengths of concrete samples taken from the pre-stressed beams produced in order to be used in Meriç Bridge are given in Table 4. 3.2 Research & Development studies regarding concrete quality As well as conducting routine activities performed with the purpose of controlling the quality of concrete, the technical staffs assigned in the construction site laboratory also carry out research 355
Table 4. Compressive strength results in the production of pre-stressed beam. Compressive strength (N/mm2 ) Rope cut Beam no.
Date of production
(Hours)
(N/mm2 )
7 days
28 days
MK 0291 MK 0319 MK 0349
06.11.15 27.11.15 09.12.15
11 18 12
48,6 46,6 52,6
63,5 65,0 65,4
79,4 80,4 78,7
and development studies in order to determine reasons of the problems in the concrete quality and find solutions. To illustrate, some of these studies performed within this scope are given below. 3.2.1 Bacterial growth in the concrete admixtures Inside the concrete produced in between 20–28.08.2014, in C25/30 class concretes 15% decrease and in C45/55 class concretes approximately 20% decrease have been determined in terms of compressive strength. At the end of the research and development studies carried out with the purpose of determining the reasons of the quality problem; it has been specified that the concrete admixture performance has decreased because of the bacteria growth in the concrete admixtures including lignin sulphonate and poly carboxyl ether therefore the water/cement ratio in the fresh concrete has increased. 7.600 kgs of concrete admixture inside the admixture tanks have been sent back to the manufacturer company and the concrete admixtures brought inside of 1 tons of plastic containers are used during the research and development studies. In the examination of the inner surface of the admixture tank used, it has been determined that due to metal corrosion, a very rough surface has occurred. A bacteria which is probably available in the admixtures including poly carboxyl ether feeds with the hydrocarbons in the admixture and becomes 2,90*1014 ea in number, at the end of 48 hours. Even though inside the admixture tank is cleaned chemically, since the bacteria may stick on the inner surface of the tank and fertilize the new admixtures for precaution against the risk of bacterial growth again, the admixture tank has been sent back and the new admixture tank have been used instead. 3.2.2 Precautions taken with the purpose of accelerating the production of pre-stressed beam Research and development studies have been carried out with the purpose of completing the beam production period in a shorter time so that the beams would be ready for installation soon. Within this scope, instead of concrete admixture including poly carboxyl ether, a more concentrated concrete chemical including the same chemical group has been used, the operation period of the mixer in the concrete plant has been increases at 50% in order to obtain the consistency of the concrete inside the mixer and with the purpose of discharging the prepared concrete inside the truck mixer easily, a surface vibrator has been mounted on the concrete plant discharge bucket. At the end of this study, as the water/cement ratio in the fresh concrete has decreased from 0,34 to 0,30 level, the unit bulk density of concrete has increased from 2.450 kg/m3 to 2.590– 2.650 kg/m3 . Since the grooves of the truck mixers are not designed according to the concrete with 2.590–2.650 kg/m3 unit bulk density, the grooves of the truck mixers have been broken and deformed. As a precaution, the current sheet grooves was cancelled and the custom manufactured grooves of 3 mm sheet grooves have been used instead. REFERENCE Uluöz, S., Düzbasan, S., Yakıt, E., Uluöz, T. 2014. Effect of Crystalline Calcite Usage in Pretensionede Precast Beam Production Over Concrete Durability, Istanbul Bridge Conference 2014, 11–13 August 2014, Istanbul – Turkey.
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Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
Design and construction of elevated viaduct at Nashik, India K. Ganesh & P. Murali L&T Infrastructure Engineering Limited, India
ABSTRACT: The 5.10 km long elevated corridor, which is part of six laning of NH-3, in the state of Maharashtra, India, has standard spans of 30 m with an overall width of 19.20 m. There are few special spans at road crossings and ramp mergers. The deck is an externally strutted segmental type, supported on spherical bearings. The segment assembly and launching was carried out by an over head launching girder. The deck is both longitudinally and transversely prestressed. Expansion joints are provided at every second span, i.e. at every 60 m, to ensure smooth riding surface. The substructure is a single pier, supported by open/pile foundation depending on the type of subsoil stratum available.
1 INTRODUCTION The Government of India has initiated the construction of six laning of National Highway No. 3 (NH-3) connecting the tourist city of Agra to the commercial capital of Mumbai, refer Figure 1. The total length of this highway is 1161 km. An important part of this highway between the places Pimpalgaon and Gonde, which are 60 kms apart, passes through the industrial city of Nashik. This elevated viaduct provides for a signal free passage through the city. The stretch of the road through the city of Nashik, was getting more and more congested due to the increase traffic and pressure of spreading the residential areas around the highway. Thus the requirement of 4 lane elevated corridor, through the Nashik city was developed and implemented. The project is implemented in Public-Private Partnership (PPP) in “Build Operate & Transfer (BOT) mode. The concessionaire of the project is PNG Tollway Limited, and the “Design & Build” contractor of the elevated corridor is L&T Construction, and detailed designers for the EPC contractor is L&T Infrastructure Engineering Limited.
Figure 1.
Location plan of highway.
357
2 GENERAL ARRANGEMENT 2.1 Development of concepts The following criteria were governing the selection of structure type. a) Contract stipulation of segmental box. The tender condition of adopting a segmental box girder, in a way helped in closing the otherwise wide construction and structural options. b) Construction friendly, including faster output. The construction period of the total highway including complete engineering and third party approvals (consisting of 5.1 km of elevated corridor, 5 flyovers, 3 VUPs, 2 Major bridges) was 36 months. So the efforts were channelized for making the standard system as far as possible, so that mass production could be achieved without any significant variation either in dimension, or in reinforcement detailing as far as possible. In-situ works were minimized and the precast structural components shall be continuous through all the seasons. There were 5 spans longer than 30 m to accommodate at-grade road crossings and almost 6 smaller spans to match the spans across the River Godhavari flowing through the city. Thus a full width precast superstructure carrying both carriage-way on a single deck was conceptualized. This would help in minimising the site works. In order to maximize the various construction resources, the number of shifting operations of equipments, especially the boring rigs, were also minimized. Since the elevated corridor passed through the busy urban area, the operational area was as minimum as possible, so that the existing traffic moves with least disturbance. Considering the above two aspects, a system consisting of a single pier with a widened piercap was conceptualized. c) Freedom for future development. In addition to the advantages given above, a single pier system gives an additional advantage of un-restricted at-grade development on both sides of the median compared to the double pier system, which was also at one point of time contemplated from the advantageous point of handling weights and enabling works. d) Aesthetics As an urban structure, that also viewed as an iconic structure in the city, shall necessarily blend with the urban backdrop. Any structure which is sleek and has less concrete in it, would be more appreciated than a massive one. So an externally strutted box with sleek features was finally chosen to satisfy the mandatory contract provisions, and all other construction and design requirements. 2.2 Layout and cross section The layout of the viaduct is as given in the Figure 2. The typical span is 30 m. However there are other odd spans such as 40 m, 21.75 m, 23.75 m, 28 m which were introduced to cater for the atgrade constraints. The geometric design of the viaduct is done for a design speed of 100 kmph. The minimum radius of curvature of the elevated corridor is 500 m. In order to ensure proper drainage, a gentle longitudinal slope of 0.3% was also provided. In order to reduce the number of expansion joints and to have a better driving comfort, tied decks were provided for two spans, and EJ at every third pier. The cross section of the flyover is as given in the Figure 3. The detailed dimension of the box girder is as given in the Figure 4. The cross section is an externally strutted box girder, with strut spacing of 3.0 m. The cross camber for the deck is given in one direction, so that the precasting of the whole box is easy and adjustment for super elevation can be achieved by rotating the whole section. This helps in simplified casting and flexibility in achieving the transition and cross slope. In general the cross slope is 2.5% and the maximum cross slope in the stretch is 5%, where the radius of curvature 358
Figure 2.
Part plan & elevation of the bridge.
Figure 3. Typical cross section of the flyover.
Figure 4.
Dimension details of box girder.
is least. The overall width of the superstructure is 19.20 m, which includes two carriageways of 8.750 m each, central median of 1.20 m and end crash barrier of 0.50 m each. The system of substructure consists of a pair of spherical bearings for each superstructure, supported on a piercap, flared from a single pier as given in above figure. Foundation system is either a group of 1.0 m dia pile or shallow footing, depending on the type of subsoil at each location. 359
3 DESIGN ASPECTS 3.1 General The design is carried out as per Indian Roads Congress (IRC) codes of practice. The following loads are considered for the design as per IRC-6. 3.1.1 Live load Live load types of Class 70R, with a single train totaling to 100 MT occupying two lanes and Class A, totaling to 55.40 MT occupying single lane, individually or in combination to give the maximum effects on various elements are considered. 3.1.2 Materials Grade of concrete for superstructure and crash barrier /median adopted is M60 and grade of steel is Fe 500. Prestressing steel is of nominal diameter 15.24 mm, conforming to IS: 14268 class II, with yield strength 1860 Mpa. 3.2 Method of analysis and design The following procedure was followed for the design of Strutted Box Girder: Since the deck is simply supported for all the gravity based loads, a regular line model analysis was carried out. 1. Longitudinal bending moments, shear forces and torsional moments were estimated on single beam model from STAAD. 2. For the moments and forces due to restraints of warping torsion at the ends, the design shear forces and bending moments in the longitudinal direction due to live load were increased by 10% and the warping moment from analysis has been found to be within this assumed value. 3. The transverse analysis was carried out in STAAD using 3D thick plate elements. 3.3 Design philosophy 3.3.1 Superstructure The spacing of struts is an important factor which will dictate the behavior of box close to the three celled box. It was observed that a superposition of the line model analysis coupled with a unit slice analysis in the transverse direction, gives, though conservative, a fairly accurate result for all practical purposes. The effect which was not obtained directly from the above analysis is the distortion, which will impart additional torsional and longitudinal warping effects, which was accounted by enhancing the corresponding unsymmetrical (in this case Live load) bending moment by about 10%, though it is seldom more than 5 to 7%, observed from rigorous analysis B.E.F analysis. External struts were also providing additional channel to transfer the torsion to the supports. A full scale plate model was done for obtaining the behavior of the strutted box in the transverse direction. The deck slab is designed for the forces obtained from the plate model directly. 3.3.2 Substructure and foundation Substructure consists of system of bearings piercap & pier. The bearing articulation on a single pier is given in Figure 5. The system of continuity provided on the deck slab is as given in Figure 6. The deck is tied together as shown, for all longitudinal forces. The joint is detailed in such a way that, moments are not transferred from one span to another. Thus the deck behaves as a simply supported beam for all the gravity loads. The bearing system was designed in such a way that all the bearings are capable of transferring the vertical and horizontal loads, including tensile forces arising due to most adverse load combination during maintenance. The stability of the structure and the bearing loads were calculated 360
Figure 5.
Bearing layout for one pier.
Figure 6.
Deck continuity.
for a maintenance condition of removal of complete wearing coat from one carriageway, while unrestricted movements of heavy traffic is allowed on the adjoining carriage way. Also enough safety factors were accounted during all the stages. Foundation system consists of either piled foundation of 1.00 m diameter or shallow footing resting on hard stratum of bearing capacity more than 400 kN/sqm. 4 CONSTRUCTION ASPECTS The construction of this long flyover on an urban stretch requires lot of planning and monitoring. Construction of foundation & substructure was taken up in stages after diverting the traffic on already completed service roads. Any utility services which were encountered on the median where piers were planned were shifted to designated locations. While this was being carried out, precasting of segments were initiated in the casting yard. Struts were precast separately and kept in position, before the box segments of 3.0 m were concreted. Short line casting method was adopted. The deck was cast horizontal corresponding to the normal camber of 2.5%, and proportionally tilted for any further camber. These segments were lifted and stacked, after cross prestressing. The design caters for two layers of stacking. The demoulding time, and lifting time were so optimized that there was a continuous supply of segments to the site. At site three over head launching girders were put to service simultaneously at various points inorder to cover the whole stretch. The weight of each segment was 40 MT. The launching girder capacity was 500 MT. 361
Each segment was transported to the site by trucks, lifted and held in position by LG. After lifting of one complete span, all the segments are dry matched and checked for the geometric accuracy. Then the segments are separated and epoxy glue is applied on the face. Temporary prestress is applied in between each segment by means of tensioning bars. After all the segments are glued, then permanent prestress is applied and released from LG to the bearing. LG is marched forward to the next span. Other finishing works were taken up subsequently. The total cost of the project is 3000 Million Indian Rupees. 5 PHOTOS OF THE COMPLETED PROJECT Some views of the finished flyover is given below.
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Conceptual design
Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
Development of a submerged floating tube bridge for crossing of the Bjørnafjord M. Reiso, T.H. Søreide & S. Fossbakken REINERTSEN AS, Trondheim, Norway
A.S. Brandtsegg & S.A. Haugerud Dr. techn. Olav Olsen AS, Oslo, Norway
A. Nestegård DNV GL, Oslo, Norway
J.H. Sekse & A. Minoretti The Norwegian Public Roads Administration, Norway
ABSTRACT: The present paper looks at Submerged Floating Tube Bridge (SFTB) configurations anchored to the seabed by groups of tethers and their response characteristics. Swell wave loads are applied on the SFTBs and the structural response is given for different cross sectional shapes in the form of box and twin tube cross sections and different number of tether groups. Both simplified modal response analyses and time-domain finite element analyses are carried out. The main findings are; 1) the added mass in vertical mode results in inertia loads being in the range 2–5 times higher for the box cross section compared to the twin tube cross section, 2) the tether stiffness needs to be adjusted to avoid large heave motions of the SFTB.
1 INTRODUCTION The Norwegian Public Roads Administration (NPRA) has been commissioned by the Norwegian Ministry of Transport and Communications to develop plans for a ferry free coastal highway (E39) between Kristiansand and Trondheim, on the west coast of Norway. The background for the plan is that studies on fixed crossings have shown the potential to develop trade and industry and thus employment and settlement in the influenced regions (NPRA, 2012). The approximately 1100 km long coastal corridor comprises today seven ferry connections, most of them across wide and deep fjords that will require longer span structures than previously installed in Norway. This implies an effort on the technical development, based on recognized experience from offshore oil and gas installations. For the deep fjords the Submerged Floating Tube Bridge (SFTB, also known as SFT or Archimedes Bridge) is regarded an attractive crossing solution. The recent attention paid to the visual impact of the structures on the surrounding landscape as well as the audible noise lead the SFTB to be a preferred solution for the crossing. The applicability of the SFTB technology has recently been proven in a feasibility study for the 3.7 km wide and 1300 m deep Sognefjord (ROO, 2012). Following the Sognefjord study, considered the most difficult and challenging fjord to cross, crossing studies are initiated for other ferry connections along the E39. One case for which a fixed link is to be assessed is the five kilometre wide and 600 meters deep Bjørnafjord between Reksteren and Os as part of the development plan for E39 Aksdal – Bergen. The assessment study for the SFTB is carried out by the design group REINERTSEN – Dr.techn. Olav Olsen – Norconsult et al. in a close collaboration with the NPRA. 365
The present paper describes the assessment study of the SFTB concept for the Bjørnafjord crossing. The focus of the paper is the global dynamic analyses of the tether stabilized SFTB solutions comparing different tube bridge cross sections as being exposed to swell waves. 2 METHODOLOGY 2.1 Environmental conditions For long slender floating strait crossings environmental loads are essential. For a Submerged Floating Tube Bridge (SFTB) supported by tethers fixed to the seabed, wind loads alone are not prone to any direct excitation of the structure. The main environmental loads are surface waves, either wind sea or swell. The wind sea is generated over a fetch length defined as the stretch of open water from the crossing to the nearest shore in the direction of the wind. Loads from wind generated waves need to be investigated as it may cause resonant response on three different time scales. First, the wind sea causes first order response at periods close to the wave spectrum peak period. Secondly, it generates slowly varying low-frequency loads. These low-frequency loads may lead to resonant response in horizontal modes of the bridge. Finally, extreme wind sea may give rise to sum-frequency wave loads exciting higher vertical eigenmodes. Swell waves are generated by distant storms in the ocean, propagating into the fjord at wave periods characteristic for the actual storm wave conditions. Hence the wave periods for swell may be in the range 12–18 seconds and may excite modes with periods in the same range. The significant wave height in a sheltered fjord for swell is considerably less than for wind sea at the same return period, typically 15–20 percent of the wave height for wind sea. Also loads due to current need to be investigated. Current introduces mainly static horizontal loads on the bridge. However, for long crossings, slow fluctuations in the current speed along the bridge may excite resonant response of the bridge. The present paper focuses on bridge response due to swell waves which is considered to represent the most critical environmental loading for the bridge, exciting primary eigenmodes (ref. section 2.2). Long duration wave measurements have not been carried out for the Bjørnafjord. Hence there is no wave spectrum available based on long term wave statistics. In lack of on-site wave statistics, open ocean swell is transformed from a location off the coast where hindcast wave data are available in terms of significant wave height Hs and peak period Tp characterising sea states with a given return period. Extreme swell with 100 year return period will be the basis for design. Due to wave refraction and wave breaking the open sea wave energy is reduced when propagating into the fjord. A 100 year significant wave height of 13.5 m is reduced to less than 0.5 m in the fjord. The swell wave spectrum is assumed to take the form of a Jonswap spectrum (DNV GL, 2014),
where σ = 0.007 for ω ≤ ωp and σ = 0.009 for ω > ωp . γ is the peak shape parameter of the spectrum. In lack of a measured wave spectrum shape, γ is chosen as 8.0. ωp is the angular frequency corresponding to the peak period (ωp = 2π/Tp ). Waves in the fjord are not uni-directional, rather the wave energy is somewhat spread. This is modelled by a directional wave spectrum which is multiplied with the Jonswap wave spectrum. The directional wave spectrum chosen is the cosn θ spectrum (DNV GL, 2014). A typical spreading parameter for swell is n = 8. 2.2 Wave frequency excitation loads The wave action on the SFTB is strongly dependent on its depth of submergence, in particular for wind sea with typical wavelength λ ∼ 30–50 m for extreme conditions. Since the Bjørnafjord is rather deep, wave action decays with depth according to exp (−kd) where k = 2π/λ is the 366
wavenumber and d is the depth of submergence. However swell waves have wavelengths in the order of 400–600 m. This means that considerable actions from swell are experienced at the depth of the tube bridge, although the wave amplitude is considerably lower than for locally generated wind sea. Swell waves are long waves compared to the cross-sectional dimensions of the tube bridge. This means that the wave excitation load can be approximated by long wave theory where the load is linearly proportional to the fluid particle acceleration vector normal to the tube bridge an . The sectional load in the horizontal (j = 2) and vertical (j = 3) directions can be taken as the inertia term in Morison’s equation,
where ρ is the water density, Cm is the inertia coefficient, Atube is the cross-sectional area of the tube bridge, and an, j is the acceleration component in direction j. In the time domain analysis, time series of fluid particle accelerations in irregular waves are generated in each node. In a frequency domain analysis the sectional loads Sf (ω) are represented in terms of load spectra proportional to the wave spectrum Sf (ω) and to the square of the load transfer function Hf (ω),
Modelling short crested waves in a frequency domain can be simplified by introducing the concept of correlation length which is a measure of typical width of wave crests in an irregular sea state and can be estimated from the wave spectrum and the wave phase function.
3 STRUCTURAL RESPONSE MODELS 3.1 Frequency domain modal analysis A simplified and conservative method to estimate the maximum response of the tube bridge due to wave loading is the modal analysis approach based on a simple linear beam equation. The maximum horizontal sway deflection of mode i of the SFTB in a sea state (Hs , Tp ) is given by
where q(ωp ) is the distributed load per unit wave amplitude and unit length and Y (ωp ) is the modal frequency response function for sway motion evaluated at the peak angular frequency ωp . L is the length of the bridge, dcorr is the correlation length to account for short-crested sea, and φi (x) are the modal shape functions for the beam with clamped boundary conditions at both ends. Ki , Mi and Ci are consistent modal stiffness (with contributions from tube bridge bending and tether stiffness), mass (including added mass and mass of tethers) and damping for the beam, respectively. The excited mode of the bridge is determined by the peak frequency of the wave spectrum. The distributed load is based on a long wave approximation as given above. The corresponding maximum moment is then given by
Similar expression is derived for the vertical deflection and corresponding moment. 367
3.2 Time domain analysis The time domain analyses of the response of the tube bridge to wave action are carried out using a general nonlinear Finite-Element-Model (FEM) framework named 3D float (Myhr and Nygaard, 2012; Nygaard, 2015). The elements are based on Euler-Bernoulli beams with 12 degrees of freedom (DOF). Cable elements with reduced bending-stiffness are used for modelling the tethers. Geometric non-linearities are accounted for by a co-rotated FEM approach, where the reference configuration is a recently deformed state. The element equations are stated in a coordinate system attached to the midpoint of the element in the reference state, and then transformed to a common component coordinate system. This allows for the utilization of small-strain elements for large global deflections, as long as the element resolution is sufficient. External loads from buoyancy, waves, and tethers are applied as distributed external loads on the structure. Loads are evaluated at Gauss points in the elements, and a Galerkin approach is used to evaluate consistent nodal loads. Since deflections of the tube bridge are small, the mean position of the geometry is used when computing wave forces. Added mass terms involving structure acceleration are added to the mass matrix, while all other loads are kept as applied loads on the right hand side of the equation system. Point forces can be applied to nodes. 4 LOAD CASES The different swell load cases are given in Table 1. The direction of the spectra is defined with zero degrees towards east and with a positive axis in the counter clock-wise direction, hence 150.0 degrees represents swell coming from the West-North West. The wave spectrum period, Tp is varied from 12 seconds to 20 seconds. Each load case (ULS 1–ULS 9) is run using 3 different seeds each of duration 60 minutes. The number of tether groups across the bridge length of 4350 m is 14, 21 and 28 for both the twin tube and box cross section. The main dimensions for the twin tube and box are given in Table 2. Their cross sectional shapes are shown in Figure 1. For all six configurations the crossing over the Bjørnafjord is straight in the vertical plane. In the horizontal plane the crossing is laid in the 1st mode shape (R = 5000 m) to suppress the first sway mode. The vierendeels are used to connect the twin tubes together in rectangular openings (rather than the more traditional triangular truss). The vierendeel is a frame capable of transferring and resisting bending moments. The tether groups are evenly distributed along the crossing path. The seabed profile is shown in Figure 2. The right-handed coordinate system is oriented with the x-axis following the longitudinal direction of the crossing, y-axis being in the tube bridge transverse direction and z-axis being in the vertical direction (positive upwards). Table 1. Swell load cases with significant wave height (Hs ), wave period (Tp ), peak shape parameter (γ), wave direction and spreading parameter n. Run id
Hs [m]
Tp [s]
γ [−]
Dir. [deg]
n [−]
ULS 1 ULS 2 ULS 3 ULS 4 ULS 5 ULS 6 ULS 7 ULS 8 ULS 9
0.4 0.4 0.4 0.4 0.4 0.2 0.2 0.2 0.2
12.0 13.0 14.0 15.0 16.0 17.0 18.0 19.0 20.0
8.0 8.0 8.0 8.0 8.0 8.0 8.0 8.0 8.0
150.0 150.0 150.0 150.0 150.0 150.0 150.0 150.0 150.0
8.0 8.0 8.0 8.0 8.0 8.0 8.0 8.0 8.0
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5 RESULTS AND DISCUSSION Normalized results are given for global displacements, moments and forces for the different tether configurations. The x-axis corresponds to the tube bridge direction, with south and north ends corresponding to x = 0 and x = L ∼ 4350 meters, respectively. All results are normalized with respect to the highest value in each displayed figure (box and twin tube cross sections normalized separately). Table 2. Main cross sectional dimensions for the twin tube and box. Parameter
Twin tube
Box
c/c tubes/width [m] Width/height [m] Wall thickness [m] Iyy [m4 ] Izz [m4 ] Vierendeel
40.0 15.0 0.96 2102 36,080 28
43.8 10.4 – 2508 28,507 –
Table 3. Eigenmodes and eigenperiods for the twin tube and box cross sections in heave and sway direction for 21 tether groups. Eigenmode
Sway, twin tube [sec.]
Heave, twin tube [sec.]
Sway, box [sec.]
Heave, box [sec.]
– 52.8 28.0 17.0 11.4 8.2
11.0 11.0 11.0 11.0 11.0 10.8
– 56.8 31.7 17.3 13.6 10.6
11.7 11.5 11.2 10.9 10.5 10.1
Horizontal layout initially in straight mode configuration (modal approach) 1 2 3 4 5 6
– 50.7 25.8 15.6 10.5 7.5
6.2 6.2 6.2 6.2 6.1 6.1
Horizontal layout initially in 1st mode configuration (FEM analysis) 1 2 3 4 5 6
Figure 1.
– 62.5 38.0 22.2 18.6 14.7
6.7 6.5 6.4 6.2 6.0 5.8
Submerged floating tube bridge cross sections; box (left) and twin tube (right).
Figure 2. Vertical seabed profile; south (left) and north (right).
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Figure 3. Normalized sway mode shape for the straight crossing (modal approach) for the box (top) and twin tube (bottom) cross sections for run id ULS 5 (Table 1), 21 tether groups.
Figure 4. Normalized sway displacement for the box (top) and twin tube (bottom) cross sections for different number of tether groups for the tube bridge crossing in run id ULS 5 (Table 1).
Table 3 gives the eigenmodes for the twin tube and box cross sections, both for the simplified modal approach and the FEM analysis. It is seen that the sway mode for the straight horizontal configuration give eigenperiods of 15.6 seconds and 17.0 seconds in 4th mode for the twin tube and box cross sections, respectively. These are in the range of the swell wave periods (Table 1) and will cause the largest responses in the sway direction, as can be seen in Figure 3 where the swell wave period is given for 18 seconds. The bottom part of Table 3 shows the eigenmodes for the tube bridge cross sections based on an initial horizontal configuration in the 1st mode. One can see that the 1st sway mode is cancelled. The remaining results in Figure 4 through to Figure 7 are based on this initial 1st mode configuration. 370
Figure 5. Normalized bending moment about the z-axis for the box (top) and twin tube (bottom) cross sections for different number of tether groups for the tube bridge crossing in run id ULS 5 (Table 1).
Figure 6. Normalized heave displacement for the box (top) and twin tube (bottom) cross sections for different number of tether groups for the tube bridge crossing in run id ULS 2 (Table 1).
For the box cross section, a 4th mode pattern is seen for the sway displacement (Fig. 4), while the tube cross section show a trend towards a 5th mode and also with some higher modes present. The sway moment for the box cross section shows a pattern in the 4th mode (Fig. 5). The twin tube cross sections show a less pronounced sway moment trend. The higher oscillation found for all three twin tube configurations correspond to the 28 vierendeels connecting the twin tubes. All results are based on the largest excitations, i.e. ULS 5 in Table 1. For the heave displacement (Fig. 6) and heave moment (Fig. 7) for both the box and twin tube cross sections it is seen that the response amplitude decreases with increasing number of tether 371
Figure 7. Normalized moment about y-axis for the box (top) and twin tube (bottom) cross sections for different number of tether groups for the tube bridge crossing in run id ULS 2 (Table 1).
groups. In fact; the number of peaks is the same as the number of spans, i.e. 14/21/28 for both the box and twin tube cross sections for the heave moment. The magnitudes of the heave displacement and heave moment are about 2.0 and 4.5 times larger, respectively for the box cross sections compared to the twin tube cross sections. This is due to the larger added mass in heave direction for the box cross section giving higher inertia loads. For the heave motion (Fig. 6), i.e. from approximately 250 meter to 3500 meter for both the box and twin tube cross sections a global single mode shape is seen. This originates from the stiffness of the tethers. The tether stiffness needs to be increased in order to reduce the overall heave motion of the tube bridges. This is achieved by increasing the cross sectional area of the tethers. 6 CONCLUSION The governing sway response mode for the box cross section is the 4th mode shape, while the tube cross section shows governing sway responses for the 5th mode shape and higher. This is in accordance with their individual eigenmodes in sway and the applied swell load. For the heave displacement it is seen that the tether stiffness needs to be increased to reduce the overall heave motion of the tube bridges. The heave magnitudes are higher for the box cross section than the twin tube cross section due to the higher inertia loads. Further works include investigating the structures for other load conditions than the pure swell. REFERENCES DNV GL (2014) DNV-RP-C205: Environmental conditions and environmental loads. DNV GL, Oslo, Norway. Myhr, A. & Nygaard, T.A. (2012) Load Reductions and Optimizations on Tension-Leg-Buoy Offshore Wind Turbine Platforms. In Proceedings of the 22nd International Offshore (Ocean) and Polar Engineering Conference (ISOPE2012), Rhodes, Greece, pp. 1–8. NPRA (Norwegian Public Roads Administration) (2012) Project Overview – Coastal Highway Route E39. NPRA, Oslo, Norway. Nygaard, T.A., J.B. de Vaal & Maus, K.J. (2015) 3Dfloat User Manual, Version 0.0. IFE, Kjeller, Norway. ROO (REINERTSEN Olav Olsen Group) (2012) Feasibility study for crossing Sognefjord – Submerged floating tunnel - Design Basis. ROO-group, Trondheim, Norway, Report 11744-ROO-R-002 Rev 01.
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Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
Three span floating suspension bridge crossing the Bjørnafjord J. Veie & S.H. Holtberget Norwegian Public Roads Administration, Norway
ABSTRACT: As part of the E39 fjord crossing project The Norwegian Public Roads Administration (NPRA) performs a feasibility study for crossing the 5 km wide Bjørnafjord. One of three alternatives for the crossing is a multi-span suspension bridge on floating foundations. The suspension bridge consists of three main spans at 1385 meters each. The superstructure is supported by two land based pylons and two floaters midfjord. The design of the floaters are based on known tension leg platform (TLP) technology from the offshore industry that has proven to be an effective concept for floating constructions at large water depths. The fjord has a depth of up to 550 m, and the seabed is dominated by marine clay in the anchor areas. These conditions are ideal for a TLP solution with anchors combining gravitational and suction abilities. The main concerns for a multispan suspension bridge of this type are addressed to dynamic response induced by environmental loads, such as wind, wave and current, and the ability to withstand accidental loads, such as ship collision. Initial analyses of the TLP concept have shown it is a feasible alternative for crossing the Bjørnafjord, and that the structure should cope with the load conditions for the site. 1 INTRODUCTION The Norwegian Public Roads Administration (NPRA) is currently working with the ferry free E39 project. A Part of the project is crossing seven fjords with a fixed link. Due to tough conditions, both environmental and geographical, traditional bridge design is not sufficient. Therefore new bridge design concepts have been developed to enable the crossing of these fjords. This paper looks at the Bjørnafjord, located just south of Bergen, and one of the concepts for crossing the fjord. The crossing site is some 5 kilometers wide, and has a depth of up to 550 meters. The width makes it not feasible for a single-span bridge solution (Klinge, 1986). Therefore some foundations in the fjord have to be constructed. Due to the depth, rigid tower type foundations are not advantageous, since these would be of an extensive size. In this paper floating foundations based on tension leg platform (TLP) technology, originally developed for the oil & gas industry, is considered. The TLP technology was developed to enable platforms to be built on greater depths than what was possible for the rigid type. The technology is based on anchoring a floater construction to the sea bed with tethers, in similar fashion of a reversed pendulum, and use the buoyancy of the floater combined with the tension in the tethers to create a stable system. The geometrical stiffness created by a lateral movement of this mechanism gives the floating structure a limited potential sideways deflection in the water plane. The vertical stiffness in the system, defined by heave, pitch and roll motions, are kept as rigid degrees of freedom. In addition to the TLP alternative, two other crossing concepts are being studied for the Bjørnafjord. One is a floating bridge, either as a strait side anchored type, or as an end anchored bucket handle type. The other one is a submerged floating tunnel. 2 SITE AND CONCEPT DESCRIPTION In the deepest part of the fjord the seabed consists of soft sediments. In the shallower parts the ground is solid rock and no soft sediments. At the north side of the fjord there is a rock plateau, 373
Figure 1. The TLP concept for crossing the Bjørnafjord.
with depths varying from 30–150 m. With a main foundation at 120–150 m depth, the main span of a suspension bridge would be about 3300 m. It is a great challenge to put one of the pylons on a foundation at 140 m depth and the span is well beyond today’s word record of approximate 2000 meters. This alternative is not found feasible for this particular crossing. Due to the width of the fjord, various floating bridge alternatives have been studied. The TLP concept for the Bjørnafjord consists of a three span floating suspension bridge with two land based towers, and two floating TLP towers in the fjord. Each span reaches 1385 meters. In total with the approach spans the length of the bridge is 4800 meters. The water depth at the TLP floater sites is respectively 550 and 450 meters. The bridge girder is made of a steel box girder type, similar to those used on The Hardanger Bridge (Statens vegvesen, 2015) and The Great Belt Bridge (Gimsing, 1998). When considering ship collision the TLP concept has a large benefit in reducing the barrier effect for ship traffic compared to the other concepts. The fact that the water plane area consists only of two floaters, and the vertical clearance from the sea level to the bridge girder, gives the concept a large navigation span clearance. 3 THE FLOATERS For the floaters several alternatives have been suggested. But after an initial screening period two different main concepts are being investigated further. One alternative is a 4-legged steel floater. The other one is a more robust single legged concrete and steel combination. Both alternatives have their benefits and disadvantages. 3.1 4-legged steel floater The 4-legged steel floater was the initial idea for the floaters in the TLP solution. It consists of four cylindrical tubes going upwards into a 4-legged pylon. Between the four legs there are truss components to create bending stiffness. The lower part/hull will consist of double bulk cylinders to give some protection towards cracking. The hull will also be ballasted so that the floater has hydrostatic stability. When it comes to installation the 4-legged steel floater has some clear advantages. It can be installed in similar fashion as a jacket, with semi submergible barges carrying the whole floater lying down out to the site. After submerging the barges and releasing the floater, the floater will be towed into position and ballasted at the bottom. This ballast will then create a momentum, erecting the floater. On the other hand the 4-legged steel floater has some weaknesses when it comes to ship collision. The steel hull is vulnerable in the case of a collision, and even the double bulk may be penetrated. Therefore the possibility of a protective structure around the floater was investigated. The concept that was developed for the protective structure was a floating ring, anchored to the tether connection points to prevent eccentricities. The floating ring would use the same principle as for the TLP, so that when the ring is pushed sideways it would submerge, and the geometric stiffness created would slow down the ship, preventing it from damaging the main structure. Initial tests have shown that the ring structure, dependent on the size, might actually stop the ship entirely, preventing it from hitting the main structure. 374
Figure 2.
Figure 3. bination.
4-legged steel floater.
Single legged concrete and steel ccom-
3.2 Single legged concrete and steel combination The single legged concrete and steel combination can be constructed in many different forms. The general idea is that the hull consists of concrete, and attached on top is a steel pylon. For the hull different designs have been investigated, and the alternative that looks most promising is a concrete cylinder attached to a concrete disk at the bottom. The disk at the bottom will consist of an inner cylinder with the same diameter as the one on top, and an outer disk divided into bulks. The disk would be submerged at such an extent that direct ship impact would be less likely since the disk would be deeper than the ships draft. This floater concept is a bit more complicated when it comes to installation. The lower concrete part would have to be cast in a dry dock and towed out to the site. Then the steel pylon on the top would have to be connected at site. The towing of the concrete hull could lead to some challenges, especially when it comes to stability. Regarding ship collision this floater alternative would be more advantageous. The concrete hull would be able to withstand a ship impact to a greater extent than the steel alternative. However a ship impact might damage the hull of the floater. Since the disk would be deep enough to prevent a ship from hitting it, the cylinder going up to the pylon would be the exposed part. A protective structure has therefore also been considered for this alternative, either as a floating ring, like for the 4-legged steel alternative, or as a fence connected to the outer line of the disk. 3.3 Tendons TLP tendons are always vertical and parallel in the initial platform position. Floater surge motion will then not give any pitch, but will be coupled to a vertical motion referred to as “set-down”. Yaw motions coupled to surge will still keep the distance between all bottom and hull anchor points the same for all tendons. Hence, no tendon will tend to loose tension. Any misalignment of the tendons will lead to loss of these ideal properties. Tolerance of tendon anchor positions is therefore an important design issue for a TLP. The Tendon system consists of three major parts: 1. Tether upper connector. 2. Tether main string body. 3. Tether bottom connector. The top connector should be able to maintain tension, but during the installation process no upward load shall be absorbed in the assembly. Connection to the threaded top of the tether, the length adjustment joint (LAJ) is achieved via threaded wedges or slips, which sit in a seat that is located on the porch structure of the hull. The load bearing ring (the seat) is resting on the porch on a flexible ring, the flex-element. Figure 4 shows a typical top connector assembly. 375
Figure 4.
Left: The top and bottom connector. Right: The mechanism of the bottom connector.
Figure 5. The aerodynamic steel box girder.
The tether structure connecting the platform (via the top connector) to the seabed (via the bottom connector) is called the main string body of the tether. The structure normally consists of steel pipe, possibly split in limited lengths and connected with special tether connectors, depending on water depth and the method of installation. Though there are two or three different systems, there is only one reliable type of bottom connection. This patented roto-latch system is applied on most of theTLPs currently in operation. The tether part of the bottom connector assembly consists of a forged steel shaft, connected via flexible connection to the bottom connector head. The casting forming the head, has lugs on the outside hooking to the receptacle, which is welded onto the foundation. The lugs also guide the head correctly into the receptacle. The latching and unlatching process is shown in the process below. The receptacle is a made of a steel forging and is welded to the foundation. Inside the receptacle slots are machined which guide the lugs to latch or unlatch the connector to the foundation. After latching locking pins are to be installed to prevent the connector from unlatching in case of accidental loss of tension. An important part of the design is the tether tension monitoring system. Other than visual inspection, during the lifetime of a tether system, there are very few ways to thoroughly inspect tethers. Therefore the continuous monitoring of the tension on the tethers is an important tool. In case of damage to one of the tethers, the change in tension will indicate possible problems. The corrosion protection system will be based on a system of anodes based on corrosion analyzers, which will be mounted along the tether system. 4 THE SUPER STRUCTURE The superstructure consists of two main cables that spans from one side of the fjord to the other. The main cables are connected to the bridge deck with hangers, and to the towers with cable saddles. The main girder is an aerodynamic steel box, with a height of 4.4 meters and a width of 34 meters. To limit longitudinal movements of the suspension bridge from live load, hydraulic buffers are placed between the pylons at axis 2 and 5 and the bridge girder to keep the girder movements within the working range of the expansion joints. The buffers allow “slow” horizontal movements of up to ±1.4 m, which should cover the total extent of movement arising from temperature change and free two-axis rotation of the bridge girder. The joint will then have a capacity of ±1.4 m. For a temperature span of ±40 degrees Celsius the movement at each joint at axis 1 and 4 will be within the limits of ±1.0 m. 376
Figure 6.
Cable clamp with the force distribution.
Figure 7.
Sea bed conditions for the Bjørnafjord and placement of the bridge.
For “faster” movements caused by traffic-, wind- and wave-induced vibrations, the buffers have a dampening effect on longitudinal oscillation of the bridge girder. It will also have an influence on the damping of the lateral slow drift motion, but the magnitude of this is still to be investigated. When movements larger than 1.4 m occur, the buffer system will act as a stop block and introduce horizontal normal forces in the girder. Each buffer should be stable for dynamic loads in the range up to 5 MN, at all positions of the piston. The design stop block force is up to 60 MN at each pylon (two buffers). An important detail is also the use of cable clamps that reduces vertical deflections in the main spans in the magnitude of 40%. At the center of the 3 main spans, the main cables and the bridge girder are interconnected by locking devices consisting of cable clamps held by trussed triangles to the bridge girder. Each clamp is a cast steel element attached to the main cable by screwed rods. The splice clamp and the steel triangle are bolted together. 5 THE ANCHORS Based on a limited initial survey with acoustic profiling and soil sampling it can be concluded that the sub bottom conditions are suitable for suction anchors (H. Systad, 2013). In addition, properties of the Bjørnafjorden clay are comparable to other deep sea clays found in the Gulf of Guinea and the Gulf of Mexico. Suction anchors are installed in both these locations, and experience from these 377
Figure 8. Combined suction and gravitation anchoring.
Figure 9. Horizontal exitation of the floater with reaction forces.
locations may be fruitful as comparison. For the initial concept study partly penetrated gravity based anchors are considered. These anchors work as combined suction and gravity structure that penetrates the bottom with the lower end. Afterwards the top part is filled and/or covered with a mass. This type of foundation was used on the Snorre-A platform (Almeland, 2013). For this foundation type the tethers are anchored by Concrete Foundation Templates (CFTs). To account for seabed topography the skirt penetration depth can be made different for each of the foundations. The cells are closed at the top with a watertight dome structure. The outer skirt walls are extended above the domes forming retaining walls for solid ballast to be placed in the compartments on top of the domes. Strict installation tolerances are required. Two docking piles for each CFT are pre-installed as guides to make sure that the CFTs are installed within the tolerances. The vertical position and the inclination are both controlled by the suction pressure in the skirt compartments. It is assumed that the upper soil layers comprise very soft clay. The average static tension CFT is counteracted by the weight of the TLP foundation and additional ballast in the top compartments in order to avoid any long term creep effects in the form of uplift. In calm weather conditions the soil resistance is not mobilized. The additional cyclic wave and wind induced loads are transferred to the soil partly by skirt friction and suction under the domes. The tension load becomes inclined as the TLP moves horizontally. The CFT may be designed to carry a horizontal load component corresponding to an inclination of approximately 7 degrees in the tethers. For 550 m water depth, this corresponds to an approximate offset of 65 m while response calculations predict an offset in the range of 20–30 m.
6 ACTION AND REACTIONS By crossing a fjord several load conditions has to be taken into consideration. Traditional loads like wind, traffic, earthquakes and accidental loads must be considered, but in addition you have wave loads, current and ship collision. These load conditions makes the design more complex than what is normal for large bridge constructions. Aerodynamics and especially hydrodynamics are generally of a non-linear nature. Coupled with a structure held by vertical cable members that are tensioned by buoyancy for the discrete points that the floaters represents and by catenary cables 378
tensioned by gravity force for the continuous bridge girder, this all together represents a complex nonlinear analytical case. Regarding design codes and regulations the Eurocode regime are used as the basis to give the correct degree of safety. Since the Eurocode doesn’t cover all aspects of designing such a type of bridge, offshore codes are utilized, but the safety level are upgraded to meet the reliability class (RC/CC) 3 according to NS-EN 1990 (Standard Norge, 2008). Suspension bridges with single box sections have been proven to be robust for spans up to 1600 meters. For Bjørnafjorden bridge spans of 1385 meters are chosen. This is only 75 meters longer than The Hardanger Bridge which has proven a robust performance after the construction was finished in 2013. The chosen spans for Bjørnafjorden is within the limits for well-known technology. Around the world more than 25 TLP type offshore structures are built and are operating in harsh offshore climate such as experienced in the North sea with wave height as high as 20 meters. In the gulf of Mexico several TLPs have proved that they can withstand hurricane conditions (American Society of Mechanical, 2004). For the Bjørnafjord, choosing a similar construction should give a high potential for a robust and durable foundation for a bridge. Forces from waves, wind and ship impact on the bridge structure will always be larger in horizontal direction than in vertical direction. Hence, by making in-plane motions (surge, sway and yaw) compliant, the largest environmental and accidental forces can be balanced by inertia forces instead of by forces in rigid structural members. On the other hand in order to make the bridge comfortable to drive on, heave, pitch and roll motions must be minimized and hence be rigid degrees of freedom. In order to realize the idea of a partly rigid and partly compliant TLP, the Eigen frequencies for in-plane motions must be below frequencies for wave energy and Eigen frequencies for out-of plane motions must be above. For an in-shore structure this means in practice that Eigen periods for heave, pitch and roll must be below 4 seconds, while surge, sway and yaw motion will be above 25 seconds. By tuning the Eigen frequencies relative to first order wave forces, higher order force components becomes important. Loads on difference frequencies (ωi − ωj ) will give slowly varying (drift) forces that together with both first and higher order wind force components might give rise to resonant in-plane motions. For such an inshore structure there is significantly more energy in the wind spectra in the range for the typical first lateral Eigen period from 70 to 130 seconds depending on the type of hull structure. Thus, wind excitation will dominate the slow drift offset. Higher order components (2ω, 3ω, …) in waves and hydrodynamic forces may give resonant out-of-plane response known as ringing. Both slow drift and ringing are considered during the design of the bridge. NPRA has been working with a combination of a suspension bridge and a tension leg floating foundation since 2011. In 2013 a concept involving moored gravity based (SPAR) type of foundations were proposed for the crossing of Sognefjorden (Jakobsen, 2013). Since the depth at the proposed location for the crossing exceeded 1200 meters a pure TLP solution were found not feasible. The size of the gravity based foundation became quite big in order to meet the SLS criteria. Later a solution combining vertical tendons and slack moorings are proposed as such a system will give large vertical stiffness and keep the structure inside approved limits concerning horizontal motions. For Bjørnafjorden it is found that the restoring forces from the tensioned tendons and the main cables are sufficient to keep the structure in place. With a total horizontal force in both main cables in the magnitude of 350 MN, the horizontal stiffness at pylon top for mode 1 (half a sine wave) is in a simplified matter found to be 350 MN/1385 m = 0.25 MN/m. Choosing the tension in the tendons to a total in the magnitude of 135 MN the horizontal stiffness are approximate 135 MN/550 m = 0.25 MN/m. This shows that the horizontal geometric stiffness for the tendons and the main cables are of an equal magnitude. For mode 2 (sine wave) the stiffness from the main cables are in a higher magnitude than from the tendons. Looking at 100 years storm a dimensioning Hs (significant wave height) of 2.7 and Tp in the range of 5 to 14 seconds for locally generated waves and Hs 1.4 for swell in the range of 14 to 24 seconds and 10 min wind speed of 29 m/s at 10 meters height gives a constant mean deflection in the magnitude of 10 to 15 meters depending on alternatives and pretension in the tethers. Looking at slow drift effects the preliminary calculations indicates a DAF in the order of 2, leading to a peak deflection in the order of 20 meters. Using 1 year return period as basis for SLS criteria the maximum amplitude will be in the range of 10 to 15 meters as Hs is 1.6 (4–14) and swell 0.7 379
Figure 10.
Horizontal deflection as half a sine wave viewed from the above.
(14–24) and v = 21 m/s. This deflection is well below a limit of L/200 = 4150/200 ≈ 21 m which is often used as a restriction. Compared to Heidrun (Teigen and Haver, 1998) a wind speed of 29 meters gives a mean vector offset of approximately 5.5 meters, while the maximum total is approximately 12 meters. Heidrun has a total tensioning of 300 MN at a depth of 345 meters, giving a stiffness in the magnitude of 300/345 = 0.87 MN/m. For the bridge case the stiffness is 2 ∗ 0.25 = 0.5 MN/m. Scaling the Heidrun displacement according to the chosen bridge stiffness gives a displacement of 12 ∗ 0.87/0.5 = 20 meters. This corresponds with initial time series analysis with low damping in the range of 0.1 to 1% of critical damping. The analysis gives a DAF in magnitude of 2, which also corresponds well with what is reported for Heidrun TLP. In general the Eurocode (Standard Norge, 2010) yields traffic loading for bridges with spans up to 200 meters. In Norway the NPRA has proposed a load case for traffic giving 9 kN/m for each lane and 2 kN/m for pedestrian lanes. In this case this sums up to a total uniform distributed load of 38 kN/m. With an optimal structural concept as described in chapter 4 a maximum vertical deflection lies in the range of 6.5 meters for an asymmetric characteristic load. 7 SUMMARY This paper presents the major characteristics of a suspension bridge supported by floating TLP foundations. The solution is one out of three alternatives for the Bjørnafjord strait crossing. By introducing a compliant system in the horizontal plane, forces from environmental and accidental loads can be balanced by inertia forces instead of by forces in rigid structural members. By keeping the degrees of freedom in the vertical plane (heave, pitch and roll) rigid the service limit state (SLS) criteria can be fulfilled. The concept has a large benefit in reducing the barrier effect for ship traffic as it has a large navigation span clearance and are flexible due to vertical clearance from sea level to the bridge girder. Although the concept represents a highly complex structural system in terms of bridge construction, it is based on combining well proven and robust technology from the offshore and bridge industry. Based on this experience and adding some ongoing research and technology qualification, the concept should be ready for construction within a few years. REFERENCES Almeland, I.-B. (2013) Tlp Technology Experiences In The North Sea Used As Foundations For A Bridge Tower. Strait Crossings 2013 – Proceedings, 918–929. American Society Of Mechanical, E. (2004) Omae 2004: 23rd International Conference On Offshore Mechanics And Arctic Engineering: June 20–25, 2004, Vancouver, Canada: Proceedings, New York, Asme. Gimsing, N. J. (1998) East Bridge, Kã¸Benhavn, Storebã|Ltsforbindelsen. H. Systad, A. S. (2013) Sub-Bottom Investigations For A Floating Structure Across Bjørnafjorden Anchoring Conditions. Strait Crossings 2013 – Proceedings, 709–719. Jakobsen, S. E. (2013) Sognefjorden Mulighetsstudie Flytebru. Klinge, R. (1986) Proceedings, Trondheim, Tapir. Standard Norge (2008) Ns-En 1990:2002+Na:2008. Standard.No, Standard Norge. Standard Norge (2010) Ns-En 1991-2:2003+Na:2010. Standard.No, Standard Norge. Statens Vegvesen (2015) Technical Facts. Statens Vegvesen. Teigen, P. & Haver, S. (1998) The Heidrun Tlp: Measured Versus Predicted Response. Applied Ocean Research, 20, 27–35.
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Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
Long railway viaducts with special spans: Part 1. Arch construction by balanced cantilever with auxiliary cables J. Manterola, A. Martínez, B. Martín, J.A. Navarro, M.A. Gil, S. Fuente & L. Blanco Carlos Fernández Casado S.L., Spain
RAILWAY ARCH BRIDGE OVER THE CONTRERAS RESERVOIR ON THE MADRID-LEVANTE HIGH-SPEED RAILWAY LINE ABSTRACT: Part of the Madrid-Valencia high-speed railway line, the Contreras Reservoir – Villargordo del Cabriel stretch represents an example of the use of the state-of-the-art construction systems, instigated by the limitations resulting from a layout that allows the running speed of up to 350 km/h, with large radii bends and gradients of less than 30% in area with rugged terrain conditions. It is against this background that the arch bridge over the reservoir stands. This is a reinforced concrete arch bridge with a 261 m span and an upper prestressed concrete deck that on the construction completion date was world record holder for a concrete railway arch bridge (Fig. 1).
Figure 1. View of the arch bridge built over the Contreras Reservoir.
381
Figure 2.
Elevation, plan view and cross sections.
1 GENERAL DESCRIPTION The bridge amounts to a total length of 587,25 m. The arch span measures 261 m and the mid-span sag is 36,944 m, which determines a span-to-rise ratio of 1/6,77 (Fig. 2). The arch is embedded in two large plinths that allow the diffusion of the load over the affected ground by means of direct foundations. It is divided in six parts with a polygonal directrix. That way the non-funicular arch is maintained while reducing the bending moments that would exist in the area between the vertical columns if the arch were perfectly curved. The cross section is a box girder with a variable depth ranging from 2,80 m at mid-span to 3,40 m at the ends. The box girder width is also variable ranging from 6,00 m in the centre of the arch to 12,00 m at the foundations, which is the width required to resist the great bending moments of the vertical axis produced by the plan curvature of the arch and the crosswind. The box girder walls range from 0,60 to 1,35 m. The upper deck span distribution is 32,625 + 12 × 43,50 + 32,625 m. The piers P-6 to P-11 are supported on the lower arch structure. The deck is made of a 3,00 m deep box girder (which determines a 1/14,50 span-to-rise ratio), a 5,00 m wide lower slab, a 6,50 m wide upper one, and series of segments that complete the total section width of 14,20 m. The web thickness is 0,50 m. The webs are thickened over the piers until reaching a total thickness of 1,27 m, to allow the anchoring of the service prestressing cables. The lower slab is 0,30 m thick. The arch has a polygonal curved directrix in the vertical plane, which corresponds to the nonfunicular of the permanent loads. In plan the arch is drawn within the circular alignment of 3 875 m radius in order to avoid eccentric forces at the points where the piers are built-in in the arch. It is made of reinforced concrete C-70, due to the great compressive forces it must bear. The variable height of the piers ranges from 3,53 to 35,38 m. All the piers are generated by one basic pier which has a rectangular box-girder cross section of a 2,60 m constant width and a variable depth ranging from 5,20 m on the upper edge, 3,20 m at the “waist” situated 5,00 m away from the upper edge, and a widening towards the base.
2 ERECTION PROCESS The construction method eventually chosen for this bridge and given its situation with respect to the water of the reservoir and the land was the cable-stayed free cantilever launching of the two semi-arches embedded in their foundations (Fig. 3). A slight modification was taken into account. The proposal was to build the arch by cable-stayed incremental launching as well, only it was to be launched from the first pier of the arch, which was to be extended until reaching the ground where it was then appropriately. At the beginning of the construction, after a particularly favorable hydrological year for this purpose, the reservoir water level was such that the foundations of the temporary pier were above water level for months on end. The construction process was carried out by first executing the approach viaduct and the deck piers using a climbing formwork for the piers and scaffolding truss for the deck. 382
Figure 3.
Cable-stayed free cantilever launching. Drawing, model and reality.
The first section of each semi-arch, between the foundation and temporary piers, is built upon a centering supported on the ground. Once the centered arch section is built, piers P-7 and P-10 are executed over the arch, to allow the advance of the scaffolding truss towards these piers. The centering is then dismantled and the advance of the semi-arches is initiated using cable stayed free cantilevers. To this end metal pylons were placed on the deck, following the vertical line of the temporary piers. From this moment on, the semi-arches advanced in free cantilevers while casted in situ using form traveler. To enable such procedure, we placed nine successive bundles of stay cables on each semi-arch. 383
Figure 4.
General definition.
One very important matter to consider is whether the use of jacks at the key in a bridge built using temporary cable staying should or should not be applied. Jacks at the arch key are aimed at eliminating the forces and strains produced in the arch as a result of the deformation provoked by shortening of the directrix due to axial compression. In this case, it was decided against placing jacks at the key, since nothing is gained from the structural point of view while it complicates the execution. This by no means implies that such a decision is to be generally applied in all cases. RAILWAY ARCH BRIDGE OVER THE TAJO RIVER IN THE ALCANTARA RESERVOIR ABSTRACT: Placed in the High-Speed Railway Line Madrid-Extremadura, the bridge has a total length of 1488 m. The span distribution is influenced by the crossing of the Tajo River, which takes place with an arch, 324 m long, and dividing the deck over it in six spans of 54 m each one. The approach spans are 60 m long, inserting two transition spans of 57 m. The emblematic element of the bridge is aforementioned arch. With curve directrix, it is formed by a hollow variable section between (4.00 m – 3.50 m high; 12.00 m – 6.00 m with). With its main span length of 324 m, it will surpass the bridge over the Contreras Reservoir, currently the largest railway arch bridge executed in Spain. 1 GENERAL DESCRIPTION The bridge has a total length of 1488 m, with a span distribution of 45 + 9 × 60 + 57 + 324 + 57 + 7 × 60 + 45 m. The deck consists on a hollow prestressed concrete section with a height of 4.00 m. This slenderness allows the structure to save appropriately the 60 m approach spans, and so the 54 m spans over the arch which, due to its flexibility, causes complementary bendings. The lower slab is 5.00 m wide, and 6.50 m the upper one, completed with cantilevers to reach a total width of 14.00 m. The web thickness is 0.50 m. Concrete HP-50 is used in the approach spans. Concrete HP-70 is necessary in the track over the arch. The approach spans have 5 prestressing tendons, between 25 y 37 ∅ 0.6 units in each web. In the spans over the arch they are complemented with upper and lower straight tendons. Due to the deck length, larger than 1200 m, the possibility of placing the fixed point of the horizontal actions in the key of the arch was studied and finally chosen as the best option. The increases of the stresses in the arch were acceptable. Using this configuration, typical expansion joints can be placed at both sides of the bridge. The arch has a curve directrix in the vertical plan. This directrix has been obtained after a detailed process of optimization of the dead load bending stresses, looking for an approximation to the 384
Figure 5.
Deck definition.
Figure 6. Arch definition.
antifunicular curve of these actions. It consists on a quasi-rectangular hollow section with variable height between the start section, 4.00 m, and the key section, 3.50 m. The width varies linearly between 12.00 m in the start section and 6.00 m in the key section. The web and slabs thickness vary to achieve an almost homogenous distribution of the compression stresses. Concrete HA-70 is used. Due to the environmental conditions of the place, and also to the singularity of the structure, a complete study about the aeroelastic behavior of the arch, during the erection and also in service state, was developed with a reduced model in wind tunnel. The conclusion of this study was that the structure was not sensitive to any accountable phenomenon of instability due to wind in the different structural configurations. The piers have a variable height between 9.60 & 71.50 m. All of them are generated from a basic pier with hollow section, 3.50 m wide and variable dimension between in the upper part, 3.20 m in the waist, and widening through the base. Spherical bearings are designed for all supports. Only in the case of the key support, it is needed to create a fixed point for the horizontal actions. Due to the need of supporting the expansion devices over the structure, box abutments with intermediate walls are projected, with maximum heights of 8.63 m at E-1 and 9.64 m at E-2. To carry out the environmental prescriptions, it is necessary to design barriers, 3.00 m high, for bird protection. The barrier is formed by steel curve tubes with 100 mm of diameter, each 0.50 m. 385
Figure 7.
Deck construction with scaffolding truss.
In their lower part, horizontal tubes with variable diameters are placed. A specific study about the aeroelastic behavior of the different shapes of barriers has been developed, resulting this one the optimum solution. 2 ERECTION PROCESS The erection of the deck is developed span by span using a scaffolding truss, placed over both abutments. The central spans placed over the arch are designed to be built using traditional formwork. To avoid excessive stresses in the arch, it has been studied a very symmetric erection of the deck. Only one span of difference is allowed between both sides during the construction. It must be taken explicitly into account in this case the deformability of the system composed by deck, arch and piers during the process of execution of the deck over the arch. The behavior as an elastic support is absolutely essential for an appropriate design of the interaction between the different structural elements that form the bridge. The erection of the arch is done using cable-stayed cantilever launching with provisional steel pylons. These towers are stayed to the foundations of the close piers, needing ground anchors to support the actions. The necessary auxiliary resources are: a tower placed over pier P-11 for the erection of one semiarch and other identical one over pier P-17 for the other one; a form traveler for each semi-arch; a cable-stay system which supports the built truck of each semi-arch anchored in the tower and other one that supports the tower anchored in the foundations. Finally, ground anchors are necessary to hold the foundations. Each tower is composed by two columns joined with a K bracing. The section of each column is a hollow steel box. These columns are separated 6.50 m to make the retention system form a vertical plan. As this separation is larger than the dimension of the pier head, the tower width is reduced in the joint between both elements. The K bracing is composed of horizontal beams placed each 4.00 m and diagonal elements. Both are steel double T rolled profiles. The form traveler is a steel auxiliary structure designed to support the formwork of each arch segment and allowing the cast in situ of that segment. The form traveler is supported in the area of the arch that has been recently erected, to prepare the casting of the next segment. These segments are about 3.80 m long, taking into account a form traveler with a total weight of 95Tons. The process has been conceived using 15 couples of stays supporting each semi-arch, and another 15 couples that hold each tower. There is a couple of stays anchored in the arch each three 386
Figure 8. Arch erection with form traveler.
Figure 9.
Current status/Final view of the bridge.
segments. The typical separation of the cables in the tower is 2.00 m. Only the first couple needs more separation, because they are almost vertical The erection process of each semi-arch begins with the casting of the first segment, which formwork is supported with a centering placed over the arc foundation. This segment needs a 387
minimum length to allow, after the hardening of the concrete, the location of the form traveler to start with the sequence of cable-stayed cantilever launching. Once the form traveler is placed in its position, the following segment is casted in cantilever. After that, the form traveler moves to the end of the second segment, to begin with the execution of the following segment. The disposal of the first family of cables requires special attention, because the pier is not braced yet by the stay system. The process begins with load steps of the 25% of the total load of the cable, following always a established sequence studied to avoid undesirable unbalances. It is necessary to control the displacements at the top of the pier. Once the first family of cables is placed, the sequence continues casting the following segment and moving the form traveler until the next cable is reached. The process enters in an iterative phase. Nevertheless, sometimes is necessary to modify the load of the cables, in order to avoid excessive stresses in the arch. The sequence finishes after the disposal of the last cables and an ultimate modification of the load of the cables, to reach the desired geometry of the arch. At that moment, one of the form travelers is removed, while the other is adapted to proceed to casting of the key segment of the arch. The process is completed dismantling the provisional stays, starting in reverse order, and removing the rest of elements of the provisional cable-stayed system.
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Long railway viaducts with special spans: Part 2. Arch construction by tilting J. Manterola, J. Muñoz-Rojas, A. Martínez & S. Fernández Carlos Fernández Casado S.L., Spain
ABSTRACT: This article presents another system of arch construction by rotation of the semiarches that has been used in several HSR bridges in Spain. It has proved to be very adequate for arched spans in the span range from 100 to 200 m. Using provisional stay cables anchored in the contiguous pier foundations, the semi-arches placed vertically on their springers over provisional hinges are rotated until meeting at the keystone, where they connected. As an example of this technique it is described a viaduct over the River Tera designed by CFC, a 645 m long structure that crosses a reservoir, where a 150 m span without intermediate supports was required. An arch made out of high-strength structural steel (S460) was used to span it. The modulation of structure length was well solved using 75 m long spans with a prestressed concrete box section girder to be built by the incremental launching. Taking into account that the river crossing length (150 m) is the double of this value, the arch solution allows for a constant span length using the arch central section as an intermediate deck support, then keeping the same deck construction solution in the entire deck.
1 BRIDGE DESCRIPTION 1.1 Design constraints The most remarkable constraint of this work is the crossing of the Tera River, in an area containing a reservoir for a hydroelectric power station that had great depth and tight limitations concerning any action that may affect the water level due to the power station operation. The construction company ACCIONA called on CFC to develop a construction variant based on its patented incrementally launched bridge technique, used in several CFC previous projects of HSR viaducts, based on a rack and pinion system for the deck pulling operation. The modulation of this work was well solved with spans 75 m long. Such length is in the current limits for incrementally launched bridges, perfectly applicable when high resistance concretes are used. In a previous work of CFC a 90 m long span had been already reached. The crossing over the river was done without supports, using an arch with double length (150 m) whose keystone serves as the deck support, enabling thus the the same construction solution of incremental launching in the entire civil work. CFC carried out the construction project of the bridge and led the project management of the entire civil engineering complex execution that took place between 2010 and 2013. 1.2 Layout The work layout is straight in plan with a constant slope in elevation. The total length amounts to 645 m with a span distribution of 60-75-150-75x5-60. There is a very steep slope on the right river margin, so a pier had to be embedded here. The slope gradually increases on the left margin. The maximum pier height reached 42 m, which makes 75 m spans suitable both formally and financially. 389
Figure 1.
General view.
Figure 2.
General elevation.
Figure 3. Typical deck cross section.
1.3 Deck The deck dimensions and proportions are those typical for railway bridges built by incremental launching: the 14 m wide platform is placed on a box section, constant depth girder, made of prestressed concrete, with a slenderness of 1/18 of the span (4.40 m). The lower slab is 5.20 m wide, while slightly slanted webs are able to achieve a width exceeding 6.50 m. Such arrangement optimizes both longitudinal and transverse performances by placing the axes of each one of the tracks as close as possible to the axes of the webs. This minimizes the transverse bending of the upper slab, and at the same time reduces the dimensions of the lower slab. The thickness values of the upper slab range from 0.30 and 0.50 m, while the typical thickness of the lower slab is 0.36 m, increasing to reach 0.76 m over the piers. The webs are 0.50 m thick. The characteristic strength of the concrete is 60 Mpa. The same as in any other incrementally launched bridge, two groups of prestressing cables are arranged. A set of straight cables provide due compressions to control variable stresses on the concrete during the launching stages. A second set of undulating cable groups, which are loaded once the launching has concluded, provide the additional capacity to resist final loads on the structure under service conditions. Construction cables providing central prestressing are arranged in both upper and lower slabs in groups connecting three segments further connected using continuous anchorages. These are loaded in the prefabrication yard, with a disposition adapted to the sectional forces during construction, 390
Figure 4.
Bending moments during launching and straight prestressing layout.
Figure 5.
Ondulating prestressing.
which as we know oscillate outside the area of influence of the launching nose, ranging from pl2/12 to pl2/24. In terms of total axial force of the prestressing this amounts to 56% of the total value. The curved service cables oscillate along the webs from one web to another. They are anchored in anchor blocks placed at the intersections with the supports. Once the launching is complete, the cables are tensioned from either ends. In order to avoid problems throughout the launching process due to the empty sheaths that reduce locally the cross section resistance, particular attention is paid to the sheaths distribution and to the local reinforcement in critical sections in the lower slab. In addition, the system usually applied by ACCIONA in its incremental launching bridges, consisting of longitudinal steel plates connected to lower slab corners that makes passing over Teflon bearings easier, contributes to the distribution of the loads on the bottom slab placed over the supports during the launching stages. 1.4 Piers The deck sits on concrete piers that have a hollow rectangular cross section whose longitudinal dimension is a constant 3.50 m, while the transverse one is variable, with a 1/40 slope running from a neck placed below the pier head and whose minimum dimension amounts to 3.20 m. The upper head is bell shaped, 5.20 m wide, decreasing lineally until reaching the neck area at the height of 7.5 m. All the supports of the bridge are sphere shaped, with one free unit and another transversely guided one in each pier at Abutment 1. At Abutment 2 both units are free. They can all be inspected through the voids designed at the pier heads that can be accessed from the deck. There are no tension problems in any of the pier supports. At the abutments, however, the supports are placed slightly wider apart in order to avoid the tension problems, therefore requiring a transverse beam which is installed in a second stage. Taking advantage of the great capacity offered by the launching device, the viaduct is fastened against longitudinal loads at Abutment 2, in order to resist the significant braking reactions of the railway. The fixing is carried out using a protuberance that originates at the abutment and enters the deck’s lower slab. The coupling of the two is carried out by means of elastomeric supports with the adequate rotation capacity. This system introduces bending moments in the deck cross section, taken into account when designing the dimensions. 1.5 Arch over the Tera River The most remarkable part of the viaduct is the 150 m long span supported on an arch which, due to its virtually straight geometry, appears to be composed of two struts. Concrete and metal 391
Figure 6.
Piers. Elevation and view.
Figure 7. Arch elevation.
Figure 8.
Steel arch cross section.
solutions were compared. The metal solution with high-strength steel (S460) proved to be more advantageous. In this situation, the strut’s dead load is of little consequence, and therefore the arch geometry generated from the antifunicular loads results mainly from the point load in the keystone. The curvature of the semi-arches is therefore very small and its front view resembles a straight element. The cross section is rectangular with beveled corners in order to create shadows bring out the edges. The typical plate attached to the walls is 15 mm thick with longitudinal T-shaped stiffeners and transverse K-shaped diaphragms placed every 4 m. 392
Figure 9.
Figure 10.
Launching nose.
Retention hydraulic frame, auxiliary frame on the semi arch, provisional hunger.
2 CONSTRUCTION PROCESS 2.1 Deck launching The constructions process of the bridge deck is carried out by incremental launching from Abutment 2. The segments are 25 m long, since the prefabrication yard had the capacity for 3 segments, which correspond to one complete span. The prefabrication cycle achieved was that of one segment per week, including launching. The metal launching nose is 45 m long, which represents 64% of the maximum span, a customary value to help optimize the forces in the deck. The connection with the deck is carried out by means of vertical and horizontal prestressing bars. 2.2 Semi-arch rotation The most remarkable part of the construction process of this work was the installation of the semi-arches, performed by rotation. The semi-arches were placed vertically over their springs over provisional hinges. Using provisional stay cables anchored in the contiguous pier foundations, the semi-arches were then rotated until meeting at the keystone, where they were duly welded together. Subsequently, the springs were blocked against the foundations and the provisional hinges were embedded in the concrete. The semiarches were divided into two segments and installed using powerful cranes. Once connected, they were then stabilized by the piers which began at the same springs using provisional rotation fastenings. The upsetting of balance in order to initiate rotation is achieved by using pairs of jacks placed at the upper portion of the pier. The back stays are anchored in a revolving frame that holds within the retention hydraulic unit. The anchoring of the stay cables in the semi-arches was carried out using auxiliary frames that were dismantled once the rotation was completed. 393
Figure 11. View of the tilting manoeuvre and the deck launching over the arch.
A control and correction system of the displacements of the arch’s twist hinges is arranged using hydraulic jacks at the provisional lower hinges. The cable is gradually released from the jacks to produce the rotation of the semi-arches until having them facing one another at midspan where they are fixed using provisional bolted embedding. The corrections of the semi-arch position are achieved in two ways: The twists of the arch cross section are corrected modifying the load in every stay cable. The other method used is the sliding system activated by the hydraulic jacks arranged at the arch springer. This system enables longitudinal displacement of the semiarches, while providing a rotation of the vertical axis. Once the position of the arch closure is determined, the fixing and welding at midspan are carried out, the hinges are locked at the arch springs and the void left on them is concreted. 3 SUMMARY SHEET Property: ADIF: Manuel Puga, Agustín Fernández, Agustín Álvarez, Construction: ACCIONA: Antonio Muñoz, Antonio Garrayo, David Higueras, José Álvaro Arch tilting: ALE HEAVY LIFTING Project: CARLOS FERNANDEZ CASADO S.L. Total length: 645 m (60-7x75-60). Width: 14.0 m. Depth: 4.30 m Deck quantities • Concrete H-60: 7529 m3 (0.83 m3/m2) • Prestressing Y1860: 651 T (72 kg/m2) • Reinforcement AP 500 S: 1694 T (187 kg/m2) Arch quantities: • Span 150 m. Sag 40 m • Steel S460: 600 T (285 Kg/m2 in 150 m deck length)
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Long railway viaducts with special spans: Part 3. Precast girders J. Manterola & A. Martínez Carlos Fernández Casado S.L., Spain
ABSTRACT: This third article completes the references of very long HSR Viaducts with special features designed by the firm CFC. It contains the description of the Viaduct over the River Guadalete, a 3221.7 m long structure. The general concept behind this work was the design of a viaduct that would fulfil all the functional and structural demands corresponding to a railway bridge while at the same time fitting in and adapting well to the specific conditions of the landscape, marked basically by a very flat floodplain.
1 VIADUCT OVER THE GUADALETE RIVER IN SPAIN 1.1 Structural concept. General description The Viaduct over the Guadalete river is located in the railway line between Sevilla and Cádiz. It has a total length of 3221.70 m. It is located in a circular alignment of 2200 m of radious. The average height of the different piers is 10 m. – The viaduct crosses twice the Guadalete River and the roads CA 2011 and CA 9023. A bridge which fulfils all the functional and structural requirements was designed from the beginning taken into account the appearance suitable adapted to the landscape which is a very flat fertile valley (Fig. 1–3).
Figure 1. Viaduct over the Guadalete river.
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Figure 2. Viaduct over Guadalete river. Typical cross section.
Figure 3. Viaduct over Guadalete river. Simply supported girders.
One of the most specific features of the bridge is the extremely length and the foundation conditions. The soil has more than 25 m deep very soft layer. These soil conditions are inadequate for resisting horizontal forces due to braking and traction railway loads. The length of the viaduct asks for a typological study taken into account track-deck interaction problems due to imposed deformations, horizontal loads and to the position of the rail and deck expansion joints along the viaduct. In this study different structural, constructional and environmental problems were considered. From the structural point of view, the strength to resist the horizontal forces due to braking and seismic loads should be compatible with the flexibility to reduce as much as possible the stresses due to thermal and long term deformations. It was decided to avoid any expansion joint on the rails in order to improve as much as possible the exploitation of the railway. A maximum length of 200 m between structural deck expansion joints was limited in order to avoid any overstressing on the rails due to track-deck interaction problems. In order to fulfill all the environmental requirements, the supports over the river beds were reduced as much as possible. A precast solution was designed to allow industrialized construction procedures suited to a very long viaduct (Fig. 4 & 6). The final solution adopted is a twin precast box girder 2.20 m deep and located below each railway axis, 2.15 m apart. The total deck width is 13.0 m. This deck allows to span the length of 396
Figure 4. Viaduct over Guadalete river. Precast arches. Table 1. Viaduct over Guadalete river. Span distribution. Stretch
Length (m)
N◦ Spans
Span (m)
1 2
823.50 207.00
3 4
810.00 207.00
5 6
547.20 207.00
7
420.00
27 2 3 27 2 3 18 2 3 14
30.50 30.00 49.00 30.00 30.00 49.00 30.40 30.00 49.00 30.00
Total
3221.70
101
30 m as a simply supported structure along most of the length of the bridge and to span 49 m with the help of two additional precast arches. This concept allows with a very repetitive structure to cross the longest spans due to the presence of Guadalete river beds and the road CA 2011. With these criteria the viaduct is split in 7 stretches with the lengths and spans shown in Table 1 (Fig. 5). The use of continuous precast arches allows to balance the horizontal forces due to permanent loads on the intermediate supports creating a well suited and new structure, Figure 6. 1.2 Track deck interaction problems During the design conception the main idea of a jointless rail bridge has been present in order to reduce maintenance problems. Two main aspects for track-deck interaction problems were studied: 1. The structural scheme to resist the horizontal forces due to braking and traction railway loads. 2. The maximum displacements due to thermal and long term deformations to avoid overstresses on the rails produced by track-deck interaction problems. By these reasons the piers under simply supported girders are able to resist the horizontal loads from each span and the expansion length of 30 m is far from producing any overstressing on the 397
Figure 5. Viaduct over Guadalete river. Span distribution.
Figure 6. Viaduct over Guadalete river. Precast arches.
Figure 7.
Displacement control under railway braking loads.
Figure 8. Track-deck interaction structural model.
rails. For the greater spans the continuous precast arches are able to resist properly the horizontal loads with an expansion length to avoid any overstressing problems on the rails. Preliminary analysis to control track-deck interaction problems were performed: the expansion lengths limitation were fulfil so the so the maximum displacement due to braking loads should be limited to 5 mm at deck level on the structural deck expansion joints. These displacements were obtained taken into account the flexibility of the soil-foundation structure using the stiffness matrix of the whole. It was assumed that all the braking loads are transmitted to the continuous deck which can be considered as an upper limit of the total load because of the continuity of the rails. A detailed model for the track interaction problems was made in order to confirm the main assumptions done in the preliminary analysis. The continuous precast arches were modelled with plane bar elements. Four additional simply supported spans are added to avoid any perturbation on the expansion joints results. The rails were included in the model as an additional structure. The connection between the rail and the deck has been done with perfectly elastoplastic elements which represent the track-deck behaviour, as it has been mentioned above. The maximum forces and stiffnesses vary according the loaded or unloaded track situation (Fig. 8). The different load cases ere done in two different models: 1. Built-in model, in which the arches are completed built in the foundation. 2. Model with soil-structure interaction: in which the soil-foundation stiffness matrix has been included. 398
Figure 9.
Figure 10.
Braking loads. Axial forces on the rail with soil structure-interaction model.
Braking loads. Axial forces on the rail built-in model.
Figure 11. Thermal action +20◦ C on the deck. Axial forces on the rail with soil structure-interaction model.
Figure 12. Thermal action +20◦ C on the deck. Axial forces on the rail in built in model.
Three load cases have been considered on each model: 1. Case 1: Railway braking load at rail level. 2. Case 2: Deck increment of temperature −20◦ C 3. Case 3: Deck increment of temperature +20◦ C The summary of the main results is as follows: 1. For the braking loads the soil structure stiffness has a relevant importance in the total displacements and on the overstressing in the rails (Fig. 9–10). 2. For the thermal actions soil-structure interaction stiffness has no influence in the results (Fig. 11–12). 3. The increment of stresses due to the combination of thermal actions and braking forces are lower than the limits 72 MPa in compression and 92 MPa in tension. 399
REFERENCES Manterola, J, Astiz, M.A., Martínez, A. Puentes de ferrocarril de alta velocidad Revista de Obras Públicas n◦ 3386 pp. 43–77 April (1999). Manterola Armisén, J.; Martínez Cutillas, A. Prestressed Concrete High-speed Railway Bridges. Bridges for high-speed railways. Ed. Rui Calçada, Raimundo Delgado, Antonio Campos e Matos. CRC Press. (2009). Martínez Cutillas, A. Track bridge interaction problems in bridge design. Track bridge interaction on highspeed railways. Ed. Rui Calçada, Raimundo Delgado, Antonio Campos e Matos, José Ma Goicolea, Felipe Gabaldón. CRC Press. (2009).
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Four spans continuous cable stayed bridges without extra cables J. Romo FHECOR Consulting Engineers, Madrid, Spain
ABSTRACT: Four Spans Cable Stayed Bridges is a particular case of multi-spans solutions. Besides the addition of extra cables it is possible to stiffen the system through the efficient use of the relationship between central span to lateral span and towers heights. Influence of deck and central pylon stiffness is analyzed for a case study. Also a parametric study carried out for a 1000 m crossing showed the importance of the ratio between lateral and central span as well as the relative height of the towers to achieve the required rigidity against unbalanced live loads. The influence of those variables and its importance in the aesthetical image of the bridge is also stated.
1 INTRODUCTION One of the main issues when designing a continuous cable stayed bridge is the way of stiffening the structure, to control the vertical deformations due to live loads. It is well known that in standard cable stayed bridge the vertical rigidity is achieved, by the backstays cables that connect the tip of the pylons to the anchor tied down piers. In the case of multi-span cable stayed bridges, the main problem is the deformation and forces created by out of balance live loads. There are several strategies in practice to achieve the required longitudinal stiffness, see Virlogeux M. 2001. A cable stayed bridge with four spans or three towers could be considered as a special case of the general one discussed in the former reference. The advantage of having only one pylon without backstays opens up a new range of solutions. In this paper a discussion of possible alternatives is presented, as well as a specific parametric study of solutions where no additional cables are been used (see Fig. 1). 2 GENERAL BACKGROUND 2.1 The classic cable stayed bridge In classic cable stayed bridges, with a central span and two lateral short spans, the vertical stiffness of the system is basically controlled by the backstays that connect the top of the pylons to the lateral
Figure 1.
Parametric study of a three tower cable stayed bridge
401
piers. The pylon’s longitudinal bending stiffness is rather small compared with the axial stiffness of the backstays. In fact the maximum shear absorbed by the pylon is normally less than the 5% of unbalanced horizontal forces transferred from the forestays cables to the tip of the pylon. The deflection of the main spans is therefore due to the elongation of the back-stays and the deformation of the forestays 2.2 Multi-span cable stayed bridges On the contrary, in continuous cable stayed bridges, the stiffness of pylons and deck play a crucial role in the behaviour of the structure. The seminal paper of Virlogeux M. 2001 includes the main concepts to explain the structural behaviour of continuous cable stayed bridges, where the importance of the relative stiffness between pylons and deck, cable arrangement as well as its connections (bridge articulation) are the key variables in the design. 3 FOUR SPANS CABLE STAYED BRIDGES Four spans cable stayed bridges is a particular case of a multi spans bridges. Examples of different solutions used in that particular case are: the Ting Kau Bridge in Hong Kong, the Queensferry Forth Crossing in Scotland, the Mazcala Bridge in Mexico and the Mersey Gatway Bridge in England. In the four cases the strategies for reaching enough stiffness are complete different, showing that site conditions as well as other key parameters as ratio between central and lateral spans play a crucial role than lead to a wide range of possibilities. To organize the discussion two main groups of solutions is discussed. First the alternatives where additional cables are used to enhance the rigidity of the system, second the solutions where the stiffness is reached by adapting the tower height’s as well as the deck’s rigidity. 3.1 Solutions with extra cables The first group of solutions includes, the use of crossing cables in the main spans and the anchoring of the central tower by extra cables, those are the schemes of the Forth Crossing and the Ting Kau Bridge. The structural behavior of those kinds of solutions is briefly discussed in the following sub-chapter. 3.1.1 Cable crossing This concept was used in the new Forth Crossing; see Hussain N. et al. 2011. In that case the overlapping of the main spans’ stay cables creates a virtual truss system (see Fig. 2). When out of balance live load is applied to one main span the tower movement causes the stays to lift the opposite main span. Over the region of the crossing stays a decompression is developed in the stays connected to the far lateral tower which is in turn tied back to the anchor piers. This system proved effective for a three-tower arrangement although the system would not be appropriate for bridges with a large number of main spans, since the length of stay cable mobilized becomes ever greater and therefore the flexibility of the system increases.
Figure 2.
Queensferry Forth Crossing Bridge.
402
3.1.2 Stiffening by long cables In the case of the Ting Kau bridge, the stabilization of the central pylons is achieved via longitudinal cables between the pylon head and the crossing point of the superstructure with the neighboring pylons, see Bergermann R. & Schlaich M. 1996. In this specific case, the additional cables contributed to the stability of the central tower during the cantilever construction of the deck. That fact is crucial due to the location of the bridge in an area with frequents typhoons. 3.2 Solutions without extra-cables Mezcala Bridge in Mexico and the Mersey Gateway Bridge in UK are solutions where no extracables are used.
Figure 3. Ting Kau Bridge.
Figure 4.
Mezca Bridge (Mexico).
Figure 5.
Mersey Gateway Bridge: Illustrative Design.
403
Figure 6.
Mersey Gateway Bridge.
In the case of Mezcala bridge, the central pylon is higher than the flanking towers, therefore the rigidity of the system is controlled basically by the longitudinal stiffness of the central tower, see Revelo C., et al. 1994. The Mersey Gateway Bridge is other example of this approach, see Sanders P. et al. From the initial client design the required stiffness was reached by reducing the height of the central pylon and by using a stiff deck. The later was the consequence of the initial idea of having a double deck for a combined use as a motorway and railway infrastructure. 4 FOUR SPANS CABLE STAYED BRIDGES WITHOUT EXTRA-CABLES 4.1 General approach As aforementioned, the problem of a multiple-span cable stayed bridge is the vertical flexibility of the system due to the action of out of balance live loads. In the case of no using extra-cables to stiffen the system, the deck and/or the central pylon should provide the required rigidity. There is no a specific rule for SLS conditions regarding the vertical deformation for the unbalanced live loads. It is a standerd practice to establish a limit based on the span/deformation ratio. For instance a normal limit is δ/L < 1:400 where δ is the deflection in load combination SLS characteristic and L the span. For single level bridges, the required stiffness could be achieved by the combination of three factors: first by using a deck with a higher stiffness compared with the deck of a classic three spans cable stayed bridge, secondly by having a stiff central pylon and third by connecting rigidly the deck to the central pylon. 4.2 Influence of deck and pylons stiffness The example shown in Figure 6 corresponds to the Mersey Bridge. In this bridge, the central pylon is connected rigidly to the deck. This is a natural way of increasing the stiffness of the bridge. Therefore the central tower attracts bending moments which could result in a huge foundation especially if the ratio moment to axil force. In that project the position of the pylons where almost fixed due to environmental conditions (see Fig. 5). The lateral span had to be in the range of 200 m which means a ratio lateral span to central span of 0.66. Figures 7 and 8 shows the structural behavior of the bridge in two situations: First when live load is applied to one of the main spans. Second, when live load is applied in one of the lateral spans. In both cases, the effect of the fixation to the central tower shows a significant benefit to control the deformation of the system. In order to evaluate the influence of the deck and central pylon stiffness, in the rigidity of the bridge against unbalance live load a parametric study was carried out. Figure 9 shows the main result of the study. There, the maximum displacement in one of the main spans due to the action of unbalance live load is plotted against the relative stiffness of the deck and central pylon. The continuous line shows the deflection when varying the relative deck stiffness from a maximum 1.00 which corresponds to the deck used in the project (4.50 m depth), to decreasing relative deck stiffness. 404
Figure 7.
Effect of the live load acting on the main span: loads, deformation and bending moments.
Figure 8.
Effect of the live load acting on the lateral span: loads, deformation and bending moments.
Figure 9 shows the relative small influence of the central pylon stiffness in the vertical deformation of the deck for the out of balance live load acting in one central span. 4.3 Influence of span’s ratio and cable disposition In order to take a wide perspective of the possibilities of this kind of solutions, a parametric study was carried for a bridge 1000 m long. Figure 10, shows the range of solutions analyzed. Main spans lengths vary from 300 to 375 m, and central pylons cables stayed from 25% to 75% of the main span (L1). There are several variables that could be considered as objectives. For instance, stays quantities or maximum bending moment in the central foundation could be establish as main parameters to define what is the optimal solution. Furthermore total cost, including constructability, is in every project the key variable. Nevertheless the cost is normally controlled by local conditions. The study was carried out considering a deck at 30 m over the level of the tower foundations, and a rigid connection between deck and central pylon. Deck and tower dimensions where similar to the ones used in the reference bridge (Fig. 9). 405
Figure 9.
Figure 10.
Parametric study influence of deck and central pylon stiffness (Mersey Gateway Bridge).
Parametric study 1000 m bridge’s length: solutions analyzed.
406
Figure 11.
Parametric study 1000 m bridge’s length.
Figure 12.
Four spans cable stayed bridges without extra cables.
As a summary of the different variables analysed, Figure 11 shows the vertical deformation due to unbalanced load plot against the percentage of the main span stayed form the central tower, which has a direct relationship with the height of the central tower. Considering a limit of the vertical deformation of L/400, Figure 11 shows the sensibility of the parameter to the ratio lateral span to central span. In all the cases the percentage of area stayed from the central tower should be less than 40–45%. That means that for this typology the central tower should be shorter than the flanking towers.
5 AESTHETICAL APPRAISAL OF CONTINUOUS FOUR SPAN’S CABLE STAYED BRIDGES WITHOUT EXTRA CABLES It is always difficult to judge or to establish general rules to assess the aesthetics of a general structural solution. In all the cases those rules have to be assed against the particular conditions of the site. From the point of view of the author, the solution without extra cables has a cleaner appearance than the ones with additional cables, especially those with cables that connect pylons to fixed points. When no extra cables are used, the central pylon should be equal or shorter than the flanking towers, to have a more visually pleasant structure. In those cases the use of three towers with similar heights would be preferred. Nevertheless, to achieve the required structural rigidity the ratio lateral to central span length should be in the order to 45% to have towers with a alike height. 407
6 CONCLUSIONS Four spans continuous cable stayed bridges has a wide range of possibilities. Solutions without extra cable are real and competitive solutions for three towers bridges. For instance in the particular case of a 1000 m crossing a disposition of spans with a ratio lateral/central span length of 0,45, without extra cables and similar tower heights could be the optimal solution in light of the parametric analysis carried out. Specific conditions play a crucial role when establish the cable and pylon arrangement. Nevertheless when it is a certain freedom in the position of central and flanking towers parametric study is highly recommendable during the conceptual phase to assess the optimal disposition. REFERENCES Virloguex M. 2001. Bridges with Multiple Cable-Stayed Spans. Structural Engineering International IABSE. Hussain N., Carter M., Kite S., Minto B. 2011. Forth Replacement Crossing-Concept Design. IABSE-IASS Symposium London Bergermann R. & Schlaich M. 1996. Ting Kau Bridge Hong Kong. Structural Engineering International IABSE. Revelo C., Alvarez C., Chauvin A., Armijo M., Paulik L., Rocha G., Arriola J. 1994. The cable-stayed bridges of the Mexico-Acapulco highway (Mezcala bridge, El Canon bridge, El Zapote bridge, Quetzalapa bridge) IABSE Conference Deauville. Sanders P., Lundorf L., Wood H., Romo J., Brennan G., Banks J. 2014. Design and Construction of the Mersey Gateway Bridge IABSE Symposium Madrid.
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Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
Particular design features for a long span cable-stayed bridge over the Harbour of Port Louis, Mauritius J. Jungwirth, J. Casper & Baumhauer SSF Ingenieure AG, Munich, Germany
ABSTRACT: Specific requirements from the contractor as well as local conditions have demanded very particular design features for the Port Luis Harbour Bridge in Mauritius. The conceptual design of the bridge has been carried out in the context of a competition for a large PPP traffic infrastructure project. The special requirements are described and the corresponding design solutions are presented. Main features are the design as a long span concrete bridge, the integral approach without joints and bearings, steel inserts in the pylon, the integral foundation as well as a pylon top restaurant.
1 INTRODUCTION In the context of a large PPP traffic infrastructure project in the area of Port Louis on the isle of Mauritius a bridge over the Harbour basin of Port Louis has been designed. The bridge connects the existing road network on the north side of the harbour to the existing road network on the south side and acts thus as a bypass for the city. The bridge features a free main span of 500 m spanning from shore to shore. For the new Harbour Bridge a cable-stayed bridge with a single-layer cable harp has been chosen as structural system. The pylon is diamond shaped. The girders of the approach bridges, as well as the main span, are single section reinforced and post-tensioned concrete box girder. The approach bridges are designed as continuous girder.
Figure 1.
Rendering of the bridge.
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As the owner intends the bridge to be valued as a landmark and tourist attraction, a restaurant has been designed on top of one of the pylons. In cooperation with an architect this additional element has been developed. After initial scepticism adding an additional element to the elegant bridge the whole team was finally convinced of the approach and also structural analysis showed that the restaurant on top of a pylon ‘fits’ pretty well to a cable-stayed bridge. Due to special requirements from the contractor as well as local conditions there are several particular design features that have been implemented in the bridge. This is namely the design as large span concrete bridge, the integral approach without joints and bearings, steel inserts in the pylon, the integral foundation as well as the aforementioned pylon top restaurant. 2 DESIGN REQUIREMENTS 2.1 Road cross-section The requirements for number and width of traffic lanes of the road cross section are given by the local road authority RDA. Both directional carriageways are situated on one bridge girder. There is a safety lane for both directions, but no footpath. The two carriageways are separated by a standard concrete median barrier and concrete parapets are provided to both sides. The widths of the lanes are conforming with RDA requirements as well as with the South African codes for traffic design: – Traffic lane. 3.75 m – Safety lane and outer margin to parapet. 1.5 m – Inner margin to parapet: 1 m Due to the specific structural design of the bridge there are two different arrangements of the directional lanes. The standard section has a spacing of 1 m between the two carriageways. The section where there is a cable features a width of 2 m to allow the anchorage of the cable and to provide a safety margin for potential impact on the cable from traffic. 2.2 Architectural design The analysis of the site has shown that site and technical requirements for the new Harbour Bridge are pretty complex. There is a large variety of existing zones interacting with each other; urban quarters, industrial districts, maritime environment, sea and mountains. Thus, the main architectural approach is to create a link between these domains with a retaining and sober structure (see Figure 2). The bridge structure acts as a link between: – north and south rim – sea and town
Figure 2. A non-homogeneous, multi-aspect environment for the new Harbour Bridge.
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– scenic landscape and modern industry – modern urban style and maritime environment – flat land and high mountains The structure should thus appear as light and transparent as possible in order to avoid any separating effect between Port Louis and the open sea. It should be a self-retaining functional structure that will not dominate the environment by avoiding that an additional strange element is added to an already quite in-homogeneous environment. However, the bridge should be representative and act as a landmark. 2.3 Arrangement of the bridge The bridge (total length 1584 m) is subdivided into the main bridge (1020 m) and a north (398 m) and south (166 m) approach bridge. The main bridge has a 490 m long main span and two back spans (2 × 265 m) on both sides. The main bridge and the approach bridges are separated by expansion joints. The approach bridges are inclined at a gradient of up to 7.5% and act thus as ramps to reach the height of the main bridge crossing the navigation channel of 50 m height. The bridge is designed to appear as one bridge even though, from a structural point of view it is subdivided in separate elements. 3 LARGE SPAN CONCRETE BRIDGE Due to the specific characteristics of the bridge structure and the lack of steel construction contractors on the isle of Mauritius as well as requirements from the PPP contractor, the bridge deck has been designed as a light concrete box girder. This choice leads to a robust and durable bridge girder. Compared to steel or steel composite girders commonly applied in mid-span of cable-stayed bridges, the concrete box girder does not entail corrosion problems in maritime areas. A steel girder has a very high maintenance for regular renewal of corrosion protection. The concrete box girder is heavier than typical steel or composite cross sections. Thus the bridge system is subject to higher loads by its dead weight, but this presents a considerable advantage in view of the structure’s sensitivity to wind. Mauritius is situated in a cyclone area. Light bridge decks could cause vibrations induced by wind. A heavy and stiff cross section such as used in this project counteracts this effect. The section of the main span is a 22.95 m wide concrete box girder optimized for cable-stayed bridges (see Figure 4). The thickness of the web, deck and bottom is reduced to a minimum. Internal steel struts distribute loads from the outside to the central suspension point. The concrete box girder distributes torsional forces in longitudinal direction to the pylons. This smart combination of materials leads to a very light and efficient structure for a concrete cross section. The standard section of the back span of the cable-stayed bridge is designed as standard concrete box girder (see Figure 5). There is both longitudinal and transversal post-tensioning. In the centre between the lanes there is space of 1m width to allow anchorage of the cables. There is a specific configuration of cross girders in the back span to enable transfer of loads from both sides to the centre where the cable is attached at the bottom of the girder.
Figure 3. Arrangement of the bridge.
411
Figure 4.
Standard section of the main span of the cable-stayed bridge.
Figure 5.
Standard section of the back span of the cable-stayed bridges.
The approach bridges are designed as standard concrete box girders (see Figure 6). The bridge deck has a total width of 21.95 m. The thickness of the edge cantilevers reduces from web to parapet to reduce the dead load and the webs are inclined to optimise the box girder. There is both longitudinal and transversal post-tensioning.
4 INTEGRAL BRIDGE STRUCTURE The main bridge girder is connected rigidly to the pylon, resulting in a very robust and durable integral framing structure. Time-dependent constraint deformations, such as temperature or creep and shrinkage, can be distributed by deformations of the 50 m long pylon legs. This structural type eliminates very difficult and expensive, maintenance-intensive bearings between bridge girder and pylon. The chosen solution is quasi maintenance free and highly sustainable leading to low live cycle costs. By being rigidly connected the two pylons act as horizontal fix points for the girder. Thus, the virtual fix point of the bridge is situated in the middle of the main span. The first piers on both sides are also directly connected to the bridge. All other piers are longitudinally free of deformation and restrained in transversal direction to use all piers during transversely load bearing. 412
Figure 6.
Standard section of the north and south approach bridges.
Figure 7.
Structural model of the integral bridge structure using SOFiSTiK.
5 CONCRETE PYLON WITH STEEL INSERTS FOR CABLE ANCHORAGE The Pylon is designed in diamond shape, offering a very good combination of high stiffness in the transverse direction and a small contact surface at the pylon foot compared to an A-shaped pylon. The pylon is divided in 3 sections: the two-piece pylon base (pylon legs), the two-piece centre part and the pylon top in which the cables are anchored. In the bend between base and centre part, a horizontal beam is designed at height of the carriageway. At this beam the carriageway girder of the bridge is connected rigidly to the pylon (see Figure 8). In the cross beam, loads from the upper part of the pylon are deviated and transferred to the lower part. In addition horizontal and vertical forces from the bridge girder are transferred via the beam to the pylon. In order to take the deviation forces in this highly loaded element, the cross beam is post-tensioned. In the upper part of the pylon, cables are anchored. A modular system made of prefabricated steel components is developed. These steel components consist of two bulkheads and anchor plates at both ends. On both sides the necessary infrastructure for further development is already integrated into the components (ladders, platforms, safety devices). Horizontal forces of opposite anchoring heads are coupled by two bulkheads. Vertical forces are distributed into the concrete by shear studs on both sides. To erect the pylon, the components are placed on the already concreted part and cast with the next concreting section. 413
Figure 8.
Diamond shaped pylon and cross section of pylon tip with inserts for cable anchorage.
6 POST-TENSIONING In addition to the cables, post-tensioning is required locally. Post-tensioning is provided in a straight line in the bottom and the top cord. Zones where post-tensioning is required are: – – – –
In the middle of the main span to compensate tension due to the horizontal forces of the cables. Next to the pylon, as there is no cable support and constraint stresses have to be considered At the very end of the main bridge where there is no cable support. At the piers’ location to avoid local stress concentration.
Furthermore post-tensioning is required for the free cantilevering along the whole main span. 7 INTEGRAL FOUNDATION The foundation of the pylons is situated in direct proximity to the shore line and in consequence groundwater level is just slightly below ground level. Therefore an integral foundation is implemented combining the temporary construction pit with a traditional pile foundation. In this way both elements contribute permanently at load bearing in final stage. In the present case of the pylon foundation, a watertight, intersecting bored pile wall is built, implemented as two intersecting circles. The wall supports itself due to vault effect. Stiffeners and back anchors are not required. The stiffening of the intersecting edge between the circles is also provided by a bored pile wall. In addition to the piles of the pile wall, traditional bored piles are planned for load distribution (see Figure 9). Within the bored pile wall, the construction pit can be excavated. To seal the pit lying directly by the sea, an underwater concrete bottom slab is cast. At level of the bottom slab the inner reinforcement for the pile wall is uncovered and connecting reinforcement is welded. The traditional piles on the inside are broken back and prepared for being cast into the pile cap. The pile caps are concreted and form a monolithic foundation in which the internal piles and the piles of the bored pile wall contribute together for load bearing of the pylons. 414
Figure 9.
Figure 10.
Pile arrangement and cross section of pylon foundation.
Selected construction stages.
8 CALCULATION PROCESS In the structural analysis are considered the final stage as well as all construction stages. This enables holistic assessment of pretension of the cables. The deformed geometry of the superstructure is taken into consideration precisely. For the calculation of internal forces the construction process of the whole structure is taken into consideration. The superstructure, the pylons and piles are modelled as beams. The stiffness of the superstructure is condensed to one beam. The bending and torsional inertia moments are exactly defined. Calculations are carried out using SOFiSTiK calculation software. The so called ‘form finding’ is the process where self-weight is introduced in order to determine the initial cable pre-tensioning forces to achieve equilibrium of forces for the non-deformed system at the “target geometry”. The geometry of the bridge deck structure is set using an iterative process during the form finding process. Pre-tensioning forces of the cables are chosen in a way that an optimal course of moments and pylon deformations is achieved at the moment of start of service. In the present case of a concrete cross section for the mid span section cable spacing has to be chosen rather small. This leads to a quasi continuously bedded beam, which in context of a rather stiff concrete girder leads to high sensitivity in the iteration process. Very little modification of strain in the cables results in high increment of stress. The standard equation solver of SOFiSTiK lead in this case to instability. Thus iteration had to be carried out partly manually. In the structural analysis, the whole construction process in accordance with the construction schedule is taken into consideration. Using the construction stage manager (CSM) from SOFiSTiK, construction stages are modelled. For each construction stage deformation and internal forces are calculated and taken from one to the next calculation step in order to obtain the induced constraint stresses and strain in final stage. 415
Figure 11.
Bridge girder bending moment. Envelop of the bending moment: max (left) and min (right).
Figure 12.
Rendering of the bridge with Sky Diamond restaurant.
Internal forces for dimensioning at ultimate limit state include the stresses from the construction stages and result from multiplying characteristic internal forces with the partial safety coefficients. All structural elements are dimensioned using the programme AQB from SOFiSTiK. 9 PYLON TOP RESTAURANT As the owner intends the bridge to be valued as a landmark and tourist attraction a restaurant has been designed on top of one of the pylons. In cooperation with an architect this additional element has been added. Located directly in front of the Port Louis city where Caudan Waterfront meets the harbour area, the Mauritius Harbour Bridge and the so called Sky Diamond restaurant will be visible to ships and yachts anchored in the harbour, air traffic as well as motorists and pedestrians in the city, making it a significant landmark for Port Louis and greater Mauritius. After initial scepticism adding an additional element to the elegant bridge the whole team was finally convinced of the approach. Also structural analysis showed that the restaurant on top of a pylon ‘fits’ pretty well to the cable-stayed bridge. Due to the very high normal force in the pylons caused by the deviation of the cables and the stabilising effect of the cable harp in one direction and the diamond shaped pylon in the other, there is quasi no influence on dimensioning from the loads of the restaurant.
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Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
A study on vehicular live load design based on actual vehicular load for a multi-span large cable-stayed bridge H. Sugiyama & H. Kanaji Hanshin Expressway Company Limited, Osaka, Japan
H. Watanabe & O. Aketa Sogo Engineering Incorporated, Osaka, Japan
ABSTRACT: In this paper, a study on the design live load based on actual possible traffic conditions and its application for a multi-span cable-stayed bridge is described. Monte Carlo simulations using vehicle models based on the vehicular load survey are conducted to obtain the maximum responses at the girder, the cable, and the tower of a multi-span cable-stayed bridge. The probabilistic distribution of maximum responses in the 100-year service period obtained from the simulations indicates that the probability of responses that exceed the design values is very small, and the distribution range of maximum responses is also very small. The live load factor of limit state design can be assigned as approximately 1.0, which assures that the bridge is sufficiently safe and significantly reduces the cross-sectional force at the lower part of the middle towers. In addition, since the tail of the vehicle-weight distribution has a large uncertainty, the small influence of the tail on the cable-stayed bridge improves the robustness of the design; this can be a factor in differentiating partial factors for small and medium bridges.
1 INTRODUCTION Recently, some multi-span large cable-stayed bridges have been constructed in the world. It is well accepted that the middle tower of such bridges is more flexible than the edge tower because of lack of anchor piers. Therefore, girder deformation at the middle spans of the bridge along with the bending moment at the lower part of the middle towers become very large under a longitudinally sectioned live load, thereby dominating the design of the section members. To address this issue, the adoption of longitudinally rigid middle towers or advanced cable systems may be considered (Hussain, et al. 2011). In current designs, the maximum vehicular live load intensity is placed only on the most unfavorable area of the influence line to produce a maximum response in the member being designed. However, when the way of loading is applied to the influence line of a multi-span large cable-stayed bridge, full loading and non-loading conditions are repeated every several hundred meters on the bridge. This loading condition is too unusual and too conservative for the design; thus, it is valuable to estimate the live load on the basis of the actual possible traffic conditions, which makes the design more rational and economical. Saitou et al. proposed more rational design live load for 4-spans continuous suspension bridge based on live load simulation (Saitou, et al. 1999). In this study, Monte Carlo (MC) simulations for a cable-stayed bridge are conducted using vehicle models and traffic congestion conditions based on the expressway vehicular load survey to obtain the maximum responses at the girder and towers in a 100-year service period. Then, to validate the obtained maximum response properties and uncertainties, the effects of the live load model parameters on the maximum response are investigated, and the results are compared with those for a plate girder short-span bridge. In addition, from the probability distribution obtained by MC simulation, the partial factors of the limit state design (LSD) method based on reliability 417
Figure 1.
Probabilistic model of vehicular live load.
Figure 2. Vehicle types.
Figure 3. Heavy vehicle ratio from the 2010 vehicle axle load survey.
evaluation are estimated using the first-order reliability method (FORM) and design value method (DVM) (ISO 2394.1998). Finally, the results of this study are applied to the design of a multi-span cable-stayed bridge. 2 PROBABILISTIC MODEL OF VEHICULAR LIVE LOAD The probabilistic model of the vehicular live load on the Hanshin Expressway, which is located in Osaka Bay Area, is developed from the characteristics of the vehicle axle load, the vehicle length, the distance between vehicles, and the traffic congestion (Fig. 1). 2.1 Classification of vehicle types The vehicles on the Hanshin Expressway are classified into five types: small-size; middle-size; large-size with two axles; large-size with tandem wheel; and trailer (Fig. 2). In addition, the weights of the cargo vehicles (large-size with two axles, large-size with tandem wheel, and trailer): can be categorized into three parts: light loaded; heavy loaded, and overloaded. Vehicle types are classified into 11 categories, as shown in Table 1. The distribution percentages of the 11 vehicle types depend 418
Table 1. Vehicle type classifications, probabilistic model parameters, and distribution percentages. Probabilistic model parameter of vehicle weight (10 kN)
Percent distribution
Vehicle type
#
Load state
number of axles
Type
Mean
S.D.
Shift
L.L.
U.L.
40%∗
60%∗
Small-size car Middle-size car Large-size with two axles
1 2 3 4 5 6 7 8 9 10 11
– – Light load Heavy load Overloaded Light load Heavy load Overloaded Light load Heavy load Overloaded
1 1 2 2 2 2 2 2 3 3 3
LOG LOG NOR LOG EXP NOR LOG EXP LOG LOG EXP
1.4 6.1 8.3 14.3 16.5 12.1 21.4 26.8 17.4 35.8 48.7
0.55 2.28 2.56 1.81 1.00 1.99 6.48 2.80 4.50 7.00 5.70
0.5 – – – 15.5 – 2.3 24.0 – – 43.0
0.5 0.9 3.5 8.3 15.5 5.8 5.8 24.0 5.2 17.4 43.0
8.0 18.5 15.5 15.5 25.0 24.0 24.0 55.0 35.8 43.0 90.0
45.00 15.00 1.69 1.91 0.40 7.21 12.41 2.38 8.23 4.87 0.90
16.00 24.00 1.27 1.43 0.30 13.77 23.69 4.54 8.82 5.22 0.96
Large-size with tandem wheels Trailer
NOR: Normal distribution LOG: Log-normal distribution EXP: Exponential distribution
Mean: Mean value S.D.: Standard deviation Shift: Offset value of origin
Figure 4. Vehicle weight relative frequency for trailers based on the 2010 survey data.
L.L.: Lower limit value U.L.: Upper limit value ∗ Heavy vehicle ratio
Figure 5. Correlation between vehicle weight and vehicle axle load for trailers based on the 2010 survey data.
on the time. The heavy vehicle ratio is used as the main parameter to represent the distribution of vehicle types. The heavy vehicle is (c),(d), and (e) in Figure 2. On the basis of the data of the 2010 vehicle axle load survey, the heavy vehicle ratio during early hours (0–6 h) is assumed to be 60%, while that during daytime (6–24 h) is assumed to be 40% (Fig. 3). The distribution percentages of the 11 vehicle types are shown in Table 1. 2.2 Model of vehicle axle load The probabilistic model of vehicular live load is developed on the basis of the data from the 2010 vehicle axle load survey. Figure 4 shows the vehicle weight histogram for trailers. The vehicle weights of trailers under light-loaded and heavy-loaded conditions are modeled by log-normal probability distributions, whereas the weight of an overloaded trailer is modeled by an exponential distribution. The parameters of the probabilistic model of the 11 vehicle types are shown in Table 1. The vehicle axle load is calculated from the vehicle weight via the correlation function between them (Fig. 5). The number of axles for the small-size and middle-size cars is assumed to be one (Fig. 2) as their weights are very light and not very influential. 419
2.3 Model of vehicle length and distance between vehicles The probabilistic model of the vehicle length and the distance between vehicles is assumed from the 2010 survey data. 2.4 Model of traffic congestion There are two types of traffic congestion on urban highways: regular traffic congestion that happens during rush hours and accidental traffic congestion caused by a traffic accident or construction work. Accidental traffic congestion also includes congestion resulting from the most unfavorable section of the influence line of the target member. Although the probability of this type of congestion occurring is very small, the response is very large. In this study, the cases of 22 traveling vehicles are considered on the basis of the combination of usual travel and the above two types of traffic congestion. Figure 6 shows the 22 traveling vehicle cases. The frequency of each case is determined by the Hanshin Expressway traffic survey data. The heavy vehicle ratios are assumed to be 60% during the early hours of the day (0–6 h) and 40% during the daytime (6–24 h; Fig. 3). The congestion duration and vehicle velocity for each type of traffic congestion are also based on the Hanshin Expressway traffic survey data. 3 TARGET STRUCTURE The target structures are a multi-span cable-stayed bridge and a typical cable-stayed bridge (Fig. 7). The influence line of the designated structural member force is calculated from the structural analysis. The length of the unfavorable principal area target member is defined as the influence line reference length Linf , which is an index that represents the characteristics of the influence line. Figure 8 shows examples of the influence line and Linf . Figure 9 shows the load distribution coefficients in the transverse direction. 4 MONTE CARLO SIMULATION 4.1 Calculation of the response forces MC simulation is employed to estimate the probability distribution of the peak response forces. The flow of the simulation is shown in Figure 10. The response force is calculated by Equation (1) Simulation year is 2000 year.
where F = the response force, Wi = the vehicle axle load at point i, ηi = the influence line value at point i, nj = the number of vehicle axles in traffic lane j, and αj = the load distribution coefficient in the transverse direction in traffic lane j. Figure 11 shows an example of the changes in the vehicle live load at intervals of t. 4.2 Estimation of the probability distribution of the peak response forces The peak response forces obtained from MC simulation are applied to the Gumbel distribution or generalized extreme value distribution (GEV) to estimate the fractile values. The probability distributions of the maximum response in the 100-year service period are calculated for 22 congestion cases. The curve enveloping their maximum values is regarded as the probability distribution of the maximum response of the target member. Maximum response is 420
Figure 6. Traveling vehicle case.
Figure 7. Target structures and structural members for the calculation of response force.
Figure 8.
Examples of the influence line.
Figure 9. Load distribution coefficients in the transverse direction.
expressed by the ratio of the calculated response to that obtained by the conventional design live load. An example is shown in Figure 12. Mean, standard deviation, and fractile values of 5% and 1% of the maximum response are shown in Table 2. The means of the maximum responses of all members except for member C is in the range of 54%–77% of the conventional design value, 421
Figure 10. The flow of the Monte Carlo simulation. Figure 11.
Movement of the vehicles live load.
Table 2. Probability distribution parameters for the ratio of maximum response in the 100-year service period.
Figure 12. Probability distribution of the ratio of maximum response in the 100-year service period.
and the standard deviations are very small. Therefore, the probability that the 100-year maximum response of any of these members will exceed the conventional design value is extremely small. The response values of the members greatly differ because of the influence line shape and influence line reference length Linf . In the load combination of the live load, the overall safety factor is 1.7 in a conventional allowable stress design. The simulation results herein indicate that the 5% fractile value of the 100-year maximum response, which corresponds to the expected value of a return period of 2000 years, is less than the conventional design response. Therefore, conventional design is assumed to be extremely safe. 5 INFLUENCE ANALYSIS OF THE LIVE LOAD AND STRUCTURAL PROPERTIES 5.1 Live load model and bridge model to be used in the comparative analysis Influence analysis focusing on the live load characteristics and structural properties is conducted. The live load characteristics are focused on large vehicle weight characteristics. The vehicle weight distributions of a large-sized car with tandem wheels and trailers are obtained on the basis of vehicle axle load survey data, which was collected using the bridge weigh-in-motion method on general roads under heavy traffic. The live load model that is formed by replacing only the weight 422
Table 3. Parameters of the probabilistic model of vehicle weight.
Figure 13.
Vehicle type Road type
Mean S.D.
Max
Large-size with tandem wheel Trailer
17.1 18.2 +6% 25.1 26.3 +5%
55 100 +81%) 90 150 +67%)
Probability distribution of trailer weight. Figure 14.
Hanshin Expwy. General road (increase rate Hanshin Expwy. General road (increase rate
5.25 7.00 +33% 10.9 13.1 +20%
Plate girder bridge.
characteristics of these two vehicle classes is referred to as the general road model. The probability distribution parameters are shown in Table 3, and the probability distribution diagram for the trailer class is shown in Figure 13. The mean values differ by about 5%, but the tail of the probability distribution, which represents the range of the heaviest vehicles, is about 1.6-times larger in the general road model than that in the Hanshin expressway model. When the structural characteristics are focused on the small-span plate girder bridge, the effect of the live load is increased. Two girder bridge influence lines with different reference lengths and numbers of lanes are considered (Fig. 14) with a focus on the bending moment at the center of the span on the outer main girder.
5.2 Calculation and analysis of the response value based on Monte Carlo simulation In addition, two plate girder bridges are employed to calculate the response value according to the general road model using the same method described in the previous section. The changes in the 5% fractile value of the maximum response of the 100-year service period due to the live load model are shown in Table 4. The rate of increase differs greatly among target members. The relationship between the influence line reference length Linf and the rate of increase of the maximum response of each target member is shown in Figure 15. The rate of increase is related to Linf ; the slope of the rate of increase exhibits greater variation in the vicinity of Linf = 200 m. If Linf is short, as in the case of two girder bridges, only a few heavy vehicles are loaded. Therefore, the maximum response occurs when the top heavyweight vehicles, which correspond to the tail of the vehicle weight distribution, are loading. A rate of increase of 58% on I-25 generally corresponds to an increase of about 60% in the tail of the trailer weight distribution. When Linf is long, the rate of increase is about 7%, which is close to the average rate of increase in vehicle weight. 423
Table 4. Fractile value of two vehicular live loads. 5% Fractile value Bridge Type
Structural Linf Expwy. General Rate of member (m) model Rd. model increase
Typical A. Girder 107 cable-stayed B. Girder 74 bridge C. Cable 188 D. Tower 586 Multi-span E. Girder 312 cable-stayed F. Cable 870 bridge G. Tower 650 H. Tower 1041 Plate girder I-25 25 bridge I-40 40
0.78 0.61 0.99 0.77 0.60 0.78 0.81 0.84 0.85 1.46
0.95 0.84 1.11 0.82 0.65 0.83 0.86 0.91 1.37 2.29
+22% +38% +12% +6% +8% +7% +6% +8% +55% +57%
Figure 15. The rate of increase between the two vehicular live load models.
6 APPLICATION TO THE LIMIT STATE DESIGN 6.1 Design equation based on the limit state design On the basis of the probability distribution of the live load obtained in this study, partial factors of the LSD formula shown in Equation (2) are estimated using the FORM and DVM (ISO 2394.1998):
Here R( ) = the resistance model function, fk = the characteristic value of the material, γ m = the partial factor corresponding to the uncertainty in material, γ b = the partial factor corresponding to the uncertainty in the member resistance model, γ i = the partial factor corresponding to the consequence of failure of the structure, γ a = the partial factor corresponding to the uncertainty in the structural analysis, SD = the dead load effect, SL = the vehicular live load effect, γ D = the dead load factor, and γ L = the vehicular live load factor. In this study, the partial factors γ i , γ a , and γ D are assumed to be 1.0, 1.05, and 1.05, respectively. Other partial factors (γ m , γ b , and γ L ) are estimated using the FORM and DVM (ISO 2394.1998). The probability distributions of the material and resistance model function are estimated on the basis of survey data for existing materials and the results of component strength tests. 6.2 Calculation of partial factors based on reliability evaluation First, a target reliability index is defined as βT = 3.8, and the characteristic value of the live load is set as the value of the conventional design. The sensitivity factor and partial factor of the member B are shown in Table 5. Since the mean of the measured value is greater than the characteristic value, and the standard deviation is small, all partial factors are 1.0 or less. Computationally, it is possible to ensure βT = 3.8 even though a safety margin is not added. The live load factors γ L for all members are shown in Figure 16. Although βT is the same in all cases, the γ L values vary greatly with the structural and live load characteristics. The difference in the values of γ L calculated by the two live load models increases as the influence line reference length Linf becomes shorter, whereas the γ L value of the general road model increases significantly with decreasing Linf . For the tail of the vehicle weight distribution, which is one of the uncertainties in the live load model, the γ L value is sensitive to the tail when Linf is short, whereas it is insensitive when Linf exceeds 200 m. 424
Table 5. Partial factor based on the reliability of member B. Probability model
Partial factor Material Member Live load
γm γb γb γL
(Uncertainty)
Mean
S.D.
Sensitivity factor α
(Yield) (Buckling) (Thickness)
1.18 1.05 1.00 0.70
0.10 0.03 0.01 0.04
0.36 0.20 0.07 −0.91
Partial factor γ 0.94 0.97 1.00 0.95
Figure 16. The relationship between Linf and the partial factor of live load γL . Table 6. Partial factor based on tower reliability. Partial factor Material Member Live load
γm,c γm,s γb γL
(Uncertainty)
Partial factor
(Compression strength of concrete fck ) (Yield strength of steel bar σY ) (Bending strength calculation formula)
1.00 1.00 1.20 1.10
6.3 The effect of application to the tower member In a multi-span cable-stayed bridge, the middle tower is constructed by reinforced concrete and is designed to resist the live load force. A cross-section conforming to the conventional allowable stress design is required (Fig. 17a). In this study, the cross-section shown in Figure 17b is designed by LSD with the partial factors given in Table 6. The results confirm that the cross-section can be significantly smaller than that in a conventional design.
7 CONCLUSIONS The present investigation leads to the following conclusions. 1. The load factor of the live load can be made smaller than that in the current design in the LSD of a multi-span cable-stayed bridge. 2. For all studied cases of a cable-stayed bridge, the 5% fractile values of the probability distribution of the 100-year service period are below or equal to the conventional design values. 425
Figure 17.
Design cross section at the basement of the middle tower.
3. The comparison with the short-span plate girder bridge shows that the effect of the tail of the vehicle weight distribution is small if the influence line reference length of the cable-stayed bridge is long. 4. The vehicular live load factor of the cable-stayed bridge, which is calculated on the basis of the reliability evaluation, is less than approximately 1.0 when the influence line reference length is long. 5. The tail of the vehicle weight distribution has a large uncertainty, and this has a small effect on the maximum response and results in the improvement of the robustness of the design. This can be a reason that a live load factor for the long-span bridge can be differentiated from that for small- and medium-sized bridges. ACKNOWLEDGMENTS The authors are grateful to Prof. Nagai, M., Prof. Okui, Y., Prof. Yamaguchi, T., Prof. Ohyama, O., Prof. Miki, T., for many useful advising. REFERENCES Hussain, N., Carte, M., Kite S. and Minto, B. 2011. Forth Replacement Crossing – Concept Design, IABSE Symposium London 2011. Saitou, N., Murakoshi, J., Nishikawa, K., 1999. A study on Design Live Load of 4-spans Continuous Suspension Bridge based on Live Load Simulation, Journal of Structural Engineering Vol. 45A ISO 2394.1998. General principles on reliability of structures. 1998.
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Comparison of variants for New Peljesac Bridge in Croatia J. Radic, Z. Savor & M. Srbic Zagreb University, Zagreb, Croatia
M. Pipenbaher Ponting Consulting Engineers, Maribor, Slovenia
ABSTRACT: The southern part of Croatia, including the city of Dubrovnik, is currently separated from the rest of Croatia by a small coastal stretch belonging to the state of Bosnia and Herzegovina. A fixed link between all parts of Croatia will be established after completion of the MainlandPeljesac Peninsula Bridge. Construction of the bridge started in 2007, but it was slowed down and finally abandoned in 2012 due to lack of funds. The client recently requested new economically more viable solutions. Two alternative bridge solutions were proposed in the new preliminary design, a continuous steel beam bridge and a multi-span extrados semi-integral bridge with hybrid deck. Finally, the multi-span extradosed semi-integral bridge with hybrid deck was chosen for further design. Both preliminary designs will be described in the paper. Keywords: box girder bridge, multi-span extradosed semi-integral bridge, steel deck, hybrid deck, deep foundations, cable stayed steel bridge.
1 INTRODUCTION The Mainland-Peljesac Peninsula Bridge is the largest civil engineering structure to be built in Croatia in the near future. A fixed link between all parts of Croatian territory will be established after completion of the Mainland-Peljesac Peninsula Bridge over a navigable sea strait, with minimum required navigation clearance of 200 × 55 m (Figure 1). This is a specific request of the neighboring country of Bosnia and Herzegovina, although the Mali Ston Bay is ecologically an extremely vulnerable area, where any larger ship traffic might
Figure 1.
Location of Peljesac Bridge.
427
Figure 2.
Continuous steel box girder bridge – longitudinal layout and ground plan.
disturb the delicate environmental balance of one of the European natural habits of oyster mussel and clam farms, and although there is no navigation channel currently chartered in this area. The bridge site is exposed to strong winds with maximum average 10-minute wind speeds of 33.4 m/s and wind gust speeds of 47.1 m/s. The bridge site also lies in the highly active seismic zone with ground acceleration ag = 0.41 g and extremely adverse foundation soil conditions. The seabed at the bridge alignment is almost level at −27 m elevation, with the stratigraphic pattern of a series of sub-horizontal layers and irregular top of the rock along the bridge. The foundation soil alongside the planned location is extremely poor, as has been confirmed by detailed geotechnical procedures and by carrying out extensive off-shore investigations, including continuous shear wave survey and geotechnical explorations from a specially equipped drill ship, drilling boreholes and taking out samples in 2004 and 2009. The distance over the obstacle amounts to approximately 2.140 m at the sea level, while the distance at the grade line elevation is 2380 m. The total length of the bridge is 2440 m. According to the Investor’s Terms of Reference, the roadway in each direction is comprised of 3.5 m wide traffic lanes, 2.5 m wide stopping lanes, and 0.5 wide marginal strips on both sides. In his additional request, the Investor required that the traffic lanes are divided in the bridge axis by safety barrier which increased the total bridge width. The construction of the bridge was already in progress when the project was abandoned due to lack of funds. The client recently requested new economically more viable bridge solutions in order to apply for EU funding, with Croatia now an EU member state. Two alternative solutions were proposed, a continuous steel box girder bridge and a multi-span semi-integral extradosed bridge with hybrid (predominantly steel) deck, both fully respecting the original road alignment and the already constructed bridge parts. 2 CONTINUOUS STEEL BOX GIRDER BRIDGE The bridge deck is continuous over 16 spans with the overall length of L = 72.0 + 96.0 + 132.0 + 3 × 164.0 + 200.0 + 256.0 + 200.0 + 4 × 164.0 + 132.0 + 96.0 + 72.0 = 2404.0 m (Figure 2). The deck structure is a continuous trapezoidal single steel box girder with constant depth d = 6.4 m, except in the main navigation span and the two neighboring spans, where the box girder depth increases parabolically to the maximum vertical webs projection of 10.6 m above piers S8 and S9. Webs are inclined at 78.7◦ . The cross section of the deck is symmetrical with double-pitched transverse roadway slope of 2.5%. The deck plate is cantilevered out symmetrical from both webs. Web spacing at the top (connection to the deck plate) amounts to 11.0 m and at 428
Figure 3. (right).
Box girder bridge – deck cross section above main span piers (left) typical deck cross-section
the bottom (connection to bottom chord) 8.44 m for the standard cross section with the web height of 6.4 m. The total steel weight of the bridge deck is 25,500.0 t. Vertical actions of the steel bridge deck are carried over to the bridge substructure by steel spherical structural bearings. Horizontal actions of the steel bridge deck in the longitudinal direction are taken over by a combination of viscous shock-transmitters and fixed bearings. There are 15 piers in total (S2–S16). The pier S2 is on land, the pier S16 is at the border between land and sea and all others piers are in the sea. All piers are of box cross section with variable dimensions. The end at the top has specially formed capital. Abutments at bridge ends are massive with parallel wings. Pier S2, S16 and part of the column S3 have been constructed in the first attempt of construction. Pier S2 on land has been constructed on shallow foundations on sound rock with foundation plate 4.0 m deep and 15.0 × 20.0 m in layout. Pier S3 is founded on 12 in sea vertical bored piles 1.5 m, about 15 m deep. Total pile lengths vary from 25 m to 27 m. The pile cap layout plan is 12.5 × 16.5 m with the depth of 3.5 m. Pier S16 is founded on 12 bored piles which penetrate 16 m into sound rock. The pile layout plan is 10.6 × 16.8 m with the depth of 3.0 m. Driven steel piles 2.0 m are utilized for foundations of 12 remaining piers S4–S15. A combination of vertical and battered piles in slight inclination of 5% has been selected. Battered piles significantly reduce horizontal movements of the pile caps and the whole bridge structure and take up an important part of horizontal actions due to wind and earthquake by their axial resistance. Piers S4 & S12–S15 are founded on 8 piles connected by a pile cap 11.0 × 18.0 m in plan, piers S5–S7 and S10–S11 on 10 piles with a pile cap 12.0 × 18.0 m in plan and piers S8–S9, which support the main navigation span, are founded on 16 piles connected to a pile cap 18.0 × 18.0 m in plan (Figure 4). The corrosion protection of steel piles is three-fold: additional wall thickness, passive cathodic protection with welded sacrificing anodes and special paint coatings. 3 MULTI-SPAN SEMI-INTEGRAL EXTRADOSED BRIDGE WITH HYBRID DECK Based on the principle of comprehensive optimization method, authors systematically developed various bridge structure alternatives, ranging from the smallest possible span (120 m) to the longest 429
Figure 4.
Pier S8 – cross section.
span (400 m). Changes of crucial criteria within the set of structural, technological, shape-related and economic parameters were analysed at the stage of an optimum bridge concept development. Authors tried to create a structurally, technically and technologically up-to-date, and financially optimal & competitive bridge, the price of which practically remains in the framework of a smaller span girder bridge, for which a greater number of pier positions significantly increases the construction price demanding deep foundations in the sea, on steel piles sometimes exceeding 120 m in length. The bridge has been conceived as inventive structure. The central system is a multi-span semiintegral bridge with six low pylons and five 285.0 m openings, so that a full symmetry in space has been achieved (Figure 5). In landscape, the bridge appears as a very light and peaceful composition. The integrally conceived bridge structure, with a hybrid superstructure, ensures seismic stability of the bridge without installation of big bearings and seismic dampers. Bearings and guiders installed on shear keys are planned only at end parts of the bridge – at abutments and piers 2-4 and 11-13. The extradosed cable-stayed deck, and 33 m high centrally placed reinforced-concrete pylons, are elastically restrained to piers so that in its central part measuring 1832 m (76% of the total bridge length) the bridge is a frame structure without bearings, which provides for an additional stability of the bridge in case of seismic action and wind gusts. The bridge deck is a continuous hybrid box structure that is suspended in its central part by stays on to six centrally placed reinforced concrete pylons (multi-span cable-stayed bridge). The steel superstructure 19.50 m in width is a three-cell box with overhangs (Figure 6). The total width of the superstructure with the wind screen amounts to 22.40 m. 430
Figure 5.
Multi-span semi-integral extradosed bridge – longitudinal layout and ground plan.
Figure 6. (right).
Multi-span semi-integral extradosed bridge – deck cross-sections; not suspended (left), suspended
Pylons are 33.0 m in height and so the pylon height to span ratio is 33.0/285.0 = 11.60. Hence the bridge can be classified as being somewhere at the limit between cable-stayed and extrados bridges. The continuous box superstructure is characterized by the systemic length of 2404.0 m which is divided as follows: 72.0 + 96.0 + 118.0 + 203.5 + 5 × 285.0 + 203.5 + 96.0 + 72.0 = 2404 m. Pylons are made of solid concrete and they measure 2.20 × 5.0 m at the top, and 2.20 × 6.2 m at the superstructure level. The high strength concrete type C60/75 shall be used for the realization of pylons. Stay-cables are basic structural and load bearing elements of the cable-stayed bridge. The superstructure is supported with stays arranged in a single plane, spaced at 12.0 m intervals. Stay-cable length ranges from the smallest 2 × 27.5 = 55 m to the largest 2 × 135 = 270 m. Each stay consists of at least 75 and of no more than 109 strands. Bridge piers 3–12, located in the sea are founded on driven steel piles 2000 mm in diameter, 55–125 m in length. At the sea level the piles are fixed to the concrete pile cap and, in this way, an appropriate load bearing capacity and horizontal stiffness of foundations will be ensured. Pile cap is 5.0 m thick with layout dimensions of 22.0 × 24.0 m. Pile caps also protect piers against direct vessel impact in case of vessel collisions. Relatively low piers S2-S4 and S11-S13 (measuring 18.5–31.33 m in height) are of box section and their external measures are constant, which greatly facilitates their realization. The piers are of octagonal cross section and they measure 4.0 m in longitudinal direction, and 10.0 m in transverse direction. The cross-sectional wall thickness is constant. In transverse direction, the walls are 0.50 m thick, and in the longitudinal direction (along the bridge length) 0.60 m thick. 431
Figure 7.
Pylons: Longitudinal section (left), Cross section (right).
Piers S5–S10, representing the bottom part of pylons, are 37.88–53.30 m high (Figure 7). The piers are of box cross-section and their external measures are constant, which greatly facilitates their fabrication. The piers are of octagonal cross section and they measure 7.0 m in longitudinal direction, and 10.0 m in transverse direction. The cross-sectional wall thickness is constant. In transverse direction, the walls are 0.70 m thick, and in the longitudinal direction (along the bridge length) 0.90 m thick. 4 CONCLUSION The construction of original Mainland – Pelješac Peninsula Bridge was abandoned due to lack of funds. It was necessary to design new economically more viable bridge solutions in order to apply for EU funding. The key challenges for the bridge design are high bridge alignment at approximately +90 m elevation, adverse soil conditions, high seismicity of the site and stringent ecological requirements. 432
Figure 8.
Computer rendering of continuous box girder steel bridge.
Figure 9.
Computer rendering of multi-span semi-integral extradosed bridge – aerial view.
Two independent design offices proposed two alternative solutions. Continuous steel box girder bridge and a multi-span semi-integral extradosed bridge with hybrid deck. In both preliminary designs the road alignment adopted in the original design, total bridge length of 2404 m and already constructed bridge substructure parts were fully respected. Due to all fore mentioned reasons, first alternative solution was design of a very light bridge. Bridge deck is designed completely in steel and most of the driven piles are designed as slightly battered to facilitate the reduction of dimensions of pile caps in plan to a minimum, and also to increase the foundation stiffness. From the architectural point of view we can conclude that the author wanted to create structure which would blend harmonically into the environment and not impose on it. In the second preliminary design, authors tried to create something unique in appearance and structurally up-to date. Multi-span semi-integral bridge is a large, impressive and excellently designed engineering structure with soul and character. Thanks to an inventive approach to the design of this extreme bridge with an integral structure, featuring a hybrid deck suspended to six pylons, the bridge meets crucial design criteria of appearance, stability, durability, usability and economy in construction, which should be reflected in subsequent maintenance costs. An independent French consultant in their preliminary study preferred multi-span semi-integral extradosed bridge alternative. Based on these recommendations the Investor decided to choose the multi-span extradosed bridge for further design. 433
Figure 10.
Computer rendering of multi-span semi-integral extradosed bridge – close-up view.
REFERENCES Hrelja, G., Radic, J. & Savor, Z. 2009. Analysis and Design of Pelješac Bridge in Croatia, Proceedings of the 33rd IABSE Symposium on Sustainable Infrastructure, Bangkok, Thailand, 9–11 September 2009: 112–113. Radic, J., Savor, Z. & Pipenbaher, M. 2013. New Preliminary Designs of the Pelješac Bridge, Structural Department of the Faculty of Civil Engineering, University of Zagreb, Croatia. Radi´c, J., Šavor, Z., Pipenbaher M. & Hrelja Kovaˇcevi´c, G. 2014. New alternative solutions for Pelješac bridge in Croatia, Report + CD of the IABSE Symposium Engineering for Progress, Nature and People, Madrid, Spain, September 3–5: 518–519. Savor, Z., & Radic, J. 2008. Main Design of the Pelješac Bridge, Structural Department of the Faculty of Civil Engineering, University of Zagreb, Croatia. Savor, Z., Radic, J., Hrelja, G., Lazarevic, D., & Atalic, J. 2009. Seismic Design of Pelješac Bridge, Bridge Structures, Vol. 5: 97–108. Savor, Z., Hrelja, G., Mujkanovic, N., & Vlašic, A. 2010. Design Loads for Pelješac Bridge, Proceedings of the Joint IABSE-fib Conference on Codes in Structural Engineering, Cavtat, Dubrovnik-Neretva County, Croatia, May 3–5 2010: 481–488.
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Gebze–Orhangazi–Izmir Motorway, Izmit Bay Suspension Bridge Nehir Güngör Chief Engineer, General Directorate of Highways, Turkey
Fatih Zeybek Suspension Bridge Project Manager, NOMAYG
ABSTRACT: The Izmit Bay Suspension Bridge consists of a three span continuous suspension bridge having the main span of 1550 m and two transition spans will becomes the world’s forth longest suspension bridge at the completion in 2016. The suspended deck is 2.907 m long and continuous between two side span piers. The side span is each 566 m. There are 120 m and 105 m transition spans on the north and the south. The navigation clearance of the bridge is 64.30 m × 1000 m. The bridge girder will be of steel box girder construction with a stiffened steel plate deck with asphalt surfacing. The transition spans between the side span piers and transition piers will be formed by single span steel box girders. The main suspension cables will be formed from pre-fabricated parallel wire strands (PPWS). The main cables are deviated at the side span piers toward the cable anchorages located below the deck of the transition spans. The top elevation of the tower is 252 m above the sea level, each tower has two steel leg and two steel cross-beams. The towers will be of stiffened steel plate construction. The anchorages will be of the gravity type and will be supported on spread foundations. The tower foundations consist of a pre-fabricated rein-forced concrete caissons placed on a gravel bed on improved soil at 40 m water depth. Soil improvement will be achieved by steel pile inclusions. The bridge is sited in one of the world’s most seismically active regions where in 1999 the 7.6 Izmit Earthquake occurred on the North Anatolian fault in 1999. This paper deals with the project overview and construction method.
1 INTRODUCTION Gebze-Orhangazi-Izmir Motorway (Izmit Bay Crossing and Connection Roads) is one of the most important transportation infrastructure projects in Turkey, having the largest tender budget so far. Gebze-Orhangazi-Izmir Motorway Project is under construction which is tendered by a Build Operate Transfer (BOT) Method . With the construction of the Izmit Bay Crossing Project, Izmit Bay which can be travelled around using the existing state roads by 70 minutes will be crossed through in just six minutes. That means 92% decrease in travel time. The Izmit Bay Suspension Bridge (IBB) will carry the new “Gebze-Orhangazi-Izmir Motorway” at the Bay of Izmit in northern Turkey. OtoyolA.S. ¸ (Otoyol Yatırım ve ˙I¸sletme A.S., ¸ a joint venture of Nurol, Özaltın, Makyol, Astaldi and Göçay) has been appointed by KGM (Karayolları Genel Müdürlüˇgü – Ministry of Transportation) under the terms of a BOT contract to plan, construct and operate a new 420 kilometer Motorway (including the 3 km Izmit Bay Suspension Bridge) between Gebze and Izmir over a period of 22 years and 4 months. 2 TENDER AND DESIGN STAGES The geotechnical investigation was conducted by Fugro between November 2010 and April 2011 under a contract from OTOYOL AS in order to address regional faults setting, geology and site conditions around the proposed bridge foundations. Construction of the Suspension Bridge was 435
contracted in July 2011 to a consortium of IHI Infrastructure System and Itochu as EPC contractor. The detailed design of the bridge, was carried out by COWI under supervision by IHI-ITOCHU and checked by HALCROW as Independent design checker. 3 DESIGN EVALUATION 3.1 Wind tunnel tests A series of wind tunnel tests was carried out to obtain the design data and to ensure the aerodynamic stability of the bridge during construction and in-service condition. The deck sectional model of a geometric scale of 1:65 was tested in smooth and turbulent flow in FORCE in Denmark, preliminary to investigate aerodynamic stability, vortex induced vibration, static load coefficient and aerodynamic derivatives on the deck section with different type of traffic barrier and with/without traffic and finally to prove the deck section to work on the deck section with the final appendices including collecting all necessary data, for construction and in-service condition .The tower aerodynamic model of a geometric scale of 1:70 was tested in smooth and turbulent flow in BLWTL in Canada to investigate aerodynamic stability, vortex induced vibration and static load coefficient for construction and in-service condition. The full bridge model of a geometric scale of 1:220 was tested in smooth and turbulent flow in Politecnico di Milano in Italy finally to confirm aerodynamic stability of the suspension bridge during construction and in-service condition. The fundamental basic wind velocity is 25.4 m/s determined based on the measured wind data at three meteorological stations near the construction site and the critical wind velocity for instability is 58 m/s for the horizontal wind. 3.2 Seismic design A performance based seismic design approach was adopted by using three different ground motions with different return periods as shown in Table 1. Nevertheless, soil-structure interaction analyses and structural analyses were conducted to evaluate the consequences of potential secondary fault deformations below the SouthAnchorage location. These analyses consider the design forced displacements of the soil in Table 2. Table 1. Seismic performance criteria. Place
Ground Motion Return Period
Seismic Performance Level
Damage Performance Level
Functional Evaluation Earthquake (FEE) Safety Evaluation Earthquake (SEE) No Collapse Earhquake (NCE)
150 years (50% in 100 years) 1000 years (10% in 100 years) 2475 years (4% in 100 years)
Immediate Access
No Damage
Limited Access
Repairable Damage No collapse, life safety Damage
Table 2. Design forced soil displacements. Design Displacement Earthquake Event
Horizontal
Vertical
SEE NCE
0.70 m 1.00 m
0.25 m 0.50 m
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3.3 Live loads and load combinations The verification was carried out for serviceability limit states and ultimate limit states respectively for different load combinations. Additionally accidental design situations such as ship impact, rupture of hanger cable, collision with vehicle restraint system, fire and tsunami are also considered. Traffic loads are determined based on EN 1991-2 by using the notional lane approach. Long loaded length approach was not used by keeping a total of 82 kN/m traffic load constant regardless of the loaded length. 3.4 Geotechnical investigation The Izmit Bay Bridge will connect the Diliskelesi peninsula to the North with the Hersek peninsula on the South. The bridge is sited in one of the world’s most seismically active regions where in 1999 the 7.6 Izmit Earthquake occurred on one of the world’s longest strike-slip trending the North Anatolian fault in 1999. The geotechnical investigation for final the foundations design included the offshore geotechnical investigation for the tower foundation area, the nearshore technical investigation for the south anchorage area, and onshore drilling for the north anchorage area was conducted by Fugro in 2011. One of important discovery was a potential secondary fault far below the ground level but at the planned location of the south anchorage as shown in Figure 2, which resulted in the south anchorage shifted by 138.25 m at the splay toward north in a safe zone between faults, the south side span pier shifted by 118 m toward north, and two tower foundations shifted by 59 m toward north. The updated bridge configuration used for the detailed design is shown in Figure 1 below. The main span was maintained as 1550 m and the suspension bridge was maintained to be symmetry. Furthermore, the separate concrete slabs in the front and in the rear of the south anchorage in the tender design was integrated into one large slab, on which the side span pier and the front and splay leg of the anchorage are located to be allowed to move together under strong earthquake as shown in Figure 2 below.
Figure 1.
General arrangement.
Figure 2.
Secondary fault at planned South Anchorage location.
437
Figure 3. Tower foundation.
4 TOWER FOUNDATIONS The tower foundations are designed as concrete caisson placed on gravel bed on improved soil at −40 m and −35 m below sea level for the north and south foundations which allow the tower foundation to move under strong earthquake not to transfer a huge seismic force into the bridge system. The soil below the caisson is improved by 2 m diameter steel pile inclusion to a depth of about 35 m below the sea bed providing the required bearing capacity and eliminating potential risk of liquefaction during earthquake. 5 m thick gravel bed separates the improved soil and the concrete caisson to provide a controlled sliding surface during seismic events. The purpose of the gravel layer is to act as a fuse under seismic loading by limiting the maximum interface shear friction and thereby reducing the seismic loads on the structure for high amplitude earthquakes. The two caissons were partially constructed in a dry dock with the base slab and outer walls fully cast in three stages, and the inner walls being partially built. They were then towed to the wet dock, where the inner walls were completed, the top slab cast and the steel shafts installed, the lower part also having its concrete cast. Since 100 years of design life is demanded, very high concrete quality is needed, with a low chloride migration ratio, low permeability and low heat of hydration. Following the completion of caissons in dry/wet dock, the caissons are towed to the final position, sunk and placed on the gravel bed at −40 m of sea level. While caisson construction was proceeding, dredging works carried out at the foundation locations by a grab dredger over an area of 87 m by 74 m and up to 5 m depth. Some 195 inclusion piles of 2.00 meters diameter and up to 35 meters long were driven by an underwater hydraulic hammer. The gravel bedding was placed using a tremie pipe mounted on a floating barge crane, and the surface of the gravel was levelled by specially manufactured hydro-mechanical levelling equipment. Submerging the caisson including steel shaft is carried out by ballasting water as the caisson is kept in position. The caisson is divided into a number of clusters. The clusters at four corners are filled with water to provide stability. The plinth and tie-beam construction is the final stage of the construction of the tower foundation and is carried out by such sequence as concrete casting of the upper part of the steel shaft, placing PC cover inside the steel shaft followed by concrete 438
Figure 4.
Sinking of caisson tower foundation; R.C.tower foundation +10.150 m.
Figure 5.
Deck section.
casting of the plinth up to +2.0 m, installation of the anchor bolts to connect the steel tower to the foundation, concrete casting of the plinth to the final height, and construction of the tie-beam. 5 THE DECK The deck is 35.93 meters wide, shaped as a closed steel box girder having a depth of 4.75 meters and carrying three lanes of highway traffic in each direction plus hard shoulders and cantilevering walkways for operational and maintenance use. Truss diaphragms at 5 m centers are used for minimum weight. The deck will be erected in 113 segments to suit the fabrication, transportation and erection capabilities available. The suspended deck is an orthotropic stiffened steel box girder of 4.75 m deep and 30.1 m wide having cantilevered inspection walkway of 2.915 m on each side as shown in Figure 5, that are suspended by the hanger ropes spaced at typically 25 m. The suspended deck has three traffic lanes each of 3.65 m and the hard shoulders of 0.5 m at the centre and 1.0 m at the side for each direction. The traffic barriers of tension rope type are provided in the central medium and at the edge of the deck. The windscreen is provided only at the tower location to avoid a sudden change of wind pressure to the vehicles running on the bridge. The water main is place at the mid-way of the inclined web to supply water for maintenance and fire fighting. The roadway lightings are provided at the edge of the deck at every 25 m in-between the hanger ropes. The thickness of the steel deck is 14 mm for all traffic lanes considering heavy traffic volume and more number of five axles articulated vehicles on middle and fast lanes. 439
Figure 6. Tower section.
6 TOWERS The 252 m tall H-shaped towers are fabricated in steel due to demands for high seismic loads and short construction time. Closed box sections are used for legs and cross beams. The towers are 241.925 m high single-cell box steel structure from the base to the cable at the tower saddle having two cross beams connecting two tower legs, separated by 30.1 m at the top and 36.2 m at the bottom, above the bridge deck level as shown in Figure 6. The overall dimensions of the tower legs are 7 m × 8 m at the base and 7 m × 7 m at the top and tapered in the longitudinal direction. Each tower leg is divided into 22 blocks to suit a lifting capacity of crane used for the erection and the tower leg blocks are connected by welding for perimeter plates and by slip resistant connection for vertical stiffeners. Mass dampers are provided to mitigate vortex induced vibration during construction and in-service condition. The towers are stiffened steel plate construction. The tower foundations are designed as concrete caisson placed on gravel bed on improved soil at −40 m and −35 m below sea level for the north and south foundations which allow the tower foundation to move under strong earthquake not to transfer a huge seismic force into the bridge system. The lower tower leg segments and the lower cross-beam were erected using a floating crane, while the tower legs segments 12 to 22 and the upper cross-beam will be erected by a self-climbing crane mounted on the support structure connected to the lower cross-beam.
7 MAIN CABLE AND HANGER CABLES The main cables are designed as being constructed by means of pre-fabricated parallel wire strand (PPWS), each consisting of 127 numbers of high strength steel wires of 5.91 mm having a breaking strength of 1760 Mpa. Per one main cable, 110 numbers of PPWS are placed between the cable anchorages (See Figure 8) and 2 extra PPWS are placed between the tower and the cable anchorage on the both sides. The hanger ropes are designed as parallel wire strand (PWS) coated by HDPE sheath and terminated to the sockets that are pin connected to the cable clamp and the hanger anchorage at the deck. Due to a large longitudinal relative movement between the main cable and the suspended deck and a large transverse movement and tilting of the suspended deck under traffic and strong earthquake, substantial number of hanger ropes are equipped with cylindrical bearing to the cable clamps and spherical bearing to the hanger anchorage at the deck. The main cables are protected by S-formed wrapping wire, paint and a state-of-the-art dehumidification system against corrosion on the entire length including at the cable saddles and in 440
Figure 7. Tower erection.
Figure 8.
Cable cross section.
the cable anchorage. The inside of the tower, deck and anchorage chambers are also protected by dehumidification system against corrosion. In order to carry out the erection of the main cable, as a first step hauling ropes were installed between 14–29 January 2015. Then an aerial walkway (catwalk) erection started on 05 Feb 2015 with hauling system supported by the tramway system. 8 ANCHORAGES The cable anchorage are designed as a gravity type anchorage consisting of the triangle shape concrete structure (upper part) where the main cable is splayed and anchored and the separate concrete slabs (lower part) on which the front and splay legs are placed on pile foundation for the south anchorage and directly on rock for the north anchorage. The transition piers are placed on 441
Figure 9.
North-South anchorage.
the rear concrete slab of the cable anchorage and the side span piers are placed on pile foundation separately built in front of the cable anchorages. The anchorages are the gravity type and are supported on spread foundations. The south anchorage is constructed on the common large concrete slab constructed on the reclaimed land on which the side span pier and the transition pier are placed to reduce a risk of relative movement between the cable anchorage and the deviation saddle on the side span pier hence a huge force introduced in the main cable under a strong earthquake. The construction of the south anchorage consist of construction of 1.0 m thick retaining wall (diaphragm wall depth to −32.0 m), excavation inside the retaining wall, construction of large concrete slab and construction of triangular cable anchorage. The overall dimension of the retaining wall is 58.0 m wide and 126.0 m long, in which the rear part is of double circle with 58 m diameter and the front part is of rectangular of 18.0 m wide and 37.4 m long. After excavation inside the retaining wall, the large concrete slab is constructed block by block. The north anchorage is constructed on land at the Dilovasi Cape of north side. Not similarly to the south anchorage, the north anchorages placed on relatively favorable rocks and no faults have been discovered in the geotechnical investigation, thus the front and rear legs of the triangular cable anchorage are place on separate concrete slabs that are connected to each other by tie-beam as shown Figure 9, and the total volume of concrete is smaller than for the south anchorage. The construction procedure is almost the same as that adapted to the south anchorage, e.g. block by block construction vertically and horizontally to suit concrete supply capacity and limit the rise of temperature in concrete below the permitted value. 9 CONCLUSION Construction of Izmit Bay Suspension Bridge has started with preparatory works such as shore reclamation of south anchorage area, construction of dry dock and mobilization activities. Then construction of caissons, anchorages, preparation of sea bed at tower locations were completed. Fabrication of tower blocks, decks, cable wires and saddles were done factories in different locations around the world, transported to site to be erected as per schedule. Construction Works of Suspension Bridge is planned to be completed at beginning of February 2016. As a result of constructing Izmit Bay Suspension Bridge, new technics of constructing suspension bridges has been found and developed. A rise is expected in the number of large span bridges and soon enough Turkey will have one more of this precious bridges, Dardanel Crossing. REFERENCES COWI, 2013. Detail Design Report of Izmit Bay Suspension Bridge Project. Cetinkaya, O.T., 2013. Izmit Bay Suspension Bridge – Overview of the Project.
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Construction of cable-stayed bridge over the Drava River on Corridor Vc, Croatia P. Sesar, M. Mašala Buhin, D. Bani´c & S. Kralj Institut IGH, Zagreb, Croatia
ABSTRACT: The bridge over the Drava River is situated on the Beli Manastir–Osijek motorway section of the road route called Corridor Vc. The total bridge span length is 2485 m, and the main cable-stayed bridge over the river measures 100 + 220 + 100 m. Approach viaducts are typical prestressed concrete structures of usual span, 35 m in length, while the cable-stayed superstructure of the main bridge is of composite construction. Selection of an optimum main span construction method, from the aspect of construction and navigability, is presented in the paper. All statically significant construction phases are presented. An overview of completed works, with support details, is also given.
1 INTRODUCTION The motorway section Beli Manastir-Osijek, in its final part passes over the 3.1 km wide Drava River flood retention area. The location itself is the most interesting part of the motorway and at the same time the greatest challenge for designers and constructors. The Bridge over the Drava River is the most significant and the largest bridge on the A5 motorway. The bridge spans over the Drava River and its left and right foreshores. The Drava River water level varies, so the high waters flood the left, Baranja shore, up to the dyke. The flood retention area on that side is swam-like and overgrown with shrubs and trees, as seen in Figure 1. The Slavonian foreshore has no dyke since the bank is significantly higher. Several options of the construction – technical part of the bridge design were analysed. Variants with different bridge lengths were analysed during preparation of the preliminary design. The problems occur in the navigation requirements and the newly defined navigational profile at the point of the main structure over the Drava River, which define the grade line position in that part. The second restrictive factor is the width of the Drava River flood retention area which is
Figure 1.
Swamp-like flood retention area on the Baranja side.
443
Figure 2.
Drava River Bridge – variant with a structure in the flood retention area.
3100 m at that location. The right bank, on the Osijek side is higher and the flood retention area ends with a natural relief. On the left, Baranja side, a dyke was constructed due to the low terrain level. This has to be crossed over by the motorway at the same time respecting the requirements set by the Hrvatske Vode Company. It was planned that a route on a structure is designed in the right foreshore because of the Vuˇcica River and the forest exploitation. On the left, Baranja foreshore variants with a bridge over the complete flood retention area were analysed, as well as the variant with a partial structure and a high embankment for the most part. The results of geotechnical investigations were used for evaluation of the different variants. 2 VARIANTS OF THE DRAVA RIVER BRIDGE The first variant analyses the option that the Drava River and its flood retention areas are overpassed by a unique structure. Overall length of the structure is Luk = 2590 m. Bridge length over the right foreshore (near Osijek) is 1170 m, and 1150 m over the right foreshore. Bridge length over the Drava River itself is 270 m. (Fig. 2). The second variant proposes construction of a bridge over the Osijek side foreshore and a bridge over the Drava River. On the Baranja side foreshore, three openings would be constructed in the foreshore, the bridge would be extended and an additional crossing for animals would be constructed in the foreshore, 75 m in length. Outside of the foreshore, near the dyke another animal crossing is planned, length 31 m. An embankment would be constructed in the Baranja side foreshore, average height 10 m, which is exposed to oscillations of the Drava River water level in the lower third. The embankment would have adequate geotechnical erosion protection measures applied. 3 SOLUTIONS OF THE CENTRAL BRIDGE PART Both variants, with the embankment or the structure in the foreshore were also analysed according to the type of structure. Three solutions for each variant were analysed , with a concrete load bearing structure, a steel structure and a composite structure. The longer and the shorter structure share the central structure which passes over the Drava River, which has also been analysed in several variants; with piers in the river and outside the navigable profile; in variants of composite steel-concrete structure, steel box structure or a concrete box type superstructure. 444
Figure 3.
Drava River Bridge – variant with an embankment in the Baranja side foreshore.
Figure 4. A view of the Baranja side Drava River bank.
While the first two variants (A and B) with a slab above the superstructure show a steady, already seen bridge structure which crosses a lowland river, variant C, with a high, inclined pylon and cable stay structure offers a new, distinct sight. The steel cross section of the central part is designed as a box type continuous girder, variable height, from 2,5 m at the bridge ends and up to 4,5 m in the centre, above the piers The upper slab is braced with 13 pieces of torsional rigid trapezoid ribs. Concrete cross section is a continuous box girder of variable height; 3,0 m at the bridge ends and in the centre, and 6,5 m above the piers. Composite cross section is a box type steel girder of variable height: 3,55 m at the bridge ends and in the centre, and 6,65 m above the piers. The upper slab is a reinforced concrete slab, variable thickness – from 0,25 to 0,45 m.
4 CONSTRUCTION OF THE DRAVA RIVER BRIDGE Construction of the motorway A5 section, from Beli Manastir to Osijek started with the construction of the Bridge over the Drava River. The adopted variant was the one where the bridge crosses the river and the left and right foreshores with a unique bridge, length 2485 m. The central structure over the river has spans 100 m + 220 m + 100 m, in order to satisfy the navigational requirements, i.e. the required navigational waterway width 50 m and height 5,25 m. 445
Figure 5. Drava River Bridge – variant A. Bridge openings: 92,0 + 136,0 + 92,0 m Bridge width: 2 × 13,2 m Overall bridge length: 271,0 m
Figure 6. Drava River Bridge – variant B. Bridge openings: 100,0 + 220,0 + 100,0 m Bridge width: 2 × 13,2 m Overall bridge length: 271,0 m
Figure 7. Drava River Bridge – variant C. Bridge openings: 150,0 + 240,0 m Bridge width: 33,4 m Overall bridge length: 390,0 m
Based on variant 3 of the preliminary design the main span was adopted consisting of a composite steel structure, 2 reinforced concrete pylons and staying cables. The load bearing steel beam is composite with a reinforced concrete carriageway slab whose cantilevers are supported by I-profiles at every 3,33 m, i.e. 3 m. Box type girders are composite with cross beams at a spacing 446
Figure 8.
Composite steel-concrete superstructure (variant A).
Figure 9.
Steel box type superstructure (variant A).
Figure 10.
Concrete box type superstructure (variant A).
Figure 11.
Selected variant of the cable stayed bridge with 2 pylons and composite superstructure (variant B).
of 3,33 m and 3 m. At the points where the staying cable enters the box, double cross beams are designed. Cable-stayed bridge was designed with 2 A pylons perpendicular to the vertical axis with staying cables, in order to decrease the bending moment caused by the self-weight of the structure and additional dead load and traffic load. Pylons are 75 m high with arm having rectangular cross section. On the front and rear side of every pylon are two rows with 10 staying cables each. The pylon piers have foundations with two groups of 25 drilled piles each, diam. 150 cm. The approach viaducts in the foreshores on the Slavonian and Baranja sides have a reinforced concrete semi-precast superstructure. 447
Figure 12.
Cable-stayed structure with a sloped pylon and steel-concrete superstructure (variant C).
Figure 13.
Construction of approach structure on the Baranja foreshore (with a part of the revetment).
The spans are 35 m, except the ones at the end, which are 28 i.e. 31,5 m and 24,5 m. The approach superstructures are supported by pairs with a circular cross section of 1,8 m. At the positions of the shore (expansion) piers, at the cross-over from the reinforced concrete to steel structure, the piers are actually walls also with foundations of reinforced concrete piles. The cross section consists of 6 precast reinforced concrete prestressed T-cross section girders, height 182 cm. Above them is a concrete slab, thickness 25 cm. The bridge construction started with approach structures from the Baranja and Slavonian sides. This was followed by construction of the pylons on the Slavonian side so that the superstructure elements could be placed. The left river bank is lined with stone revetment for app.1.5 km, symmetrical on both bridge sides since the erosion is very prominent on the banks, and the line moves to the shore during time. The river navigation will be ensured by construction of regulation facilities on the opposite side, i.e. the right river bed side, also during the bridge construction stage. 448
Figure 14.
Construction of the Slavonian and Baranja side pylons.
5 END OF CORRIDOR VC IN THE CROATIAN SOUTH The continuity of traffic flow shall be enabled only when Corridor Vc is completed, i.e. its construction through Hungary, Croatia, Bosnia and Herzegovina and at the Croatian farthest south up to the Port of Ploˇce. South section bears the mark Motorway A10 and is planned to be the final, southern section of Corridor Vc. Analysing the wider zone of the future motorway border crossing, a new eastern entrance into the town and port of Ploˇce was also included into it, as well as the ˇ Ceveljuša Interchange on the Adriatic motorway. Several high serviceability motorway corridors are planned for construction at the farthest Croatian South, in the Dubrovnik – Neretva County. One of these corridors is the A1 Motorway, Zagreb – Bosiljevo – Split – Dubrovnik, an integral part of the Adriatic – Ionian motorway. The second corridor is the A10 Motorway, and the third is the new express road connecting the Plo´ce Interchange and the Port of Ploˇce. This road has a very complex route, overcoming a very demanding terrain configuration. It connects the motorways A1 and A10 with the town and port of Ploˇce as well as with the State road D-8 (Adriatic motorway). The Port of Ploˇce is one of the most important Croatian ports (second according to national significance). It is located half way between Split and Dubrovnik, and is the exit to the sea for the European Corridor Vc, offering the shortest route to the main centres of Bosnia and Herzegovina which give over 80% of the port’s overall traffic. In this respect, the express road Ploˇce Interchange – ˇ Ceveljuša Interchange – Town of Ploˇce becomes a crucial connecting route for the overall traffic in this area, with special influence on the Port of Ploˇce. Its construction is a pre-requirement for free ongoing of all traffic flows in this region. 449
Figure 15.
Motorway route on Corridor Vc within the Croatian motorway network.
6 CONCLUSION The Motorway A5 on Corridor Vc is an important transport connection between Croatia and the neighbouring countries; Hungary and Bosnia and Herzegovina Corridor Vc importance for Croatia is based on better connection of the Slavonia – Baranja area with the Adriatic or coastal area with a pre-requirement of economic, tourist and cultural interference. It is an important part of Bosnia and Herzegovina road network and it allows Hungary a link with the Adriatic, i.e. the Port of Ploˇce. This is why the link between the motorway A5 through Slavonia and Baranja, Corridor Vc in Bosnia and Herzegovina and the motorway A 10 Mali Prolog-Ploèe is so important. When this link between Pannonia and the Adriatic is finished a quicker economic, demographic and political development will be possible for the whole region. REFERENCES Crnjak M. et al. 2007. Design and build of highway on Corridor Vc in Croatia; 1. BiH road congress. − Highway Beli Manastir – Osijek – Svilaj, Section: Beli Manastir – Osijek, Main design, Institut gradevinarstva Hrvatske, d.d., Zagreb, 2008. Highway Vc. Section Beli Manastir – Osijek: Preliminary design bridge over the Drava River on corridor Vc. Highway Vc. Section Beli Manastir – Osijek: Main design bridge over the Drava River; Institut IGH, d.d., Zagreb, 2013. Highway Beli Manastir – Osijek – Svilaj, Section: Beli Manastir – Osijek, Basic documents for obtaining location permit, Institut IGH, d.d., Zagreb, 2003.; 2014. Highway Vc. 2014. Section Beli Manastir – Osijek: Main design 1. faze autoceste; Institut IGH, d.d., Zagreb. Highway Vc. 2014. Section Beli Manastir – Osijek: Implementation design bridge over the Drava River; Institut IGH, d.d., Zagreb. − Hrvatske, d.d., Zagreb, 2007. Institut gradevinarstva Kralj S. et al. 2007. Highway A5, project elements section Osijek-Ðakovo; 4. Croatian road congress. Sesar P. et al. 2007. Highway Beli Manastir-Osijek with analysis of the possibilities of Drava River crossing; 4. Croatian road congress.
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Strait crossing of the Thermaikos Gulf with a mixed long-span bridge and subsea tunnel system M. Malindretou-Vika Aristotle University of Thessaloniki, Thessaloniki, Greece
P. Spyridis Institute of Structural Engineering, University of Natural Resources and Life Sciences, Vienna, Austria
ABSTRACT: The realization of the proposed connection, spanning 8.5 km of open sea, is the response to the needs and requirements of the metropolitan center of Thessaloniki and to the necessities of its inhabitants. While becoming part of the freeway “Egnatia Odos” the project aims to improve the highway infrastructure system of the city. This connection serves as an answer to the growing needs of the modern commuter of Thessaloniki thus completing and improving it’s external ring road. The new highway, will by-pass the city centre offering immediate relief of traffic, while ensuring faster overall travel times. The design is guided by the requirements and the limitations of the project area, while taking in consideration the social, political and economic status of the city of Thessaloniki, respecting its historical importance and contemplating on the preservation of the gulf’s ecosystem. 1 INTRODUCTION 1.1 General Information Thessaloniki is the second largest Greek city, situated at the North of the country, with a population of approx. one million (1,000,000) inhabitants in its’ metropolitan area. Historically speaking the city has been structured over multiple layers of different cultures and ethnicities and has always been the evolutionary hub of the region as well as the Balkan Peninsula in general. Between the late 1960’s and early 1980’s, because of the political and economic turmoil in the country, the city had to deal with an unprecedented amount of uncontrollable growth which has left most infrastructures unable to cope with the advancing populations’ needs. Therefore in recent years the acute traffic congestion has become an ordinary phenomenon. 1.2 Triantafillids proposal In 1964 Triantafillidis (1966), professor at the Urban Planning Department of the Aristotle University of Thessaloniki, was asked to perform a study on the possible implications and solutions regarding the city’s spanning over the next 50 (fifty) years. His results, presented in 1966, included various solutions, based on a four point plan, which would aid Thessaloniki in becoming a modern European metropolis. His proposals included the creation of a European transport Hub through the city’s port, the relocation and expansion of the Makedonia Airport, the relocation of the International Fair from the city’s center to its’ outskirts and the construction of a railway passage over the Thermaikos Gulf, by which the western and eastern parts of the metropolitan area would be connected, without aggravating downtown traffic. The results of this urban study since then became the point of reference for every proposal regarding Thessaloniki’s future in recent years. The aforementioned railway connection was supposed to undergo construction between the years 1991–2016. Even though the city’s population far exceeded Triantafillidis’ expectations, the 451
Figure 1.
General area map and diagram of Thessaloniki’s sprawling (L) and the proposed connection (R).
evolution of the city, East and West, did not. The Eastern and Western parts of the city came to grow separately creating two distinct urbanized areas—in need of a point of reference—which still today meet at a narrow piece of land, the city’s center, geographically constrained by one side from the sea and by the mountains from the other unable to cope with the needs of the growing population. The uncontrollable city sprawling resulted in sparse low-rise housing in the suburban areas and an increased construction rate of high-rise buildings in the city’s center. Therefore, unlike Northern Italy’s paradigms of multimodal “diffused cities”, as identified by Indovina (1990), Thessaloniki was forced to rely heavily around a unique node, resulting in intensified traffic, increasing gas emissions and noise pollution. The sporadic proposals upheld throughout the years, have persistently failed to address this major issue lowering quality standards and increasing the populations’ dismay (Fig. 1). 1.3 Area of interest The surrounding region presents numerous problematics in need of confrontation in order to present a sustainable solution both ecologically and urban planning-wise. The project area is situated close to the Axios River’s Delta, one of Europe’s most important wetlands, featuring over 100 species of birds. Moreover 93 species use the riverbanks as an intermediary migratory station. Three other rivers empty into the gulf (Aliakmon, Gallikos, Loudias) therefore creating an unstable seabed (the mean total volume of the sediments in the Delta rises up to 14,260,000 m3 per year), while also contributing to the humid climate and ever-present foggy atmosphere during the autumn and winter months. Due to the complex ecosystem that has been created by these four rivers the wetland area, covering almost 12.000 hectares, the site is under the protection of the 1971 Ramsar Convention on Wetlands “The site is an extensive coastal zone formed by the three rivers and includes the lower riverbeds of the rivers and their estuaries with predominant salt marshes and extensive mudflats. Natural vegetation areas are criss-crossed by the drainage ditches delineating the arable land. In spring and summer the neighboring extensive rice fields are flooded, creating a unique landscape.” (Ramsar Sites Information Service 1998) The Ramsar Convention forbids any kind of human intervention inside the protected zone (Fig. 2). Adding to the complexity of the site, the Northern Greek region is prone to intense seismic activity, including numerous earthquakes above the 4 Richter scale over the past 20 years (Fig. 3). 1.4 Objective The proposed highway connection over the gulf of Thessaloniki will become part of the Egnatia Odos highway system , deviating the traffic from the Euzonoi node to the recently designed Aggelohori-Nea Moudania highway, serving the city’s commuters East to West and vice versa bypassing the center. Furthermore, ongoing traffic towards the Makedonia InternationalAirport and 452
Figure 2. The area protected by the Ramsar Convention (L). Seismic activity for the past 20 years (R).
the tourist region of Chalkidiki will burden no more Thessaloniki’s already congested metropolitan area, liberating it from all the indirect traffic. 2 METHODS AND SOLUTIONS 2.1 Connecting the gulf In order to deviate traffic the proposed connection must be made by joining the two points located at the outskirts of the metropolitan area, both east and west. The new highway will become attached to the existing infrastructure towards the west entrance of the city, by reconstructing the Euzonoi intersection on top of the Athens-Thessaloniki highway, which connects Greece’s two largest cities. On the opposite side of the gulf the Municipality of Aggelohori, which stands on the verge of the peninsula, will become the eastern node of the project. The deviation will have a total length of 15 km and a travel time of 12 minute (at a traveling speed of 80 km/h) instead of a 58 km route and a travel time of 78 minutes for the same destination using the current infrastructure. 2.2 The project After the reconstruction of the Euzonoi intersection in order to accommodate the new highway the project traverses 6.5 km of the agricultural buffer zone towards the wetlands. Before entering the protected area the road turns into an underground tunnel so as to leave undisturbed the preserved by the Ramsar Convention flora and fauna. Due to the fact that the aforementioned limitations extend for several miles inside the sea basing, the underground tunnel travels for a further 4.5 km before exiting the zone of limitations, emerging 1.35 nautical miles of the coast inside the Thermaikos Gulf. In order to accommodate the connection the construction of an artificial island, 500 m long and 250 meters wide, is necessary so as to provide a connection point between the tunnel and the bridge section. From the artificial island onwards the rest of the remained distance spans 4 km above sea before reaching, the other side of the connection, the Municipality of Aggelohori (Fig. 1R). 2.3 The bridge The first step towards designing the bridge section was the selection of the specific typology suitable for the characteristics of the area. Because of the proximity of the Makedonia International Airport (at 13 km of distance) the total height of the structure would need to be maintained reasonably low. For this reason a suspended or cable-stayed type of bridge would be unsuitable, due to the direct air-traffic towards and from the airport.Considering the airport’s needs the bridge as colored with a white opaque coating, easily visible from a distance without creating any reflection, while the illumination of the deck was thoroughly thought avoiding to create a “runway effect”, by directing the light downwards.The only solution available was that of an arch type bridge with a 453
Figure 3. The typical arch (L), and construction detail (R).
suspended deck, which would not only eliminate the need for extremely high pylons but also satisfy an unobstructed view of mount Olympus as observed by the city. Since the region is constantly battered by intense seismic activity, with at least three major earthquakes stronger than 6.5 grades on the Richter scale during the past century, the arch type construction also meets the specific anti-seismic requirements. 2.3.1 The arches In order to obtain a fluid and elegant design the decision was made to construct the main part of the bridge using continuous arches supporting the main deck, instead of proceeding with a widely diffused girder bridge technique. The continuous arches bear considerably better the tensions caused in case of an earthquake as opposed to simple vertical piers. Subsequently the form of each arch was studied; the objective was to obtain a transparent and light structure, which would not obstruct the view. For the aforementioned reasons the arches design would need to have considerably long spans, while the double arched form are heavily influenced by the fluid forms of Italian engineer Sergio Musmeci. In order to cover the needed distance above sea sixteen (16) continuous arches with variable sections have been designed. Their spans vary from 168 m up to 228 meters while their respected rises varying from 16.4 m up to 42.8 m above sea level (Fig. 2). 2.3.2 The main arch The main arch where the suspended deck reaches a 52 m of clearance has a span of 350 m and a rise of 105 m. The arch, with a variable rectangular section, passes through the two separate decks, while suspending them by 52 cables (each with a 15 cm diameter). The port authorities requirements covered 30 meters of clearance for the current conditions, a height that was almost doubled during the design phase, accommodating any possible future needs. 2.3.3 The deck The choice of the two distinct decks (one per each direction) was undertaken in order to relieve the pressure created by the intense wind that blows in the area, reaching the speed of 120 km/h. For that same reason the hollow deck section is formed like an inverted airfoil generating a downward force that keeps it attached to the supporting structure. The deck is 38,5 m wide and consists of two driving lanes (4,5 and 4 m respectively), an emergency lane (2,5 m) and a walkway (2 m). The tubular section has a maximum height of 4,3 m. The two decks are constructed in modules 20 m long, welded together by means of a C-beam, that 454
keeps them 6,7 m apart. The total weight of each module rises to 234,444 kg (234 ton). For the first 100 meters, following the exit from the tunnel, the deck is covered by an elliptical roof, so as to prevent and protect the passing cars from the possible high waves during rough weather (Fig. 3). 2.3.4 The foundations The gulf of Thermaikos reaches a maximum depth of 1400 m. On the proposed project area the maximum perceived depth is 28 m, which does not create particular problems regarding the laying of foundations for a structure so massive as the one previously described. Despite the large sediment volume produced by the near rivers, the consistence of the seabed does not present insurmountable difficulties. In order to create a stable foundation the use of numerous micro-piles needs to be employed, up to a depth of 30 m below the seabed. The micro-piles serve as a stabilization method, creating an ideal base for supporting the weight of the over-structure. A similar method was employed a decade ago for the Harilaos Trikoupis Bridge (at Rio-Antirrion) in the Peloponnese region, with similar seabed and seismic characteristics. Regarding the earthquake protection, the foundations need to be free to move in a controlled manner, thus absorbing the eventual vibrations. Therefore, a thick layer of gravel is poured over the micro-pile structure, approx. 2 m thick, on which the foundation base is lowered upon, held in place by its own weight. 2.3.5 Structural design basis The structural function of the arch construction is based on the following features: – The arcs should mainly bear axial forces and transfer them to their base “pushing” each other in the horizontal direction, while the abutments bear mainly vertical loads. – The main arc is a so-called arc and tie bridge, which incorporates a tie between two opposite ends of the arc. The tie is capable of withstanding the horizontal thrust forces, which would normally be exerted on the abutments of an arc bridge. – The tubular deck is appropriate to bear torsional and bending stress, which mainly appears due to traffic loads. The material used for the main construction is Steel S355/Fe510, and for cables and tendons high strength steel. Regarding the load typesin this preliminary design phase, gravity loads are supposed to represent the most critical static loads of the structure. The following actions have been taken into consideration: – Permanent Loads: Self Weight; Equipment; Road – Pavement Weight – Variable Loads/Traffic Loads: Uniform Load – UDL; Concentrated Load TS The response of the bridge under wind and seismic loads is subject to special dynamic analyses. Special caution is suggested for the design under impact loads and furthermore to a robust design against collapse. Temperature effects should be limited by use of special constructional joints. The response of the ground together with the formation of the foundations is also extremely critical. Stability under second order effects and robustness against collapsing should be particularly investigated. Breaking actions are not covered. The checks are done for the Ultimate Limit State but a design study should also focus to the Serviceability and Durability (Fatigue) Limit States. Finally, a crucial part is the construction design, on which the realization of the project is highly dependent. 2.4 The tunnel The underground part of the highway consists of a tunnel 4.5 km long. This shall be realized using Tunnel Boring Machines (TBM), specialized equipment that allow for the complete automation of the digging process of the galleries, including the coating and finishing layers. These machines can excavate through rocky surfaces, as in the example of the San Gottardo passage in Italy, or even dig under the seabed (i.e. Napoli’s underground railway). The above technique is to be employed 455
Figure 4. The cross section of the tunnel and the section of the artificial island. Table 1a. General data. Total bridge length Total tunnel length Island dimensions Main arch span Adjacent arches span Main arch span/rise Road section Boat clearance main arch Boat clearance sec. arches Max. capacity/direction Estimated cost
4000 m 4500 m 500 × 250 m 350 m 228 m 1/3 2 lanes + shoulder. 52 × 280 m 40 × 100 m 3000 vehicle/h 1 billion €
Table 1b. Arches structural data. Material Main arch rise Arches rise Arch cross section Weight of typ. arches Weight of main arch
Steel S 355 108 m 16.4–42.80 m Variable rectangular 57,190 ton 6070 ton
throughout the whole length of the tunnel, including the underwater part, since it is impossible to use a submerged type of tunneling (a widely used method in similar cases) because of the restrictions sustained by the Ramsar Convention. The proposed underground passage consists of two twin galleries, measuring 15.5 m in diameter. Each of the galleries has two driving lanes, an emergency lane and a walkway. For safety purposes, a fire-shelter (refuge) connects both passageways, every 300 m. Right below the road surface a secure passageway can be used for evacuation purposes (Fig. 4). 2.5 The artificial island The underground tunnel reaches its end at the proposed artificial island, approx. 1.35 nautical miles off the coast. This elliptical man-made surface is 500 m long by 250 m wide. For its construction the land reclamation technique will be employed. Famous examples using this type of construction method are the International Kansai Airport (in Osaka, Japan), the International Airport of Hong Kong, the Burj Al-Arab Hotel, as well as The World and Palm Island complexes in Dubai. The aforementioned technique consists of tap into the seabed’s resources for the necessary materials and consequently placing those using GPS navigational systems. Once the sand layer is in place, construction continues using rocks up to 70 kg each. When the desired height has been reached a 456
Table 1c. Deck data. Material Total length of the deck Length of the suspended deck Total width of the deck Width of the deck per direction Weight per meter Total weight
Steel S355 4 km 280 m 38,6 m 16 m 11.75 ton 52.875 ton
Table 1d. Suspension system Material Number of cables Cables’ dimension Cable Type No. of strands per cable No of wires per strand Total weight of cables
Figure 5.
HSS steel (1200 MPa) 52 D = 11 cm DG-P12 12 7 cold galvanized wires 107.5 ton
Rendered image of the main arch of the bridge and the suspended deck.
raised kerb made of 3000 kg rocks is put in place in order to contain the rocky layers below and to serve as a wave breaker, where necessary, preventing the island from collapsing. Between each layer special geotextiles are used, to reinforce the stability of the island and contain the material in place. In order to build new structure on the island the terrain needs to be compacted using heavy machinery, inserting a series of vibrators that are used to solidify the sand, thus creating an ideal base for any type of construction (Fig. 5). 2.6 Construction The construction needs to take into account the sensibility of the area in which the project is to be constructed and preserve the environmental impact as low as possible. Therefore, the assembly pieces need to be manufactured on dry land, subsequently transported in place and mounted upon the foundations using specialized heavy machinery. Above in Tables 1a–d the most important technical data of the project are listed. 3 SUMMARY After the completion of the Egnatia Odos, the traffic in Thessaloniki’s metropolitan area has become an issue of great importance. The excessive growth of the number of vehicles that pass through 457
the area every day is the origin of great distress and discomfort for the citizens of the region. Even though a number of solutions have been proposed in recent years, none of them succeeded in resolving the problem in a satisfactory manner. This proposal aims to not only resolve the problem, by actually deviating all the traffic not directed to Thessaloniki’s center, but to do so by creating a new point of interest for the whole area. In order to achieve the goal, the deviation has to take place outside the limits of the metropolitan area. The new highway from west is attached to, the already in use, Athens-Thessaloniki highway at the Euzonoi intersection, then crosses the Thermaikos Gulf to connect to the already planned highway from Aggelohori to Moudania. The project described in this paper takes into consideration all aspects of this complicated area (with a high environmental value, being an intensive seismic zone, with existing infrastructures, and variable climate conditions) one by one and resolves them in the most thorough way, through intensive research and innovative architectural solutions. The outcome of this research is not only an effective deviation of traffic but also a project of high architectural value (Fig. 5). REFERENCES A.A.V.V. Rivista, Rassegna, “Il mondo dei ponti”, editrice Compositori, Brescia 1997 A.A.V.V., “Atlante del’acciaio”, UTET, Torino 2001 Albenga, G. I ponti. Torino: Unione Tipografico Editrice Torinese. Arcila, T.M. 2002. Ponti. Gribaudo. Arici, M. & Siviero, E. 2005. Ponti e viadotti. Palermo: Dario Flaccovio Editore. Ballio, G. & Mazzolani, F.M. 1987. Strutture in acciaio. Milano: Hoepli. Ballio, G. 1994. La progettazione in acciaio. Massa: Consorzio CREA. CNR 10011, “Costruzioni in acciaio: istruzioni per il calcolo, l’esecuzione, il collaudo e la manutenzione”, Poligrafico dello Stato di Roma, 1985 Dani, F. 1988. Il libro dei ponti. Pomezia: SARIN. Ermopoulos, I.X. 2000. Steel and Composite Bridges, Kleidarithmos. Eurocode 1: Basis of design and actions on Structures Eurocode 3: Design of Steel Structures Ferrari, P. & Giannini, F. 1994. Geometria e progetto di strade. Torino: UTET. Furiozzi, B. & Messina, C. & Paolini, L. 1978. Prontuario per il calcolo di elementi strutturali. Brescia: Le Monnier. Indovina, F. 1990. La città diffusa, Venezia: Daest. Jodidio, P. 2005. Piano: Renzo Piano Building Workshop 1966 – 2005. Cologne: Taschen. Karageorgis, A. 2001 General description of the Axios River catchment and the Gulf of Thermaikos. Athens. Karageorgis, A. & Anagnostou, Ch. & Georgopoulos, D. & Albuisson, M. 2000. Distribution of suspended particulate matter determined by in – situ observatoins and satellite images n the NW Aegean Sea. Athens. Leonhardt, F. 1979. I ponti: dimensionamento, tipologia, costruzione Milano: Edizioni di scienza e tecnica. Nicoletti, M. 1999. Sergio Musmeci: Organicità di forme e forze nello spazio. Torino: Collana Universale di Architettura no 54. Pertangeli, M. 1997. Progettazione e costruzione di ponti. Milano: Mason. Polano, S. 1996. Santiago Calatrava: Opera completa. Milano: Electa. Siviero, E. 1999. Il tema del ponte. Bologna: Compositori. Siviero, E. & Benedetti, A. 2002. La concezione strutturale nel progetto di ar chitettura. Bologna: Editrice Compositori. Spyridis, P. 2006. Analysis and Design of A Three-Span Composite Bridge with Open Girder Sections (Diploma Thesis). (in Greek) Torroja, E. 2002. La concezione strutturale. Torino: Città studi edizioni. Triantafillidis, I.D. 1966. City Planning of Thessaloniki. Thessaloniki. Thessaloniki: Ministry of Public Works. Troitsky, M.S. 1992. Orthotropic Bridges. Cleveland: James F. Lincoln. “Axios, Loudias, Aliakmon Delta.” Ramsar Sites Information Service. Web. 15 Nov.2011. . “Egnatia Odos S.A. – Project Identity” Ramsar Sites Information Service. Web. 12 Jan.2015. .
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Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
Segmental prestressed concrete multispan large bridges V. Barata, J.P. Cruz & P. Pereira Civil Engineer, LCW Consult, Lisbon, Portugal
ABSTRACT: For the past 30 years, a vast network of new transportation infrastructures has been constructed in Portugal, implying many significant crossings over wide rivers, deep valleys and large plains. These challenging circumstances provided the best opportunity to look for different structural concepts and new technologies, using segmental construction. This paper presents some cases of segmental prestressed concrete bridges designed by our company.
1 INTRODUCTION Throughout the following items, an overview focused on the structural concepts is made – beamand-slab, box girders, arches and cable stayed bridges – and in the construction technologies – precast beams, span-by-span construction, incremental launching, cantilever construction and stay cables. Some considerations are made about the cross sections and also about the seismicity and the correspondent anti-seismic strategy. The bridges presented are made of concrete elements or segments, assembled by reinforcement and post-tensioning. The choice between precast and cast-in-place solutions depends on site conditions, the project size, access constraints, and precast facilities in the close vicinity to the construction site, schedule construction and equipment available. 2 BEAM-AND-SLAB BRIDGES 2.1 Precast post-tensioned T-beams N10 – Santa Iria da Azóia Viaduct [traffic opening 1998]: located in Lisbon, parallel to Tagus River, has a total length of 1890 m, in a high seismic region. The viaduct allows the crossing of two water courses, North Railway line and road N10. Its width varies between 24 m and 42.65 m. The viaduct incorporates 525 post-tensioned T-beams (50000 m2 of deck) mounted by crane and connected by a cast-in-situ slab, poured above pre-slabs. Longitudinally, it is divided in seven continuous and independent structures with a length of about 300 m each, in order to solve the problems caused by temperature variation and seismic actions. 2.2 Precast pre-tensioned T-beams Vale Flores Viaduct [traffic opening 2006]: located at A10 motorway, in Lisbon region, with a total length of 1180 m. Due to the relevance of the seismic actions and to the geologic-geotechnical characteristics of the foundation, it was considered convenient to build two separate structures: the West Viaduct, 459
Figure 1.
Santa Iria da Azóia Viaduct.
Figure 2. Vale Flores Viaduct.
about 592 m long, and the East Viaduct with a length of approximately 587 m, both with current spans of 35.2 m. In the transverse direction there are two parallel decks, with a total width of 36.13 m. Each deck consists of five precast pre-tensioned reinforced concrete T-beams, mounted by means of a travel crane, which continuity is established over the supports by post-tensioning, applied on the slab of the deck. This 0.25 m thick deck slab was cast over pre-slabs. At both abutments viscous anti-seismic devices connected to the deck were installed. Ribeira Chã Viaduct [traffic opening 2011]: located at the road N1-1A in São Miguel Island, Azores, in a high seismic region. It has a total length of 310 m with current spans of 40 m. It is 18 m wide and is formed by 5 pre-tensioned T-beams with a span of 40 m, connected by a 0.25 m thick upper slab. Precast beams are mounted by means of a launcher. Structural continuity is assured by reinforced concrete over the supports. Ductility criteria has been used to detail the reinforcement of the piles. 2.3 Precast pre-tensioned U-beams Leça Viaduct [traffic opening 2005]: located at A41 motorway at North, spans Leça River and a road. The total length is 650 m and the current spans are 30 m long. The transverse platform of 34.20 m is formed by two similar decks, each one with two Ubeams connected by the upper slab. Beams are mounted by means of a travel crane. Pre-tensioning was applied in the first phase for the purpose of transport, mounting and supporting total deck self-weight. Post-tensioning ensures the structural continuity of the deck, in the second phase. 460
Figure 3.
Ribeira Chã Viaduct.
Figure 4.
Leça Viaduct.
Figure 5.
Bridge over Ribeiro da Ponte.
2.4 Span-by-span with launching gantry Bridge over the Ribeiro da Ponte [traffic opening 2006]: located at A7 motorway nearby Vila Pouca de Aguiar. The bridge is made of two independent structures, each carrying one carriageway. The extrados structure has a length, between end supports, of 338 m, while the intrados structure has a length of 424 m. The main span is 40.5 m long and each deck consists of two prestressed concrete beams, with a constant depth of 2.85 m, connected by a reinforced concrete slab. Diaphragms were placed at support sections to interconnect the longitudinal beams and center the sections over the piers. The deck is supported by fixed pot-bearings at the 3 highest shafts and by unidirectional sliding pot-bearings at the other piers and abutments. The decks were built, “in situ”, in a single phase using a launching gantry. Louredo Viaduct [traffic opening 2011]: located at A32 motorway nearby Oliveira Azeméis is 510 m long, with current spans of 44 m and 38.8 m wide. 461
Figure 6.
Louredo Viaduct.
Table 1. Beam-and-slab bridges/viaducts. Structural Concept Bridge/Viaduct Beamand-Slab
Lenght Width Main Transverse (m) max (m) span (m) section
N10 – Sta. Iria Azoia 1888
42,8
38
A10 – Vale Flores
1180
36,1
35,2
N1 – 1A – Rib. Ch˜a S. Miguel Island (Azores) A32 – Louredo
310
18,0
40
510
38,8
44
A7 – Rib. Ponte
424
30,1
40,5
A41 – Leça
650
34,2
30
postensioned T-Beams pretensioned T-Beams pretensioned T-Beams
Deck Sismicity/ construction Antiseismic method strategy crane
High/D
crane
High/D
launcher
High/BI
π – shaped deck span-by-span Low/EB launching girder π – shaped deck span-by-span Low/EB launching gantry pretensioned crane Low/EB U-Beams
BI – Base isolation D – Ductility VD – Viscous damper EB – Elastic behahviour
The overall width is achieved by two decks, each one with two webs, cast-in-place. The deck is built span-by-span, by means of a launching girder. 3 BOX GIRDER BRIDGES 3.1 Incremental launching The Bridge over the Águeda River [traffic opening 1998]: located at national road N1 crosses the Águeda valley at a height of about 60 m, requiring the construction of a relatively tall and long engineering structure since the valley is fairly open. The bridge is 855 m long between the axes of supports at the abutments and the deck is 15.3 m wide with a 2% slope. The current spans are 51.5 m long. The bridge is curved in plan (R = 3999 m) and in longitudinal section (R = 30000 m). The deck consists of a prestressed concrete box-girder, with a height of 3.7 m and was erected using the incremental launching method by lift-push. For the launching stage, the prestress is bonded and consists of tendons and bars into the slabs. In the second stage, external prestressing cables are installed, the necessary for the bridge service. 462
Figure 7.
Bridge over Águeda River.
Figure 8.
Bridge over Alcarache River.
The deck was built in standard segments of 25.75 m, concreted in a fixed shuttering located behind the right abutment. Piers are 60 m high, in general. 3.2 Span-by-span with launching girder The Bridge over the Alcarrache River [traffic opening 2000]: at municipal road M517, it has a total length of 965 m and crosses the reservoir of the Alqueva Dam, with current spans of 62.5 m. The deck consists of a prestressed concrete box-girder with 3.25 m at a constant height and 11.5 m wide. The piers have a maximum height about 50 m. When the reservoir of the dam attains its maximum level the higher piers will be submerged in more than 35 m high. The deck was constructed span-by-span by a lower launching girder. In one of the abutments two viscous anti-seismic devices were installed. 3.3 Balanced cantilever construction The Bridge over Ceira River [traffic opening 2014]: located at A13 motorway, near Coimbra, it crosses the valley through two distinct structures: the bridge itself and the access viaduct, in a total length of 930 m. The bridge is a continuous portal frame with a prestressed reinforced concrete deck, between abutment E1 and pier P4, with an overall length of 582 m and a main span of 250 m. The bridge deck is a single box girder with two inclined webs, with a total height of 14.5 m over the main piers, reducing to 5.5 m at mid central span. The top slab, prestressed transversally, is 26.40 m wide. The lower slab has a maximum thickness of 1.80 m over the main piers, reducing to 0.25 m at mid central span. Each one of these piers is about 100 m high, consisting of two shafts with a box section, 20 m apart between axes, the walls being rigidly connected to the diaphragms of the deck. The top of the main piers is 0.50 m eccentric to the interior of the curve, in reference to the bottom of the lower slab of the deck, in order to reduce the effects of the deck curvature. In order to reduce the effect of the permanent deck torsion, which produces bending moment on the main piers, vertical prestressing is adopted in the outer walls of these shafts, anchoring on the top end in the deck slab and on the bottom end in the micropile cap, where it is used also as a resistant tie. 463
Figure 9.
Figure 10.
Ceira River Bridge. Geometry control of cantilever construction.
Ceira River Bridge.
Table 2. Box girder bridges/viaducts. Structural Concept
Brridge/ Viaduct
Box Girder N1 – Águeda
Deck Sismicity/ Lenght Width Main Transverse construction (m) max (m) span (m) section method 855
15,3
51,5
M517 – Alcarrache 965
11,5
62,5
A13 – Ceira
26,4
250
930
Anti-seismic strategy
box girder Incremental Medium/D launching box girder span-by-span High/VD launching girder box girder balanced cantilever Medium/EB
BI – Base isolation D – Ductility VD – Viscous damper EB – Elastic behaviour
The main structure was built using the balanced cantilever method with pairs of segments concreted “in situ”, in each side of the main piers, the geometric control of which is shown. The foundations of the main piers are hybrid, distributing the load by the spread footings at surface and in depth through micropiles. These foundations are transversally eccentric in reference to the piers, ensuring an approximately uniform stress distribution in the soil. 464
Figure 11.
Zêzere River Bridge.
Figure 12.
Corgo River Valley Viaduct.
4 ARCHES 4.1 Cantilever construction with auxiliary cables and towers The Bridge over Zêzere River [traffic opening 1994]: at national road N348, it crosses the reservoir of Castelo do Bode dam by means of an arch bridge of reinforced concrete about 100 m high from the bed of Zêzere River. With a span of 224 m and a rise of 44 m, the arch girder of 6.0 × 3.75 m overall cross section, is a twin cell box and supports the superstructure by means of piers with centers spaced 21.5 m apart. The 10.5 m wide deck is a π-shaped cross section, 1.8 m deep and 385 m long. The arch construction started with a segment 7.0 m long cast over scaffolding. From the cast “in situ” extremities of the arched girder, nineteen 6.0 m long segments were cantilevered out from both sides – using two 30 ton travelling formwork carriages – and were joined by a short closing segment in the center. During construction the weight of the arched cantilevered segments was supported by temporary cables attached to rock anchors embedded in the hillsides. The spans of the access viaducts superstructure were cast at the same time. After the closing of the two halves of the arch, the cables used during construction were taken down and the deck was built. 465
5 CABLE STAYED BRIDGES 5.1 Balanced cantilever construction with definitive cable stays The Viaduct over the Corgo River Valley [traffic opening 2013]: at the A4 motorway, near the city of Vila Real, it stands, at its maximum height, 230 m over the river bed, with its deck describing a concave curve with vertical radius of 10.000 m and slopes of approximately 5%. It is a prestressed concrete viaduct with a total length of 2796 m, divided in 3 continuous Sub-Viaducts: West, Central and the East, measuring 855 m, 768 m and 1167 m, respectively. The Central Sub-Viaduct is a cable-stayed solution with central suspension in a half-fan shape, symmetrical in relation to the masts. In the masts, saddles are used, for deviating cable stays spaced 6.0 m apart along the deck and 1.20 m along the masts. The central external suspension is made possible due to the internal suspension of the box-girder webs by pairs of structural steel ties. The Central Sub-Viaduct has a central span of about 300 m, balanced by adjacent spans of 126 m and continuous spans extending through each side, resulting in a total length of 768 m. The decks have a central box girder 3.5 m high, with 0.60 m thick webs, 9.40 m apart between them, with overhangs supported by struts spaced at regular intervals of 3.0 m. The 28.0 m wide top slab is prestressed transversally, which complements the internal suspension action, helping the transfer of forces from the stays to the box-girder. The foundations of the towers are hybrid, distributing the load over the surface soil by spread footing and in depth through micropiles. 6 FINAL REMARKS In conclusion, it is possible to state a few main remarks regarding the segmental pre-stressed concrete large bridge, in the following terms: – It is an efficient, extensive and economical method for a large range of span lengths and structure types; – Allows a reduction of construction schedules, especially in the case of precast construction, where segments may be manufactured while the substructure is being erected, and rapidly assembled afterwards; – Requires an effective geometry control during the construction, based in a high level design taking into account not only the distinct stages but also the long-term effects on the materials; – Reduces the inconvenience to the population, and the traffic interferences during construction; – Increases workers’ safety.
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Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
Experience of some long multi-span bridges in Queensland, Australia (Part 1) J.A. Hart & E. Kittoli AECOM, Brisbane, Queensland, Australia
ABSTRACT: This is the first of two papers covering some long multi-span bridges in both urban and rural locations in Queensland, Australia. The New Gateway Bridge, renamed the Sir Leo Hielscher Southbound Bridge, is a major concrete box girder bridge, with a 260 m cast-inplace main span with navigation clearance of 55 m and a total length of 1,627 m, with typical 71 m approach spans. The development of the concept design is discussed, particularly with respect to the need for a similar appearance to the adjacent existing bridge while also introducing the contractor’s preferred construction methods. The new bridge has major structural differences in terms of cross section, some span lengths, articulation, and the erection method for approach spans. The bridge is notable for the specification of a 300 year design life, which led to a major investigation into corrosion mechanism theories, material properties, and construction variability, variability, as well as an education campaign for the construction workforce. The second paper will deal with viaducts in both rural and urban settings which have multiple spans of precast beams.
1 BRIDGE ENGINEERING IN QUEENSLAND This paper and its companion review AECOM’s experience in recent years on long multi-span bridges at three sites in Queensland, Australia. Queensland is about the size of western Europe but with 1.5% of the population. The large distances and low population density present one challenge in financing rural transport infrastructure. About 40% of the population live in or adjacent to the capital city, Brisbane, which has the traffic issues typical of moderately large cities. The climate ranges from temperate to tropical. Tropical cyclones occur in summer and bring high winds, storm surges, and flooding rains. In 2011 and 2013 widespread severe flooding followed initially cyclonic events. In terms of structural durability, the climate is quite benign. There is no need for de-icing of roads in winter. It is usually only coastal environments that present severe durability issues. Nevertheless, the Queensland Department of Transport and Main Roads (TMR) puts an emphasis on durability, with strong specifications for concrete structures and a disinclination to use steel bridges except in special situations. Australia is geotechnically mature and lacks high mountains, and Queensland was never glaciated. There are few rivers navigated by ocean-going ships. Queensland does not present many sites demanding long-span bridges, as opposed to long multi-span bridges. Australia is a developed country with high labour costs, boosted in Queensland in recent years by a boom in the construction of mining infrastructure. There is a strong preference for using precast pretensioned beams cast off site in standard forms. This context shapes the nature of the local bridge engineering industry. The influence of the above factors as well as functional requirements, location, loading, geotechnical and hydraulic conditions, and design life are explored with respect to the concept designs of three example projects, including the choice of span lengths and structural form in each case. The subject is presented in two papers. Part 1 concerns the New Gateway Bridge, and Part 2 the Yeppen Viaducts and the Logan Motorway Interchange Viaduct. 467
2 EXISTING GATEWAY BRIDGE The Existing Gateway Bridge (EGB) is a prestressed concrete box girder bridge carrying the Gateway Motorway across the Brisbane River, with a main span of 260 m. When completed in 1986, this was a world record span, and still ranks in the top ten. The bridge has 18 spans and a total length of 1627 m, with a navigation clearance of 53 m and a deck width of 21.9 m. The three main spans were cast in place in travelling formwork, and the approach spans were precast in full width segments, suspended from an erection truss, using the span-by-span method with 300 mm wide cast-in-place joints between segments. 3 NEW GATEWAY BRIDGE (SIR LEO HIELSCHER SOUTHBOUND BRIDGE) EGB was duplicated as part of a more extensive motorway upgrade project through a design-andconstruct contract. The client was Queensland Motorways Limited (now Transurban Queensland), on behalf of the Queensland Department of Transport and Main Roads (TMR). The contractor was the Leighton Abigroup Joint Venture assisted by VSL, and the designer was AECOM SMEC Joint Venture with subconsultant Cardno and verifier AAS Jakobsen. Coffey was the geotechnical designer. The tender commenced in 2005, the contract was awarded in 2006, and the New Gateway Bridge (NGB) was completed in 2010, when both bridges were renamed in honour of Sir Leo Hielscher, a distinguished Queensland public servant. The spans arrangement and articulation are shown in Figure 1. 4 DESIGN REQUIREMENTS One tender requirement was to match the appearance of the adjacent existing bridge. However, for NGB: – – – –
the deck is wider to include a shared path, or potentially a future lane the current SM1600 traffic loading is much heavier than the previous T44 loading the piers are protected against collision from specified ships design standards have changed in many ways
Figure 1.
Spans arrangement and articulation.
468
– 2000 year average recurrence interval for ultimate wind and earthquake loads – the design life for durability is 300 years. These are significant differences and, combined with the contractor’s construction method preferences, led to major differences in the structural form of the two bridges, while preserving their outward similarity. 5 SUPERSTRUCTURE At the concept design stage, cross section, construction method, span lengths, and articulation were strongly inter-related and they cannot be treated individually. The NGB main span superstructure depth varies from 15.7 m at the piers to 5.2 m at midspan, matching EGB very closely. A single cell as in the EGB main spans was found to be inefficient for the 26.9 m width of deck between barriers on NGB. Therefore the bottom flange was widened from 12.0 m to 15.0 m and a third web introduced. Inclined webs were considered, but it was found that formwork complications countered material savings. Inclined, variable depth webs result in variable width of the bottom flange, which was considered to be an undesirable and unnecessary visual difference from EGB. The deck slab is transversely stressed. The extra web and wider web spacing enhance shear and torsion capacity, reducing the demand for vertical web prestressing, and providing more anchorage locations for the main tendons. The shear and torsion design was based on the AASHTO LRFD 2006 code because the Australian code does not deal well with this aspect of box girders. The cross section is shown in Figure 2. Cast-in-place cantilever construction is the only practical method for spans of this scale. Each cantilever comprises 30 segments, with equal numbers of 3.0 m, 4.0 m, and 5.0 m long segments, so that the form traveler capacity was utilized efficiently. The top flange, being governed by the transverse action of the deck and the need for prestressing anchorages, was constant along the bridge. The bottom flange and the webs were stepped down in thickness with distance from the piers. Longitudinal prestressing was arranged so that the new segment on one side of the cantilever was anchored in the previous segment on the other side. Longitudinal reinforcement was designed to enable launching of the form traveler before prestressing of the just cast segment. Both these features contribute to shortening of average cycle time. The governing limit for longitudinal prestress was generally the requirement for zero tension under permanent loads plus 50% of liveload. Cantilever prestress comprised tendons of 17 or 19 15.2 mm strands, with six tendons typically anchored at each segment. Midspan prestress was anchored at bottom flange blisters in the main span. The contractor had a strong preference to erect the typical 71 m long approach spans by the balanced cantilever method. EGB had been erected span-by-span and included an 88 m long final
Figure 2.
Main spans cross section.
469
Figure 3. Approach spans cross section.
approach span which had required a temporary pier. There was a very neat solution by simply making the final approach span 71 m and increasing the main side span from 145 m to 162 m. Then the balanced cantilevers almost met with just enough gap for a cast-in-place diaphragm segment connecting them. This decision was made with some trepidation because it was not strictly conforming, but it expanded rather than diminished the main spans and was accepted. The approach span piers of EGB have pinned bearings top and bottom, with longitudinal restraint only at the abutments and the river piers, and the only movement joints being between the approach spans and the main spans. Such pinned piers require temporary fixing during superstructure erection, and eventual replacement of the bearings is rather difficult. For NGB the piers are monolithic top and bottom and there are movement joints at halving joint segments in the final approach spans adjacent to the main side spans and at an intermediate span on the longer northern approach. The halving joints were temporarily fixed to enable balanced cantilever erection. Interestingly, the final approach spans and piers become continuous with the main spans, following release of the halving joints. This was a complication for design, but facilitated both construction and maintenance. The result then is a bridge which consists of four multi-span continuous frames separated by movement joints. It follows that shrinkage, creep, and thermal strains will induce stresses in the indeterminate structure, as well as govern the range of movement at the joints. For the approach spans, the cross section comprises two match-cast segments subsequently connected by a stitch pour, as shown in Figure 3. Match casting was originally prohibited on durability grounds because of some problems encountered at that time on some older match cast bridges in the United Kingdom. Match-casting offered significant program advantages and was offered as a non-conforming alternative. (The conforming offer was for a cast-in-place balanced cantilever scheme.) Match-casting was eventually accepted subject to minimum residual compression at the joints of 1 MPa in service, the detailing of a small recess at the top of the joint which was filled with epoxy from the deck to ensure there was no gap in the squeezed epoxy, and the provision of a second layer of waterproofing over the joint. The typical approach span depth is 3.3 m, rather than the 3.0 m of EGB, to enable the minimum residual compression to be achieved. The dual segment cross section, as shown in Figure 3, and the dimensions of the segments were influenced by the contractor’s possession of sets of formwork from a previous project. The smaller segment size also was favourable for erection by crane and erection gantry. Again, design complications were encountered in order to achieve these construction advantages. To align the outer webs of the approach spans and the main spans, the precast segments are strongly asymmetric, with an inclined neutral axis and non-uniform stresses across the flanges from self-weight and prestressing. Once the stitch pour was completed, the overall section became symmetric for subsequent loading. This is an example of a design outcome being tuned to the particular preferences of a contractor and which does not necessarily apply in general. Table 1 summarises the differences between EGB and NGB. 6 SUBSTRUCTURE The river piers have twin blade columns, typical of balanced cantilever bridges, to provide stability for cantilevering and flexibility for the continuous structure in service. The pier head is monolithic 470
Table 1. Comparison of Existing Gateway Bridge and New Gateway Bridge. Design or Construction Aspect Main spans lengths Main spans cross section Bridge Width Main spans construction method Approach spans lengths
Approach spans cross sections
EGB
NGB
145 m, 260 m, 145 m Single cell box with 12 m wide bottom flange 23 m Cast-in-place balanced cantilever 60 m (abutment end) 71 m (typical internal) 88 m (adjacent to main spans, requiring temporary pier) Twin-cell precast box with wide, reinforced cast-in-place joints.
162 m, 260 m, 162 m Two cell box with 15 m wide bottom flange 28 m Cast-in-place balanced cantilever 60 m (abutment end) 71 m (remainder)
Approach spans construction method Longitudinal fixity Movement joints
Span-by-span construction with concrete infill joints Abutments and river piers 2 No. (in main side spans) with load transfer by steel needle beams)
Bearings
Abutments, top and base of all approach piers, needle beams)
Two single cell match-cast precast boxes with epoxy joints and a longitudinal stitch pour. Balanced cantilever with match cast epoxy joints All piers 5 No (at abutments and in 3 No, approach spans) with load transfer through halving joints. Abutments, halving joints
with the superstructure. The blade columns and the girder webs extend through the pier head as diaphragms and walls. Prestress is provided in three directions. The EGB river piers have pilecaps set low in the water, which were built in cofferdams. The contractor saw this as an expensive and slow operation. In parallel, the question of how to protect the piers against ship impact was leading to providing a submerged island of rockfill around the piers. These two aspects merged to produce a scheme with piles and pilecaps installed on rockfill islands above water, with causeways to the river banks, so that the whole operation became terrestrial rather than maritime. The extra rockfill used during the NGB pier construction was subsequently displaced to form part of the submerged protection island for EGB. There was thus a substantial saving in temporary works for the river piers. Cast-in-place bored piles were appropriate for the river piers. Preliminary designs were produced for pile diameters ranging from 1.5 m (as used on EGB) to 2.5 m. Taking local plant availability into account, the selected diameter was 1.8 m, with 24 piles per pier and maximum ultimate pile load of 35 MN. The design intent was to set out the piles as directly under each blade column as practical to reduce the load distribution action and the dimensions of the pilecap, even though greater pile depth was needed for the same pile capacity. The pile spacings are 3.6 m along the bridge centerline and 3.3 m transversely. The piles pass though alluvial sediments to socket into interbedded siltstone and sandstone at depths of 25 m to 55 m. The client initially required that the pile construction be verified by a geotechnical engineer descending into the pile to inspect the socket and the base, as had been done on EGB. The contractor had strong resistance to this, based on workplace health and safety legislation, requiring hazards to be avoided if practical. This was a major topic of negotiation during the tender assessment period. Eventually the absence of direct inspection was accepted subject to multiple measures to assure pile capacity: – a borehole being undertaken on the centreline of every pile – more conservative socket and base properties adopted in geotechnical design – sacrificial test piles tested to near failure by the Osterberg method 471
– visual inspection of every pile by a video camera apparatus specifically set-up for that purpose and able to operate underwater, and – testing of one production pile at each pier to 1.2 × design load. The testing and inspections confirmed the adequacy of the design. The approach piers have blade columns with three internal cells, and most are supported on large groups of precast pretensioned 550 mm octagonal piles (up to 45 in number). These piles are a standard component in Queensland. The four southernmost piers are founded on spread footings because the rock is high. Piers 14 and 17 are the end piers in a length of continuous superstructure. These piers are not sufficiently high for the columns to provide the desired flexibility to accommodate shortening strains in service, and they are therefore supported on a single line of cast-in-place piles. With pile lengths typically in excess of 35 m, it was necessary to splice the piles. The accepted splice detail in Queensland has dowel bars cast in the toe of the upper pile segment fitted into holes formed in the top surface of the already driven lower segment, with an encircling steel collar and all gaps filled with epoxy. 7 300 YEAR DESIGN LIFE A 300 year design life was specified, following a study by TMR which indicated that such a life would maximize economic benefit for a major piece of infrastructure. This challenging specification is believed to be unprecedented. By necessity the design process to address this unique requirement went beyond the deemed-to comply approach of the codes, where materials performance is assumed, and the strength of the materials and the concrete cover is defined by an adopted exposure classification. The process adopted took a first-principles and deterministic approach to modelling the environmental influences and material performance. This was supplemented by a probabilistic approach using Monte Carlo simulation of the rate of deterioration based upon published data on the variability of material properties and construction quality. There was much recourse to the disparate research literature The design process for addressing durability is summarised as follows: – – – – – – –
Define the characteristics of the environment Identify the potential deterioration mechanisms in that environment Determine the likely rate of deterioration Assess the material life Define the required material performance Take a probabilistic approach to the variability of the relevant parameters Assess and define the need for further protection
The bridge site is 7 km from the mouth of the tidal Brisbane River. The water is virtually seawater, and chloride ingress was a particular concern for the durability of the main pier pilecaps. Testing of core samples from pilecaps on EGB indicated surface chloride concentrations of 0.4 to 0.5% by weight of concrete after 20 years. A worst case surface chloride concentration of 0.65% by weight of concrete was used in the analysis and is similar to published data for marine splash environments. Testing was performed on trial mixes of the proposed concrete to establish a chloride diffusion coefficient. Modelling of chloride ingress (as shown in Figure 4) was subjected to the probability analysis with the finding that ordinary reinforcement required a cover of 150mm to achieve a 90% probability of no corrosion initiation before 280 years. The pilecap was dimensioned and the structural design carried out on this basis. Such a large cover would lead to wide surface cracks. To control such cracking, a relatively light mat of LDX 2101 stainless steel reinforcing was placed at 75 mm cover in the main pier pilecaps. A 50 MPa ternary blend concrete consisting of 30% fly ash and 21% blast furnace slag was used in these elements. The piles of the river piers were cast in sacrificial steel liners. Such liners are standard on TMR projects in Queensland because they promote durability by preventing collapse of soil strata into 472
Figure 4.
Chloride diffusion curve before variability analysis.
the bored hole of the pile and by facilitating accurate installation of the reinforcement cage. The liners are not relied upon for strength because they are liable to corrode. This principle was adhered to on NGB, but the liner was designed to be much thicker than usual, and its design life contributed to the total life of the piles. In order to avert alkali silica reaction, a minimum of 25% flyash was provided in all concrete. A disadvantage of fly ash is increased carbonation rate for concrete with equivalent water/cementitious material ratio and total cementitious content. Published data for similar mix designs to those proposed was used to estimate the carbonation rates for predictive purposes. The rates selected were 4.5 and 3.0 mm/yr0.5 for S40 and S50 concrete with 25% fly ash, respectively. An atmospheric CO2 concentration of 0.04% was assumed. The depth of carbonation predictions suggested that 55 mm cover to exterior surfaces of superstructure elements would be acceptable in preventing premature carbonation and corrosion initiation provided an S50 concrete or other mix with a maximum carbonation rate of 3.0 mm/yr0.5 was used. It was recommended to monitor the structure for carbonation after 250 years and to apply a barrier coating if necessary. In addition to reinforcement corrosion due to chloride ingress or carbonation, the durability assessment for the second Gateway Bridge also considered sulphate attack, alkali aggregate reaction and acid sulphate soils. The northern approach substructure concrete was isolated from potentially aggressive soil or groundwater, by epoxy coating of driven piles to a certain depth and removing and replacing soil adjacent to the pilecaps. To provide a further option for future durability enhancement, the substructure reinforcement cages were made electrically continuous with access terminals to enable monitoring of electric potentials and installation of a cathodic protection system. The 300 year design life would not be achieved without control applied by the construction team to realise the design assumptions of the durability plan. Trial concrete mixes were tested to ensure the chloride and carbonation diffusion coefficients were achievable, and the concrete cast was tested during production to verify that the properties were achieved in the structure. Apart from using high performance concrete, the single most important element in achieving durable structures is to ensure the correct thickness of high quality cover concrete. The design must ensure the detailing is practical enough to prevent reinforcement congestion, but the construction must ensure that the requisite cover is achieved, the compaction of the concrete achieves a good layer of dense cover concrete, and the concrete is sufficiently cured to prevent any early age cracking or incomplete hydration. 473
Figure 5.
3 C’s logo.
The contractor undertook an extensive education program with a logo of 3 C’s to represent Cover-Compaction-Curing as well as the Roman numeral for 300 to encourage commitment from all workers to achieve the 300 year life. 8 SUMMARY NGB is a case study of the design-and-construct process, where the concept design is heavily influenced by the contractor’s construction method preferences as well as the client’s mandatory requirements. The duplication of an existing bridge imposed constraints but also created a situation where the advances in both design and construction over a period of 25 years could be appreciated. The process also offered the opportunity to propose alternative features to the client which could be shown to satisfy technical requirements and to provide value. NGB looks similar to EGB, but is structurally very different, particularly in the approach spans. Design complications were taken on in order to make the construction efficient in cost and time. NGB was designed and constructed in less than four years. NGB is unique in having a 300 year design life, which led to a major investigation into corrosion mechanism theories, material properties, and construction variability, as well as an education campaign for the construction workforce. ACKNOWLEDGMENTS The contributions of the various entities involved in the project, as listed in section 3, are acknowledged. This paper expresses the opinions of the authors and does not necessarily represent the views of any of those entities. REFERENCES Connal, J, & Berndt, M. 2009 Sustainable Bridges – 300 Year Design Life for Second Gateway Bridge, Austroads Bridge Conference, Auckland. Connal, J, Wheeler, K, Pau, A, & Mihov, M. 2009 Design of the Main Spans, Second Gateway Bridge, Brisbane, Austroads Bridge Conference, Auckland. Hart, J, Connal, J, & Berndt, M. 2011 The Influence of Concrete Material Behaviour on the Design, Construction, and Durability of the Sir Leo Hielscher Bridge, Concrete in Australia, Vol 37, No 4. Hart, J, Muccillo, J, & Connal, J. 2009 Design of the Approach Spans to Second Gateway Bridge, Brisbane, Austroads Bridge Conference, Auckland.
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Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
Experience of some long multi-span bridges in Queensland, Australia (Part 2) J.A. Hart & E. Kittoli AECOM, Brisbane, Queensland, Australia
ABSTRACT: This is the second of two papers covering some long multi-span bridges in both urban and rural locations in Queensland, Australia. The first paper dealt with the New Gateway Bridge (renamed the Sir Leo Hielscher Southbound Bridge), a major concrete box girder bridge. This second paper covers two projects. The Yeppen Floodway Upgrade Project comprises three viaducts with a total length of 2.6 km crossing a flood plain near Rockhampton in Central Queensland. The Logan Motorway Interchange Viaduct is on the Springfield commuter railway line on the outskirts of Brisbane. All of these viaducts have superstructures comprised of pretensioned precast deck units or T-beams, each of a form which is economical and widely used in Queensland. Still there were choices to be made amongst these standard forms, affected by alignment, hydraulics, loading, and substructure options, as well as cost and availability. 1 INTRODUCTION This paper and its companion review AECOM’s experience in recent years on long multi-span bridges at three sites in Queensland, Australia. The subject is presented in two papers. Part 1 describes some general aspects of bridge engineering in Queensland and covers the New Gateway Bridge. This Part 2 paper provides some further information on loading and typical bridge types and then describes theYeppen Viaducts over a major floodway in Central Queensland and the Logan Motorway Interchange Viaduct for a railway on the outskirts of Brisbane. 2 TRAFFIC LOADING SM1600 has been the Australian highway traffic loading since the publication of the Australian Standard AS 5100 in 2004. The loading has different cases for stationary and moving traffic, each comprising combinations of four triple axle loads plus a uniformly distributed lane load. Both cases amount to a total load of about 1600 kN over a length of about 25 m. The ultimate load factor is 1.8. This loading does not represent a particular type of truck but is intended to produce an envelope of load effects which is appropriate for a range of spans and a range of very heavy trucks. History shows that truck loads have increased over time, and the intention is that this loading is an upper bound of feasible loading that might remain valid through the 100 year design life of current bridges. By international standards, it is a heavy loading. Major highway bridges are also designed for HLP 400, a heavy load platform totaling 4000 kN on 16 axles over a length of 27 m, with an ultimate load factor of 1.5. 3 STANDARD STRUCTURAL COMPONENTS FOR ROAD BRIDGES The current practice for the design and construction of typical road bridges to Queensland Department of Transport and Main Roads (TMR) standards is to select from the following structural components and tailor the design to the particular project. 475
Figure 6. Transversely stressed deck units.
Figure 7.
Deck unit with deck slab.
Figure 8. T-beams with composite slab.
3.1 Transversely stressed deck units These are rectangular pretensioned precast elements with a constant width of 596 mm and variable depth from 300 mm to 1050 mm covering spans up to 25 m. Gaps between the units are grouted and transverse stressbars are installed at 2 m spacing. A typical bridge cross section is shown in Figure 1. This bridge type, though used throughout the state, is particularly suited to remote areas to minimize the site concrete work. 3.2 Deck units with deck slab As an alternative to transverse stressing, the deck units may be connected by a deck slab which acts compositely for subsequent loading. The overall structural depth is increased and the cost may be higher, but this bridge type, as shown in Figure 2, is commonly used where a higher performance barrier is required, the deck width varies within a span, the alignment is more than gently curved, or durability is an issue. The typical range of spans is as for transversely stressed deck units. 3.3 T-beams with Deck Slab These are pretensioned precast beams, most commonly having an open-topped section with inclined webs and cantilever flanges, as shown in Figure 3. Standard depths are 1200 mm, 1500 mm, and 1800 mm, used for spans of 25 m, 30 m, and 35m approximately. The bottom flange and webs are standard; the flange widths are adjustable so that the flanges of adjacent beams abut and can be 476
sealed with tape for casting the deck. A sacrificial form is placed between the webs before casting the deck. 3.4 Precast Prestressed Concrete (PSC) Piles These are pretensioned precast octagonal piles with a dimension between opposite faces of 450 mm, 500 mm, or 550 mm. The largest section is most common. The longest length of pile which can be lifted and transported is 27 m. Where longer piles are needed, they are spliced with a stainless steel collar and dowels. The pile capacity derives from both shaft friction and end bearing and is confirmed by pile driving analyser (PDA) testing. 3.5 Cast-in-Place (CIP) Piles CIP piles are reinforced concrete piles. TMR specifies that they be constructed in permanent sacrificial steel liners extending to a founding stratum which is suitable for rock socketing. The liner prevents hole instability, ensures the full cross section of the pile, and enables accurate placement of the reinforcing cage. Shaft friction on the liner is neglected in design because of the possible eventual corrosion of the liner. CIP piles are commonly used with diameters up to 1500 mm, with occasional use of larger diameters. CIP piles can carry large axial and bending loads and can often be extended directly into a headstock eliminating the need for a pile cap. There is construction risk related to rock socket roughness, cleanliness of the base, and the placement of concrete in wet holes. 4 YEPPEN FLOODWAY UPGRADE PROJECT The Bruce Highway is the principal highway that connects the major towns and cities along the east coast, north of the Queensland capital, Brisbane, which lies in the far south. The city of Rockhampton in Central Queensland is built on the banks the Fitzroy River, approximately 600 km north of Brisbane, and has a population of around 83,000. Immediately south of Rockhampton, the Bruce Highway crosses over the Yeppen floodplain and the Yeppen Lagoon, part of the Fitzroy River catchment. This section of the highway is approximately 3 km long and provided a single lane in each direction on a low level embankment with culverts and short bridges across theYeppen Lagoon and the prominent water channels. It had flood immunity for a 15 year Average Return Interval (ARI) flood event and was inundated during major or moderate flood events resulting in closure of the highway. For instance, this section of the Bruce Highway was closed for fourteen days in the 2011 major flood, and for eleven days during the 1991 flood, causing freight, economic and social disruptions. In addition, Rockhampton Airport and the North Coast Railway also become inoperative during such events leaving the city and north Queensland isolated from the south. To minimize this disruption during such flood events, the Yeppen Floodplain Upgrade Project is being undertaken by TMR primarily to provide immunity for flood events with up to 100 year ARI and to duplicate and upgrade this section of the highway. The project is being undertaken in two phases and delivered through two separate projects: Yeppen North and Yeppen South. 5 YEPPEN NORTH This first stage of the Yeppen floodplain upgrade involved duplication of the existing Yeppen Lagoon bridge with a higher level bridge and upgrade of the Bruce Highway, south to the Yeppen Roundabout and north to Port Curtis Road. The existing bridge over the Yeppen Lagoon is a 200 m long, two-lane bridge with 10 m spans, consisting of transversally stressed deck units with PSC piles. The existing bridge now forms the northbound carriageway of the Bruce Highway and the 477
new Yeppen Lagoon Bridge the southbound carriageway. There is a roundabout intersection with the Capricorn Highway 130 m south of the bridge and it was necessary to maintain existing levels at the roundabout. To improve the capacity of the roundabout, a southbound slip lane is provided to bypass the roundabout, requiring widening of the southernmost 120 m length of the bridge. The slip lane is initially blocked until it is connected to the slip lane bridge being constructed in the next phase of the highway upgrade. A hydraulic study showed that a bridge length of 420 m is required, with the bridge soffit above the 100 year ARI flood level. Road alignment considerations placed the bridge approximately 14 m downstream of the existing road bridge and 120 m upstream of an existing railway bridge. It is therefore hydraulically efficient to align the new piers with the existing road bridge, which pointed to practical span lengths of 10 m or 20 m using deck units or 30 m using T-beams. Geotechnical investigations showed the typical soil profile at the site could be divided into two layers: an upper layer, 5 m to 8 m thick, of surficial clay and a lower layer of granular strata consisting of sand, sand and gravel, and clayey sand and gravel, generally becoming stronger with depth. Bedrock was located at a depth of around 45 m. This geology lends itself to friction piles, and PSC piles and CIP piles were considered. PSC piles were preferred over CIP piles because the soil profile is suitable for driven piles and because the CIP piles would be much deeper and entail construction difficulties at this location . In particular, driving the steel liner to the deep rock and sealing against water ingress would be difficult due to the dense but permeable materials overlying the rock. The superstructure type, span length, and substructure type were explored in the preliminary design, taking into account flood performance requirements, hydraulic assessments, site geology, and interaction with the existing bridge to determine the optimal bridge span and substructure type arrangement. The superstructure depth was limited by vertical alignment required to tie into the low level roundabout and the bridge soffit level necessary to provide the waterway area to achieve the 100 year ARI flood immunity requirement. A 30 m span satisfied the structural depth requirement but substructure loads were too high for an efficient piling layout. hence the 20 m deck unit was adopted, with 21 spans. Construction costs are lower for transversely stressed deck units (Fig 1) than deck units topped with a reinforced concrete slab. However, spans 1 to 6 require a concrete deck (Fig 2) to accommodate the varying deck width due to the slip lane. The piers are comprised of reinforced concrete headstocks, two circular columns (three for the wider section) and pile caps supported on 550 mm octagonal PSC driven piles. The abutment substructures are comprised of reinforced concrete headstocks on raked PSC driven piles. Both abutments have spill through batters with reinforced concrete embankment protection. The strength of the ground material, particularly the upper weak materials at shallow depths has an impact on the articulation of the bridge. The stiffness and strength of the substructure consisting of columns, pile cap and piles allows the longitudinal movement and loads to be managed. Movement joints are provided at every third pier, with longitudinal continuity of the superstructure provided at the other piers. A pile cap is not required at the abutments as the pile height above ground and span loads are lower than that at the piers. The longitudinal movements and load are accommodated with raked PSC piles at the abutments. Construction of the Yeppen North Project was completed in November 2013. 6 YEPPEN SOUTH The Yeppen South Project is for the further upgrade of the Bruce Highway across the floodplain south of the Yeppen Roundabout and is currently under construction. The project includes the construction of two bridges: Yeppen South Bridge, which forms the northbound carriageway of the Bruce Highway, and the Slip Lane Bridge, which connects to the Yeppen Lagoon Bridge bypassing the low level intersection at the Yeppen Roundabout. The north and southbound carriageways are connected through a high level gated transition embankment north of the Yeppen South Bridge which can be opened during a flood to form a bi-directional roadway across the floodplain. The 478
Yeppen South Bridge is adjacent to three small existing bridges on the Bruce Highway, all of which are 20 m span deck unit bridges. The North Coast Rail line runs parallel to the Bruce Highway, along the eastern side adjacent to the embankment for the southbound carriageway. The centreline of the railway is between 15 m to 125 m from the centreline of the southbound carriageway. Both bridges have a deck width of approximately 11.5 m and carry two lanes of traffic. The Yeppen South Bridge has a stopping bay near mid-length where the deck width increases to 15 m. The Slip Lane Bridge has a higher performance concrete barrier for 120 m from the southern abutment because of the closeness of the railway line. Further hydraulic analyses for the Fitzroy River catchment area were undertaken to determine the required bridge length, positioning, and deck soffit levels to achieve the requited immunity and minimise the afflux at critical locations such as the airport and the North Coast Rail. The outcome is a bridge length of 1645 m for Yeppen South Bridge and 545 m for the Slip Lane Bridge. The hydraulic analysis also determined parameters required for the bridge design: the flood levels, flow velocities, and scour. Geotechnical investigations show the ground generally consists of an upper layer of peat or loose silty clay underlain by predominantly granular strata with clay and gravel layers which overlie the sandstone bedrock at depth. The thickness of the granular strata and depth to rock vary significantly along the lengths of the bridges, between 15 m at the southern end of the Yeppen South Bridge to 35 m at the northern end of the Slip Lane Bridge. At some locations in the floodplain, boreholes were drilled to a depth of 69 m without encountering competent rock. Superstructure and foundation options were investigated to arrive at the optimal solution. Options for short and medium span length arrangements were considered. Short span options included deck unit spans up to 25 m and T-beam spans up to 35 m. These systems are labor intensive due to the construction on-site of the RC deck on the precast beams and there is delay while the deck concrete gains strength before applying construction loads, possibly leading to increased construction time. However, the beams can be erected using large cranes. Medium span options were box girders, either whole span or segmental, which can be manufactured ahead of erection and stored on site for the period required for developing concrete strength. The constructed box girder spans may be used for delivery of remaining girders at the deck level upon completion of post tensioning and grouting. Box girders have deeper structural depth, resulting in a raised vertical alignment and higher approaches. They are suited for bridges with a high vertical alignment, require specialist construction, and incur high setup costs for casting yards and erection gantries. Generally they are cost efficient for bridges of a scale which justifies the establishment costs. The following foundation options were considered for both bridges: CIP piles, PSC piles and composite piles. CIP and PSC piles are as discussed previously. Composite piles consist of a steel liner with a concrete core and shear keys to transfer the loads between the liner and concrete core. They do not need a rock socket and therefore may result in shorter pile lengths than CIP. For the Yeppen South bridges there were a number of risks associated with the composite piles: formation of a soil plug at locations where rock is shallow, hard driving, and corrosion of the structural steel liner limiting the design life of the bridge. Also composite piles required submission of a detailed proposal to TMR before they could be incorporated, which posed a risk to the very short design and construction program if they were not accepted. Since the ground conditions varied with bed rock at a significant depth, PSC piles were selected over CIP piles and composite piles as they would be shorter, cost less, better accommodate the risk from variable ground conditions, and their capacity could be confirmed through dynamic testing. The other benefit of the PSC piles is ease of construction over than their CIP counterparts. The length of the pile segments is limited to 27 m for transportation, and at some location piles are spliced. The bridge options were assessed for the cost implications and taking into account the associated risks, such as ground conditions and the potential for flooding at the bridge site as well as the whole of life cost. The 35 m span option was found to be the lowest cost solution. It provides the maximum span that can be constructed using standard forms of construction and erection and the optimum 479
Figure 9. Logan Motorway Interchange Viaduct.
balance between the cost of foundations and of superstructure. It differs from Yeppen Lagoon because the span length is not constrained by an adjacent bridge. The same structural form was adopted for both bridges to facilitate structural detailing, precasting of elements, and construction over the floodplain. The Yeppen South Bridge is 1645 m long consisting of forty-seven 35 m spans (Fig 3). The bridge has cast–in-situ kerbs and a steel post and rail barrier. The Slip Lane Bridge is 545 m long consisting of fifteen 35 m spans and one 20 m span. The superstructure for the 35 m spans on both bridges consists of 1800 mm deep T-beams with a 200 mm thick RC composite deck slab. Six beams are required for the Yeppen South Bridge with one more beam at the stopping bay. Seven girders are required for the Slip Lane Bridge to provide additional capacity for the concrete barrier and variable beam widths for the deck curvature. The 20 m span of the Slip Lane Bridge consists of transversely stressed deck units. The remainder of the bridge has cast-in-situ kerbs and a steel barrier. The piers for both bridges will be comprised of reinforced concrete headstocks, columns, and pile caps supported on PSC piles. The abutment substructures will be comprised of reinforced concrete headstocks on PSC piles, with concrete embankment protection over the spill through batter. The bridge articulation utilizes finger type expansion joints at a maximum of every five spans to accommodate the resulting range of movements. The joint system adopted incorporates as a standard detail underside draped PVC trough drainage seal to drain to a collection point provided at the low side of the deck which will prevent water ponding at the bearing shelf leading to corrosion damage of bearings and restraints and staining of piers. 7 LOGAN MOTORWAY INTERCHANGE VIADUCT 7.1 Background The Richlands to Springfield railway extends Brisbane’s suburban passenger train network to the outer growth hub of Springfield. The project was undertaken by the Trackstar Alliance, comprising Queensland Rail (QR), Thiess, AECOM, and Aurecon. TMR was a major stakeholder. The two-track railway crosses the interchange between the Centenary and Logan Motorways on a viaduct over 800 m long (Fig 4). The Centenary Motorway and the interchange are planned for future expansion, and the pier positions have to cater for both the current and future layouts of the interchange. Furthermore, there is a 69 degree skew crossing of a roadway at one end of the bridge. 7.2 Superstructure QR has a range of standard precast beams, but the maximum length of 25 m could not always span the gap between zones of valid pier positions, and the transversely stressed beams could not accommodate the necessary skew. A box girder option was considered but found to be too costly. 480
Figure 10.
Railway viaduct with T-beams.
Figure 11.
High skew crossing – plan.
On account of the restricted traffic, the railway loading was not the typical A300 loading to AS 5100, but A200, comprising multiple groups of four 200 kN axles. As it happens, the A200 loading is quite similar to the SM1600 loading for road bridges for the desired span lengths. The solution was to adopt TMR style T-beams, normally used for highway bridges, for this railway bridge. A typical span of 33 m was thus achieved (Fig 5), and this was stretched to 36 m where necessary by means of closely spaced beams supported on inverted-T headstocks. An unexpected issue arose with the design requirements for compression fatigue of the flanges of the precast T-beams. In the Australian code, there is a fatigue limit based on the peak compressive stress and not affected by the stress range. This differs from European and American practice. Consequently, the precast flanges were made thicker. The Australian code is currently being revised and suggestions made on this topic have been included in the latest draft. The high skew crossing, where vertical clearance was also a constraint, was achieved by means of a span with reduced skew of 45 degrees and excess deck width (Fig 6). The structure has deck units and a deck slab, with wide gaps between some of the deck units (Fig 7). Short, trapezoidal and triangular spans were used to resolve the difference in skew between the end span and the typical viaduct spans. The solution is not an elegant structure, but it is functional and economical. 7.3 Substructure Typical piers have single circular columns with a diameter of 2400 mm and height up to 20 m. The ground conditions are variable, and typically comprise 15 m to 20 m of mixed alluvial soils and extremely low strength mudstone and sandstone overlying very low strength sandstone. The horizontal constraint of adjacent roadways pointed to the choice of a compact group of a few CIP piles rather than a larger group of PSC piles, which were also at risk of pulling up with insufficient penetration at some piers. A group of four 900 mm diameter CIP piles was chosen. QR accepted 481
Figure 12.
High skew crossing – section.
piles that were designed for polymer fluid to stabilize the excavation before concreting. This was an economical solution which permitted shaft friction to be activated at a higher level than for a fully lined CIP pile. Pile integrity testing by a low impact stress wave method did not reveal significant defects in any of the piles. Two special portal frame piers were needed where no single pier position could satisfy both current and future road layonts. These piers were detailed to facilitate extension of the headstock, the construction of a new column, post-tensioning of the headstock, and the demolition of one of the initial columns. 8 SUMMARY The four viaducts demonstrate the range of standard bridge components and the site-related reasons for the choice of which system to adopt. The typical layouts were compared with some non-standard alternatives during design development and were found to provide the best value. ACKNOWLEDGMENTS The contributions of the various entities involved in the projects are acknowledged. This paper expresses the opinions of the authors and does not necessarily represent the views of any of those entities.
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Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
Multi-span bridges: The first Chilean experience and future challenges M.A. Valenzuela, M. Márquez & I. Vallejo Public Works Ministry, Santiago, Chile
ABSTRACT: The traditional typology of Chilean bridges mainly consists of multi-span bridges with girder superstructure. This type of bridge is economical and quick to build, becoming the most used typology because of favorable geographical conditions of the north and center south of the country, where there aren’t deep and extensive spans. Suspension bridges has little development in Chile, there is only small and medium span suspension bridge. Currently, a asymmetrically multi-span suspension bridge project in the Chacao channel is developed, with two main spans of 1155 and 1055 m each. This paper describes and compares the current state of art of the Chilean bridges, including suspension bridges and the Chacao multi-span suspension bridge. Also, this paper develops an analysis of potential sites for the design of single long-span bridges and multi-long-span bridges in the south of the country.
1 INTRODUCTION 1.1 Traditional chilean bridges Most of the bridges that are in Chile, are managed by the Roads Department of the Ministry of Public Works. Their is mainly one type of bridge that is constructed in Chile, which considers isostatic beams in the superstructure. These beams can be of concrete, pre-stressed and post-stressed, or steel, with a maximum span of 50 m (Vallejo, I., 2014). These types of beam bridges are cheaper and faster to build, making them the most used typology due to favorable geographical conditions in the north and center south of the country, where there is no need to cross deep (eg. Amolanas Bridge) and large valleys (eg. Juan Pablo II Bridge). The Amolanas Bridge (Fig. 1) is the highest road bridge in Chile and is located on the La Serena stretch – Los Vilos, Route 5 North in the Coquimbo Region. It has a length of 268 m and a height of 101.3 m. This bridge has its own characteristics for seismic protection, such as slide bearings on piers and visco-elastic dampers in abutments acting longitudinally (Moroni, O., et al. 2004). Juan Pablo II Bridge crosses the Biobío River in the city of Concepción. Currently, it is the longest bridge in the country (Fig. 2). It is 2430 m and is structured in 75 spans of 32.4 m each, with piers-piles and pre-stressed beams. Since 1995, the typology of bridges expands from simply supported beams (with expansion joints in the abutments for beams), to continuous beams and slab bridges, reducing points of failure by eliminating expansion joints to improve their seismic behavior and reduce maintenance cost. From 2002 to 2012, the continuous beam bridges were built with seismic isolation (eg: Marga Marga bridge, Figure 3), and continuous slab on piers and abutments with a maximum span of 60 m. The seismic isolation is performed through elastomerics plates in the support beams, which, by the elastic behavior of the material, the superstructure has a degree of independence from infrastructure, reducing damage during an earthquake (Campos, R., 2012). The new tendency in bridges is to design it without expansion joints (“jointless bridges”), being implemented by the Road Department, highways and bridges in developed countries such as the 483
Figure 1. Amolanas Bridge, Region of Coquimbo, Chile. (Morales, C., 2014).
Figure 2.
Juan Pablo II Bridge in Concepción, Biobío Region, Chile.
Figure 3. Marga-Marga Viaduct, reinforced concrete bridge with 10 spans with 54 m of maximum height, Region of Valparaíso, Chile. (Morales, C., 2014).
United States and currently recommended in “the new Seismic Criteria Chilean”. The benefits obtained for these structures are cheaper and safer, functionality and performance is improved in relation to disaster response, especially in cases of earthquakes (Campos, R., 2012). Table 1 shows the main structural elements and their types, of a traditional Chilean bridge. The following illustration shows the main elements of a traditional Chilean bridge. 484
Table 1. Elements of a traditional beam bridge. Element
Type 1
Type 2
Type 3
Materiality
Piers Abutments Deck
Wall Wall Beam
Pier-piles Pier-piles Slab
Pile Spill-through
Reinforced concrete Reinforced concrete 1. – Reinforced concrete 2. – Steel
Figure 4. 2013).
Main elements of a traditional Chilean Bridge (Hightway Manual, Volume 3, Road Department,
Table 2. General characteristics of the bridges President Ibáñez and Augusto Grosse. Bridge
Presidente Ibáñez∗
Augusto Grosse∗∗
Typology Span Pylon Typology Pylon Height (m) Pylon Materiality Deck typology Deck Width (m) Deck Height (m) Cable typology Hangers Corrosion protection
Suspension 210 Portal 27 Square steel profile Truss 9 3,1 8 parallel locked coil Locked coil Paint
Suspension 138 Portal 14,5 Circular steel profile Truss 6,2 2,1 5 parallel locked coil Steel bars Galvanized
∗ Mondorf,
P., 2014; Dominguez, J. & Flores, J., 1988 and MOP, s/f (b). ∗∗ MOP, s/f (a).
2 SUSPENSION BRIDGES IN CHILE Suspension bridges constructed in Chile are small and medium spans which are single suspension span (Fig. 7). These bridges no need to have a thorough study of geographic or weather conditions, or have a significant environmental impact compared to major bridges (Vallejo, I., 2014). To illustrate this typology, the main features of the bridges President Ibáñez (Fig. 5) and Augusto Grosse (Fig. 6) are presented in Table 2 to show detailed illustrations of them. The President Ibáñez Bridge cables pass through saddles (Fig. 7) on top of the piles and end in small anchor block. The hangers are connected to the main cables by clamps with pin (Fig. 8). Augusto Grosse Bridge presented fixed main cables on top of the piles using friction plates and hangers are connected with the main cables using clamps friction plates (Fig. 8). Augusto Grosse Bridge presents two different anchor blocks. One is founded on rock, so it is smaller with a depth of 3 m, a width of 7 m and a length of 3 m. The other anchor block has a length of 15 m and a depth of 5 m (Fig. 9). 485
Figure 5.
Presidente Ibáñez Bridge, Aysén Region, Chile. (Campusano, J., 2012).
Figure 6. Augusto Grosse Bridge, Aysen Region, Chile.
Figure 7.
Detailed drawings of cable connections of President Ibañez Bridge. (MOP, s/f (a)).
Figure 8.
Detailed drawings of cable connections of Augusto Grosse Bridge. (MOP, s/f (b)).
3 CHACAO BRIDGE The Chacao Bridge project links the Island of Chiloé with the mainland through a longitudinally asymmetric multi-span suspension bridge. With this bridge, Route 5 South would link in their last 200 kilometers, providing greater social value to the island by reducing travel time (Fig. 10) (Vallejo, I., 2014). 486
Figure 9.
Detailed drawings of the anchor blocks of the Augusto Grosse Bridge. (MOP, s/f (a)).
Figure 10.
Geographical map of the location of the Chacao Bridge. (MOP, 2012 (a)).
Figure 11.
Outline of Chacao Bridge, Chile. (MOP, 2012 (b)).
Currently, the bridge is in a phase of design and it is expected to open in 2020. For this project, basic engineering were performed a detailed studies such as topography, bathymetry, marine hydraulics, soil mechanics, wind, geophysics, geology and seismicity. The Chacao Bridge project structure consists of a 2750 m length, with an access viaduct 140 m on the south side, and two main span of 1155 m the north and 1055 m side south (Fig. 11). Currently, it’s the longest multi-span bridge in the world. The bridge has three reinforced concrete pylons. The south pylon, located on the island, reaches 157 m in height with direct foundations. The North and Central pylons are 199 m and 175 m in height respectively, which have deep foundations in the channel with piles of 2.5 m in diameter. The central pylon has a “Y” inverted shape (to balance the asymmetric own weights). The north and south pylons were portal type with two horizontal bracing beams, one on the top and one at the height of the deck. The two anchor blocks of the main cables are located on the banks. These blocks transferring the loads from the cables to the ground. 487
Figure 12.
Saddle housing and saddle of the Storda Bridge, Norway. (Márquez, M. & Valenzuela, M., 2013).
The bridge superstructure is composed of an aerodynamic orthotropic box beam steel deck and a system of cables. The deck has a width of 25 m, where two lanes each direction with safety parapets are arranged longitudinally at the ends and in the middle of the cross section. The bridge deck is continuous from the south pylon to the north abutment. At the ends of the deck, will be installed expansion joints and a longitudinal support system. Hydraulic buffers will also be installed to dampen the longitudinal seismic loads. The cable system is composed of two main cables and the hangers. The main cables are formed by strands, which are precast steel wires with resistance of 1860 MPa and hexagonal cross section. These strands are installed strand by strand from one anchor block to the other (PPWS). On the other hand, the hangers are the cables between the two main cables and the deck. The saddles are located on the tops of the pylons inside of protection housing. They are fabricated of steel and connecting the pylon with the cable (Márquez, M. & Valenzuela, M., 2014) (Fig. 12). This bridge will have access to the entire structure for inspection and its maintenance. Inside the pylons there will be elevators and stairs. The deck will have an inspection car inside it and other outside it. The main cable will have inspection cars, that moves along its length. The corrosion protection system consist in dehumidify inside all the structural elements (main cable-anchor blocks-deck-housing saddle) and paint all the steel elements of the bridge. Finally, for the entire bridge will be implemented by BMS, including monitoring system for structural control, maritime, air and vehicular traffic, in addition to the weather conditions, providing security for users and for the bridge itself. 4 INDUSTRY For the Chacao bridge project, the industry can carry out the shallow and deep foundations. The larger diameter pile built by a Chilean company is 2 m, so if for some project a larger diameter pile is needed, MOP will have to hire an expert foreign company in deep foundations, which have the machinery necessary for piles of large diameter. Chile has the equipment, experience and ability to construct pylons through climbing method and anchor blocks, which require a large volume of concrete. For the steel elements, Chile does not manufacture the steel plates required for the fabrication of deck sections. In Chile, only longitudinal elements are manufactured as reinforced concrete. If the deck will be of truss, Chile would have the industrial capacity and experience to manufacture the sections of the bridge deck. Chile does not have the industry for the manufacturing of the main cables nor for their installation (Vallejo, I., 2014). 5 FUTURE SITES FOR PROJECTS OF LARGE SPAN BRIDGES IN CHILE In Chile, the access to Chilean Patagonia by land is through the “Carretera Austral” (Route Ch-7), which has 1240 km of length and ends in Villa O’Higgins in the Aysen Region. A stretch of the 488
Figure 13.
Maps of potencial sites for long-span bridges connecting the Austral road.
Figure 14. Magallanes and Chilean Antártic Region, Chile. (Military Geographical Institute).
Austral road that does not have a road yet (see dashed line in Figure 13), where the passage from Hornopirén to Caleta Gonzalo is with ferries. In this section there are two fjords, the Quintupeu and Cahuelmo, in which they planned to build bridges for connecting the course of the road (Fig. 13). The bridge project for the fjord Quintupeu should be 300 m in length. While, the bridge project that will cross the fjord Cahuelmo will be 1300 m. There is still no charted road to joint Villa O’Higgins with Puerto Natales in the Magallanes Region (Fig. 14). Currently, study the feasibility of linking these two cities through a highway that linking islands, crossing channels and fjords with the possibility of construct large bridges with single and multi-span. On the other hand, the Road Department is studying the connection of the all country even the most southern city. This route will also need a huge-span bridges (length of the Magallanes Strait, 3.7 km aprox.).
6 CONCLUSIONS The bridges in Chile have evolved into a more continuous design and seismic protection. It has remained the preference for girder bridge, but the construction method or the beam materiality, makes the difference. 489
Table 3. General differences between Chilean suspension bridges and the Chacao Bridge. Item
Chilean suspension bridges
Chacao Bridge
Bridge typology N◦ of pylons Materiality of pylons Geometry of pylons Deck typology Main cables Fundations Detailed basic engineering studies Corrosion protection
One single suspension span 2 Steel profile Frame typology Truss Parallel locked coil Shallow footings No Painting, galvanized
2 Suspension spans 3 Reinforced concrete Frame typology and “Y” inverted Aerodynamic box beam Hexagonal strands (PPWS) Footings and piles of 3 m in diameter Yes Painting, galvanized dehumidification system
The suspension bridges built in Chile are of small span, because of this, they have large differences with which it is projected for the Chacao channel or for futures large bridges. The most significant differences are shown in Table 3: In Patagonia area there are places where it can project multi-span suspension or cable-stayed bridges. For this, it is more feasible to import the steel elements, while for concrete elements, the Chilean companies have the experience and the capacity to build them. BIBLIOGRAPHY Campos, R. & RCQ Engineering, 2012. “Design of Bridges and Viaducts”. The Engineering that coming, V Congress AICE (ed.). November 10, 2012, Santiago. Campusano, J., 2012. “Historical Evolution of Bridges in Chile”. Seminary: Challenge for the Design and Construction of Chacao Bridge, Roads Department (coord.), Ministry of Public Works, October 16, 2012, Santiago, Chile. Dominguez, J., & Flores, J., 1998. Suspension Bridge. Memory for title of Civil Engineer Civil Works, University of Santiago de Chile, Santiago. Garrido, M. L., 2013. “Chacao Bridge, Integrating Territory”. Road Department, MOP, Seminar: Challenges for Design and Construction of Chacao Bridge. Road Department, October 16, 2013, Santiago, Chile. Márquez, M & Valenzuela, M., 2013 “Chacao Bridge Technical Team visit to the Scandinavia”, Technical Report, Highway Directorate, Ministry of Public Works, Santiago, Chile. Mondorf, P., 2014 “President Ibáñez Bridge Repair 2011, Hangers Renewal Deteriorated by Corrosion and Fatigue”. ACCT Chile – MOP, First International Conference of Bridges – Chile 2014, Future Challenges: Design, Construction and Maintenance. September 24 to 26, 2014, Santiago, Chile. MOP, 2012 (a). Chacao Bridge Project: Background Referential. Roads Department, Ministry of Public Works. MOP, 2012 (b). Chacao Bridge Project: Technical Challenges. Roads Department, Ministry of Public Works. MOP, 2013. Highway Manual, Volume 3: Instructions and Design Criteria, Edition 2013. Ministry of Public Works. MOP (s/f (a)). Detailed drawings of the Augusto Grosse Bridge. Road Department, MOP. MOP (s/f (b)). Detailed drawings of the President Ibañez Bridge. Road Department, MOP. Morales, C., 2014. “Seismic isolation system for the viaduct Las Cruces, Route 160”. AICE – Association of Structural Civil Engineers (ed), III Conference: Structural Engineering Sample, June, 2014, Santiago, Chile. Moroni, O., Sarrazin, M., Benavides, C., & Dìaz, A., 2004. “Dynamic characteristics of Chilean bridges with seismic protection” in Revista Sul-Americana de Engenharia Estrutural, V. 1, n. 2, p. 31–54, 2004. Valenzuela, M. & Marquez, M., 2014. “Chacao Suspension Bridge: Special Structural Singularities”, ACCT Chile – MOP, First International Conference of Bridges – Chile 2014, Future Challenges: Design, Construction and Maintenance. September 24 to 26, 2014, Santiago, Chile. Vallejo, I., 2014. Suspension Bridges of large span: Design and Construction applied to Chile. Memory for the degree of Civil Engineering, University of Los Andes, Santiago, Chile.
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Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
Optimization of cable weight in multi-span cable-stayed bridges. Application to the Forth Replacement Crossing A. Baldomir, E. Tembrás & S. Hernández School of Civil Engineering, University of Coruna, Spain
ABSTRACT: An application of the optimum design methodologies will be presented in order to minimize the amount of steel needed for the full set of stays for the Queensferry Crossing (formerly the Forth Replacement Crossing). The optimization problem is formulated considering the cross sectional area of each individual cable as design variable. An approach based on simulating the prestressing forces by temperature increments in cables is performed to solve at the same time the optimization problem and compute the initial cable tension forces under dead weight. A multimodel optimization approach is used in order to consider two stages of the bridge construction. Finally, a modification of the spatial distribution in cables for this bridge project is proposed and the results are compared after the optimization process.
1 INTRODUCTION The experience accumulated over the years by engineers is the most important source of knowledge to address a new bridge design project. The conventional methodologies to design structures are essential to set a starting point when a new project is on track. However, it seems profitable to include as many tools as possible to improve the final result. One of the disciplines that is wellknown in an academic environment is the structural design optimization, as well as in aeronautical and automotive industry. Structural optimization is not yet a common tool in civil engineering due in part to the lack of this subject in the academic curriculum in most of the civil engineering degrees. Some references of optimum structural design can be found in Haftka (1990), Hernandez (1990), Vanderplaats (2001) or Arora (2004). Cable stayed bridges, being challenging structures, represent a very fruitful and convenient case for the application of structural optimization methodologies. They have been studied in depth by many authors as Walther et al. (1988), Gimsing (1997), Virlogeux (1994), among others. This bridge typology can be considered expensive in terms of budget, complex in terms of engineering design and relevant in terms of social impact. Therefore they deserve to be designed using the best up to date technologies. Some authors have made interesting applications of these techniques in cable-stayed bridges as Negrão and Simões (1997); Sung et al. (2006); Baldomir et al. (2010) or more recently Martins et al. (2014) where time-dependent effects are added to the optimization problem. The main goal in these works is to minimize the material cost of the bridge under displacement and stress constraints. In this work, an optimization methodology is proposed to minimize the system cable volume in a multi-span bridge. Two phases of the construction procedure are considered in the optimization process by applying a multi-model optimization technique. Nevertheless, the methodology is general so any number of construction stages can be added to the formulation. A commercial optimization tool is employed to solve the problem where temperature variation in cables are used as design variables to compute the prestressing cable forces under dead load while satisfying initial geometry requirements. 491
Figure 1.
Map of bridge location (left). The three towers under construction (right).
Figure 2.
Geometry of the finite element model.
The new projected bridge over the Firth of Forth has been chosen as application example of this approach. The reason for choosing this bridge is to know the influence of crossing cables in the optimization results. Finally, a modification of the actual scheme about the crossing cable length is proposed in order to improve the bridge design.
2 THE FORTH REPLACEMENT CROSSING: BRIDGE STRUCTURAL MODEL The Forth crossing bridge will be the longest three tower cable-stayed bridge in the world and it will be completed in 2016 according to Transport Scotland publications. This structure is under construction and it was projected to replace the Forth Road Bridge because of corrosion problems in the main cables. The new bridge is 1950 m long with two main spans of 650 m and three towers around 200 m high. In Figure 1 a map of bridge location is shown and a photograph of the towers construction. From the point of view of the structural layout, this bridge has a special feature which consists of overlaying some of the families of stays in the center of the two main spans. As explained by Gimsing (1997) this cable system allows to project slender towers because of the extra horizontal bracing effect. In this case the bridge is projected with a total of 18x6 cables where a subset of 16 stays are overlaid in the center spans. A simple 2D finite element model was used for the structural analyses. It consists of 108 rod elements for cable modelling and 158 bar elements for the deck and pylons. The data to build the finite element model have been obtained from the reports published by Transport Scotland (2009) and a presentation of Arup in the congress Bridges Middle East (2009). The supporting conditions have been assumed considering the usual practice in similar bridges. The deck is simply supported at both ends and at the intermediate supports of the approach spans. Material data and mechanical properties are shown in Table 1. Cable areas are design variables in the optimization problem. 492
Figure 3.
Cross section of the deck. Source: Transport Scotland.
Table 1. Mechanical properties of bridge model.
E (GPa) υ Area (m2 ) Moment of inertia (m4 )
Tower (below deck)
Tower (above deck)
Cables
Deck
35 0.3 45 450
35 0.3 32 220
200 0.2 var. –
200 0.2 2 7.6
Two load cases are considered: dead load and dead load plus live load (4 kN/m2 ). The dead load is applied to the full bridge but the live load is only applied to one of the main spans due to the symmetry considered and because this location produces higher bending moments. 3 OPTIMIZATION APPROACH 3.1 Formulation of the optimization problem The objective is to find the minimum volume of steel cable required by the stays while satisfying the following constraints: 1) Cable tension must be lower than 800 Mpa 2) Vertical displacements of the deck under dead weight must be lower than L/10000 where L is the main span length. 3) Horizontal displacement at the tower top under dead weight must be lower H /10000 where H is the central tower high (206.89 m). 4) Vertical displacement of the deck under dead and live load is limited to L/500 5) Horizontal displacement at the tower top under dead and live load is limited to H /250 Mathematically the problem can be expressed as:
where Ai = cable cross sectional area (full set of design variables in this problem); li = cable length; σ i = cable tension; wj = vertical displacement of node deck j; uj = horizontal displacement at the tower top k; subscript dindicates under dead load and l under dead and live load. During the optimization procedure the cable areas are modified every cycle and therefore a new set of prestressing forces need to be calculated in order to satisfy the geometry constraints under dead load. These forces can be obtained by solving the linear system of equations based on the 493
Figure 4.
(I) Bridge model before attaching the cantilevers. (II) The full bridge model.
unitary load method. Thus, this stage should be made before continuing with the next cycle of the optimization procedure. In order to avoid this stage that can be cumbersome a modification of the initial problem is proposed by adding the cable forces as design variables. Therefore there will be two design variables per stay: the cross sectional area and the prestressing force to accomplish the desired geometry under self-weight. The commercial optimization software Altair Optistruct v11 (2013) was used to solve the problem but internal forces in cables are not available as design variables. Then prestressing cable forces were introduced as thermal loads using the thermal expansion coefficient as design variable. 3.2 Multi-model optimization Two stages of the bridge construction were included in the optimization procedure supposing a cantilever construction method. It was selected the critical cantilever stage and the full bridge model (Figure 4). An updated optimization problem can be formulated including displacement and stress constraints over the incomplete configuration of the bridge. The new optimization problem can be defined as:
where the superscripts (I), (II) indicates the bridge model configuration according to Figure 4. The above mentioned problem can be solved numerically in different ways depending on how the structural responses are going to be obtained. The structural responses may be obtained by parallelizing the analysis of both models. But a better option is to arrange both structural models in a unique file that is computed many times in the optimization process. This strategy is usually coined as multi-model analysis. Then, only one analysis is thus required to get all the structural responses. With this approach the phase of data transferring from the various models at each iteration to the optimization algorithm is not necessary. Three sets of design variables were defined: 1. The cross sectional area of each stay. This set is shared by both models. 2. The thermal expansion coefficients (prestressing forces) for model I. 3. The thermal expansion coefficients (prestressing forces) for model II. 494
Figure 5.
Cable steel volume.
The objective function is the cable steel volume of model II as stated in Equation (2) and the constraints of both models are simultaneously evaluated at each iteration of the optimization process. 3.3 Optimum design considering identical cross sectional area for each stay A first optimization problem considering only one value of cross sectional area for the full set of stays was done. The objective is to know the minimum volume of steel cable in a more conservative design and then comparing it with the improvements when allowing a different cross sectional area for each stay. A better comparison criterion could be to check the amount of material of the actual design versus the optimized one but in absence of actual data the previous criterion has been chosen. The convergence is achieved after 15 iterations with a final cross sectional area of 0.06706 m2 and the total volume of steel in cables is 1762.094 m3 . At the final design only the vertical displacements of 6 nodes of the deck are active constraints for the dead load condition. 3.4 Optimum design considering each cable cross sectional area as a single design variable This case corresponds to the problem formulated in Equation (2). The problem converges after 67 iterations with a final steel volume of 1170.774 m3 . Figures 6–7 show the evolution of the optimization process. The sizing is about 33% lighter than the previous model where only one cross sectional area of stays may be used. The distribution of the cable areas along the bridge are shown in Figure 8. Only half of bridge length has been drawn. There are 46 active constraints at the final design divided in 20 active constraints for the dead load condition and 26 for the dead and live load case. More specifically in Figure it was shown the active constraints for dead load case (in green color) and the active constraints for dead and live load case (blue color). The circular points represent the node of deck where vertical displacement constraint is active while the blue lines the cables where stress constraint is active. It is noticeable that at the final design there are active constraints of both configurations which indicates that it is necessary to consider a multi-model approach in order to obtain only one design that satisfies the constraints of several bridge configurations. 4 A PROPOSAL DESIGN OF THE FORTH BRIDGE PROJECT In order to improve the final design a modification of the original bridge is proposed by increasing the crossed cables length from 154 m to 330 m. Figure 9 shows the new bridge configurations like previous case. 495
Figure 6.
Evolution of the cross sectional areas.
Figure 7.
Evolution of the objective function.
Figure 8.
Cable steel areas along the bridge deck.
496
Figure 9. Active constraints.
Figure 10.
Design variation for the Forth Bridge.
Figure 11.
Cable steel areas along the bridge deck in the proposed model.
The optimum steel volume for this configuration was 1067.448 m3 after a multi-model optimization process with 118 iterations. The values of cross sectional areas of stays are shown in Figure 11. For this case 30 stress constraints are active, horizontal displacement at tower top nodes and 40 vertical displacements on deck are active. This design is 8.8% lighter than the optimum design obtained in previous section. 497
5 CONCLUSIONS Several conclusions can be drawn from this work: 1) Some finite element models representing adequately the Queensferry Crosssing have been developed to identify a more efficient distribution of material in the stay scheme. 2) A formulation of the optimization of the amount of stays weight considering the completed bridge and a model of it before completion has been carried out including stress and displacement constraints and the cross area of the cables as design variables. 3) In order to speed up the optimization procedure two strategies have been developed: one consisted on including all structural models in a single file to diminish data transfer and another one was to add as design variables the initial prestressing forces of the stays with the aim of avoid solving a system of linear equations at each iteration of the optimization problem. 4) A first group of numerical results showed the material savings that can be obtained when each stay has different cross area with respect to the case when all of them have equal cross section. 5) Additional reduction of material can be obtained if the length of the span where the stays overlaps is enlarged. 6) In overall, the work demonstrates the usefulness of optimization methodologies for dealing with real engineering problems. REFERENCES Altair Hyperworks v11.0, 2013. Altair Engineering Inc. Arora JS 2004. Introduction to optimum design. 2nd ed. Elsevier Academic Press. Baldomir A, Hernandez S, Nieto F, Jurado JA 2010. Cable optimization of a long span cable stayed bridge in La Coruña. Advances in Engineering Software 41: 931–938. Gimsing NJ 1997. Cable supported bridges. 2nd ed. John Wiley and Sons. Haftka RT, Gürdal Z, Kamat MP 1990. Elements of structural optimization, Second Revised Edition. Kluwer Academic Publishers. Hernández S 1990. Métodos de Diseño Óptimo de Estructuras, Colegio de Ingenieros de Caminos, Canales y Puertos. Hussein N 2009. Bridges. Boutique Footbridges to Major Crossings, Bridges Middle East, Abu Dhabi, UAE. Martins AMB, Simões LMC, Negrão JHO 2014. Cable stretching force optimization of concrete cablestayed bridges including construction stages and time-dependent effects. Struct Multidisc Optim. DOI 10.1007/s00158-014-1153-4. Negrão JHO, Simões LMC 1997. Optimization of cable-stayed bridges with three-dimensional modelling. Computers & Structures 64(1–4): 741–758. SungYC, Chang DW, Teo EH 2006. Optimum post-tensioning cable forces of Mau-Lo Hsi cable-stayed bridge. Engineering Structures 28: 1407–1417. TRANSPORT SCOTLAND 2009. Forth Replacement Crossing DMRB Stage 3 Scheme Assessment Report Part 2: Engineering, Traffic and Economic Assessment, Vanderplaats GN 2001. Numerical optimization techniques for engineering design. Vanderplaats Research & Development, Inc. Colorado Springs. Virlogeux M 1994. Erection of cable-stayed bridges, the control of the desired geometry. International Conference AIPC-FIP, Deauville. Walther R, Houriet B, Isler W, Moia P, Klein JF 1988. Cable Stayed Bridges. Thomas Telford.
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Design parameters of suspension bridges: Updates of state of art and its application on multi-span typology I. Vallejo, M.A. Valenzuela & M. Márquez Civil Engineer, Public Works Ministry, Santiago, Chile
ABSTRACT: Suspension bridge typology among other types of bridges is the most optimal for spans higher than 1100 m. In recent years, suspension bridges have had a breakthrough in technology, materials and methods of design and construction, which has led them to be bigger and longer. Furthermore, with the emergence of the new typology as multi-span bridges, such as Taizhou Bridge in China and will be the Chacao Bridge in Chile with two asymmetric spans of 1055 and 1155 m each. This paper analyzes and updates the main design parameters of this bridge typology, based on a thorough study on the state of art of suspension bridges until 2014. This study has as its conclusion the growth of some design parameters such as the main span, while others parameters ratios are kept within a certain range, such as the width of the deck and his deep. 1 INTRODUCTION Suspension bridges have been used from the beginning to cross large spans in valleys, ravines or between two banks separated by rivers or sea. Suspension bridges are the typology of optimal bridge in relation to costs among all types of bridges, for larger spans 1100 m. In smaller spans competes with cable-stayed bridges [Gimsing, N. & Georgakis, K., 2012]. In even smaller span, up to 500 m, it competes with other typologies such as arch bridges, truss bridges and others. This typology have had in the last two decades a great development and technological advancement in their methods of design, construction and materiality [Vallejo, I., 2014]. Therefore, this type of bridge has an upswing in their construction from the 90s, becoming the most used typology for spans greater than 1000 m (Fig. 1). Today there are 28 suspension bridges with spans higher than 1000 m, while the cable-stayed typology has only 3: Russki Bridge with 1104 m main span (Russia, 2012), Sutong Bridge with 1088 m main span (China, 2008) and the Stonecutters Bridge 1018 m main span (China, 2009). The first multi span suspension bridge opened to traffic the year 2012, the Taizhou Bridge in China. This configuration provides an optimal solution for greater distances (2000 m) using a larger number of pylons and suspended spans. The benefits of this configuration compared to a traditional are: 1) decrease the amount of piers on the approach viaducts, 2) keep the maritime routes open, 3) reduce of the risk of foundations in water, 4) reduce loads on main cables and in the anchor blocks, 5) reduce the dimensions of these elements and their costs. 2 GENERAL PARAMETERS OF SUSPENSION BRIDGES DESIGN 2.1 Main span With the evolution of suspension bridges in the last two decades, the amount of suspension bridges that exceed 1000 m main span has increased (Gimsing, N., & Georgakis, K., 2012). In 1931, the 499
Figure 1. Number of suspension bridges greater than 483 m of span (Brooklyn Bridge), built by decade (the bridges under construction are included).
George Washington Bridge in New York, United States, is the first suspension bridge to exceed the 1000 m, with 1067 m of main span. Their main spans are also increasing, with Akashi Kaikyo Bridge, Japan, holding the record, opened in 1998, with 1991 m main span. The 25 longest suspension bridges in the world by main span are shown in Table 1. Table 2 and Figure 2 show the graphic of the bridges that marked a record in the length of the main span, as well as the chronological evolution of the main span. 2.2 Deck The Severn suspension bridge constructed in 1966 in the UK (Fig. 3) is the first to innovate in the design and construction of a deck of box girder. The first metallic deck was constructed of truss girder. Currently, the trend is build the deck of box girder, with exceptions such as the Second Tacoma Narrows suspension bridge for aesthetic reasons (Fig. 4) and the Chinese bridges, Balinghe and Aizhai, for reasons of construction method. The box girder deck has better aerodynamics behavior and is lighter and economical (Myerscough, M., 2013). Another advantage of this type of deck is it’s ease to maintain, because it is sealed so it can be dehumidified, process that dries the air and maintains a relative humidity of 40%, which prevents corrosion inside the deck. The dehumidification system in the deck was first used in the Little Belt Bridge, Denmark, in 1970, and due to its good performance, today, it is common to design the deck with dehumidification system and that some already built bridges have integrated this system as corrosion protection (Bloomstine, M., & Sorensen, O., 2012). The suspension bridge with the longest continuous deck in the world is the Great Belt East Bridge, Denmark (Fig. 4), which doesn’t have vertical supports from one anchor block to another (2694 m). Suspension bridges with wide traffic lanes require very wide decks which compromises their aerodynamic behavior, so in the 90s, a double box girder deck was designed as Vierendeel typology. 500
Table 1. Longest suspension bridges in the world. Deck Typology
N◦ Bridges
Country
Year
Length Span
1 Akashi Kaikyo 2 Xihoumen
Japan China
1998 2009
960 + 1991 + 960 578 + 1650 + 485
3 Great Belt 4 Izmit Bay 5 Yi Sun Sin
East Denmark Turkey South Korea
1998 2016∗∗ 2012
535 + 1624 + 535 560 + 1550 + 560 357.5 + 1545 + 357.5
6 Runyang 7 Nanjing 8 Humber 9 Jiangyin 10 Hardanger 11Tsing Ma 12Verrazano Narrows 13 Golden Gate 14Yangluo 15 Höga Kusten 16Aizhai 17 Mackinac 18 Chacao 19 Ulsan Harbor 20 Huangpu 21 Minami 22 Fatih Sultan Mehmet 23 Balinghe 24Taizhou 25 Ma`anshan
China China U.K. China Norway China U.S. U.S. China Sweden China U.S. Chile South Korea China Japan Turkey China China China
2005 2012 1981 1999 2013 1997 1964 1937 2007 1997 2012 1957 2020∗∗ 2015∗∗ 2008 1988 1988 2009 2012 2013
470 + 1490 + 470 417 + 1418 + 357 280 + 1410 + 530 336.5 + 1385 + 309.3 1380 455 + 1337 + 300 370.3 + 1298.2 + 342.9 342.9 + 1280.2 + 342.9 250 + 1280 + 440 310 + 1210 + 280 1176 548.6 + 1158.2 + 548.6 212 + 1155 + 1055 + 324 303 + 1150 + 355 1108 274 + 1100 + 274 210 + 1090 + 210 1088 320 + 1080 + 1080 + 320 360 + 1080 + 1080 + 360
∗ Ss = Single
Truss girder Double box girder Box girder Box girder Double box girder Box girder Box girder Box girder Box girder Box girder Box girder Truss girder Truss girder Box girder Box girder Truss girder Truss girder Box girder Box girder Box girder Truss girder Box girder Truss girder Box girder Box girder
span, Ms = Multi span; ∗∗ In construction
Table 2. Longest main span of suspension bridges. Bridges
Country
Year
Main Span (m)
Union Menai Fribourg Wheeling Cincinnati Clifton Brooklyn Williamsburg Bear Mountain Delaware Ambassador George Washington Golden Gate Verrazano Narrows Humber Akashi Kaikyo
UK UK France USA USA USA USA USA USA USA USA USA USA USA UK Japan
1820 1826 1834 1849 1867 1869 1883 1903 1924 1926 1929 1931 1937 1964 1981 1998
140 176 266 308 322 386 486 489 497 533 564 1067 1280 1298 1410 1991
501
Typology∗ Ss Ss Ss Ss Ss Ss Ss Ss Ss Ss Ss Ss Ss Ss Ss Ss Ss Ms Ss Ss Ss Ss Ss Ms Ms
Figure 2.
Graphic of the chronological evolution of the main span.
Figure 3. Left: deck of aerodynamic box girder of Severn Bridge, UK (Gordon, S., 2010). Right: the two suspension bridges Tacoma Narrows, United States (Márquez, M. & Valenzuela, M., 2013).
Figure 4.
Box girder deck of Great Belt East Bridge, Denmark (Weight, A., 2009).
Examples of this deck are Xihoumen, China (2009) and Yi Sun-Sin Bridge, South Korea (2012, Fig. 5), which has a better aerodynamic stability with this deck. The ratio between the width of the deck with the main span or the height of the deck with the main span, are kept within a certain range (Table 3). The detail is showed in Figure 6. 2.3 Pylons Suspension bridges began to be constructed with pylons of masonry, but after the San Francisco earthquake in 1906, USA, it became evident that this construction method had great flaws, since it was a very fragile system, and thus it wasn’t earthquake – resistant. Thus, the pylons were began to be made of steel, providing ductility and flexibility, which could better release the energy of 502
Figure 5. Double box girder deck of Yi Sun Sin Bridge in South Korea (right) and Xihoumen in China (left) (Valenzuela, M., 2013). Table 3. Ranges of ratio width/main span and height/main span.
Width/main span Height/main span
Figure 6.
Truss girder deck
Box girder deck
1/41–1/66 1/157–1/272
1/27–1/67 1/300–1/471
Ratio height/main span and width/main span.
earthquakes. The peak of the steel lasted only 50 years, mainly because of the high cost of this building material. In the 70s, reinforced concrete was began to be use as a building material due to its ductility, its capacity to resist earthquakes and for being more economical than steel. The multi span bridges are consider concrete pylons (Chacao Bridge), steel pylons (Taizhou Bridge) or composite pylons (Ma’Anshan Bridge). The Humber Bridge, UK, opened to traffic in 1981 was the first bridge to have their 180 m height pylons made of reinforced concrete (Vallejo, I., 2014). The Figure 7 shows the chronological evolution of the height of the pylons. 503
Figure 7.
Evolution of the height of the pylons.
Figure 8. Table 1).
Ratio between the main span and height of the pylons (the numbers correspond to the bridges of
Even if the pylon has been increasing consistently its height, due to the growth of the main span, these two elements have maintained a relationship within an almost constant range between 5 and 8 (Fig. 8), except the Fatih Sultan Mehmet Bridge, Turkey, which has a ratio of almost 10 and the Aizhai Bridge, China, with a ratio of 9, because their pylons have a low height since they are on top of hills. The maximum pylon height of the multi span are 200 m in the Taizhou Bridge (central pylon), Ma`anshan reached 175 m (all pylons) and 199 m for Chacao Bridges (north pylon). 2.4 Cables The basic unit of the main cables used for this type of bridge is steel wires. These steel wires have the property of having a high tensile strength while maintaining ductility compared to other steel structure. A typical wire used in a suspension bridge has a circular section and a diameter of 5 to 5.5 mm (Gimsing, N., & Georgakis, K., 2012). Within the parameters that have been set within a fixed range is the sag ratio. This ratio has been maintained between 1/9 and 1/10.5 in all major suspension bridges longer than 1000 m. The main cables of a suspension bridge can be constructed using two methods: the “Air Spinning or AS” method, where the erection of the main cable is by installing wire per wire from one anchor block to the other, and the “Prefabricated parallel-wire strand or PPWS method”, where the erection of the main cable is by installing strand per strand, which are prefabricated parallel wire cables that come with the exact length to reach from one anchor block to another. This method is used in most of the bridges of Asia and has the advantage of reduced installation time. The decision of which method to use, depends on the experience of the builder with each method, because both methods have advantages and disadvantages. The latest innovation in technology of the main cables was developed in South Korea. This innovation is the increased tensile strength of the steel wires of the main cable to 1960 MPa, which was applied for first time to the Ulsan Bridge, which is still in construction (Fig. 9). 504
Figure 9.
Evolution of the wires resistance of the main cables.
3 STRUCTURE With multi-span suspension bridges, more rigid pylons begin to be designed longitudinally because the central pylon must compensate the differential loads when the main spans are loaded asymmetrically. The Taizhou Bridge, China, is the first multi-span bridge, that has two main spans of 1080 m each and a central inverted “Y” shaped pylon that has a base of 4 supports, which acquire great stability to resist the displacements on top of the pylon due to the longitudinal horizontal forces of the cables. In 2013, the Ma’anshan Bridge was opened to traffic in China, which is also a multi-span bridge that has two main spans of 1080 m each. In this bridge, the central pylon is a portal frame like traditional pylon shape of a suspension bridge, but it was made with concrete until the lower cross beam and then is was made of steel, which helps resist the movement on top of the pylon. Currently, the Chacao Bridge in Chile, is being designed and it is projected to open to traffic by 2020. This bridge will be asymmetric multi-span bridge, because their main spans have 1155 m and 1055 m. Those asymmetric spans generate asymmetric dead loads which will be offset by the heights of the pylons. The north pylon, for the longer span, has a height of 199 m, the central pylon has 175 m height, and the south pylon, for the shorter span, has a 157 m height. The central pylon also has the inverted “Y” shape and will be built with reinforced concrete. This pylon will have the flexibility to resist displacements due to asymmetric loads between the two spans. The critical load case is when a main span is fully loaded and the other not, and to counter this disadvantage is that the saddle atop the central pylon should have greater anti-slip friction with the cable. The central pylon, by allowing a moderate displacement at the top, avoids generating unwanted twists and deflections in the cable structure and the deck. By allowing these displacements at the top of the pylon, it should have a large support base or foundation, to counteract the large bending moment that is generated. It is for this reason that the design of the Chacao Bridge 2001 central pylon had an “A” shape, which gives it a great basal stiffness which completely restricts the displacement on the top of the pylon (Fig. 10). Today, the central pylon has “Y” inverted shape design with a greater restriction of cables on top of the pylon, which gives the bridge a better structural behavior (Marquez, M. & Valenzuela, M., 2014). 505
Figure 10.
Central pylons of Chacao (2001) and Chacao (2014) projects.
4 CONCLUSIONS 1) The fast progress on the suspension bridge typology in all its aspects, in design and construction, as well as the materials that they are constructed and his maintenance. 2) There are parameters that continue increasing like as the main span, being the world’s longest 1991 m, and the tensile strength of the main cables wires reaching 1960 MPa. Other ratios between design parameters of these bridges are kept in a certain constant range, such as the length of the main span with the height of the pylon, also with the width of the deck or his deep. 3) Due to this progress there is a tendency to increase the length of the span and as an optimal solution, the multi-long-span bridges were born. The multi-span suspension bridges emerge as an optimal solution, but they should be studied in detail: the feasibility of the construction of the central pylon, the load distribution on main cable, pylons, anchor blocks and central saddle. BIBLIOGRAPHY Bloomstine, M. & Sorensen, O. 2012. Dehumidification of suspension bridge main cables. En S. &. Frangopol (ed), Bridge Maintenance, Safety, Management and Life-Cycle Optimization (pp. 1724–1731). London: Taylor & Francis. Gimsing, N. & Georgakis, C. 2012. Cable Supported Bridges: Concept and Design, Editorial Wiley, Denmark. Gordon, S. 2010. Critical Analysis of the First Severn Bridge. Bridge Engineering 2 Conference, University of Bath, 2010, Bath. Márquez, M. & Valenzuela, M. 2013a. Chacao Bridge Technical Team visit to the US, Technical Report, Highway Directorate, Ministry of Public Works, Santiago, Chile. Marquez, M. & Valenzuela, M. 2014. Chacao Suspension Bridge: Special Structural Singularities, ACCT Chile – MOP, First International Conference of Bridges – Chile 2014, Future Challenges: Design, Construction and Maintenance. Santiago, Chile. Myerscough, M. 2013. Suspension Bridges, Past and Present. The Structural Engineer, 12–21. Valenzuela, M. 2012. Chacao Bridge Technical Team visit to China and Korea, Technical Report, Highway Directorate, Ministry of Public Works, Santiago, Chile. Vallejo, I. 2014. Suspension Bridges of large span: Design and Construction applied to Chile. Memory for the degree in Civil Engineering, University of Los Andes, Santiago, Chile. Weight, A. 2009. Critical Analysis of the Great Belt East Bridge, Denmark. Department of Architecture and Civil Engineering (coord.). Proceedings of Bridge Engineering 2. University of Bath, Bath.
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Comparative study of prestressing consumptions in 7 different constructive methods for 75 m multi-span box girders A. Ferreira & B. Lima STRENG Engenharia de Estruturas, Porto, Portugal
F. Lopes FEUP, Porto, Portugal
P. Pacheco BERD/FEUP, Porto, Portugal
ABSTRACT: In the last few years new constructive methods have been developed and applied in multi-span prestressed concrete decks, with medium-large spans (70 to 80 m), where, typically, box girders cross sections are adopted. Each constructive method has its own field of application and frequently decisions are taken after a comparative analysis of different options, comprising main issues like Safety, Risk, Economy and Time. Present paper presents a comparative study of prestressing consumption on 7 constructive methods for 75 m multi-span concrete box-girder decks. Thus, being mainly useful for the cost evaluation of alternative methods (Economy). The methods considered in this study are: (1) span by span, cast in situ, with constructive joints over the piers; (2) span by span, cast in situ, with constructive joints at 0.2 of the span; (3) precast segments – span by span (simply supported); (4) precast full span segments (simply supported); (5) precast full span segments with post continuity; 6) in situ balanced cantilever and finally (7) precast segmental balanced cantilever. The study considers a unique width of a typical highway cross section and Eurocode design criteria and materials are adopted. All solutions are performed with the same volume of concrete per span and with the same height, but with different geometries (optimized for the conditions of each case) in order to obtain a uni-parametrical study on prestressing consumption. A typified intermedium representative span (in a multi span deck) is considered in all cases. Common prestressing layouts are adopted. The materials’ time-dependent behavior, the influence of the construction stages and the internal forces redistributions are taken into account according to standard procedures. Stresses are analyzed during construction and service stages. All the design optimization procedures are performed by means of genetic algorithms. The prestressing consumptions for mentioned 7 cases are presented and results are very briefly discussed. 1 INTRODUCTION It is known that main factors of decision, while selecting a Constructive Method are: Economy; Time and Safety, not necessarily by this order. Economy is normally evaluated by the sum of Materials consumption; Main Equipment; Man Power; Secondary Equipment; Site Preparation; Indirect Costs and Others. In what regards to Material Consumption one of the main factors is the Prestressing Consumption. It is known that for the same deck cross section and for the same spans, the prestressing consumption may have relevant variation. In this paper a study of uni-parametric variation of prestressing consumptions is presented for 7 different constructive methods as per Figure 1. Present approach is to be clearly understood us an exercise to give tendencies and never exact, relative or absolute, values. Indeed, it is known that the design always depend on multiples factors. For example, it is possible to achieve significantly different results if the optimization procedures 507
Figure 1.
Constructive Methods and Structural Systems Characterization.
also comprise concrete volume optimization or reinforcing steel optimization. And, in fact, common design procedures always consider simultaneously multiple factors. Thus, the propose of this approach is no more than sharing “what would be the tendency variation of prestressing consumption, for different constructive methods considering the same concrete consumption and the same deck height for all (Lopes 2015) 1.1 Main Assumptions In order to perform a uni-parametrical evaluation of different prestressing consumptions following assumptions were adopted: – All assessments performed according to Eurocode (EC 0a 0b, 1 2, 2003, to 2007),; – Equal concrete volume and height of the deck for all 7 evaluated solutions, comprising different geometries, optimized for each case; 508
– Equal height and width of the cross section in all cases; – The deck’s geometry definition and the prestressing layout/dimensioning performed simultaneously by means of a genetic algorithm (Ferreira & Lima (2015)). In order to perform this optimization, the relation used between costs for concrete, prestressing and reinforcing steels is 100:2:1. Other secondary assumptions were adopted. Some of them may influence the absolute results, but as they are the same in all cases, there influence on the relative values is to be reduced: – Infinite number of spans; – Service stages analyzed at the end of the construction and 10,000 days after it; – The stresses redistribution due to creep and shrinkage determined based on the simplified formula of Trost & Wolff (1969); – Prestress strands of 0,6” (1,5 cm2 of cross section area), with fpuk = 1860 MPa, and stressed at 0.75fpuk ; – 10% prestress immediate losses, constant throughout the cable length; – Prestress time dependent losses considered between the end of the construction and 10,000 days after it; – Prestress stress at the age of 10,000 days assumed equal to 1000 MPa; – C40/50 concrete class at 28 days, and equivalent concrete class of C30/37 at stressing; – No stresses considered due to pier/foundations-constraints to imposed deformations. 2 LOADS AND COMBINATIONS 2.1 Loads In all cases the loads are evaluated according to Eurocode and comprise: – Concrete self-weight; – Prestress; – Dead loads, due to the presence of new-jerseys and bituminous layer, of 36.35 kN/m (longitudinally); – Road bridge traffic loads - Load Model 1; – Temperature gradients of +10◦ C and −5◦ C; – Movable scaffolding system rear reaction (valid for span by span construction) considered equal to 1.60 the rear reaction for the weight of the wet concrete, to take into account the scaffolding system self-weight; at a distance of 2 m of the construction joint section; – Form traveler’s (valid for balanced cantilever construction) self-weight considered equal to half the weight of the heaviest segment (SETRA (2007)); resultant force located at the middle of the segment in construction; 2.2 Load Combinations and Design Assessments Following combinations of actions were adopted: Table 1. Load combinations. Combination Construction Stages Characteristic Frequent Quasi-Permanent
Expression Gk,j “+” P Gk,j “+” P “+” Qk1 “+” ψ0,iQk,i Gk,j “+” P “+” ψ 1,1 Qk1 “+” ψ2,i Qk,i Gk,j “+” P “+” ψ2,i Qk,i
509
Following concrete stress verifications were considered in table 2: At joints of precast elements a more demanding stress verification is required, as shown by the table 3:
3 DECK CHARACTERIZATION / PRESSTRESING LAYOUTS 3.1 Cross Sections The Cross section follows a typical prestressed concrete box girder design, where the indicated variables are defined together with the prestressing amount, by means of mentioned Genetic Algorithm.
3.2 Longitudinal Section / Prestressing Layout The longitudinal sections, with a 75 m span length, are as follows:
Table 2. Stresses verifications. Combination
Maximum compressive stress
Maximum tensile stress
Construction Stages Frequent Quasi-Permanent
0.60 fck (t) – 0.45 fck
fctm (t) fctk,0,05 0 (zero)
Table 3. Stresses verifications at joints.
Figure 2.
Combination
Maximum compressive stress
Maximum tensile stress
All combinations Characteristic Quasi-Permanent
0.60 fck 0.45 fck
0 (zero) -
Mid Span Cross Section.
510
4 PRESTRESSING CONSUMPTIONS Following prestressing consumptions were achieved table 5:
Figure 3.
Support Cross Section.
Table 4. Longitudinal section.
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Table 5. Prestress consumptions.
Figure 4.
Construction Method
Prestress consumption kg/m2
(1) Cast in Situ; Span by span, Joints over the piers (2) Cast in Situ, Span by span, Joints at 0.2 of the span (3) Precast segments, span by span (simply supported) (4) Precast full span segments (simply supported) (5) Precast full span segments, with post continuity (6) In Situ, Balanced cantilever (7) Precast Segments, Balanced cantilever
20.4 15.1 29.7 23.4 19.2 18.0 32.0
Prestressing consumptions tendency.
5 CONCLUSIONS The performed study allows to establish the tendency variation of prestressing consumptions, for the different constructive methods presented in figure 4. Cast in situ, span by span with joints at (0.2x L), reveals to be the method with less prestressing consumption. The segmental methods have the higher consumptions. REFERENCES Lopes, F. 2015. Comparative study of prestressing consumptions in different constructive methods in different mediul/large span bridge types, in Portuguese, MSc Thesis, FEUP, Porto Eurocode 0a, 2002, Basis of structural design (BS EN 1990:2002), CEN. Eurocode 0b, 2005 – Annex A2, Application for Bridges (EN 1990 – Annex A2), CEN Eurocode 1, 2003, Actions on structures, Part 2: Traffic loads on bridges (EN 1991-2), CEN, Eurocode 2, 2005, Design of Concrete Structures, Part 2 – Concrete Bridges – Design and Detailing Rules (EN 1992-2:2005), CEN Precast Concrete Products, 2008 – Bridge Elements (BS EN 15050:2007), BSI. SETRA, 2007 – Service d’Études Techniques des Routes et Autoroutes,. Design Guide - Prestressed Concrete Bridges Built Using the Cantilever Method. Lima, B. & Ferreira, A. 2015. “Optimized bridge deck design using a genetic algorithm”, Proceedings of the Multi-Span Large Bridges Conference, CRC Press. Trost, H. & Wolff, H.J. 1969. Zur wirklichkeitsnahen Ermittlung der Beanspruchungen in abschnittsweise hergestellten Spannbetontragwerken. Hannover: Techn. Univ., Lehrstuhl und Inst. für Massivbau
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KaTembe Bridge over Espírito Santo Estuary, in Maputo T. Mendonça, V. Brito & M. Monteiro General Manager at BETAR Consultores Lda, Lisbon, Portugal
ABSTRACT: The crossing Maputo/KaTembe has been planned since colonial regime as an important investment to develop economic trade and to reach several communities in the South of Maputo Province and further South Africa and Swaziland borders. The crossing earned a new impulse in 2009 when an alternative corridor stablished the bridge over the narrows, at the Estuary’s mouth. As constraints of different stakeholders were being cleared different solutions for the crossing were studied, namely a cable stayed option and a suspension bridge, depending on the main span dimension. The access viaducts are composed of multi-span bridges. Also restraints at insertion points on the two banks of the Estuary dictated two different types of structures. At the North side the access viaduct will be constructed by the cantilever method. At the South side the deck will be cast with a launching girder.
1 SCOPE The crossing Maputo/KaTembe has been planned since colonial regime as an important investment to develop economic trade and reach several communities in the South of Maputo Province. Furthermore, a deep-water port was being study for cape Dobela and so road network to South Africa, Swaziland and South Mozambique would have to be improved. Although the great extension to access the bridge at Matola towards KaTembe, as initially planned, the crossing over Espírito Santo Estuary would not interfere with navigation near Maputo harbor. But the construction didn’t start. After independence this older ambition was renewed. In 2008 another location for the bridge was studied linking Maputo at N1 road junction to KaTembe Norwest, along 4.3 km over the Estuary. This solution would have many piers in the water and was similar to Armando Emílio Guebuza Bridge, with multiple spans, the central ones with 210 m length (Figure 1). Despite the location far away from Maputo Port area of influence – that is accessed by very large ships – the piers would interfere with docking movements around Matola Port and so security issues were discussed.
Figure 1.
Preview of a solution linking Maputo to KaTembe Northwest with a multi-span bridge.
513
Figure 2. The two corridors proposed for the crossing: the left one (the one that crosses the word “Estuary” in the image) around Matola and the right one over the narrow and directly to KaTembe.
The study of the bridge insertion at KaTembe and the road connection to the South towards Bela Vista was further developed leading to variants of this route. In 2009 a new impulse was given for the crossing when an alternative corridor was considered over the narrows, at the Estuary’s mouth (Figure 2). This new solution has evident repercussion in the urban development of the South bank as it provides a major opportunity to improve KaTembe municipal district. 2 MAIN CONSTRAINTS 2.1 Navigation For the navigation channel it is required a minimum vertical clearance of 60 m and a clear span of 200 m. And the docking maneuvers impose 100 m wide in addition. Harbourage activities have also conditioned the location of the bridge foundations and have imposed a clear corridor of 150 m next to the main pier. A cable stayed bridge then became the natural (and more economical) structural solution for this crossing. 2.2 Geologic and geotechnical conditions The geotechnical survey embraced the execution of boreholes for visual assessment, SPT surveys (Standard Penetration Test) and LDP tests (Light Dynamic Penetrometer test). At the North bank the resistant layer is located 30 m deep at brown clays. Near the harbor the deep is smaller up to 11 m. However, in the South bank the alluvial deposits are very high and vary from 15 m up to 30 m. Under it there are other layers with poor resistance. The ground layer of sand is located from 33 m to 52.5 m (Figure 3). 2.3 Hydraulics The estuary’s bed consists of mud and alluvial material subject to the flow and its erosion action. The scour phenomena results from localized erosion due to pylon existence and generalized erosion due to lower bed evolution. 514
Figure 3.
Geological profile at the bridge alignment.
To prevent local scour a rockfill layer will be laid down around pylons. An extensive study of floods and tides of Maputo Bay and KaTembe Bangoloene swamp was performed. It defined the road alignment minimum elevation at the south bank. 2.4 Aeronautics and flight restrictions The Maputo International Airport (MIA) is located in the North of Maputo and its terminal building is 6 km faraway from the Estuary. It has two aeronautic corridors that are prepared to be used in poor visibility conditions. Civil Aviation Convention rules impose a volumetric limitation for the obstacles that are aligned to the conical surface for the approaching of landing strips. Therefore bridge pylons maximum height should stay under that imagine line. In this case they cannot have the same height (Figure 4). 2.5 Environment An extensive environmental impact study was conducted across Maputo and Maputo Province, including the districts of KaTembe, Boane and Matutuine. The global project concerns the crossing over the Estuary and also 192 km of roads to Ponta do Ouro (and South Africa) and to Boane city as it is indicated on Figure 5. That highway to the south passes through Maputo Special Natural Reserve (also known as Elephant’s Reserve) and the river Futi corridor. This canal is used by wildlife in their migration to South Africa Reserve. 2.6 Rail traffic At Maputo railway central station three international lines (Ressano Garcia to South Africa, Goba to Swaziland and Limpopo to Zimbabwe) are connected in multiple lines that enormously enlarge the railyard. The piers cannot be placed next to the rail lines. 2.7 Road traffic Nowadays, the connection between Maputo and KaTembe is assisted by ferry. That service level is very poor. There is low frequency of vessels, low reliability and high prices, frequent accidents occur on the docking platform at KaTembe side and consequently the connection between the banks 515
Figure 4. Aeronautic restrictions – plan view and side elevation of the conical surface.
stops. There is also a long road that crosses Matola through Boane ending in KaTembe Southwest. Despite its extensive length this road is narrow and unpaved. The new bridge then becomes an urgent option. Its road alignment and elevation were defined to ensure an excellent level of service on the link of the two banks, quickly and safely (Figure 6). The road profile is high enough to guarantee the necessary navigation clearance. The platform comprises two sides rigidly separated with two lanes each. 2.8 Water supply and others The water supply of KaTembe region that is needed for its development was planned to be provided by a water pipeline coming from Maputo. The region cannot take anymore boreholes as the water is becoming salty. The water pipeline should have a diameter of 1200 mm. More recently state authorities of water resources management decided to provide water from KaTembe south-west. At the banks there are innumerable restraints for footing the piers regarding the current usage of the ground – railyard, harbor facilities, boat dock, services, etc. The transition piers between the main structure and the access viaducts and also the retention ones required particular attention and conditioned the structural solution in either the spans dimension or the constructive process to be adopted. 516
Figure 5.
Maputo Province map and the global road project.
Figure 6.
Road cross section.
3 THE BRIDGE 3.1 The main bridge Betar was involved in the global project embracing the bridge and its transition points on the two banks. Once the bridge will allow Maputo urban expansion – that is currently smothered by poor construction in the North – the global project included as well a Master Plan and Environmental Studies for KaTembe region to stablish over 400.000 inhabitants for the next 30 years. 517
Figure 7.
Crossing Maputo/KaTembe preview for the cable stayed solution.
Figure 8.
Main bridge – deck and pylon cross-sections.
At the conceptual design phase two structures were studied as the constraints of different stakeholders were being cleared – a cable stayed bridge and a suspension bridge at last. The cable stayed solution is the most economic. The first concept design placed the two pylons on the Estuary. It was a symmetrical cable stayed structure with a central span of 350 m long and four traffic lanes. Due to docking problems and to future port expansion concerns the span was widened up to 423 m. But the towers grew differently to fulfill the aeronautic requirements that now joined the constraints list. The bridge is aligned with the clear space for landing included in Maputo’s airport approaching corridor. The main bridge was 815 m long. The deck was a composite cross section with two longitudinal concrete girders at each side of the platform and connected with a steel grid supporting a thin concrete slab. The stays were arranged in a semi-harp pattern with two inclined planes (Figure 7 and 8). The towers had the maximum height possible and were unequal. Therefor they had different number of stays and they suspended different lengths of the platform. These lengths were optimized in accordance to an economic relation height-span. Thereby the total extension of the main span could fulfill the total navigation clearance. 3.2 Access viaducts The proposed structure for the northern approach viaduct was a concrete box girder bridge with variable depth, 911 m long (inserted in the cable stayed solution) with multiple spans up to 135 m (Figure 9). It was planned to be executed by balanced cantilever method due to several constraints at the ground. 518
Figure 9. Viaducts deck cross-sections at piers and mid-span – North bank.
Figure 10. Viaducts deck cross-sections at mid-span and over piers – South bank.
Figure 11.
Layout for the cable stayed solution.
The southern viaduct was 1193 m long (inserted in the cable stayed solution). The deck was a concrete box girder and was conceived to be constructed with a launching girder (Figure 10). The final layout of the cable stayed solution is shown on Figure 11 and 12. However, the solution was still conditioned to additional security features to protect the South pylon located on water and also logistic difficulties at the North bank pylon implanted inside the Port. 519
Figure 12.
Crossing Maputo/KaTembe preview for the cable stayed solution, shown through West.
A suspended solution was then studied since the towers were already at the maximum elevation level. The new solution is 30% more expensive. The central span expanded to 675 m long. The deck became a three-dimensional steel truss and the viaducts were redesigned. 4 FINAL REMARKS The conceptual design and also its complementary studies were a very challenging work carried out by Betar either as a global manager or as a specific designer. The dimension and complexity of the global project embraced multi-disciplinary teams regarding several studies for traffic, growth prediction, roads, structures, hydraulics, environmental, social, economics, financial, juridical, etc., among others specialties.
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The South Approach Viaduct of Izmit Bay Crossing Project Nehir Güngör Chief Engineer, General Directorate Of Highways, Turkey
ABSTRACT: Gebze-Orhangazi-Izmir Motorway (Izmit Bay Crossing and Connection Roads) is one of the most important transportation infrastructure projects in Turkey. The South Approach Viaduct is located on an important part of Gebze-Orhangazi-Izmir Motorway Project and south of Izmit Bay Suspension Bridge. The project location crosses North Anatolian Fault and it lies on the secondary fault zone because of this, different approaches are implemented for the design and the construction of the Viaduct. This paper outlines briefly the design criteria, the construction methodology and application stages of project; which is quite different from the traditional methods.
1 INTRODUCTION Gebze-Orhangazi-Izmir Motorway (Izmit Bay Crossing and Connection Roads) is one of the most important transportation infrastructure projects in Turkey, having the largest tender budget so far. Gebze-Orhangazi-Izmir Motorway Project is under construction which is tendered by a Build Operate Transfer (BOT) Method. The Izmit Bay Crossing is the critical link of the 420 km GebzeIzmir Motorway. KGM awarded the project through BOT model to the OTOYOL A.S. in 2009. Nurol Insaat ve Ticaret A.S. was awarded by OTOYOL A.S. for the design and construction of the on-going the South Approach Viaduct of the Izmit Bay Crossing in 2012 (Figure 1). The structural design was carried out by Wiecon, geotechnical, geological and seismological evaluations of the proposed project region was performed by Fugro-Sial. The Izmit Bay Crossing Project; that is composed of a 3 km-long suspension bridge and a 1379 m long South Approach Viaduct (SAV) will connect the Diliskelesi peninsula on the North with the Hersek peninsula on the South. SAV is located along the western part of Hersek peninsula. SAV is located at one of the most seismically active places in the world. The Project construction site which has the potential to experience significant earthquakes associated with the relative motion accommodated on the North Anatolian Fault is underlain by deep deposits of soft soils and areas of unstable and liquefiable soils. SAV will experience significant earthquakes because of one of the world’s longest strike-slip trending the North Anatolian Fault Zone (source of the 1999 Magnitude Mw 7.6 Izmit and Mw 7.2 Duzce earthquakes). The deck is orthotropic steel deck. The superstructure of SAV consists of two steel box sections connected by a cross beam system which is arranged at a 4 m spacing. The total width of the deck is 35.93 m and the longitudial slope of the viaduct is 3.5%. Piers are flexible steel columns (two leg frame). Slurry walls are implemented as a box type deep foundation under SAV of the Pier foundation. 2 DESCRIPTION OF THE PROJECT The South Approach Viaduct is exactly 1378.92 meters long, from Km:7 + 081.76 to Km:8 + 460.68; has a straight alignment of 299.11 m, from Km:7 + 081.76 to K:7 + 380.87, a 899.19 meters long section with constant radius of curvature R = 2000 m between Km:7 + 380.87 to Km:8 + 280.06 and a clothoidal section from Km:8 + 280.06 to the end of the viaduct. The 521
Figure 1.
Location and alignment of the SAV.
Figure 3.
Cross section of superstructure.
Figure 2.
General layout of SAV.
following section has a constant radius of curvature, out of defined limits from SAV. The clothoid parameter of these two section is A = 670. The longitudinal profile includes a radius with salient angle R = 20 000 m from Km:7 + 081.76 to Km:7 + 277.00; a constant gradient of 3.5% from Km:7 + 277.00 to Km:8 + 397.26, and then a radius with a re-entrant angle R = 10 000 m. The structure of deck consists of 12 spans; the South side span is 72 m long from South Abutment A12 to pier P11, from the South, the 10 main spans vary from 100 to 136 meters long, each one longer by 4 meters than the previous one, the North side span between P01 and transition pier of Suspension Bridge P00 is 126.92 meters long. 2.1 Structure type The entire structure above the foundations upward consists of steel. The deck section is a twin box orthotropic steel structure continuous for the entire length of the project. The two box sections are linked by a series of transvers steel beams which are spaced at 4 m. All enclose spaces such as the inside of the piers and cap beam as well as the two boxes of the superstructure are protected against corrosion primarily by dehumidification system. SAV is designed to be launched from A12 to P02, it is necessary to overcome the geometric alignment problems between A12 and P02 in order to make the launching method feasible. The method employed on this project to make the deck suitable for launching is keeping the two box sections on a constant horizontal radius all the way from A12 to P02, the deck is fabricated in section behind the abutment in 40 m long sections and after completion of each segment it is launched by means of strand pullers SLU. The two longest spans will be constructed by heavy lifting of the entire span due to the fact that the geometry in that portion of the deck is not suitable for launching. The weight of the heavy lifting sections is 4 750 tons. The total weight of the deck is 26 000 tons. 522
Figure 4.
One quarter of a module with all parts shown.
Figure 5.
Section of pier and deck.
Piers are flexible steel box (4 m × 4 m) columns. The reason of the selection of the pier structure type and shape is to reduce mass and to decrease the structural stiffness which results in a period shift. An increased structural period generally results in a reduction of acceleration. The piers consist of two vertical brackets, composed of steel box girders with sections of roughly 4.00 × 4.00 meters in dimension. The height of the brackets varies from 6.55 m in Pier P11 to 47.739 m in pier P01 and the thicknesses of the plates composing the piers vary depending on the height of the pier. Total weight of the piers is 7 000 tons. The brackets are rigidly linked to the footings. The rigid link is made by steel anchorage blocks which are embedded into the concrete cap of the slurry wall foundation (letting the top reinforcement of the cap pass through the embeded plates). 3 GEOTECHNICAL EVALUATION SAV is located at one of the most seismically active places in the world. SAV construction site crosses which has the potential to experience significant earthquakes associated with the relative motion accommodated on the North Anatolian Fault, is underlain by deep de-posits of soft soils, and areas of unstable and liquefiable soils. SAV is within 5 km from the North Anatolian Fault and within a zone of secondary deformation around the primary trace of the North Anatolian Fault. Due to the very active tectonic environment of the project, the project design criteria suggest that Soil-Structure Interaction (SSI) analyses be performed to evaluate key issues that influence the response of the structure. These include (i) the strong ground motions originating from the fault rupture on the North Anatolian Fault to the south of the structures, and (ii) fault rupture 523
Figure 6.
Section of foundation.
effects originating from the relative motion accommodated on the secondary faults in the southern approach structures area. The ground motions including the fling effect considers the tectonic deformation due to fault near viaduct. The ground motions with secondary fault effect consider that faulting is not at a specific foundation location but faulting between any two piers. Numerical analyses were performed to evaluate the demands on a pier foundation in terms of displacements and rotations due to fault rupture through a foundation. 3.1 The foundation system The slurry wall was selected for the pier foundations due to its more satisfactory performance in reducing the risk of overturning under the earthquake loads and limiting the foundation rotations under fault rupture demands. The length of the slurry walls vary from pier to pier based on the soil profile and the earthquake loads at each pier. Due to the close proximity of the viaduct to the North Anatolian Fault, the size of the foundation system is controlled by demands from earthquake loads rather than static loads. A performance based design of the foundation system was adapted for this project. The slurry wall system geometry and plan view dimensions (8 m × 21 m) were selected for each Pier location. The slurry wall construction required a special technique for providing the continuity of horizontal reinforcement, special reinforcement cages in different shapes and dimensions for each panel type were used in accordance to the dimensions of the excavated slurry wall panels. Each steel reinforcement cages was assembled horizontally on the ground in a single longitudinal section. Each cage was stiffened, using convenient stiffening elements placed between the main reinforcement bars, to provide the cage rigidity needed to avoid deformations during lifting and lowering into the trench. The maximum weight of each cage was about 40 tons. Concreting works were executed by using tremie method.
4 DESIGN EVALUATION 4.1 Design philosophy SAV is located in a highly seismic area, with the added complication of a potential fault rupture zone traversing the alignment of the viaduct, it was required to come up with a special seismic isolation solution for this project in order to ensure that this structure will withstand the specified EQ. In Detail Design; the flexible structural system was selected which has a significant contribution to the overall isolation philosophy. 524
Seismic design philosophy for seismic protections is light–flexible structure, energy sharing system and energy mitigation. Energy mitigation takes advantage of both seismic isolation and energy dissipation and is therefore the most effective means to achieve an economical design with relatively large factors of safety. By using this design approach it is possible for structures to withstand large earthquakes without major damage and thereby ensuring the continued use of the structure immediately after such an event. Seismic isolation is chosen for: i) decoupling of the structure, ii) limits energy that is transferred to the structure, iii) period shift reduction – reduction of spectral acceleration during seismic event. 5 SEISMIC ANALYSIS For the design of this viaduct great emphasis is placed on the seismic design as the earthquake event essentially controls the design of the substructure of this project. Three-level Design History Motion for earthquake evaluation: Functional Evaluation Earthquake (FEE) Safety Evaluation Earthquake (SEE) No Collapse Earthquake (NCE)
50% probability of occurrence in 100 years 10% probability of occurrence in 100 years 4% probability of occurrence in 100 years
FEE level ensures the essential elastic response in substructures for more frequent or expected earthquake. SEE level allows the stress to exceed the yielding stress at the extreme fibers of the section of piers. The criteria shown given in the AASHTO LRFD code are bases on this SEE level earthquake. NCE level governs the limits on the inelastic deformation in substructures and the design displacements for the supported isolators of the superstructure for maximum rare earthquake with high magnitude or maximum considered earthquake (MCE). A three level design is part of the entire design process in the viaduct seismic design. The seismic provisions in AASHTO-LRFD form the basis of the design. AASHTO Guide Specification for LRFD Seismic Bridge Design, 2012 is taken as complementary reference to the main AASHTOLRFD code. The seismic design is carried out by using a force-based approach with an appropriate modification factor. The factor, R reflects the non-linear energy dissipation capability of the piers, the increase in natural period, damping and most importantly, the ductility and redundancy of the structural components. The ductility of the substructure is mainly in the long steel piers. The pier cap and piers comprise a ductile moment-resisting frame. In this project, the pier heights vary from 46 m to 6 m and isolators shall be installed between the superstructure and substructure at each pier location. The taller piers are much more flexible and therefore undergo much larger deflections than the short piers and the isolator at short piers would therefore be much more effective for energy dissipation and this is the reason why the short piers have been fitted with twice the number of dampers that the tall piers. For the taller pier, a plastic hinge is allowed to develop based on AASHTO Code. In order to develop the high ductility of the tall piers previously mentioned modification factor will be used. 6 ISOLATION SYSTEM FOR SAV SAV is located in a region which exposes this structure to severe seismic conditions. To make conditions even more challenging it is required that the SAV be able to withstand a fault displacement event with movements up to 1000 mm in a direction perpendicular to its alignment. In order to be able to design a structure to experience only limited damage under the maximum earthquake event of 2 475 years, it will be required to incorporate a seismic isolation system into the design, which would effectively be able to isolate the SAV and thereby reduce the forces and displacement that this structure will be exposed to, during the occurrence of the design earthquake. 525
Figure 7.
General view of SAV (02.15.2015).
Seismic isolation takes advantage decoupling of the structure, limits energy that is transferred to the structure and period shift reduction – reduction of spectral acceleration during seismic event. In order to reduce the accelerations that the SAV will be exposed to during the specified seismic event, it has been decided that a combination of mechanical devices will be utilized to mitigate the forces and accelerations to levels that will ensure that the structure will not collapse and not result in loss of life, during and immediately after the occurrence of such a catastrophic event. The isolation system chosen for this viaduct had to fulfill several critical conditions one being the potential large displacement as a result of fault rupture and the second being to ensure and guarantee adequate isolation of the superstructure, to ensure that it performs in accordance with the project seismic performance requirements. The isolation system chosen essentially consists of a two part system: the bearings and dampers. The isolation system for energy dissipation that fulfilled all of the requirements essentially consists of LRB bearings. They have vertical load carrying capacity and re-centering ability. In conjunction with a finely tuned damper system that effectively isolates the superstructure form the substructure during an extreme seismic event, at the same time dissipating large amounts of energy, which is essential for the survivability of this structure. 7 CONCLUSIONS In general, when designing structures in seismically active areas, critical structures are typically located away from known fault. However, for long structures, a fault maybe unavoidable, and fault rupture risk is impossible to preclude. In the subject project interpreted geotechnical and geophysical data collected during site investigation revealed numerous traces of the secondary fault zone on the entire area near the South anchorage of the main suspension bridge and SAV. However, in the light of the complex projects, the South Approach Viaduct is being built by expertise that can even stand against very intense earthquakes. This is a very promising step in the engineering sector in Turkey. This project might be considered as a good model in Turkey; which verifies that major iconic structures can be constructed at one of the most seismically active places in the world. REFERENCES Fugro-Sial, 2012. Additional Site Investigation Results for South Approach Viaduct. Kızılkaya, A., Gökalp, A., Yıldız, N., 2014. Continuous Horizontal Reinforcement Diaphragm Wall Application, Istanbul Bridge Conference. Wiecon, 2014. Detail Design Report of the South Approach Viaduct of Izmit Bay Project.
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Effect of hangers disposal on the steel consumption for bowstring arch bridges M. Daraban & I.R. R˘ac˘anel Strenght of Materials, Bridges and Tunnels Department, Technical University of Civil Engeneering of Bucharest, Romania
ABSTRACT: Bowstring arch bridges defines a structural system where the compression force in the bow is balanced by the axial tensile force in the tie, which is also called stiffening girder. They are based on a Swedish concept implemented for efficient bridges in terms of steel consumption for spans until 120 m, aesthetics and the advantage of low construction height. The optimization of this type of structures was discussed since 1940, the subject being treated at IABSE Congress Association Internationale des Ponts et Charpentes. This structural system is used for almost a century, allowing a variety of shapes, which is expressed both in bow geometry, in structural system chosen for tie and in technology (hangers, anchors) and erection systems. The variety of architecture inevitably led to different hangers disposal based on aesthetic or design, disposal which has a significant effect on the consumption of steel and structural response, which is treated in this paper.
1 BOWSTRING ARCH BRIDGES FOR MULTI-SPAN LARGE BRIDGES Bridges have always been considered complex artwork. Their evolution shows the level of knowledge of the laws of nature, their greatness intended to inspire power, security and trust. If we analyze the top openings of arch bridges, first is Chaotianmen bridge form Chonging China, which is built in 2009 with 552 m span. This bridge is longer than Lupu bridge from Shanghai China, who held the record for the structure with the largest opening until 2003. All arch bridges with span over 250 m, constructed until this time, have hangers arranged in Langer system because this system allows the minimum ratio f /L(rise/span). The field of spans under 250 m is currently owned by bowstring arch bridges. This type of structures is based on a Swedish concept, implemented to design efficient bridges in terms of steel. As a basic principle, the moments must be taken by the bow and by the tie directly proportional to their flexural stiffness. This principle was signaled for the first time by Prof. M. Ritter at IABSE Congress (Association Internationale des Ponts et Charpantes) in 1940. Optimal distribution of steel between the two elements should be considered and for maximum efficiency, the bow is so designed as buckling resistance is fully used, but without neglecting its agreeable shape. The severity of the Swedish design rules, combined with high ratio between engineers salaries and materials cost, cause the implementation of this type of structure, which is a very efficient bridge structure. Thus, engineers have designed structures or bridges unspectacular with not many different structural elements. An eloquent example of a multi-span large bridge realized with multiple bowstring arch is Porsi bridge, over the Lule river in Laponia, built in 1960. In this paper three solutions for hangers disposal are discussed: Langer, Nielsen and radial disposal. The radial hanger disposal is a recent architectural solution for bowstring arch bridge. We can mention two important bridges, one built in 2005 over Danube in Bratislava with 231 m span, called Apollo, and the other one, buit in 2007 on the A1 motorway Torino-Milano in Reggio Emilia, designer by Santiago Calatrava with 220 m span. 527
The best solution of the three mentioned in terms of static efficiency and low costs is Nielsen solution, patented in 1920. The major advantage of Nielsen system is the reduction of bending moments on the bow and tie and the disadvantage is the difficulty of calculation. For some structural conformation, ascending hangers may unload under live loads. The optimization of such type of Nielsen structures involve complicated analysis of second order. This disadvantage of the solution does not appear in Langer system, that’s why Langer solution is often preferred by designer to Nielsen solution, neglecting the financial implications.
2 THE AIM OF THE PAPER. DESCRIPTION OF THE ANALYZED STRUCTURES The aim of this paper is to establish the effect of hangers disposal on the steel consumption for bowstring arch bridges. To achieve this proposed target, a Nielsen bowstring arch bridge was designed. The span of the bridge is 100 m and the rise is 22.5. The design was conducted so that no hanger unloading under live loads occurs. If this condition is accomplished, then the calculations may be performed in linear elastic domain, using a first order analysis. The angles between bows and the tie is 76 degrees. The bows are provided with top bracing. The tie consists of an orthotropic plate. The hangers support the tie girder at each 10 m. The entire structure is made of steel S235 J2. The cross girders are spaced equally at 2 m. Keeping the conformation described above, the hangers were changed to those shown in figure 4, then to those shown in figure 5, resulting a Langer arch bridge, respectively an arch with hangers radially arranged. A comparison of bending moments on bows, ties, cross girders and on the longitudinal ribs at ULS for the three arrangements of hangers described above is presented. The resulted cross section for each structural element mentioned above, for each of hangers disposal solution is also shown. It is important to mention that the tensile stresses caused by local forces induced into bows and ties by hanger weren’t considered
Figure 1.
Bridge over Lule river, Laponia, picture from
Figure 2.
Cross section of the bridge (structural details for cable anchorage from )
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in sizing calculation. The values of these stresses depends on the chosen construction system for attaching cables and shouldn’t modify bows and ties section along whole length. 3 CONSIDERED ACTIONS The actions considered for sizing the structural elements were: the permanent loads and live loads LM1, specific for a highways traffic, choosing the coefficient:
The combination of actions considered at ULS:
Figure 3. The disposal of hangers in Nielsen structural system.
Figure 4. The disposal of hangers in Langer structural system.
Figure 5. The radial disposal of hangers.
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4 STUDY RESULTS The static calculation was performed using a finite element software for structural analysis pro. 4.1 Cross beams Static analysis confirmed that the arrangement of the hangers has an insignificant influence on sectional efforts in cross beams. 4.2 Longitudinal ribs The variation of bending moments in longitudinal ribs according to the hanger disposal is due to the stiffness of the elastic supports provided by hangers in their attachment point to the tie.
Figure 6.
Moment envelopes for longitudinal ribs at ULS 146.
Figure 7.
Comparison between the dimensions of longitudinal ribs for the three solutions analyzed.
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4.3 Bows and tie
Figure 8.
Moment envelopes for bows and ties at ULS.
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Figure 9.
Comparison between sections and internal forces in bows for the three solutions analyzed.
Figure 10.
Comparison between sections and internal forces in ties for the three solutions analyzed.
5 CONCLUSIONS The current trend is to design Langer bowstring arch bridges, with a small ratio f /L (rise/span) which allows a circular shape for bows, avoiding the parabolic shape. From the technological point of view, the difficulty involved by erecting a parabolic bow or a circular one is the same. A small f /L ratio for bows has the following implications: axial forces and bending moments in bows and ties are higher, the hangers must be disposed in Langer solution, the impossibility to provide a bracing on top. Arches without bracing on top require higher values for the moment of inertia in the plane perpendicular to the plane of bows to ensure the lateral buckling stability of the bows. As a conclusion, the Nielsen bowstring arch bridges are more efficient that bowstring arch bridges with Langer or radial disposal of hangers. The biggest problem of Nielsen solution is the hangers unloading which can be solved adopting a higher headway in small steps, till the hangers will be in tension in all load cases. To achieve a pleasant shape for bows, a parabolic curve can be chosen, which has furthermore the advantage of reducing bending moments at springers. REFERENCES ; ; ; .
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Strategy for durability of structural concrete in Mega-Sealinks in tropical sea-waters V.K. Raina World Bank, Bridge Branch, Department of Roads, Ministry of Physical Infrastructure and Transport, Kathmandu, Nepal
ABSTRACT: Tropical sea-waters, generally, are highly charged with the attacking chlorides, sulphates and sometimes even with molluscs – the tripple killers of structure–durability, given the warm temperatures and high humidity. The technical requirements for design of structures in such waters, are a class of their own that call for clinical attention. 1 DURABILITY AGAINST WHAT? The structural concrete here is located in severe marine enviromment with extreme exposure to high water salinity and temperatures, temperature gradients, airborne salt spray and sometimes possibly even the limestone-devouring marine borers (also called: molluscs or piddocks). This exteme environmental exposure over a service-life of at least 100 years, calls for precise attention during design, construction, operation and maintenance – a very demanding task that cannot be fulfilled with traditional and simple materials and procedures. The basic methodology is to define exposure conditions for a number of relevant exposure classes, and then use these classes as basis for a ‘strategy’ to be consistently incorporated in the design and specification. For each class the relevant deterioration mechanisms must be defined, and a ‘multi-stage protection strategy’ be developed and applied. A multi-stage protection strategy comprises two or more selected barriers that each: – prevents or slows down transportation of aggressive chemicals/pollutants from the environment into the concrete structure, or/and – passivates the subsequent potential deterioration mechanism. 2 EXPOSURE CONDITIONS 2.1 Sea water and ground water pressure on foundations and substructures This zone comprises all foundations and piers and abutments up to level below splash and tidal zones. Sea water with salts enters the concrete by hydraulic pressure and salt diffusion. 2.2 Water ‘splash’ The splash zone varies and can extend from almost near land level and up to about +4 m above high tide level (HTL). The concrete surface is subjected to alternate wetting and drying (which causes capillary suction), resulting in repetitive salt concentration build up again and again. 2.3 Water ‘spray’ Likewise the spray zone also varies and extends from a few metres above land level and up to about 6 to 7 m above HTL. The concrete surface is thus subjected to alternate wetting and drying (capillary suction), again causing salt concentration build up. 533
2.4 Sea water composition, chlorides, sulphates, other salts, and pH The shallow areas are characterised by very saline water with up to about 4–4.5% salts (40,000– 45,000 ppm). The salinity can reach even about 6%. Out of the salinity, more than half the salts are chlorides. 2.5 Temperature The sea water temperature can vary between about 17◦ C and 37◦ C with a ground water temperature of about 27◦ C. 2.6 Molluscs (marine borers) One group of molluscs (also called piddocks or marine borers) of the family: Pholadidae, has an ability to excavate for itself depressions in calcareous rocks and some types of concrete. Hence concrete with limestone (calcareous) aggregate can be the victim. Concrete with granite and gabbro stones can only be destroyed by molluscs which rasp the material off – which is unusual. 2.7 Airborne chlorides, sulphates and sea water spray These cause Salt concentration same or higher than in the sea water. 2.8 Windborne dust and sand Dust and sand storms can occur, bringing material which is contaminated with chlorides (NaCl and CaCl2 ) and sulphates (specially gypsum). 2.9 Humidity This can be very high, even 100% during many months in a year, and this eases movment of chloride and sulphate ions that use mositure as the vehicle in travelling through the pores in concrete in reaching steel and cement, respectively. 2.10 Temperature The air temperature in the shade can vary between low to about 45◦ C. The maximum temperature on irradiated surfaces exposed to direct sun can reach about 50◦ C, and even more. 2.11 Oxygen and carbon dioxide The ambient atmosphere contains oxygen and carbon dioxide, the former enhances the corrosion by chlorides and the latter enhances carbonation (reacting with calcium Hydroxide, forming calcium carbonate which lowers the pH, thus increasing acidic ambience). 3 DETERIORATION MECHANISMS 3.1 Chloride attack Steel reinforcement corrosion caused by two different actions: – Chloride-induced corrosion due to High concentration of chlorides within concrete matrix, and – Corrosion due to inward penetation of Carbonation-front from concrete surface. Chloride induced corrosion: a) A high chloride content causes dissolution of the protective ‘passive’ ironoxide layer on the steel, resulting in active corrosion, typically as pitting corrosion. Pit corrosion is usually a localised corrosion which reduces the rebar cross-section significantly over small areas and can be very dangerous. 534
b) The chloride concentration which is required to cause corrosion, the so-called threshold value, is not a constant, but varies with many parameters, of which the most important are: – relative humidity in the concrete pores – cement type – cement content – additions, such as silica fume and fly ash, etc. – water/cement ratio – availability of oxygen – chloride contamination (from surrounding water and soil, and from the aggregates, the mixwater, and the curing-water) – denseness of concrete c) The main factor deciding whether or not chloride induced corrosion takes place is the chloride concentration around the steel (reinforcement and prestressing cables). The following general guidelines apply in normal ambience: Chloride concentration, % by weight of concrete Below 0.05% 0.05–0.10% Above 0.10%
Risk of corrosion Low Corrosion does not normally take place, but there are severe exemptions to this Medium Active corrosion is normally decided by rest of the above mentioned parameters High Corrosion will normally take place
Corrosion caused by carbonation-front penetrating inwards from surface to steel. Corrosion may be initiated when the concrete cover gets completely carbonated. The penetration of carbonation front from the concrete surface depends mainly on: a) b) c) d) e) f) g)
relative humidity in the concrete pores cement type cement content additions such as silica fume and fly ash, etc. water/cement ratio availability of carbon dioxide denseness of concrete.
The maximum rate of carbonation is usually found when the relative humidity is in the range 60–80%. Carbonation does not take place in saturated concrete and very slowly in near-saturated concrete because water-filled concrete pores prevent ingress of carbon dioxide. Cementiteous combinations which form a high calcium hydroxide content during hydration will cause a slower rate of carbonation e.g. use of CEM I (ASTM cement Type I) without any pozzolanas. Combinations with ASTM-CEM III or pozzolanas will give denser concrete and develop a lower penetrability of carbon dioxide and thus a lower rate of carbonation and lower amount of calcium hydroxide formation during the hydration. Corrosion, caused by loss of the protective passive alkaline layer (reduction of high pH) around steel due to carbonation of the concrete cover, is usually the so-called general corrosion which results in a relatively uniform reduction of the rebar cross section in large areas. 3.2 Sulphate attack Expansive sulphate reactions are seen in presence of sulphate ions when CEM I with a moderate to high C3A-content is used in cement. The expansive reactions will result in cracking and ultimately 535
disintegration of the concrete structure (due to formation of sulpho Aluminate crystals which demand and wish to occupy more volume than available and hence cause expansive bursting pressure). Sulphate attacks are usually not seen in concrete with ASTM-CEM-V (because it has low C3A content). 3.3 Seawater attack The combination of chloride and sulphate salts in seawater usually will not cause much deleterious effects of chemical reactions in the concrete if the concrete is dense, has low water/cement ratio, and C3A-content is well below 8% (about 3% as in CEM V, called Sulphate Resisting Cement SRC). 3.4 Salt weathering (alternate wetting and drying of concrete) Alternate wetting and drying of concrete in the splash zone and spray zone causes a build-up of salts within the concrete pores in consequence of ingress of seawater and Evaporation of the water leaves precipitated (damaging) salts in the concrete pores. During subsequent wetting the salts are re-hydrated, generating a physical expansive pressure in the concrete pores. Such pressure may lead to cracking and spalling of the cement skin and later the surface layer of the cement mortar. This effect reduces the thickness of the concrete cover and thereby the protection of the rebars. This deterioration mechanism is specially effective in the Arabian Gulf where the salt content in the seawater is very high and the drying effect between wettings is rapid due to high temperatures. 3.5 Alkali–aggregate reactions Alkali-silica and alkali-carbonate reactions are chemical reactions between alkalis (sodium and potassium ions) in cement and reactive silica or carbonate aggregate particles (e.g. Opal). The reactions take place in a moist alkaline environment only. The reaction products, upon drying, demand and occupy a larger volume than the dissolved parts of the aggregate particles and thereby create bursting stresses within concrete, leading to cracking and disintegration of the entire concrete matrix. 3.6 Leaching Concrete exposed to water with a low carbonate content may be subjected to gradual dissolution of first the calcium hydroxide and later the calcium silicated hydrates, thus exposing the matrix of concrete. This effect is called ‘leaching’. 3.7 Temperature Difference between coefficients of thermal expansion of coarse aggregate and cement mortar may lead to internal cracking in the concrete matrix when the concrete is exposed to large temperature and moisture variations (differential thermal movements). As an example, we note that a concrete mix with 370 kg cement per m3 made with a dense limestone aggregate may have a coefficient of thermal expansion of about 6.2 × 10−6 per ◦ C, whereas the same concrete made with e.g. gabbro coarse aggregate may have a coefficient of thermal expansion of 9.3 × 10−6 per ◦ C. The temperature sensitivity of a concrete structure is therefore reduced significantly by using different Coarse Aggregates. 3.8 Weathering from abrasion (mechanical action of airborne dust and sand) Weathering of concrete surfaces from abrasion with airborne dust and sand is normally only observed on inland structures in open landscape where the wind can carry dust and sand unhindered to the lower surfaces of concrete structures. Structures with high surface strength are normally not significantly affected. 536
3.9 Non-structural cracks Non-structural cracks in concrete comprise: – Plastic settlement cracks – caused as the Concrete ‘settles downwards’ around rebars, thus ‘breaks its back’ over the rebars (creating cracks in cover); also creates water bellies under the rebars which causes loss of bond when this water dries up! – Plastic shrinkage cracks – caused by lack of ‘evaporation-protection’ during the first few hours after placing the concrete – if the rate of surface-evaporation is faster than rate of bleeding, the concrete surface tends to dry up and shrink and – being plastic – hence cracks at surface and the crack traverses downward until its propagation stops due to initial setting of concrete. – Thermal cracks from heat of hydration – caused by large differential temperatures between newly and previously cast concrete or between different locations in newly cast concrete. – Drying shrinkage cracks – caused by drying out of the concrete matrix when effective curing is stopped too early. 3.10 Structural cracks The most important structural cracks are: Cracks caused by restrained thermal movement – Such cracks should be hindered by (1) providing a sufficient quantity of closely spaced small diameter reinforcement close to the concrete surface and (2) by dividing the structure into units which can contract and expand relatively freely. Cracks caused by structural tensile stresses – Reinforced concrete structures are designed to crack in tensile zones. The design should, however, ensure that the spacing of the cracks and the associated crack widths are both sufficiently small (crack width control). Cracks caused by mechanical damage and impact – Minimise the risk and the consequences of impact from vehicles on the bridges and vessels passing through the bridges. 4 EXPOSURE CLASSES The structures may be divided in to following different Exposure Classes: Exposure Class Structures SUBM-A (buried, submerged)
SUBM-B (in tidal zone)
Exposure conditions
• Water temperature 17–37◦ C • Up to 20–30 m water pressure by sea-water/ groundwater • Cl and SO4 in sea water/soil • Waterborne sand • Molluscs Portions of • Air temperature in shade: Structures in 0–47◦ C tidal zone, up to • Max. temperature on Design High Water irradiated surface exposed DHW-level to direct sun: 80◦ C • Alternating wetting and drying with seawater • Cl and SO4 in seawater • Atmospheric O2 and CO2 • Waterborne sand • Molluscs
Foundations and other structures buried in the ground, Submerged, marine structures
Deterioration mechanisms • Chloride induced steel reinforcement corrosion • Expansive sulphate attack • Abrasion • Mollusc attack • Chloride induced steel reinforcement corrosion • Salt weathering • Carbonation induced steel reinforcement corrosion • Abrasion • Mollusc attack
(continued) 537
Exposure Class
Structures
Exposure conditions
Deterioration mechanisms
SPLA (in splash and spray zone)
Portions of Structures in splash and spray zone
• Chloride induced steel reinforcement corrosion • Salt weathering • Carbonation induced steel reinforcement corrosion • Abrasion
SUP (above splash and spray zone)
Portions of structures above splash and spray zone
• Air temperature in shade: 0–47◦ C • Max. temperature on irradiated surface exposed to direct sun: 80◦ C • Occasional wetting with seawater spray • Cl and SO4 in seawater • Atmospheric O2 and CO2 • Waterborne sand • Air temperature in shade: 0–47◦ C • Max. temperature on irradiated surface exposed to direct sun: 80◦ C • Airborne Cl and SO4 from seawater • Atmospheric O2 and CO2 • Waterborne sand
• Chloride induced steel reinforcement corrosion • Salt weathering • Carbonation induced steel reinforcement corrosion • Abrasion
5 ELEMENTS OF MULTISTAGE PROTECTION STRATEGIES 5.1 Structural design For all structures it would be a general requirement that good run-off is facilitated, accumulation of dust and water is prevented, and steps taken to slow down sulphate, carbonation and chloride ingress. This will require proper geometric shapes such as rounded corners, sloping self-draining surfaces, etc. The structures shall be designed with adequate consideration to minimising the formation of cracks from temperature effects. The risk of formation of cracks from tensile stresses exceeding the acceptable tensile strength of the concrete shall be minimised by adequate amount and distribution of smaller dia reinforcement close to the concrete surface, avoidance of large member thicknesses, abrupt changes of cross sections leading to large stress gradients, etc. 5.2 Concrete materials and composition 5.2.1 Concrete materials The concrete shall be made of materials and with a mix-design which ensure a potentially dense and durable concrete without built-in weaknesses (e.g. high built-in chloride and sulphate content) and an efficient barrier (cover) between the surface and the reinforcement. a) Cement and pozzolanas Several types of cement and combinations of cements and pozzolanas can be used. – Concrete with CEM I (Type I ASTM Cement) with a low C3A content unless made very dense, should only be used for un-reinforced structures. – CEM I with a moderate C3A content, alone or in combination with silica fume or fly ash, may be used for dense concrete in marine structures. – Another option is e.g. CEM III B, the blastfurnace-slag cement. 538
Evaluation of the adequacy of a binder or binder-combination should consider certain important parameters e.g. the rate of and the total heat development,risk of non-structural crack formation, resistance against chloride penetration, threshhold value for steel corrosion, etc. b) Coarse aggregate (CA) High quality coarse aggregates are available as limestone aggregates as well as gabbro aggregates. The limestone aggregate has sufficient strength, and wear on equipment in contact with the aggregate. Thermal movement is lower for limestone aggregate than for the higher strength CA. The latter has a specific gravity 5–10% higher than that of the limestone aggregate. Both types have a low absorption of 0.3–0.6%. Higher strength CA is safer against molluscs. Unit weight of concrete made with such CA is slightly higher. c) Fine aggregate (FA) – Fine aggregates are usually reclaimed from the sea or from inland deposits and require very thorough washing. – In both cases, the grading of a source tends to be single sized, thus requiring blending of FA from two or more sources to provide a suitable grading. – Usage of marine sand requires careful washing and draining to remove most of its chlorides. Many of the marine sand sources consist to a large extent of weak limestone or broken sea shells. It is generally difficult to wash out the chlorides from the calcium carbonate sands because some of the chlorides are apparently trapped inside the particles, requiring a long diffusion time for removal. – Fines from the coarse aggregates can be used to blend with natural fine aggregate to obtain a better sand grading. d) Water Water for all usage in and in contact with concrete shall be of potable quality with very low sulphate and chloride contents (not just potable water – because humans can drink even slightly salty water!). e) Admixtures – Water reducing admixtures and ‘high range’ water reducing admixtures are normally required to make workable concrete mixes with low water/cement ratios. – A retarder may be required to ensure sufficient initial setting time for placing, compacting and finishing the concrete as well as for being able to compact newly and previously cast fresh concrete together without creating a ‘cold joint’. f) Corrosion ‘inhibitors’ Are available for mixing into the concrete with the aim of ‘passivating’ the chloride ions which penetrate the concrete cover on way to the rebars. It is critical that the inhibitor is still active after many years (e.g. 50 years), when the need for it arises. There is at present some uncertainty about the long term performance of corrosion inhibitors. 5.2.2 Concrete composition (w/c ratio, Grading, Admixtures) A critical parameter for a low permeability and highly dense concrete mix is a low water/cement ratio and low water/binder ratio where additions such as pozzolanas are used. Capillary continuity (non-denseness) is normally unavoidable even in well cured concrete if the water/cement ratio is above about 0.50. For High Performance Concrete (HPC) a maximum water/cement ratio between 0.40 and 0.35 is normally specified. The lower the water/cement ratio the lower will be the permeability if the concrete can be placed, compacted and finished well! The structural design and reinforcement should allow for usage of upto maximum of 20 mm to 40 mm coarse aggregate size in most of the structural elements if possible in order to increase the coarse aggregate content (and low shinkage, etc.) and reduce the cement content. 539
A high bulk density of the combined coarse aggregate should be obtained by suitable combination of the coarse aggregate fractions; similarly by suitable combination of coarse and fine aggregates. Due consideration should be given to obtain a cohesive mix with a suitable workability with a high aggregate content (good Gradation). Admixtures and Admixture dosages should be so selected as to give a robust fresh concrete with low tendency of segregation and bleeding, which has a low workability loss (slow and not abrupt) giving sufficient time for placing, compaction and casting together with previously placed fresh concrete. 5.3 Concrete production All storage and manufacturing facilities for concrete constituents as well as batching and mixing plants shall be of such design, types and quality as to ensure continuous production of very uniform concrete mixes. Hot weather concreting precautions shall be taken to meet the fresh concrete temperature requirements. Cooling with chilled water, flaky ice or liquid nitrogen (in water) are methods of reducing the fresh concrete temperature. During the hot weather season, scheduled ‘secondary dosage’ of water reducing admixture and/or ‘high range’ water reducing admixture to be added on site immediately before discharge may be required to maintain a workable concrete mix for a longer time. 5.4 Concrete ‘execution’, including ‘curing’ The main aim during execution of concrete for a concrete structure is to ensure development of the required potential properties of the concrete. Cracking should be kept to a minimum or avoided using partial prestressing if required – where possible. Blow holes and other surface defects should be minimised by e.g. using ‘self compacting concrete’ or ‘controlled permeability formliner’. Application of controlled permeability formliner (CPF) will result in a densification of the outer few mm of the cover which will be significantly less permeable. The concrete cover shall be sufficient to prevent chloride concentrations exceeding the threshhold value at the reinforcement level within the specified service life. The minimum cover should be at least 70 mm, but should not be more than 80–90 mm (risk of cracking of a thick un-reinforced cover layer). The cover could be reduced if stainless steel or non-metallic reinforcement is used. Another element in providing an impermeable concrete cover is to minimise or eliminate the risk of formation of thermal cracks. This can be achieved by application of binder combinations with a low heat development combined with conformity with temperature difference limits obtained from experience. 5.5 Reinforcement One of the main effects of the requirement of a dense, impermeable concrete and its cover is to protect steel reinforcement from corrosion and inhibit ingress and movement of pollutants through any capillaries in the concrete matrix. If the concrete cover protection is not considered sufficient, the steel reinforcement could be cathodically protected right from the beginning or the structure prepared for cathodic protection by ensuring rebar continuity andas yet another option to build-in the anodes for future-use when required. (But large scale cathodic protection in the water-field is neither easy nor dependable). An element in a multistage protection strategy could be to use reinforcement which does not corrode: a) Epoxy coated steel has been used for many severely exposed concrete structures over the past more than 50 years but with varying degrees of success. In some cases (e.g. in many road bridges 540
b)
c) d)
e)
in the U.S.A.) corrosion has started long before expected. As a consequence, several states in U.S.A. have prohibited use of epoxy coated steel. The main cause for premature corrosion of epoxy coated steel is practical defects in the coating, either pin holes or holes formed during handling, tying and construction, and/or its defective bond to the steel. In both cases chloride ions can get access to small areas of the steel causing severe pit corrosion and a quick, dramatic reduction of the rebar cross section. Use of epoxy coated steel is not recommended for concrete structures in the severely aggressive environment under discussion here. Stainless steel. Stainless steel is available in grades that will not corrode in a chloride contaminated environment. Use of stainless steel reinforcement may eliminate any durability related measures needed against chloride ingress and steel corrosion. The long term performance of stainless steel in concrete structures is well documented with examples of more than 70 years excellent performance. Stainless steel reinforcement can be used as replacement for all reinforcement or in selected, specially severely exposed structural elements or parts of structural elements; stainless steel may not be used in combination with ordinary black steel (owing to possibility of corrosion because of presence of dis-similar metals). Non-metallic reinforcement: bars, sheets, fibres. Many types of non-metallic materials are available as alternative or supplementary reinforcement for concrete structures. Some types, e.g. fibres used as supplementary reinforcement, result in more evenly distributed stresses in the structure with a reduced wide crack formation. In addition to relatively high prices, unfortunately few of these materials have a long track record, hence not recommended.
5.6 Coating the concrete surfaces up to a certain elevation Coatings may be applied on concrete surfaces as an element (one barrier) in a multistage protection strategy to provide protection against: a) Water absorption b) Ingress of chlorides and sulphates in seawater c) Ingress of carbon dioxide and oxygen All surface protection systems should, as a minimum, be water repellent and prevent ingress of chorides and sulphates. Impermeable systems will also protect against ingress of carbon dioxide and oxygen, whereas water vapour diffusion open systems may allow penetration of these gases. Each system’s ability to bridge across cracks in the concrete, its bond strength, flexibility, mechanical resistance and UV-resistance should be assessed. Impermeable epoxy systems can be used for high performance concretes and these systems dry out due to self desication during the hydration. When applied correctly, impermeable systems, e.g. epoxy coatings, have good long-term records from marine structures in the Arabian Gulf. 6 SELECTED ‘DURABILITY STRATEGY’ 6.1 General In the preparation of the durability strategy for the structures under reference here, it is logical to search for and study major structures subjected to similar exposure conditions. The obvious example is the King Fahd Causeway (KFC) – the Sealink constructed in the first half of the 1980s between Saudi Arabia and Bahrain. The strict specification for the concrete structures of KFC has resulted in an excellent performance with one minor exception – in that: – Scaling to a depth of 7–8 mm on the vertical faces on the north side of the southern girders. This scaling is generally only seen on the low bridges, but extends to a height of about (levels) +6 to 541
+7 m in some locations on the high bridges. The total extent of scaling is estimated to be about 1% of the total concrete surface area. A TEST PILE, constructed similarly and placed close to this sea-link prior to construction of the sea-link, has been inspected and tested regularly. The pile is not epoxy coated. From measurements of chloride penetration in the test pile in 1994 and 1999 it has been calculated that the time to initiation of corrosion of the reinforcement is about 600 years!! Although this figure cannot be accurate, it no doubt gives an indication of the magnitude of slow timeframe with which this important deterioration mechanism is progressing. The bridge structures were all epoxy coated in the precast yard up to +4 m level, and the chloride penetration behind this coating is nil. During construction, the epoxy coating was extended in-situ up to +8 m level on all piles. The girders are not epoxy coated. KFC is the oldest major bridge structure in this part of the Arabian Gulf and, with its excellent durability performance, it constitutes an important example. Based on the current condition of KFC concrete structures and the estimated remaining service life from chloride penetration tests, it is obvious that KFC should form a significant part of the basis for the durability strategy for the subject concrete structures. Despite the fact that there has been significant development within concrete construction of durable structures in aggressive exposure conditions over the last 40 years, the only changes which represent significant and very well long-term documented improvements should be considered as deviations to the materials, methods and procedures applied for KFC. 6.2 ‘Overall’ durability strategy The most important deterioration mechanism for the subject concrete structures is chloride induced steel reinforcement corrosion. When normal steel is used, the main worry is to prevent chlorides from penetrating the concrete cover to the location of the steel rebars, as also the corrosion of steel from chlorides within the concrete matrix. The ‘Strategy’ for ‘Durability’ is therefore ‘Summarised’ as follows: a) Provide a dense, impermeable concrete and a concrete cover of minimum 70 mm thickness to steel rebars. b) Alternatively, use stainless steel locally (in areas with the most severe exposure, such as the splash zone & areas where the 70 mm cover cannot be obtained). c) The general use of stainless steel could also be considered as the most effective and safe method of securing a long service life. The concrete cover can be reduced to a minimum of 35 to 40 mm where stainless steel is used. d) Apply measures for prevention/reduction of crack formation in fresh, hardening and hardened concrete. e) Use a low/slow heat cement; optimise mix design through trial mixes and full scale castings. f) Ensure compliance with temperature requirements during hardening. g) Use of Controlled Permeability Formliner (CPF) on all formed surfaces – which gives a highly densified surface layer of concrete (to 3–5 mm depth). h) Apply epoxy surface coating on all concrete surfaces up to 2 m above splash zone. i) Use higher strength aggregate (e.g. like Gabbro) in all foundations and substructures upto a level of 2 m above HTL to prevent mollusc (manirne borer) attack. The selected strategy has multi-stage barriers against chloride ingress and carbonation for the substructure from: a) Epoxy coating of concrete surfaces up to the suggested elevation. b) Densified surface layer, 3 to 5 mm, by application of CPF (Controlled Permeability Formliner). c) Dense and low permeability concrete and concrete cover with minimum cracks. 542
The Selected Strategy has Multistage Barriers against Chloride Ingress and Carbonation for the superstructure from: a) Densified surface layer, 3–5 mm, by application of CPF. b) Dense and low permeablility concrete and concrete cover with minimum cracks. Protection against subsequent steel reinforcement corrosion – should chlorides or carbonation eventually reach the rebars – could comprise use of stainless steel rebars or establishment of rebar continuity to allow for a future cathodic protection system (although the efficacy of the latter is more dependable in Laboratory work than in a large sea-structure project in the field!). The protection against mollusc attack is provided by using the higher strength coarse aggregate which is considered a sufficient barrier against these marine borers (_they can eat into limestone, generally not into the higher strength CA like Gabbro). 6.3 Design For all structures it would be a general requirement that: a) b) c) d)
good drainage run-off is facilitated, accumulation of dust and water is prevented, and steps are taken to slow down carbonation, and sulphate and chloride ingress, requiring proper: – geometric shapes such as rounded corners, sloping surfaces, and – highly dense concrete of low w/c ratio (High Performance Concrete).
The structures shall be designed with adequate consideration to minimise the formation of cracks from temperature and other effects. The risk of formation of cracks from tensile strain exceeding the tensile strength of the concrete shall be minimised by: a) adequate amount and distribution of the small diameter reinforcement at close spacing, close to the concrete surface, and b) by avoidance of large thicknesses and abrupt changes of cross sections that lead to steep stress gradients. 6.4 Concrete materials and composition 6.4.1 Concrete materials a) Cement and Pozzolanas – Blastfurnace slag cement with a slag content of 66 – 80% by weigh of cement, CEM-III/B 52.5 SR, shall be used for all reinforced and prestressed concrete. (This cement type, but a lower strength class, was used for KFC). – No pozzolanas will be used. – Silica fume shall not be allowed, except for underwater concrete where it may improve the cohesiveness of the concrete to prevent washing out of fine material. b) Coarse Aggregate CA – Higher strength CA (not Lime stone) shall be used for all substructures (and foundations) to prevent attack from molluscs (marine borers). – Limestone aggregate shall be used for all superstructure in order to obtain lower thermal movements. – The lower specific gravity of the limestone aggregate will reduce the concrete density by 3–5%. The wear on all equipment in contact with concrete with this coarse aggregate is significantly reduced. c) Fine Aggregate The fine aggregate will probably have to be formed out of two or more raw sands from inland and marine sources, thoroughly washed. 543
d) Water Water for all usage in and in contact with concrete shall be of potable quality with very low chloride and sulphatecontents (not just the potable water). e) Admixtures – ‘Water reducing’ and ‘high range water reducing’ admixtures must be used with very low w/c ratio concrete with a normal workability and also if a self- compacting concrete is sought.(There is insufficient experience available with the latter.) – No other admixtures are expected to be required, except for a retarder to allow enough time for concrete placement and compaction. 6.4.2 Concrete composition a) The following seven classes of concrete for reinforced and pre-stressed concrete and for blinding can be adopted (ref. section 5. earlier):
Exposure class
Concrete class
SUBM-A SUBM-B
A40 A20 S40 S20 B40 B20 C (Blinding layer)
SPLA SUP
Maximum water/cement ratio (by wt.)
Minimum cement content kg/m3
0.40
380 400 400 420 380 400 –
–
Coarse aggregate type Non– calcareous (Gabro) Calcareous (Limestone) –
Maximum size of coarse aggregate mm 40 20 40 20 40 20 20 r 1 0
Minimum characteristic compressive cube strength (28 days) MPa 60
12
b) The above seven concrete classes for the suggested exposure classes for the indicated structureportions, shall be as follows: – Class A40 concrete – shall be used for all structures in exposure classes SUBM-A and SUBM-B, except for concrete deposited underwater and when the usage of 40 mm nominal coarse aggregate size is not permitted – Class S40 concrete – shall be used for concrete deposited underwater, except when the usage of 40 mm nominal coarse aggregate size is not permitted. – Class B40 concrete – shall be used for all structures in exposure classes SPLA and SUP, except when the usage of 40 mm nominal coarse aggregate size is not permitted. – Class B20 concrete – shall be used for all other structures in exposure classes SPLA and SUP. – Class C concrete – shall be used for blinding for all structures. Coarse aggregate size shall be selected in accordance with blinding thickness. c) Other concrete composition ‘requirements’: – Concrete workability shall be determined by the contractor taking into consideration complications in placing and compacting the concrete in different structural elements. Self compacting concrete may be a useful alternative for certain structural elements, but there is insufficient experience available in this as yet. – The ‘total acid-soluble chloride content’ in the concrete mix, from all sources, shall not exceed 0.10% by weight of cement for all reinforced and prestressed concrete in and on the sea, and along the coast (The Engineer to decide as to up to how far inland). This limit is difficult to meet, but is not un-realistic. 544
– The ‘total acid-soluble sulphate content’ as SO3 in the concrete mix, from all sources, shall not exceed 4.0% by weight of cement. – To eliminate any risk of alkali-silica or alkali-carbonate reactions, the maximum content of equivalent Na2 O shall be 3.0 kg/m3 concrete with a mortar content of 60%. – The concrete temperature shall not exceed 32◦ C at the point and time of placing. – A high bulk density of the ‘combined coarse aggregate’ shall be obtained by suitable combination of the coarse aggregate fractions. Similarly, shall the high bulk density of the ‘combined aggregate’ be obtained by suitable combination of coarse and fine aggregates. Due consideration shall be given to obtain acohesive mix with a suitable workability using a high aggregate content and low w/c ratio. – Admixtures and Admixture-dosages shall be so selected as to give a robust fresh concrete with low tendency of segregation and bleeding, which has a reliable workability loss-slow and not abrupt, giving sufficient time for placing, compactionand casting together with previously placed fresh concrete. 6.5 Concrete production All elements of Multistage Protection Strategy, mentioned above, shall be implemented as required. 6.6 Workmanship and construction methods The Specifications shall define limits for methods and workmanship, with due concern for the variation in exposure classes. This would include basic requirements, including: a) the use of slip-forming would not be preferred; b) controlled Permeability Formliner (CPF) shall be used on all ‘formed’ surfaces in order to minimise blow holes and other surface defects and to further densify the outermost 3–5 mm of the concrete cover; c) curing requirements would be strict and comprehensive, steam curing would not be allowed (possibility of cracks from temperature differentials); d) restrictions would apply to total and differential temperatures during casting and curing; e) restrictions would apply to ensure careful handling during transport and erection of structural elements; f) strict QA and QC procedures would be required. Concrete cover – the concrete cover to any steel – also including stirrups and links – shall comply with the following:
Ordinary, “black” steel Stainless steel
Minimum cover to any steel mm
Minimum thickness of spacer mm
Nominal cover mm
70 35
75–80 40–45
75–85 40–50
Spacer blocks shall be of cement mortar of similar quality as the mortar of the concrete. Stainless steel spacer may be used if the contractor can demonstrate complete bond with the surrounding concrete. 6.7 Reinforcement Reinforcement shall be ordinary (black) steel (HYSD) or stainless steel. Ordinary and Stainless steel may not be mixed in the same structural element and need not be electrically isolated (using dis-similar metals can encourage corrosion). Epoxy coated steel reinforcement shall not be used. 545
Application of non-metallic reinforcement, whether bars, sheets or fibres would not be allowed for a concrete structure in this aggressive environment with a 100 years service life requirement. 6.8 Coating of concrete surfaces up to a certain elevation The 3-layer (primer in 2 coats) solvent free, two component, epoxy coating on King FahadCauseway (Sea-link) has proved to perform excellently in preventing chloride ingress and already has been durable for about 30 years so far. The spray zone on KFC extends up to 6 to 7 m above MSL. In order to ensure that the coating protects all surfaces in risk of being subjected to salt spray, it is recommended that all reinforced and prestressed concrete structures up to a certain elevation be epoxy coated. No need to epoxy-coat the superstructure, unless subject to salt spray. 6.9 Quality system To achieve the set Durability target in the structures in the harsh environment, it is very important that the selected contractor establishes and strictly follows Quality Assurance (QA) and Quality Control (QC) systems. It is assumed that the selected contractor will operate such quality system and will be ISO certified. Additionally, all aspects of the project, right from the preliminary and detailed design, through method statements, work procedures, inspection-and-test-plans, etc., must be covered by and be implemented under strict compliance of written Quality Assurance (QA) and Quality Control (QC) plans. 6.10 Testing of finished concrete structure Critical properties of the finished concrete (in the structure) shall be documented as part of the required quality documentation prior to contractual handover. Such documentation shall include measurement of: a) Concrete cover; b) Chloride diffusion , and c) may also include measurements for: Accelerated carbonation depth, Electrical resistivity, Capillary absorption or ISAT (initial surface absorption test). These sets of test results will provide the base-line quality of the truly achieved performancerelated qualities in the finished structure from which, later, residual-service-life updating can be made. 7 ‘OPERATION AND MAINTENANCE’ REQUIREMENTS TO BE CONSIDERED IN THE DESIGN 7.1 General The structure shall be designed and built with due regard for future inspection and maintenance, based on the strategies and decisions adopted in design, specifications and construction methods. Advanced structural monitoring systems for early warning of risk of chloride induced steel corrosion shall be incorporated at strategic locations as automatically monitored corrosion and reference cells. An inspection and maintenance programme for the life-time of the structures should be established at the design stage itself as an integrated part of the strategy for achieving the required durability. 546
Also organisational aspects of the future ‘operation and maintenance’ should be taken into account at an early stage, ensuring that all parts of all sea-link structures will be maintained to the same standard, following the same pre-planned procedures and providing consistent records in the same formats. The planning should facilitate easy access and easy repair/replacement of special items like Expansion Joints, Bearings, wearing courses and surface protection, crash barriers, mechanical and electrical installations, and other items where the recurring or potential needs for maintenance, repairs and replacement can be foreseen. 7.2 Planning for ‘access’ and ‘inspection platforms’ The planning for access will comprise the design of the structure (working space, stairs, inspection openings) and also the provision of the necessary permanent inspection platforms (gantries), which must be tailor-made to suit the inspection and maintenance of piers and girders in the various bridge structures (viaduct and high bridge portions). The inspection gantry should be easy to transport and install at any location along the structure, with the ability to let the inspection crew safely reach all parts of the piers and girders. Permanent rails under the bridge superstructure should be provided for easy movement of the gantry, and the gantry should be able to pass the piers when travelling on such rails along one bridge in its entire length. The gantry should, for safety reasons, be operated directly by a properly trained and fully selfreliant crew placed on the travelling gantry platform itself, not by persons on the bridge deck above. During periods between inspections, the gantry should be stored and safely maintained in the maintenance yard. REFERENCES AASHTO, 2001. LRFD Bridge Design Specifications – S.I Units. 2’nd Edition, Interim Revisions, May. AASHTO, 1998. LRFD Bridge Construction Specifications, First Edition. AASHTO for Concrete Durability Design, 2002. Guide to the construction of reinforced concrete in the Arabian Peninsula, CIRIA and The Concrete Society. CEB, 1989. Durable Concrete Structures, CEB Design Guide, Second Edition, Bulletin No. 182, Lausanne. Neville, A. 1995. Properties of Concrete, Fourth Edition, Addison Wesley Longman Limited, England. Raina, V.K. Concrete for Construction – Facts and Practice. Shroff Publishers and Distributors, Mumbai.
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Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
Project Westgate – Lekki Beltway Bridge, Lagos, Nigeria C.M. Bednarski Studio Bednarski (Architects) Ltd, London, UK
A. Adaõ da Fonseca Engenheiros Consultores, Lda., Portugal Faculty of Engineering, University of Porto, Portugal
ABSTRACT: Reflecting on sustainable design and search for a 21st century urban bridge design paradigm, this paper presents a 10.5 km multi-span bridge designed for Lagos, reported in 2014 to have a metropolitan area population of 21 million, making it the largest metropolitan area in Africa and one of the world’s biggest megacities. Traffic congestions in Lagos, the ultimate chaos-city, immortalised by Fela Kuti in his “Go-Slow”, are already world-famous.
1 GENERAL 1.1 The Lagoon This Lagos Lagoon is located at the southern part of the Lagos metropolis, linking theAtlantic Ocean (to the west and south) and the Lekki lagoon (to the east). Its surface area is about 6354.708 sq.km and its perimeter 285 km. The lagoon provides places of abode and, to a much lesser degree than it would be the case in the developed world, recreation. While it provides the means of livelihood and transport, it also serves as dumpsite for residential and industrial waste and discharge. The lagoon is polluted, heavily in some areas. The Lagos Lagoon consists of three main water zones: Lagos Harbour, Metropolitan and Epe Division. Interestingly, water in the Lagoon is very warm, probably a factor of its shallowness. According to Oyenekan, J. (1988) “the bottom water of the lagoon has high temperatures which were relatively constant throughout the year. The temperatures varied between 32.7◦ C in December 2002 at the entrance of Ogun River near Ikorodu and 27◦ C in October 2003. The temperatures fluctuated only narrowly throughout the year. The annual temperature range was only 7◦ C. During the rainy season (May to November), the influx of riverine water and the heavy cloud cover in the sky resulted in a gradual fall of the temperature to a minimum of 26◦ C”. Differential salinity in the Lagoon is the result of the impact of the Atlantic Ocean. It fluctuates both seasonally and semi-diurnally. These fluctuations are greatest in the Lagos Harbour area because of the direct link with the Atlantic Ocean. 1.2 Lekki Peninsula Lagos is located on southwest of Nigeria, on the Gulf of Guinea and comprises the islands of Lagos and Victoria and the outer Lagos Mainland and part of Lekki Peninsula. Lekki Peninsula has been singled out as the main direction for urban expansion of Lagos. The new Lagos airport is located in Lekki and a substantial new city is being developed there. Based on the proposed land use plan the New City will be divided into 5 linear development zones. At present the only practical link between Lekki and the main metropolitan area of Lagos is by way of Lagos Island and the existing 11 km Third Mainland Bridge. All routes out of Lekki go via the Lagos Island with its high levels of traffic congestion. It is thus critical that a new rapid transport link, with minimal levels of congestion, and bypassing the Lagos Island, is provided. 549
Figure 1. Aerial view with Lekki Beltway Bridge outline.
2 LEKKI BELTWAY BRIDGE GENERAL Lekki Beltway Bridge has been conceived as a large volume multi-modal by-pass for Lagos. Normally such gathering ring roads are land based. In case of Lagos, it will be built on water. This is the direct result of the scarcity of land in Lagos, land development and urbanisation patterns to date, and the land values, believed to be the highest in Africa. As the Lagos Lagoon is shallow and cannot support navigation of large vessels, there is no justification for any ‘dramatic’ bridge solutions that would call for large spans and/or high clearances. This also makes this bridge significantly more cost effective than a similar by-pass bridge would have been over navigable waters. The bridge carries eight lanes of road traffic, two cycle lanes, statutory service lanes and monorail, reaching up to the new airport. It is also intended as a symbolic structure – a green avenue snaking along the Lagoon with a row of palm trees along its whole length. The bridge’s paired down structural solution in some ways draws from the modernist architectural icon the Farnsworth House by Mies van der Rohe, with its steel structure clearly and legibly articulated. The main design effort went into making it not only efficient as a structure, and easy to build, but also refined in proportions and detail. With its elevated monorail beams, the bridge feels almost classical in scale and articulation. The north-west landing point of the new bridge, on the existing Third Mainland Bridge, is located north of an existing junction island in order to disperse potential traffic congestion hotspot and to simplify the connection and the time that it would take to construct. The bridge is sinusoidal in plan to make the crossing more enjoyable than a relentless straight line would be.
3 MONORAIL Elevated monorail tracks are located above the bridge deck, along both edges of the bridge, so as to permit long views of the Lagoon from the train and views of the Lagoon from the cars. Were they placed at the deck level the tracks would have to be fenced off with heavy protection both for the sake of general public safety but also to prevent vandalism and terrorist attacks. Such 550
Figure 2.
Outside view of the bridge.
Figure 3.
Farnsworth House.
fencing would negate the key idea of the bridge as a green avenue with panoramic views of the Lagoon, as the views from the bridge would be totally obscured by the monorail fencing. The final monorail system will be selected based on functionality studies and cost analysis. The presently considered options include a levitating maglev monorail, like that linking Shanghai to its airport, or a wheeled system, like for example the monorail that links Haneda airport to Tokyo. 4 STRUCTURE Construction in a marine environment demands a high durability concrete. For that, concrete compactness is a first requisite. In addition, accurate detailing and positioning of reinforcement must be secured. Furthermore, minimization of construction costs of a long bridge requires repetition of procedures. All that “mise en œuvre” is best achieved with precast concrete or with a repetition of in situ long spans. Available geotechnical information leaves no doubt about need for pilling but length of piles is not well determined. Correlation of pilling costs and deck span costs, together with elevation 551
Figure 4. Aerial view of the bridge.
Figure 5. View from monorail carriage.
equipment costs, will determine the best dimension for the bridge spans and the type of structural cross-section. One-first decision is to go for either one single deck almost 50 m wide (Figure 7) or for two twin parallel decks, each almost 25 m wide. One-second decision is to opt for several parallel span-by-span “U” or “box” precast shaped beams, or to opt for two parallel wide box-beam decks built with precast segments or in situ concrete. Up to 40 m spans, “U” and “box” precast 552
Figure 6.
Inner view of the bridge.
Figure 7.
Concept cross-section.
beams are best, but larger spans going up to 100 m, or even more, are being built with movable scaffoldings equipped with OPS technology (Pacheco et al. 2002). Figure 8 shows the construction of 50 m multi-spans with precast segments. A new generation of movable scaffoldings equipped with OPS technology is seen in Figure 9 for 70 m in-situ concrete multi-spans (Pacheco et al. 2011). The newest movable scaffolding system with OPS for spans up to 100 m is shown in Figure 10, and it is due to be used in the construction of a 90 m multi-span viaduct in Turkey. The final resolution of the bridge structure will be arrived by further cost and engineering studies after more information is available about local geotechnical conditions. 553
Figure 8. Laguna Bay Bridge, Santa Catarina, Brazil – 50 m spans (credits to Camargo Corrêa, Aterpa, M. Martins and Construbase).
Figure 9.
River Cabriel Bridge, Valencia, Spain– 70 m spans (BERD courtesy).
5 LIGHTING DESIGN This will be an unusual bridge – with no lampposts. The monorail structure will support streetlights so there will be no need for lampposts on the bridge. In addition to functional lighting required by the highways authorities, architectural lighting will form a critical part of the visual expression of the bridge and change it into an object of art by night. Colour changing lights creating waves of 554
Figure 10. 3-D model of the movable scaffolding system M1 – for up to 100 m spans cast in situ – first unit under construction (BERD courtesy).
Figure 11.
Creative uses for the bridge.
coloured light at night will be enjoyed from the Banana Island and the Lagos Island, and from the northern banks of the Lagoon. This linear light-based art would also greet air passengers coming to land at the new airport in Lekki. To be sustainable, the lighting design will consider environmental impact as a part of the design process. Lighting design, wherever possible, will use energy efficient light sources such as LEDs and long life fluorescents rather than incandescent light sources. The bridge will be illuminated using high quality optically efficient luminaires that direct light onto the intended area of the structure with minimal residual light spill. Reflectors, louvers, cowls and other beam shaping accessories will be used to control and direct light throughout the bridge deck area.
6 ‘SPOT’ ISLANDS As the two bridge branch locations that would link the Lekki Beltway Bridge to the Banana Island and to the Lagos Island would require two level junctions, they could also support ‘water born’ property development. The exclusive Burj al Arab hotel in Dubai, standing on its own little island, could serve as an example, as indeed could the old lighthouses that sometimes stood away from the shores. 555
Table 1. Cost estimate of the Lekki Beltway Bridge.
7 OTHER CREATIVE USES FOR THE BRIDGE Considering the substantial width of the bridge, international or local annual events, like running races or cycle speed races (Figure 11), could be organised using one half of the bridge and putting the remaining four lanes on the other side of the line of palm trees to temporary two-way road use. 8 COST Cost estimate for the construction of the bridge and its multilevel junctions is performed with 2015 unit costs in Table 1. REFERENCES Pacheco P. & Adão da Fonseca A. 2002 Organic Prestressing. Journal of Structural Engineering ASCE: 400–405. Pacheco P. Coelho H. Borges P. & Guerra A. Technical Challenges of Large Movable Scaffolding Systems. Structural Engineering International (IABSE). Vol. 21 Number 4: 450–455.
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Innovative construction methods
Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
High productivity in bridge construction – the OPS effect P. Pacheco Porto University, BERD S.A., Porto, Portugal
H. Coelho, A. Resende, D. Carvalho & I. Soares BERD S.A., Porto, Portugal
ABSTRACT: The growing dissemination of industrialized solutions has made productivity demand become a commonplace of bridge construction process nowadays. There is not a unique formula since different solutions have proved to be efficient. The choice for the most adequate solution is not linear and depends on several factors of quite different nature besides cost efficiency, such as bridge geometry, deck section, span length, span arrangement and bridge total length, environmental and climacteric conditions, site logistics, local construction and design traditions, constructor experience and preferences and overall project schedule. In this paper, a general reflection regarding productivity is substantiated by a bridge construction process practical case. Throughout the text, particular reference is made to Organic Prestressing System (OPS), an actively controlled prestressing system. Besides increasing structural efficiency and safety, OPS confirmed a positive impact in productivity in its recent applications to bridge construction equipment. Keywords: Bridge Construction, Productivity, Movable Scaffolding Systems (MSS), Organic Prestressing System (OPS)
1 INTRODUCTION The demand for productivity in bridge construction may potentially derive from some very distinct causes: – Impact of construction time in operation and construction yard costs; – Interest of the concessionaire in opening the bridge to traffic as soon as possible in order to anticipate the capital return; – Limited time for construction due to inadequate planning and control of bridge design and extended tender processes – leaving limited time for construction process preparation. This often results in narrow chance to make key adaptations in bridge design in due time. – The impact of bridge construction on existent road network (traffic interference during construction and increased traffic flow after bridge opening); – Political promises and marketing – pressure to integrate inauguration ceremonies with electoral events. Regardless of the context and case by case restrictions, the approach for achieving high productivity in bridge construction should be based on the following general guidelines: – Preparation of construction process from an early stage of bridge design. Simultaneous development of bridge design and construction process enables a more efficient design of both bridge and construction equipment, often leading to more productive and less expensive solutions; 559
– Adequate and timely planning of operation tasks, including preparatory works, assembly of construction equipment, current operation cycle and disassembly. In some cases, operation planning may justify changes or adaptations in construction equipment design; – Bridge construction process analysis shall cover all operations to be done on site. The productivity on bridge construction strongly depends on adequate preparation and simplification of each task to be done on site. Repetitive tasks such as steel reinforcement placement and formwork adjustments (using the example of cast in situ construction) may be largely simplified by an adequate design and by adding adequate auxiliary means to construction equipment; – Reduction of in situ tasks to a minimum – for example, by prefabricating complete reinforcement cages; – Use of specialized labor force in all stages: bridge design, bridge construction process design and also bridge construction – sometimes the labor force during bridge construction is disregarded and consequently the productivity is much smaller and safety is impaired. Operation crew must receive an adequate training prior to beginning of site activities; – The bridge construction process shall always include a specific risk analysis for all tasks to be done on site. All tasks shall be designed to have an acceptable risk level. Most commonly, a safe and intuitive task tends to be a more productive one; – Coordination and efficient communication between bridge designer, construction process designer and operation crew; – Monitoring and continuous improvement: during the construction of a bridge, the first cycles are often slower (especially in cases in which operation crew is not familiarized with construction technology or construction process). Lessons learned during first spans may lead to implementation of very effective improvement measures. 2 ORGANIC PRESTRESSING SYSTEM (OPS) The present chapter, which appears to be disconnected from the previous, gives a brief and overall overview of Organic Prestressing System (OPS) to all readers not yet familiarized with this technology. This introductory chapter is important for a comprehensive reading of practical case described next. OPS is an automatically adaptive prestressing system which has the ability to increase or decrease prestressing forces according to load variation. It was first developed using as an inspiration the behaviour of nature formed structures (biomimetics), more specifically the muscle behaviour. One can describe it as a prestressing system in which the tension applied to cables is automatically adjusted to the actuating loads, through a control system, in order to reduce the structural deformations and minimize tensions (Pacheco, P. 1999). Using bridge construction equipment as an explanatory reference, the OPS main elements are 1) the actuator and the active anchorage, 2) the unbonded cables, 3) the sensors, 4) the electronic controller in the girder control unit, 5) passive anchorage and 6) deviation shores (please see Figure 1). The OPS control is ruled by an algorithm that adopts the girder mid span deflection as main control variable. Mid span deflection is monitored by pressure transducers (sensors) continuously transmitting signals to the control unit (PLC). The control algorithm computes actuation decisions (hydraulic cylinders stroke variations) which consequently affects tension on prestressing cables. Actuation decision is based on mid span deflection changing tendency, filtering instantaneous deflection noise due to vibration. To ensure an adequate reliability level, OPS is provided with distinctive and redundant sensors with measures that are permanently compared to guarantee that the algorithm decision is always based on accurate information. If any inconsistency or incoherence is detected, there are several alarm combinations (buzzer and color light) warning the operator to check the data – always available on real time on intuitive touch screen interface (on girder control unit). Operator must then confirm if the data is correct and 560
Figure 1.
OPS main elements and layout.
Figure 2.
M60-I on concrete pouring phase.
take immediate action so that regular operation may be re-established and potential risk situations prevented. Granting information to the operator about the bridge construction equipment behavior, the OPS increases the safety factor of the structure. Application of OPS to bridge construction equipment allows the design of lighter and more efficient structures. An important indirect impact on productivity and operational costs is due through considerable weight reduction and load capacity increase, especially since these equipments have to endure frequent launching, assembly, disassembly and transport operations. 3 CASE STUDY – BERD M60-I CORGO RIVER 3.1 Bridge overview The bridge over Corgo river in Vila Real (north of Portugal) is a concrete bridge with a total length of 2796 m, including approach viaducts and the 552 m long cable stayed main bridge. By the time of conclusion, this was the 2nd highest cable stayed bridge in Europe. The East viaduct, with a total length of 1278 m comprising 22 spans with a maximum length of 60 m was built with M60-I (please see Figure 3). 561
Figure 3.
Bridge over Corgo river elevation.
Figure 4.
Deck cross section.
The deck cross section is a single box with a total width of the crossing platform changing from 25 m in approach viaducts to 28 m in cable stayed bridge. Even though the box is quite wide (10 m between external web external surfaces), the continuous cantilever is supported by precast concrete struts positioned every 3 m (please see Figure 4).
3.2 Projects challenges The development of East viaduct construction solution presented some interesting challenges: – M60-I design started being the bridge design almost finished. Construction works had already started and piers were actually partially constructed. To fulfill Contractor’s schedule the M60-I had to be prepared to build one 60 m span every 10 calendar days (9 working days) and equipment assembly was scheduled to start in less than 10 months. Furthermore, the interfaces between bridge piers and M60-I had to be defined 4 weeks later to avoid interruption of bridge piers construction; – East viaduct comprises four distinctive types of piers with significant variation of width (4, 4.5, 5 and 6.5 m). Pier width variation means increased complexity for design of movable scaffolding system (MSS) supports (especially in underslung equipment) and for definition of interfaces between bridge piers and MSS (already with tight schedule in this specific case); – The bridge is located over a deep valley and great part of construction was to be carried out in winter period. The Contractor required M60-I to be prepared for work under wind velocities greater than usual (launching operation with wind velocity up to 70 km/h). To accomplish Contractor’s schedule it was necessary to work even in adverse climatic situations (usual at winter time in that region); – Due to the significant height from the deck to the ground (pier heights up to 113 m in East viaduct) and also because of the irregular topography it was not possible to have auxiliary equipment – mobile cranes – to assemble the machine supports. This means that the M60-I needed to be self-supporting and completely autonomous; – Equipment assembly and disassembly in difficult conditions and with short preparation time. Topography was very difficult for assembly operation (there was only 18 m of free space behind the abutment) and also for disassembly operation (on last span the MSS was around 75 m from the ground); – East viaduct was actually constructed in 3 different stages (first two by MSS – see Figure 3 – and a final stage comprising part of cantilever and the precast struts was built by auxiliary equipment). First stage included construction of bottom slab and webs. During construction of top slab in second stage, the “U” shaped deck section built in first stage was already rigid but still not prestressed. Interaction between M60-I main girders and “U” shaped deck section assumed an increased importance – main girders needed to be rigid enough so that the stresses in non prestressed 1st stage concrete were kept within allowable levels; – The deck section built by M60-I was significantly heavy, with the deck weight varying from 338 kN/m in mid-span to 531 kN/m near the pier section. 562
Figure 5.
Figure 6.
M60-I elevation and components identification.
M60-I front support on fixed phase.
Figure 7.
M60-I rear support on fixed phase.
3.3 Equipment description The solution developed by BERD to build the East viaduct was an underslung MSS – M60-I. The main components of the M60-I are: (1) Main Girder and Noses; (2) Rear Support Ring; (3) Support Brackets; (4) Transversal Structures; (5) Bogies; (6) OPS Elements and (7) Front Support Ring (please see Figure 5). Basically the M60-I operates in 2 distinctive phases with different structural configurations: 1) Fixed Phase and 2) Launching Phase. The Fixed Phase includes all operations of span (n) deck construction and the Launching Phase comprises all operations in-between the M60-I fixed on span (n) and fixed on span (n + 1). During Fixed Phase the MSS Front Support is performed directly on the pier and is done by a hydraulic cylinder placed inside a recess in the pier (please see Figure 6). The MSS Rear Support is the rear support ring which suspends both girders and transfers the loads to the previously built deck cantilever by metallic shims (please see Figure 7). During Launching Phase the MSS Front Support changes from the hydraulic cylinder positioned in the pier recess to bogies’ wheels positioned over support brackets fixed to the pier by prestress bars (please see Figure 8). The MSS Rear Support changes from the metallic shims to wheels on metallic rails previously assembled over the deck (please see Figure 9). 3.4 Working cycle The Contractor demand for 10 days working cycle markedly affected the conception of M60-I, from design to operation. Working cycle operations were designed to guarantee the predefined schedule while regarding all safety measures – period in which M60-I is unbraced and most affected by wind exposure (Phase 2) was reduced to minimum. 563
Figure 8.
M60-I front support on launching phase.
Figure 9.
M60-I rear support on launching phase.
Table 1. Representation of M60-I working cycle.
As aforementioned, the construction of deck box section by M60-I was performed in two stages. The first stage comprises the construction of the bottom slab and the vertical webs (“U” shaped section) being that the top slab is poured in the second stage. This decision simplified the inner formwork and the reinforcement steel preassembly, increasing concrete pouring productivity and deck prestressing productivity as well – deck prestressing operations could start 15 to 20 hours after finishing concrete pouring (starting on “U” section cables). Simplification of inner formwork was particularly advantageous in transition spans, in which the rearrangement of blisters relative position (different prestressing layout caused by span length variation) and deck width variation (transition from viaduct to cable stayed bridge section) required significant changes in internal formwork. Like in many other span by span cast in situ deck construction processes, the most consuming task in terms of time and manpower is reinforcement steel assembly and placement. The M60-I was equipped with two independent movable cranes specially designed for the rebar operations (cages were preassembled on construction yard). Supporting brackets relocation operation includes brackets disassembly, transversal sliding to achieve clearance from piers, transport to next pier and reassembly procedure. Relocation of 30 tonne weight supporting brackets was surely one the riskiest operations of M60-I operation cycle, strongly affected by climatic conditions – mainly the wind velocity. Being the M60-I directly supported on the pier and deck cantilever (front and rear support, respectively) throughout all Fixed Phase, two main advantages arise: (1)Avoiding concrete loads on brackets allows a significant 564
Figure 10. Auxiliary cranes placing rebar.
reduction of their weight and consequently the weight and complexity of all equipment and auxiliary structures used for its relocation – this is particularly significant in face of challenging East Viaduct −5% longitudinal slope; (2) Given that the Brackets are unloaded during almost the entire cycle it is always possible to choose a period with good climatic conditions to perform the relocation – this operation was never in the cycle critical path. When the concrete of both stages reaches the resistance predefined by bridge designer, the deck is prestressed allowing the MSS to move forward to the next span. Prior to the M60-I longitudinal movement it was necessary to lower the girders, by that unloading the formwork and then move the girders transversely. In this particular project the deck lower corner demanded previous stripping of lateral formwork panels before lowering the girders. After transversal movement of the formwork lateral panels the girders are lowered and moved outside towards launching position, guaranteeing compatibility with bridge pier position and geometry (please see Figure 8). In this particular application, the M60-I longitudinal movement relied on gravity, since the deck longitudinal slope was −5,0% for almost the entire viaduct. The movement is controlled by 4 hydraulic winches on the rear support ring. During the launching operation the MSS girders are connected by the rear support ring which confers greater stability regarding lateral overturn (to achieve Contractor’s requirement for launching operation with wind velocities up to 70 km/h). The rear support ring is also equipped with 4 hydraulic jacks positioned between the rollers and the metallic structure, allowing rear reaction control, and consequently the mitigation of imposed displacements effects throughout launching operations – this was particularly advantageous for last spans, in which longitudinal slope varied significantly. Longitudinal launching duration is around 1 hour, which is exceptionally fast for an underslung MSS. The M60-I was able to build several spans in less than 10 days, being 9 calendar days and 7 working days the fastest working cycle, clearly exceeding expectations around cycle productivity. The East viaduct construction was finished on Contractor schedule and, most important, without accidents. 3.5 OPS impact Besides advantages previously referred, related with structural performance, deformation control and structural monitoring, three specific advantages may be pointed out, regarding OPS application to this particular project: – When the deck top slab was concreted the “U” section previously constructed was already hardened. By controlling the MSS deformation to “zero”, it was possible to limit the stresses in the “U” section during the second concreting stage – there was no need for reinforcement due to construction process; 565
– The OPS deformation control incurs in great importance to avoid time loss due to pre-camber works. This was particularly relevant in the final spans of the Corgo river Bridge project due to their variable length and longitudinal slope variation; – During the bridge construction, while the MSS disassembly project was being developed, a collision between the MSS front nose and the form traveler used in the construction of the cable stayed bridge deck was detected. There was a need to disassemble the front nose during last launching and before concreting last span. The front nose removal meant a considerable increase of main girders’ self-weight imposed positive bending moment (between MSS supporting sections). By changing the OPS parameters it was possible to introduce an initial negative bending moment that counteracted the effect of nose removal, therefore guaranteeing the structural safety of the MSS during last span concreting. 4 CONCLUSION As introduced before, the demand for productivity is a common requirement of bridge construction process nowadays. The specific challenges of each project may justify the need to adopt special solutions, different concepts or even strengthen technical coordination between different entities engaged in the same project. Simultaneous and coordinated development of bridge design, construction process design and erection equipment design is strongly recommendable but not always possible which, per se, represents a major challenge. Besides description of general design approach to achieve productivity goals, particular attention was given to the effect of OPS application in construction equipment. The design of lighter structures and continuous structural monitoring are advantages almost taken for granted. Furthermore, the use of OPS allied to careful, timely and detailed preparation of all tasks to be done on bridge site has shown great advantages on productivity – contributing to the achievement of working cycles that appeared to be unrealistic, plus reinforcing structural and operational safety – many times the accidents occur on tasks improvised on bridge site. Ultimately, safety is a synonym of productivity per se, since accidents are one of the most common causes for significant delays in bridge construction activity. ACKNOWLEDGMENTS The authors wish to thank all the BERD team who worked in these Projects, to LCW Consult (bridge designers), ConstruGomes (equipment operator) and FCC, Soares da Costa, RRC (bridge builders). REFERENCES CNC, Confederación de la Construcion, 2007. Manual of Self launching scaffoldings, in spanish, 1a ed., CNC, Madrid, 2007. EUROCODE 1: Actions on structures – Part 3: Actions induced by cranes and machinery, 2005. Pacheco, P., Adão da Fonseca, 2002, A. Organic Prestressing, Journal of Structural Engineering, ASCE, pp. 400–405. Pacheco, P., Guerra, A., Borges, P., Coelho, H., 2007. A scaffolding system strengthened with organic prestressing – the first of a new generation of structures, in Structural Engineering International, Journal of the International Association for Bridge and Structural Engineering, Vol. 17, Number 4, November 2007, pp. 314–321(8). Pacheco, P., H. Coelho, P. Borges, A. Guerra, 2011. Technical Challenges of Large Movable Scaffolding Systems, Structural Engineering International (IABSE), Vol. 21, Number. 4. pp. 450–455. Rosignoli, Marco, 2013. Bridge Construction Equipment, Thomas Telford, (pp. 6–7).
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Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
FlexiArch-Stress Ribbon combination for multi-span pedestrian bridges A.E. Long, D. McPolin & S. Nanukuttan School of Planning, Architecture and Civil Engineering, Queen’s University Belfast, UK
A. Gupta & D. Robb Macrete Ireland Ltd, Toomebridge, UK
ABSTRACT: The potential benefits of combining the elegance of the stress ribbon concept with the robustness and speed of construction of the FlexiArch is discussed. In combination, multi-span pedestrian/cycle bridges which are innovative, highly durable and have optimal full life cycle costs can be produced with lengths of over 100 m. As the stress ribbon system is well known, the main emphasis of this paper will be on the FlexiArch. Since 1900 few arch bridges have been built, but with the development of the innovative FlexiArch this trend can be reversed as they can be installed rapidly, are cost competitive, have all the attributes of masonry arches and are very sustainable. Thus the FlexiArch represents a very attractive alternative to heavily reinforced cast in situ arches currently used in combination with stress ribbon deck systems.
1 INTRODUCTION Strength, stiffness, durability and minimal maintenance are some of the attributes of arch bridges acclaimed by structural engineers throughout the world. In addition their aesthetic qualities are recognised universally; so much so that there are hundreds of thousands of arch bridges in the world (some over 2000 years old). Thus in the early 1990’s when the first author was asked ‘Why is it that so few arch bridges have been built since the 1900 as they are aesthetic, strong, durable, and require little maintenance?’, he decided to respond to this perceptive question which acted as a catalyst for research targeted at this challenge. The need for centring and accurate voussoirs for arches meant that they could not compete in terms of speed of construction with prestressed concrete/steel beam and slab systems which rose to prominence in the 1950’s and 1960’s and are still widely used. However, many of these beams and slab bridges, even though their specified design lives were 120 years, have deteriorated after only 30–40 years and a significant number have already had to be replaced. Where aesthetics was of paramount importance rigid precast concrete arches, heavily reinforced so that they could be safely lifted into position, were often adopted. Like beam and slab bridges these are also vulnerable to reinforcement corrosion and they do not have the high levels of durability associated with unreinforced masonry arches. In this context the UK Highways Agency (2004) recommends the use of the arch form and that ‘consideration should be given to all means of reducing or eliminating the use of corrodible reinforcement. Basically the challenge was to develop an arch system with all the attributes of an unreinforced masonry arch but one that: – – – – –
Can be installed as quickly as alternative types of bridges. Eliminates the need for centring – expensive to construct/install and often difficult to remove. Uses existing well accepted methods of analysis/design for conventional masonry arches. Is cost competitive and suitable for construction off-site. Uses precast concrete for the voussoirs instead of stone voussoirs. 567
Figure 1.
Stress ribbon and arch combinations (Strasky 2005).
In this paper the concept of the patented ‘FlexiArch’ system (Long 2004), developed to meet this challenge, will be described and case studies will be presented for three specific applications of this versatile system – chosen as a representative sample from over 50 Flexi Arch bridges already in service in the UK and Ireland. Whilst stiffened deck suspension bridges are widely used for long span road and rail bridges, the extremely elegant Stress Ribbon system is primarily utilised for pedestrian bridges. Evidence of primitive forms, based on ropes from liana or bamboo, have been found in tropical jungles but Professor Jiri Strasky (2005) has attributed the initial one built in the 1950’s to Ulrich Finsterwalder (1891–1988). Nevertheless the major development of the Stress Ribbon system over the last 40 years is due to Strasky’s design input, research and persistent promotion to structural engineers. In his highly acclaimed book (Strasky 2005) many examples are shown of long span pedestrian bridges over the rivers of his native Czechoslovakia. The combination of strong minimally sagging cables with slender precast concrete deck units which could easily be pulled into position has resulted in extremely elegant pedestrian bridges with little environmental impact. However, like flat arches, the high horizontal forces generated at each end, whether single or multi-span, represented a real challenge to foundation designers. Strasky’s response was to combine the attributes of a stress ribbon with an arch so that the resultant horizontal forces on the foundations were eliminated or minimised. Such an ideal system is shown in Figure 1 where the foundations only have to resist vertical forces – eliminating raking piles and making them suitable for estuarine locations. Perhaps because of the relatively long spans involved Strasky has adopted cast in-situ reinforced concrete arches in combination with stress ribbon decks. Nevertheless as both stress ribbon decks and FlexiArches benefit from the use of precast concrete units the speed of construction of the combined system could be greatly enhanced and steel corrosion risks in the arch eliminated. Thus possible combinations are considered later in this paper. 2 THE FLEXIARCH SYSTEM 2.1 FlexiArch concept and method of manufacture The ‘FlexiArch’ is constructed and transported to site in flat pack form using polymeric reinforcement to carry the self-weight of the arch unit during lifting but once in place it behaves as a conventional masonry arch. The preferred method of construction of the arch unit is shown in Figure 2 (see Long 2013 for more detail). For the manufacture of each arch unit the tapered voussoirs 568
Figure 2.
Method of construction: FlexiArch.
are precast individually then they are laid contiguously with the top edge touching, in a horizontal line with a layer of polymeric reinforcement placed on top. In-situ screed, approximately 40 mm thick, is placed on top and allowed to harden so that the voussoirs are interconnected. When lifted gravity forces cause the wedge shaped gaps to close, concrete hinges form in the screed and the integrity of the unit is provided by tension in the polymeric reinforcement and the shear resistance of the screed. The arch shaped units are then lifted and placed on precast footings at the bridge site and all the self-weight is then transferred from tension in the polymeric reinforcement to compression in the voussoirs, i.e. under load it acts in the same way as a conventional masonry arch. Experience of using this method of manufacturing (Figure 1) has shown that it has a number of advantages over traditional methods. The voussoirs can be accurately, quickly, and consistently produced with the desired taper in relatively simple shuttering. High quality concrete can be utilised for the individual precast voussoirs to enhance the durability of the arch and greatly reduce the variability in the physical properties normally associated with natural stonework. 2.2 Lifting & installation The primary function of the polymeric reinforcement is to provide sufficient tensile strength so that the FlexiArch units can be lifted safely; both from the flat casting bay on to a flatbed lorry and from the lorry in its designated arch form into position on the precast sill beams at the bridge site. With safe working being of primary importance, carefully designed tests, which accurately simulated the boundary conditions, were carried out to ascertain the strength of the polymeric reinforcement (Long 2013). Provided the recommended procedures are adopted during installation individual units can be accurately located on site every 15 minutes. 2.3 Validation testing and design A wide range of static loading tests have been carried out to validate the performance of the system. These have included model tests in the laboratories (at fifth, quarter and third scale) with granular or concrete backfill where it was possible to achieve the ultimate capacity (Long 2013). However as model tests are not considered to be reliable at predicting behaviour at serviceability loads a number of full scale tests were carried out at Macrete. These included tests on 5 m span × 2 m rise and 10 m span × 2 m rise FlexiArches with lean mix concrete backfill and a test on a 15 m span × 3 m rise with lightweight concrete backfill. At full scale the strengths of the arches were significantly higher than the maximum capacity of the loading rigs. However, the maximum loads applied (equivalent wheel load of 320 kN – or lane loading of over 1000 kN) were still in excess of the maximum for the factored loads imposed on road bridges. Thus the tests confirmed that like conventional masonry arches, which have enormous reserves of strength, the FlexiArch system (with uniformly high strength voussoirs) more than satisfied the stringent requirements for highway bridges. 569
Figure 3.
Sheinton Bridge, Shropshire.
Once constructed the FlexiArch behaves as a conventional arch and as a consequence standard design/analysis tools for arches have been used in the design process e.g. Archie software analysis system (Obvis 2007), Ring software (Limitstate 2009). Both approaches give comparable estimates of strength for the system with granular backfill but were found to be conservative relative to the model tests in the laboratory (around three times stronger). As anticipated the strengths of FlexiArches with concrete backfill were very much higher (around ten times stronger than predicted). In order to gain a better understanding of the behaviour of the system with concrete backfill FlexiArches with a range of geometries were analysed. Using two dimensional finite elements in conjunction with relevant material properties it was found that the deflections predicted were in good agreement with those measured in the full scale tests (Long 2014). 3 FLEXIARCH EXEMPLARS Over 50 bridges have been constructed in the UK and Ireland over the past eight years. In this section details of three different applications of the system are given. Relevant photographs are available of all three projects (Macrete 2015). 3.1 Flexiarch replacement bridge, Sheinton, Shropshire In 2009 the old bridge in the small village of Sheinton, Shropshire was irreparably damaged by flooding. Thus engineers from Shropshire Council decided to replace the three span bridge with a much longer single span arch to greatly reduce the risk of flood damage in the future. As they wished to reduce construction time to a minimum and avoid the use of a bridge with corrodible reinforcement they selected a 13.7 m span × 2.7 m rise ‘FlexiArch’ and ordered eight one metre wide units from Macrete Ireland, Ltd. Each 13t unit, shown in Figure 3(a), was placed on the precast sill beams in a matter of 10–15 minutes. Once all the ‘FlexiArch’ units had been located and the precast spandrel wall installed the bridge was ready for the lean mix concrete backfill (Figure 3(b)). The spandrel walls were then finished in stonework (Figure 3 (c)) to produce an aesthetically pleasing solution. 3.2 Bridge widening, Bouthray, Cumbria In the devastating flooding experienced in Cumbria in 2009 severe damage was caused to a two span arch bridge at Bouthray. Cumbria County Council engineers decided to replace the upstream face (over 2 m wide) with an arch system without corrodible reinforcement. The main span 6.6 m × 1.77 m rise and side span 3.72 m × 1.2 m rise ‘FlexiArch’ units selected were supplied by Macrete and transported to site by lorry before being lifted (Figure 4(a)). Figure 4(b) sows an overall view of the site after the precast concrete spandrel walls were located. Local stone was utilised to face the spandrel walls and the finished bridge (Figure 4(c)) is not only aesthetically pleasing but should have a design life of over 120 years. 570
Figure 4.
Bridge widening Bouthray, Cumbria.
Figure 5. Tameside bridge strengthening.
3.3 Bridge strengthening, Tameside, Manchester Tameside’s 78 year old Jubilee Bridge, which spans National Cycle Route 66 in Manchester had been weakened by extensive reinforcement corrosion and spalling. Replacement was unacceptable due to the disruption to services and a key transportation corridor. Sprayed concrete, used for repair in 1974, was clearly not a long term solution. Aware of other arch bridges over the linear cycleway, the consultants suggested using the Macrete FlexiArch. Thus in December 2012 fourteen FlexiArch units (1 m wide) were installed (Figure 5(a)), the first ever application for bridge strengthening. The 7.4 m span units were placed on lightly greased laterally extended sill beams along each abutment. Then they were pushed horizontally in pairs beneath the bridge using two hydraulic jacks (Figure 5(b)). When all 14 units had been located, spandrel walls were constructed and then the gap between the FlexiArch unit and the original deck soffit was filled with foamed concrete. Tameside Council now have an aesthetically pleasing bridge with a design life of over 120 years (Figure 5(c)). 4 FUTURE DEVELOPMENTS The three exemplars give an indication of the versatility of the ‘FlexiArch’, however the authors firmly believe that the system has yet to achieve its full potential. For example: – The maximum span could be increased to 25–30 m for highway loading and even more for pedestrian bridges. – In the construction of new or the replacement of existing multi-span bridges where the lateral forces can be minimised so that very slender intermediate piers can be utilised, Figure 6. 571
Figure 6.
Plan view of Multispan FlexiArch bridge with slender piers.
Figure 7.
Bridge Olomouc, Czech Republic (Strasky 2005).
4.1 Stress ribbon/arch systems A significant number of pedestrian bridges made up of Stress ribbon decks supported by an arch have been built in the Czech Republic and other parts of the world. One fine example, illustrated in the book by Strasky (2005), is the bridge over the freeway near the city of Olomouc which was built in 2007. The two span stress ribbon deck (Figure 7) is supported by an arch. The 79 m long deck is an assembly of precast concrete segments, subsequently prestressed, which is supported by a slender 64m span cast in-situ reinforced concrete arch. The geometry of the bridge, the load and the level of post-tensioning are designed such that the horizontal forces in the stress ribbon and the arch are equal so that overall it functions as a self anchoring system. Thus the footings have, in essence, to be designed only for vertical loads. The completed bridge at Olomouc is exceptionally slender, elegant and aesthetically pleasing, but good ideas often attract the attention of designers/engineers in other countries – the Phyllis J Tilley Bridge in Fort Worth, Texas USA is very similar to Strasky’s at Olomouc – in this case a steel arch is utilised. 4.2 Possible stress ribbon/FlexiArch system The use of a FlexiArch as opposed to a rigid reinforced concrete or steel arch in combination with a slender stress ribbon deck has obvious advantages as both are based on the use of precast concrete 572
Figure 8.
Stress ribbon/FlexiArch system with length up to 105 m.
elements. Taking advantage of the fact that the FlexiArch is ideal for multi-span bridges and that maximum spans for such bridges could be 30 m–40 m the following concept could have potential (Figure 8) ie a total length of some 80 m–105 m for a two span arch system. Again this could be designed to be self anchoring so as to widen the potential use of the FlexiArches where ground conditions would otherwise be unsuitable. 5 CONCLUDING REMARKS A self anchoring combination based on a stress ribbon deck and FlexiArches has great potential for application to pedestrian bridges for the following reasons: – As both systems utilise precast concrete elements, the speed of construction could be comparable to other more conventional RC/PC beam systems. – Since the FlexiArch is highly durable (no corrodible reinforcement), the combined system could have a design life far in excess of alternatives. – As the FlexiArch can be placed in a matter of hours existing transportation routes being crossed would have minimal closure times (eg pedestrian bridges over freeways and railways). With the support system in place the stress ribbon cables and deck could then be placed, quickly thus minimising congestion. ACKNOWLEDGEMENTS Finance provided by the ICE R&D Enabling Fund, KTP Scheme, Invest Northern Ireland, DRD Roads Service (NI) and the Leverhulme Trust is gratefully acknowledged as is the input by K. McDonald, B. Rankin and I. Hogg. REFERENCES LimitState 2009. LimitState: RING 2.0 software. See http://www.limitstate.com/ring. Long, A.E. 2004. Queen’s University Belfast, Concrete arch and method of manufacture. International Patent, Publication 27 May, No. WO 2004/044332A1. Long A., Kirkpatrick J., Gupta A., Nanukuttan S. and McPolin D. 2013. Rapid Construction of arch bridges using the innovative FlexiArch. Proc. ICE, Bridge Engineering, Vol. 166, Issue BE3, Sep, pp. 143–153.
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Long A., McPolin D., Kirkpatrick J., Gupta A. and Courtenay D. 2014. FlexiArch: from concept to practical applications, The Structural Engineer, July, pp. 10–14. Macrete 2015. Macrete FlexiArch Projects-Ashton Bridge, Bouthray Bridge, Sheinton Bridge, macrete.com (accessed 15/2/2015). Obvis Ltd. 2007. Archie-Msoftware. See http://www.obvis.com (accessed 04/04/2013). Strasky J. 2005. Stress ribbon and cable supported pedestrian bridges. Thomas Telford Publishing, London, 232 pages, ISBN: 0 7277 3282 X. UK Highways Agency 2004. BD 91/04, Un-reinforced masonry arch bridges. Design Manual for Roads and Bridges, Vol. 1, Section 3, Department of Transport, Highway and Traffic.
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Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
Balanced lift method for the construction of bridges with two spans S. Foremniak, W. Weiss & J. Kollegger Vienna University of Technology, Vienna, Austria
ABSTRACT: Conventional construction methods for building bridges with one tall central pier over deep valleys are the balanced cantilever method and the incremental launching method. In a research project at the Vienna University of Technology a different approach for the construction of bridges with one tall pier, called the balanced lift method (Kollegger 2006), was investigated. The case study for the construction of an alternative design according to the balanced lift method for the San Leonardo Viaduct, a bridge originally built using the balanced cantilever method, will be discussed in this paper. Furthermore several large-scale tests and the first application of the balanced lift method are outlined in this paper. 1 BALANCED LIFT METHOD 1.1 The underlying idea behind the balanced lift method The balanced lift method is a bridge construction method for bridges with tall or short piers, which was developed at the Vienna University of Technology. The underlying idea is to assemble the key elements of the bridge in a vertical position and then rotate the bridge girders from the vertical into the horizontal position with the aid of compression struts, as shown in Figure 1. The inventors’ objective was to create a construction method that could accelerate and simplify the construction of bridges with very specific boundary conditions. Two different designs for bridges with completely different boundary conditions are described in this paper. Even though both constructions of the bridges start with fragile elements consisting of ulta-thin precast elements, monolithic bridges with solid cross-sections are obtained in the end. A span range from 50 m to 250 m is suggested by Gmainer (2011) for the application of the balanced lift method. 1.2 The importance of ultra-thin precast elements For bridges built with the balanced lift method, the weight of all the elements that are put in motion during the lifting or lowering operations is of utmost importance. The amount of force that is needed
Figure 1. The balanced lift method for bridges with tall and short piers.
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Figure 2.
San Leonardo Viaduct. Near termini-imerese in Italy.
for the rotation process during the bridge construction depends, if one sets friction and eccentricities aside, on the weight of the bridge girders and the compression struts. To minimize the weight of these elements the Institute for Structural Engineering at the Vienna University of Technology started developing cross-sections out of ultra-thin precast elements that would be filled with in-situ concrete after their installation on the construction site. The application of these elements would allow a combination of the positive attributes of precast elements and in-situ concrete. The precast sections are made of hollow wall elements or ultra-thin precast elements or of a combination of both, allowing the production to be very economical and fast. 2 BRIDGES WITH ONE TALL CENTRAL PIER 2.1 The original bridge and the new design The original design of the bridge is a cantilever beam joined at the pier, with supports at the ends (see Figure 2) The total length of the San Leonardo viaduct is 105 m. The width of the viaduct is 23 m and the viaduct thickness is from 3 m to 10.60 m. The cross-section consists of a steal box beam, which is resistant to torsion. The beam is formed of a lower slab, two vertical walls and an upper slab connecting these walls and acting as deck for both carriageways. The span ranging from 40 to 125 m is suggested for the application of the balanced lift method (Gmainer 2011). The San Leonardo Viaduct would have therefore been a perfect candidate for a construction with the balanced lift method. The compression struts would have had positive effects on the height of the bridge girders, creating a slimmer and more elegant girder, reduced span lengths and created a more interesting look of the bridge. The design according to the balanced lift method proposes the construction of two balanced lift bridges and then connecting them with the deck slab. The cross-sections of one half of the bridge are shown in Figure 3. The spans are divided into lengths of two times 54.40 m and two times 50 m. The height of the bridge girder is held to 3.38 m. A comparison of the thicknesses of the viaduct of both designs is shown in Figure 3. Half of the deck slab (one carriageway) would have a width of 11.25 m. The compression struts and the bridge girders are made of hollow wall elements and ulta-thin precast elements during the first construction phases until they are filled with cast in-situ concrete. Although the construction of the bridge starts, due to the application of 576
Figure 3.
Sections of the design according to the balanced lift method with the original design in grey.
thin-walled precast elements, with quite fragile elements, a monolithic concrete bridge with solid cross-sections is obtained in the end. 2.2 Construction phases The construction phases of the San Leonardo Viaduct built with the balanced lift method are shown in Figure 4. The two 86.70 m high piers are to be built using climbing formwork. Figure 3 shows the two different cross sections of the piers. Auxiliary piers are needed in order to ensure that the end parts of the bridge girders do not touch the sides of the valley. A minimal weight of the compression struts and the bridge girder is very important for the lifting process, therefore these construction elements are made out of precast elements, which are filled, once the balanced lifting process has been carried out, with in-situ concrete (see Section C-C and Section D-D in Figure 3). The compression struts consist of three segments. Once both compression struts have been assembled in a vertical position, the bridge girders can be put together. Each bridge girder consists of five segments. After the bridge girder elements out of hollow wall elements and ultrathin precast elements have been assembled in a vertical position, the two bridge girders can be 577
Figure 4.
Construction phases for the San Leonardo Viaduct built according to the balanced lift method.
578
connected together by a tendon. This tendon has to carry the tensile force that develops during rotation of the bridge girders. The next construction phase consists of lifting the compression struts and bridge girders by approximately 20 m with the aid of strand lifting units. In the next step, the lifting of the lower points of the compression struts leads to a rotation of the bridge girders. Once the base points of the compression struts have reached the desired height, the middle ends of the bridge girders can be lowered to reach their final position. After adjusting the position of the elements in-situ concrete is placed in the joints. Post-tensioning tendons are stressed and the compression struts and bridge girders can be filled with concrete. Finally, the deck slab is added to complete the structure. 2.3 Large-scale tests Large-scale tests were used to provide valuable insight into the building process and stability of lightweight box-sectioned bridge girders out of concrete as would be used for the San Leonardo Viaduct if built with the balanced lift method. Two box-sectioned bridge girder, with a height of 2.89 m and a width of 6.00 m, were built (Figure 5) and tested. The webs of the lightweight box-sectioned bridge girders where made out of hollow wall elements, which were manufactured fully automatically on a rotary production. The deck and bottomw slabs of the lightweight girders were made out of precast elements with a width of 70 mm. In order to assess the performance of the bridge girders during transport and lift operations, loading tests were performed on both test specimen. Fot the tests the box-sectioned girders were placed on 3 bearings, the load was applied with four hydraulic jacks and the deflection of both the bottom and deck slab were measured as described in Foremniak (2014). 3 BRIDGES WITH ONE CENTRAL SHORT PIER 3.1 The design Even though the method was primarily invented for bridges over deep valleys, i.e. bridges with tall piers, it also – with minor changes- can be applied for bridges with shorter piers. The design of bridges with short piers according to the balanced lift method will be described using the example of the bridges over the Lafnitz and Lahnbach rivers of the S7 motorway in Austria, whose start of construction has been set for the summer of 2015. In contrast to the bridge designed instead of the San Leonardo Viaduct using the balanced lift method, the spans and cross-sections of the Lafnitz and Lahnbach bridges are quite small, with lengths of 120 m and 100 m respectively and widths of 14.5 m. The area where the bridges are to be built is part of the ecologically sensitive “Natura 2000” nature reserve, therefore an encroachment on the natural habitat through the erection on falsework is not accepted by the Austrian motorway operator ASFINAG. The construction site should be as small as possible and limited to the central pier and abutments. To meet all these criteria, the balanced cantilever method, incremental launching or the balanced lift method are the only construction options. It could be shown that the erection costs based on a design applying the balanced lift method (Figure 6 design based on the balanced lift method shown in dark grey) were 30% lower than the original design by incremental launching of steel bridge girders (Figure 6 original design shown in light grey). The balanced lift design was also over 50% slimmer than the steel bridge girders, with 2.0 m and 4.6 m, respectively (Figure 6 Section A – A original design shown in light grey, design according to the balanced lift method shown in dark grey). It was proposed to build the central section of the webs using the balanced lift method as shown in Fig. 6, to install the end sections of the webs with mobile cranes placed beyond the abutments and to build the deck slab similarly to the original design using a formwork carriage. In the course of preparing the alternative design, the abutments and the locations of the central piers were rotated through 30◦ on plan with respect to the longitudinal axis of the bridge as a response to the 579
Figure 5.
Box-section girder out of hollow wall elements and ultra-thin precast elements.
Figure 6. Comparison of the original steel-concrete composite bridge design and the design based on the balanced lift method for a post-tensioned concrete bridge over the River Lafnitz.
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Figure 7.
Large scale test of the designed bridge over the Lafnitz river.
location of the riverbed and in order to provide an improved design for the passage of deer. These changes resulted in a bridge design with two equal spans. After ASFINAG became convinced of the benefits, both financial and aesthetic, a detailed design of the two bridges for crossing the Lafnitz and Lahnbach rivers was commissioned. 3.2 Cross-section The S7 motorway is designed with two separate carriageways, therefore each bridge (Lafnitz and Lahnbach bridges) has two carriageways that are to be erected detached from another, making a total of four bridges to be built. The cross-section of the bridges of the Lafnitz and Lahnbach rivers is a doubled webed t-beam with a width of 14.5 m (Figure 6 Section B–B). During the construction phase and the erection using the balanced lift method only the out walls of the webs of the double webed t-beam cross-section (u-shaped bridge girders out of thin-walled precast elements) are existent. Since two bridges are needed for the crossing of the two rivers and each bridge crosssection consists of two bridge girders a total of eight separate balanced lift method operations are to be performed during the construction of the S7 motorway bridges. 3.3 Large-scale tests In the course of research projects and the design of the S7 motorway bridges it became possible to carry out a large scale test of the balanced lift part (72 m in length) of the bridges described in the previous section. The large scale test was based on a 70% scaling of the designs for the S7 motorway bridges, and therefore had a total length of 50.4 m (Kollegger 2010). After four days of construction, the bridge was ready for the balanced lift (in this case lowering) part of the construction process. The lowering process was carried out by slowly and simultaneously lowering the two top points of the bridge girders with the help of two mobile cranes. After the lowering process the geometry of the structure was checked and the nodes at the pier were filled with concrete in order to provide a higher resistance against horizontal wind loads. The large scale model is shown in Figure 7. 4 CONCLUSION By applying the balanced lift method, the spans of bridge girders are reduced by the connections with the compression struts, thus enabling considerable savings in construction material. The assembly and erection operations of the bridge are concentrated at the pier and the rotation of the bridge girders can be carried out much faster than by horizontal launching or the cantilever method. The small space requirements and the high construction speed might be of advantage when an obstacle 581
such as a railway or a busy motorway has to be spanned by a bridge and traffic interruptions have to be kept to a minimum. For bridges with high piers the fast construction period through the application of precast elements is most advantageous. ACKNOWLEDGEMENT The financial support by Österreichische Forschungsförderungsgesellschaft (FFG), ASFINAG, ÖBB Infrastruktur AG and Vereinigung Österreichischer Beton- und Fertigteilwerke (VÖB) is gratefully acknowledged. The good cooperation with Schiemmta Consult GmbH during the detailed design of the bridges for the S7 motorway, and with Franz Oberndorfer GmbH & Co KG during the construction of the large-scale test structure is also gratefully acknowledged. REFERENCES Foremniak, S. 2014. Bridge girders out of hollow wall elements and ultra-thin precast elements. The 10th fib International PhD Symposium in Civil Engineering. Quebec, Canada. Gmainer, S. 2011 Brückenklappverfahren – Untersuchungen zur Entwicklung eines praxisnatauglichen Bauverfahren. Dissertation, Vienna University of Technology. Kollegger, J. 2006. Brückenklappverfahren. German patent No. 102006039551. Kollegger, J. & Gmainer, S. & Wimmer, D. 2010. Maintaining Balance. BRIDGE Design & Engineering 61 (4): 30–31. Weiss, W. 2015. Herstellung von zweifeldrigen Talbrücken mit dem Brückenklappverfahren. Master Thesis, Vienna University of Technology.
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Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
An innovative system of precast segmental span-by-span construction for span lengths of above 100 m J. Muñoz-Rojas, S. Fernández & C. Iglesias Carlos Fernández Casado S.L., Spain
P. Pacheco, H. Coelho & A. Resende BERD, Porto, Portugal
ABSTRACT: An innovative solution has been jointly developed by the companies BERD and CFC in order to push the limits of the span-by-span construction with precast segments to values above 100 m, which are to date exclusively in the realm of the balanced cantilever systems. The particularity of the proposed erection system is that, in order to reduce the weight and length of the launching gantry, the construction is performed in two stages. In the first stage, a 1/2L portion of the deck is erected by extending the precast segments symmetrically on either side of the support section in symmetrical cantilevers which are then prestressed with a set of cables located in the upper slab. Next, the gantry places the central segments that are connected to the parts already built with another set of prestressing cables located in the lower slab. The paper presents two cases for which this system has been proposed, a 3.5 km long viaduct in Cartagena (Colombia) with a classical parabolic variable-depth PC box section and 102 m long spans, and a 10 km long viaduct in India with a 124 m long extradosed structure. 1 INTRODUCTION Prestressed concrete construction with precast segments is a very competitive solution for large and medium span viaducts. Segments may be assembled either span-by-span or applying the cantilever method. In the former case launching gantries can be used; alternatively, should the conditions underneath the deck allow, segments may be laid on provisional supports. Launching gantries can also be utilized when cantilever method is applied; other options include using hoist frames placed at the advance end or erecting the segments by means of cranes from below in case the conditions underneath the deck should allow this procedure. We will now focus on the launching gantries system: this is the solution providing the best of performances, regardless of the conditions underneath the deck, while at the same time allowing easy access to the advance forefront of the installation. The maximum lengths of the span-by-span solution or those of the cantilever method are obviously determined by the capacity and cost of the auxiliary assembling machinery. The usual equipment that can be found nowadays on the market places the current cost-efficiency range of the span-by-span construction around 50 m in competition with in-situ construction with a movable falsework. In the cantilever method solution, however, where only one or very few segments that need to be suspended at the same time, the spans are therefore larger, ranging about 70–90 m and even reaching over 100 m in some cases already put into practice. As with all productive activities, technological evolution and industrial competition tend to inexorably and constantly push or modify the existent limits further. One of the systems that emerged in recent years and opened the door to such developments is the OPS system of the firm BERD. By using active prestressing in the launching gantry, the systems allows the reduction of need for steel in the metallic structure of the girders. At the same time, among other advantages, it provides a more effective and active control of the deformation during the assembly process. 583
Figure 1. Cantilever and span by span segmental construction with launching gantries.
In cooperation with BERD, CFC has explored the application of this system to different projects carried out with precast segments and built span-by-span, with special spans or requirements. The first of these experiences was the New Pumarejo Bridge in Barranquilla (Colombia). In this case, span-by-span construction was envisaged for the approach viaducts. This case is extraordinary not only because of its typical span (70 m) but also because of the heavy load that is to be supported by the girder, given the exceptionally wide deck (38 m). Comparative studies showed that this proposal was cheaper than the cantilever method or the in-situ construction with a movable falsework. The second experience presented in this paper explores the potential of the system for tackling large spans, as long as 125 m, built to date by applying only the cantilever method (L > 80 m). The solution reached is an intermediate assembling procedure half-way between the cantilever method and the span-by-span construction. It is the result of a compromise between the pursued goal of achieving assembly pace and performance similar to the span-by-span method as well as reaching an optimal equipment cost. 2 PROPOSAL FOR THE “GREAT VIADUCT” OF CARTAGENA DE INDIAS (COLOMBIA) 2.1 Project description The first application of the system was developed in cooperation with the firm OHL for the public tender called for a large viaduct in Cartagena de Indias (Colombia). This is to be a viaduct 3.5 km long, with over 30, 102 m long spans. The deck cross section is a conventional box girder of a variable depth, 5.10 m deep at the supports and 2.40 m deep at midspan. The box girder’s upper slab is 11.70 m wide, while the lower is 5.80 m wide. To optimize the equipment and the segment handling elements, the cantilevers are to be executed in a second phase, one or two spans out of sync with the advance forefront. 2.2 Description of the planned execution procedure The central question here was to find a solution to execute a deck of nearly 50.000 m2 in a project where the spans were originally to be built applying the free cantilever method with in-situ segments. However, the deadline was an important factor given that we are dealing with a concession contract. The advantages of the industrialization involving prefabricated construction quickly proved to be the most suited solution. Since the overall span was impossible to modify, we had to think of a solution to span the 102 meters as efficiently as possible. As mentioned above, the execution of one span is carried out in two phases: – In the first place, a symmetrical “T” is built from the pier, with cantilevers whose length amounts to 1/4 of the span. – The central portion is then completed with the remaining segments, which are connected with one cantilever on either side. 584
Figure 2. Typical cross section.
Figure 3. Elevation of the launching gantry.
Figure 4.
Erection of starting segments.
The segments are put in place by means of an upper, lattice launching gantry whose length is approximately 1.5 times that of the span. Given the features of the project, the gantry is supported at the front directly on the foundations, while in the back it rests on one end of the previously built cantilever. To optimize the prestressing of this segment, the support is carried out 20 m away from the pier axis. The launching gantry has a couple of binary supports at the front end, used exclusively during launching operations. The segments are lifted from the already built deck by means of hoisting units and are thereupon brought to its definitive position with the help of a winch that passes along the upper chord. They are finally suspended by adjustable elements. 2.3 Detailed sequence The building sequence of a span begins by placing the girder, supported at its front, binary supports. The initial segment is then built upon the pier. This segment is the one that has the greatest depth, so in order to lighten its weight its length is reduced to 3.0 m. For the same reason the concreting of the inner diaphragm of this segment is carried out in a second phase, once it is installed in its definitive position. 585
Figure 5. Working position of the launching gantry.
Figure 6.
First stage. Erection and prestressing of segments near pier.
Figure 7.
Second stage. Erection and prestressing of mid-span segments.
– Once the deck girder is placed the launching gantry is moved forward and placed on definitive supports in order to build the segments that rest upon the front foundations and upon the rear cantilever of the already built deck. – The remaining segments are then placed in the following, two-phase sequence: In the first place, a “T” symmetrical to the piers is built with cantilevers 23.5 m long (7 segments). These segments are connected using a first set of upper cables. – The rest of the central segments are built thereafter (11 5.0 m segments). It is at this point that the greatest stresses at the rear end of the cantilever are produced due to the support given to the device from which the segments are suspended. The prestressing installed in the previous phase is designed so that it may adequately endure this. – Once the segments are in place and the geometrical adjustments performed with the help of two, in-situ, 0.50 m joints on either end, the second phase of prestressing cables installation begins. This phase is made up of two sets of cables, those located in the lower slab, which connect the central segments, plus an additional set of upper cables at the sections close to the piers whose function is to provide a reserve of additional compressions in these sections to help resist the final loads of the structure. 586
Figure 8. Typical cross section.
2.4 Technical details Great Viaduct (lenght 3.5 km, span 102 m). Cartagena de Indias. Colombia. Project CFCBERD/OHL Building Contractor. (Gonzalo García-Villalba, Mauricio Aguirre). 2014 Main deck values: Concrete 0.66 m3 /m2 Prestressing: 33 kg/m2 Passive reinforcement: 125 kg/m2 3 PROPOSAL FOR “GREEN FIELD SIX LANE EXTRADOSED CABLE BRIDGE” 3.1 Project description The second project in which the application of this construction procedure was studied was an extremely long-span viaduct, almost 10 km long, over the Ganges for an international call for tenders, which is currently suspended. The required modulation dictated 123.5 m long spans along the entire viaduct, to be carried out with an extradosed stay-cables. In this case, the great width of the cross section, 29.40 m, made us propose a two-phase execution. The first phase consisted of the execution of the central core with prefabricated, match-cast segments with epoxy-filled joints, while the second phase contained the building of the lateral cantilevers made up of in–situ slabs supported on struts. From the point of view of the longitudinal configuration of this civil work, it is divided into 8 separate structures. On each stretch the deck is built-in at the two central piers in order to resist longitudinal actions, whereas in the rest of the bridge it is left unrestrained with the help of sliding bearings while transversely fixed at the piers. Each stretch that begins at a pier consists of one initial segment, 3.50 m long and 12 5.00 m long segments. The 9 central segments are axially cable-stayed to a 15 m high tower. The configuration of the cable-staying is extradosed. This allows constant depth all along the deck in spite of the 123.5 m long span. The cast is, therefore, one and the same for all the segments of the entire bridge. 3.2 Execution procedure description There were two main guiding ideas in the design of this viaduct, almost 10 km long and 30 m wide, is to be built: – Firstly, the section dimensions have to be divided into small parts, mutually as independent as possible, in order to be built independently and minimize the critical path. – Secondly, a maximum prefabrication must be achieved. This is why we decided to apply a two-phase procedure similar to the one developed in the previously described project: a precast segmental construction performed span-by-span with a t 587
Figure 9.
Figure 10.
Elevation of typical span.
First stage. Erection and prestressing of segments near pier.
launching gantry as light as possible. In a first stage symmetrical cantilevers beginning at the pier are erected. Thereafter the construction of the central part is done. The box girder is optimized in order to lighten its weight, bringing it down to a width of mere 11 m. Once the continuous girder with this central core is put in place, the lateral cantilevers are concreted resting upon prefabricated struts. The most innovative aspect introduced in this project is the way of compensating the tensions of the segments’ joints during construction. In an extradosed bridge the required compressions levels are provided by the combination of prestressing cables and the extradosed cable-staying system, which also evens out a portion of vertical reactions through the cables’ slant. The customary procedure in these bridges is to gradually compensate the loads from the start by means of stay cables applying the free cantilever method aided by the prestressing. Thereafter, once the structure is completed, the remaining prestressing cables are installed and these contribute to resist the rest of the dead and live loads that will act upon the structure. In this work the roles of the cable-staying and the prestressing are inverted. This big advantage facilitates both the execution of the tower and the installation of the stay cables without slowing down the cycle. The weight segment central core and the reactions of the launching gantry during erection are compensated by conventional prestressing, Once the continuous girder with the segment core is placed the remaining self-weights (lateral cantilevers resing upon prefabricated struts) and the dead load are resisted by the stay cables. 3.3 Detailed sequence – Launching of the metallic scaffolding, hanging of the segments and their assembling in groups of 13 at the symmetrical cantilever of the front pier. As an approximate order of magnitude, the average weight of the segments is about 125 T. – Prestressing of the segment group of the front pier’s “T” at the advance forefront. – Hanging and assembling of the segments in groups of 6 as the current span completion is accomplished. – Prestressing of the segment group as the current span completion is accomplished. – Shifting of the scaffolding to the following span. Return to point 2, and repetition of the operations of segment launching and hanging 80 times until completing the structure. 588
Figure 11.
Second stage. Erection and prestressing of mid-span segments.
Figure 12. Third stage. Erection of prefabricates pylons.
Figure 13.
Fourth stage. Installation of stay cables.
– Transport and assembly of the extradosed cable-staying towers utilizing a crane. The towers are 16 m high and their weight is 150 T. – Cable-staying (phase 1) of the deck portion composed of the central core. – In-situ construction of lateral cantilevers with transverse form travelers. – Cable staying (phase 2) of the deck portion composed of the complete cross section. 3.4 Technical details Green Field Six Lane Extradosed Cable Bridge near Kachhi-Dargah on NH-30, close to near Bidupur in Dist. Vaishali on NH-103 (project currently suspended). Project CFC-BERD/ISOLUX. (Javier Encinas, Pedro Vizeu). 2014 Main deck values: • • • •
Concrete 0.68 m3 /m2 Prestressing: 33 kg/m2 Stay cables: 7.1 kg/m2 Passive reinforcement: 120 kg/m2
4 CONCLUSIONS. ADVANTAGES OF THE SYSTEM The described system for span-by-span construction with prefabricated segments in two phases differs from the usual span-by-span procedure construction. However, this solution enables the tackling of such spans as have to date only been built applying the balanced cantilever system. The advantages of this innovative solution can be summarized as follows: – Lower investment in launching girders given the shorter span to be loaded in each phase (less structural steel, length reduction). – Greater approximation of the sectional forces in the structure during construction to those produced in the definitive structure with the resulting optimization of the prestressing amount and distribution. 589
– Greater geometrical control during assembly. This is due to the fact that, on the one hand, the system allows to correct the alignment of the initial “T” , and on the other, adjust the assembling of the remaining segments by way of two closing joints. – The two-phase sequence does not affect the cycle significantly. According to various studies, one span can be built in seven days in the case of the Cartagena viaduct (102 m long span), while in the case of the extradosed, 125 m long bridge, the pace of two spans in three weeks can be accomplished. REFERENCE Pacheco, P., Coelho, H., Resende, A. , Soares I. 2014. High productivity in bridge construction – the OPS effect. 9th International Conference on Short and Medium Span Bridges, July 15–18 2014. Calgary, Alberta, Canada.
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Launching of fully welded steel long span bridges: Bogibeel bridge A.K. Mathur, S.S. Shukla & J. Gupta RITES Ltd., A Government of India Enterprise, Gurgaon, Haryana, India
ABSTRACT: Railways is constructing a Rail cum Road Bridge at Bogibeel, over River Brahmaputra in Assam (India) connecting Dibrugrah town on the South bank in Lower Assam to Upper Assam and Arunchal Pradesh bordering China through Sisibargaon on North Bank. Brahmaputra, one of the world’s largest rivers, having length of 2890 km & width between 3 to 22 km, divides the transport network of Assam Valley into Northern and Southern parts along the river banks. Khadir width at location of bridge is 10 km with a bridge length fixed as 4943 m based on hydraulic model studies. To efficiently construct such a long bridge over the highly dynamic river, a unique methodology was envisaged for round the year construction independent of the river hydraulics. This paper describes development of the Launching Scheme and process employed to erect in position fully welded 125 m span truss girders using Incremental push launching.
1 INTRODUCTION The Bogibeel Bridge superstructure is divided into three parts viz. the Viaduct portions on North bank and South Bank supporting the roadway and the Main Bridge consisting of two Tier Rail cum Road composite bridge with two track railway on the lower level and three lane roadways on the top level. The main features of the Bogibeel Bridge are – – – – – – – – – – – – –
South Viaduct – 833.973 m Main Bridge 1 × 33.4 + 39 × 125 m + 1 × 33.4 m North Viaduct 833.973 m Superstructure; Through Warren type triangulated fully welded composite Simply Supported steel truss girder Roadway deck width 9.075 m with 0.5 m crash barrier and 1.5 m footpath on each side. Height of Truss 11 m (Centre to Centre) Centre to Centre of truss 10.6 m Total Structural Steel Quantity – 68000MT Substructure; Double D shaped well foundations of size 16.2 m × 10.5 m Design Discharge – 73000 cumec Protection Work – North Guide Bund −2792 m, South Guide Bund – 2043 m Erection Method – Incremental Push Launching Design as per IRS Bridge Rules for MBG Loading, Steel Bridge Code and IRC codes (70R/Class A Loading). Compliance to Euro-Codes for Fatigue criteria.
2 DESIGN SERVICE STAGE The superstructure is designed as a simply supported through type Warren triangulated fully welded composite steel truss girder. 591
3 CONSTRUCTION STAGE For the construction/erection of the main truss girder (125 m) alternative schemes such as Cantilever construction of truss girders by incremental erection and in situ connection of components, Erection of individual truss girders by using floating cranes or pontoons, erection by rolling completed truss girders over temporary launching trusses, use of two level launching truss i.e. rolling completed truss girders over already erected and positioned trusses, were considered. The relative merits and challenges were considered and finally Incremental Push Launching scheme using the final truss girder itself as part of a continuous consist (upto 10 spans weighing approximately 17000 t at a time) for the launching with minor strengthening arrangements was selected based on overall economy in use of the final truss girder itself in lieu cost of fabrication of additional launching trusses/Pier head brackets etc. 4 CONSTRUCTION AND ERECTION/INCREMENTAL PUSH LAUNCHING Structural analysis of the incremental push launching scheme was conceptualized using a series of truss girders with launching nose and tail attachments, made continuous and nomenclature as a “consist” (1 consist =10/9 truss girders in a series), through welding using temporary connection pieces at the ends of each truss girder. The system of consists being push launched was designed using individual simply supported truss girders (designed for service Stage stresses) and checked as a continuous structure during launching (construction/erection loads). The unique and most complex analytical check was at the stage of lowering of the truss girders onto their final positions on the bearings. At this stage all locked in stresses in the continuous girders due to the effect of continuity are to be released by a systematic procedure of jacking up and jacking down of some of the supports. The stresses in the connections are brought to minimum values and under such reduced stress levels in the connections, the connections are cut with flame cutting and the ends of the trusses are finished to the required specification. Since the launching levels are higher than the required final levels of trusses, all the trusses are jacked down systematically to lower the trusses on final bearings. 5 THE LAUNCHING PROCESS Prior to launching, all components of the truss like Nodes, Bottom chord, Top chords, Vertical, diagonals, cross girders, cross bracings, stringers, brackets etc are fabricated and transported to Lay Down Assembly sheds. The components are placed accurately on the floor to make modules of about 30 m each. Each of these modules (weighing approximately 110 to 135 T) are transported to vertical assembly yard and joints between the modules are welded to form the truss supported on sledges in actual cambered profile. The other components like Cross Girder, bracings etc. are connected to form whole truss (Fig. 1). A prefabricated launching nose (52 m long) is fixed to the front end of the truss to reduce the cantilever bending moments and reactions on bearings due to cantilever action during launching. A specially designed launching bearing (Fig. 4) will be mounted on each pier. The truss girder is pulled forward with the help of a pulling device consisting of two 1000 t capacity multi strand jacks mounted on Pier P1 of main bridge and is continued till the nose touches down on the launching bearing mounted on subsequent support. Launching nose is profiled upward so that the nose lands freely on the launching bearing in spite of cantilever deflection of approx 1500 mm in order to negate the deflection of the nose tip. The girder is continuously monitored for any lateral shift and corrective actions is done by side shifting of truss using guide/jack arrangement. At this stage the second truss is moved forward and connected to the rear end of the first truss with connection pieces welded between the chords of the two trusses. The process is continued further in a similar operation after moving the anchor frame to the first panel abutting the End 592
Figure 1.
Schematic view of launching stages.
bottom cross girder of the newly attached Truss Girder. This process is continued till the entire consist (10 girders) are moved onto the piers and the rear end is attached with a launching tail. While a new set of pulling device/arrangement is set up on the 11th pier for further pulling the first consist forward towards the 20th per and onwards to the end, a new set of girders forming the next consist are erected and launched in a similar manner. Once the first consist reaches its final position on the 40th pier, the Nose and tail of the first consist is dismantled. Since the previously launched consists are already in place, the nose of subsequent consist cannot cross beyond the receiving pier cap. To facilitate further launching, a pier bracket is attached to every such 10th Pier and a launching Bearing is mounted on this pier bracket. Part of the nose beyond the launching bearing on Pier bracket is dismantled. After dismantling part of the nose, further launching is done in stages and simultaneously part of the nose is dismantled. This process is repeated till the rest of the trusses are launched. 593
Figure 2.
Schematic arrangement showing cutting and positioning of truss.
Figure 3.
Bottom chord resting over launching sledge and launching track.
6 FINAL POSITIONING OF TRUSS GIRDER ONTO PERMANENT BEARINGS The girders forming the consists are continuous during launching and individual truss girders need to be separated for lowering onto the final position on bearings for service stage. Due to the effect of continuity there are locked in stresses at the connecting members of the truss girders which need to be released. By a systematic procedure of jacking up and jacking down of some of the supports (Fig. 2), the stresses in the connections are brought to minimum values and under such reduced stress levels in the connections, the connections are gradually cut with flame cutting and the ends of the trusses are finished to the required specification. Since the launching levels are higher than the required final levels of trusses, all the trusses are jacked down systematically to lower the trusses on final bearings. 7 ENABLING WORK REQUIRED FOR LAUNCHING 7.1 Launching track & launching sledges The completed trusses are moved from the assembly yard to launching position over a track consisting of steel girders with stainless steel top plate (Fig. 3). The truss is supported and erected on “Sledges” having a stainless steel backed PTFE plate which slides over the launching track. 7.2 Sliding plates During launching process, sliding plates are inserted between the bottom flange of the superstructure and the top of launching bearing/Sledges. The sliding plates consists of steel backing plate with PTFE. The PTFE is chambered in the steel plate and fixed with countersunk bolt. 594
Figure 4.
Launching bearing.
Figure 5. A-struts.
7.3 Launching bearing Rocker type launching bearings (Fig. 4) are arranged under both sides of the bottom chord of the truss at each pier/support. Launching bearings at a support are connected with each other by two separate beams 2000 mm apart. Two lateral guiding arrangements are located in the line of the cross beams. The lateral guides take the lateral forces due to jacks employed to correct lateral shift of girders under wind, transverse forces due to eccentric launching forces, temperature, differential settlement or tilting of bearings etc and transfers the forces to the bearing pedestals. Forces in longitudinal direction due to friction are transferred to the launching bearing using a stainless steel interface and further to the pedestal through neoprene pads. 7.4 Launching nose/tail In order to reduce the reactions and bending moments in the members of truss during full cantilever stage of launching, a nose of 52 m length (Fig. 8) is attached to the front and rear end of the of the leading and trailing truss of the consist respectively. During maximum cantilever stage, the nose will deflect considerably, estimated at around 1500 mm hence the bottom of the nose is given an upward taper in the front to ensure that the nose rides over the launching bearing even after deflection at time of touch down. 7.5 A-strut To reduce the stresses due to the local bending of the bottom chord vertical temporary beams has been provided between the bottom chord and top joint of both sides of the truss. The A – struts (Fig. 5) are placed by bolting in the zones of the maximum vertical forces due to launching in the first and last trusses of each segment and removed after final positioning of girder. 595
Figure 6. Anchor arrangement at first cross beam.
Figure 7.
Pulling arrangement and strand jack at pier.
7.6 Pulling systems To pull consist of truss girders a pulling device (Fig. 9) consisting of multi strand jacks placed at top of pier P1 (Figs. 7 & 8) and properly anchored to the substructures (piers), is used. The anchor beam is located at the first cross beam (Fig. 6) of the truss. These multi strand jacks, located at pier P1 pulls the strands which are anchored to the anchor beam. To limit the strand length the anchor beam is shifted after each launching stage of 125 m. 8 MEASURES CONSIDERED DURING LAUNCHING 8.1 Smoothening of bottom plate of chord Contact surface of Bottom Chord should be properly smoothened to ensure minimum friction between supporting PTFE and Bottom Plate. We are required to limit the maximum friction coefficient between PTFE and Steel as 5%. Increase in coefficient of friction will transfer increased force in Cross girder, Launching bearing through jack and hence in Pier. If the friction coefficient exceeds above 5% then number of truss in consist can be reduced to restrict the force on in the pier within permissible value. 8.2 Continuous observation on PTFE over Sledge and launching bearing PTFE plates with thickness 5.5 mm and allowable compression ≥15 N/mm2 have been used for sliding material this PTFE will slide on the stainless steel plate. Due to continuous operation 596
Figure 8.
Picture showing pulling arrangement and launching nose at pier P1.
Figure 9.
Pulling jack at pier P1.
under heavy loads PTFE is likely to deform over time. Hence PTFE surface should continuously be monitored for abrasion both in sledges and launching Bearing, to achieve desired friction for operation. 8.3 Lateral movement of truss and side pushing Train of trusses may shift laterally during launching due to various reasons like settlement of one support launching bearing with respect to the other, difference in friction coefficient of one side of PTFE and other, imbalance force transfer through Jack etc, wind force or temperature differentials etc. To control the side shift, continuous observation are required at all launching bearing location and shifts recorded. A 20 mm gap between side guide arrangement and bottom plate of Bottom chord of truss has been provided which is maintained. A Jack arrangement providing lateral force has been provided in all launching bearing for the sideways repositioning of truss from either side. 8.4 Slack in strand of jack All the wires of the pulling cables should be slack free and uniformly stressed to avoid imbalance in forces transfer to the truss girder. 597
8.5 Monitoring of deflection of launching nose Continuous measurement of the deflection of the tip of nose needs to be done with the help of leveling staff installed at the nose for reference to the design values. A 1500 mm upward taper has been provided in the launching nose to counter the deflection of the tip of nose in maximum cantilever position. 8.6 Installation of strain gauge The Strain Gauges are installed at critical stress locations (as determined from design) of the leading and trailing end truss girders of the consist. Three numbers of Strain Gauges are installed at each of the above mentioned critical locations identified on the Bottom and top chords. The values measured shall be checked with the design values. In case the strain measured is more than the permissible limits; it shall be referred to designer for further analysis. 9 CONCLUSION: BENEFITS OF ADOPTING INCREMENTAL PUSH LAUNCHING 9.1 Executability Proposed launching scheme enables complete fabrication and assembly of truss under conditions of controlled environment to achieve high quality of welds and paint, minimizing requirement of on-site operation. 9.2 Optimum use of material strength The proposed launching scheme optimally utilizes material strength of the final truss by providing continuous span arrangement during launching. By this method launching forces get distributed and structure has capacity to handle more erection loads. 9.3 Damage to corrosion protection system Damage to the corrosion protection system is negligible as all relative movement is through PTFE surface only and there is no exposure to clamps, fixtures or other lifting/handling devices. 9.4 Better control over structure to be launched As there is no handling and lifting of individual trusses etc during the process of launching there are no risks of damage to members or abrasion on painted surfaces, breakdown of truss or submergence in the water etc. Due to continuity in the consist during launching and inbuilt redundancy chances of failure/dislodgement of individual truss is eliminated. 9.5 Non dependency on river hydraulics The Launching is being done from above the pier level and there is no dependence/hindrance on/due to river hydraulics. The scheme envisages a very Limited use of Pontoons for lifting/dismantling of nose/tail segments. 9.6 Cost effectiveness The proposed launching scheme is cost effective as it involves minimum amount of enabling works, tools and plants. It also allows continuity of launching during rainy/or peak flood seasons thus reducing overall construction period and related idling/storage cost.
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Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
Swivel lowering operation of the viaduct over the River Tera F.J.M. López, M.B. García, M.M. Cañueto, J.M.G. Parejo & M.T. Serrano Acciona Ingeniería, Alcobendas, Madrid, Spain
ABSTRACT: The deck is a concrete box girder which shall be launched from one the abutments and has a total length of 645 m. It is simply supported on concrete piers spaced 75 m. However, because of the wide riverbed, one of the theoretical piers has been replaced by a pair of inclined steel semi-arches. At their lower end they are fixed to the foundation of those adjacent concrete piers. A distinct feature of this viaduct lies in the special construction method used for the erection of the inclined semi-arches: they have been lifted into a vertical position and, by means of a swivel lowering operation and external cables, erected and joined at their upper end. This method involved the provision of pins at the base of the semi-arches. In addition, the hinges were designed to be able to allow the adjustment of the semi-arches during the final phase of the erection.
1 DESCRIPTION OF THE VIADUCT The Viaduct over the River Tera has a total length of 645 m. It has eight spans of 60 + 4 × 75 + 150 + 75 + 60 m span lengths. It is located in the North-Northeast high-speed Corridor connecting Madrid and Galicia, in the Zamora-Lubian section, subsection Otero de BodasCernadilla. The deck is a prestressed concrete box girder that has been launched from one of the abutments and has a total width of 14.00 m where two railway lanes two pedestrian walkways are located. The viaduct main span, see Figure 1, is 75.00 m long in relation to the deck but 150.00 m long in relation to the substructure. This is because the theoretical center pier was replaced by two inclined semi-arches of S460 steel. The angle to the horizontal is about 30◦ . Those semi-arches have an approximately rectangular cross section with variable width and depth, ranging from 6.00 × 3.00 m at the base to 5.40 × 2.50 m at the crown. Each one has a length of 81.00 m and their total weight is 3000 kN. The crown element has a parallelepiped shape and supports both the definitive bearings as well as the temporary guiding bearings. For maintenance reasons both semi-arches as well as the crown are accessible in safety conditions. 2 ERECTION METHOD Basically, the erection method of the semi-arches involved its assembly in a basically vertical position and a subsequent rotation with the aid of guiding cables, detent hydraulic cylinders and the force of Gravity. The semi-arches are rotated in relation to a pin located at the base and the guiding cables are anchored to the adjacent footings. See Figure 2. In the initial position the two semi-arches were slightly slumped against the piers. The semiarches were anchored to the piers by means of quick fastener devices to ensure safety during the assembly of all the elements needed for the operation. To break this static equilibrium, it was necessary to introduce into the system loads with horizontal components in order to rotate the semi-arches, causing them to slump over the riverbed in a controlled manner so that the guiding cables progressively increase their tension. The balance breaking loads were generated by a set of horizontal hydraulic cylinders located on top of the main piers. 599
Figure 1.
Main span.
Figure 2.
Layout of the guiding cables, pins and anchors.
Once the balance was clearly broken the operation progressed itself simply by controlling the length of the guiding cables by means of the hydraulic cylinders. The guiding cables increase their stress as the semi-arches rotate when the guiding cables increase their length. Thus, from a structural point of view, the swivel lowering operation is a sequence of intermediate equilibrium positions between the semi-arch, its guiding cables and the actions of self-weight, dead loads, and wind acting on the system. It is a clearly isostatic system. See Figure 3. Once contact was made between the semi-arches the resulting geometry was evaluated by topographic means. Everything was as planned. Schematically, the stages of the swivel lowering operations were: – – – – – – – – – – – – –
Placement of anchors in the adjacent footings. Installation of the pin in the main piers. Erection of the concrete piers. Assembly and welding of the sections of semi-arches in a substantially vertical position. Placing of guiding cables and hydraulic cylinders. Placement of balance breaking horizontal hydraulic cylinders on the top of the main piers. Tensioning of guiding cables. Balance break. Rotation of the semi-arches. Crown joint. Adjusting of geometry. Detensioning and complete removal of guiding cables and auxiliary elements. Final fix and concreting of the base of the semi-arches. 600
Figure 3.
Swivel lowering movement.
3 SPECIALLY DESIGNED ELEMENTS FOR THE SWIVEL LOWERING OPERATION This unique process of erecting the semi-arches has involved the design of a set of auxiliary elements as described in this section. 3.1 Anchor of the cables in adjacent footings As explained above the guiding cables modified their length to allow the rotation of the semi-arches. It was decided that a passive anchor be placed to the guiding cables at the back of the semi-arches. Consequently, the anchor to the adjacent footings was active. See Figure 4. The active nature of the anchor is achieved with the installation of detent hydraulic cylinders. See Figure 5. 3.2 Pins at the base of the semi-arches These elements were located at the base of the semi-arches. There were two pins on every semiarch. These elements are responsible for the rotation of the semi-arches. They are not mere pins in the classic sense of Strength of Materials because they were capable to displace longitudinally in a controlled way. See Figure 6. These elements were designed with the ability to slide differentially in the longitudinal direction. Acting in a coordinated way in the bearings of the two semi-arches it was possible to have additional degrees of freedom for the accurate settlement of the geometry resulting from the swivel lowering operation. Thus, each pin had small horizontal hydraulic cylinders to enable differential longitudinal movements. Additionally, they had to be designed to withstand the longitudinal loads and lateral and transverse overturning moments due to the self-weight of the semi-arch and the wind. 3.3 Passive anchor frames on the back of the semi-arches As described above the passive anchor of the guiding cables was placed on the back of the semiarches. See Figure 7. 601
Figure 4. Anchor of the guiding cables to the adjacent footing.
Figure 5.
Detent hydraulic cylinders.
3.4 Elements for balance breaking These elements were required to kick start the rotating movement of the semi-arches. They were to be able to overcome the slight slump that semi-arches presented against the piers in the initial position. Also they were to withstand the possible necessary reaction to maintain balance due to a potential gust of longitudinal wind on the semi-arch. They were designed to rotate along with the semi-arches in the first stage until the angle reached by them was sufficient to stabilize the moment induced by the self-weight of the semi-arch and an eventual overturning moment in the opposite direction induced by the wind. See Figure 8. 602
Figure 6.
Pins at the base of one semi-arch.
Figure 7.
Passive anchor frame.
Figure 8.
Balance breaking elements.
603
Figure 9.
Final position of the balance breaking phase.
4 LOADS TAKEN INTO ACCOUNT FOR THE SWIVEL LOWERING OPERATION The main loads present during the swivel lowering operation were self-weight and dead loads of the semi-arches and auxiliary elements, and wind. It is generally easy to maintain control of the action of self-weight. In the case of the wind, this is not so clear. Thus, although the assessment of the effects of wind action has been performed according to Ministerio de Fomento (2010), during the two months prior to the completion of swivel lowering movement itself an anemometer was installed to continuously record wind speed at the top of the main piers. The different wind speeds that have been considered were as follows: – Design speed for the design of the elements: 42 m/s. It is equivalent to 150 km/h. Return period of 4 years according to Ministerio de Fomento (2010). – Design speed during the swivel lowering movement itself: 20 m/s. It is equivalent to 72 km/h. This value was agreed with the Owner. – Top speed recorded during the last two months: 16 m/s. It is equivalent to 58 km/h. Figure 9 shows the angle for which a longitudinal wind of 20 m/s and direction from the semiarch to the pile produces a destabilizing moment that exactly balances the stabilizing moment of self-weight and dead loads. Figure 10 shows the angle for which a longitudinal wind of 42 m/s and direction from the semiarch to the pile produces a destabilizing moment that exactly balances the stabilizing moment of self-weight and dead loads. The angles of the semi-arches are different because the crown is fixed to the left one, so the centers of gravity of both semi-arches are not symmetrical. 5 EXECUTION OF THE SWIVEL LOWERING OPERATION The swivel lowering movement itself lasted for two full days, from sunrise to sunset, and a few hours of the third day. It is therefore a slow operation. The balance breaking is the slowest phase. It was necessary during this phase to deal with the force of the balance breaking horizontal cylinders, the measured force in the detent hydraulic cylinders and the minimum allowed tension of 5 kN per strand. This minimum value was necessary to ensure they do not slip in respect of their respective anchor wedges. Once the balance breaking phase was completed the rotation of the semi-arches progressed at a rate of 5.00 m per hour measured as guy cable length. Thus, during the first day it was only possible to break the balance in both semi-arches. They were left overnight in the so called long-term safe position shown in Figures 10 and 11. 604
Figure 10.
Long-term safe position.
Figure 11.
Position at the end of the first working day.
The second working day ended once contact between the semi-arches was made. The operation was halted because of the high temperatures recorded that distorted the topographic measures that were taken to verify the geometry of the crown. See Figure 12. During the morning of the third day, before the temperature increase of the steel plates of the semi-arches, topographical checks were performed and it was determined that their position was correct, so it was not necessary to act on the hydraulic cylinders located on the bearings for adjustment. The welding of the crown joint started immediately and, days later, the guiding cables were definitely loose according to plan provisions. Once the auxiliary elements were removed and the pins were fixed, the launching of the deck continued, as shown in Figure 13. 6 IMPROVEMENTS TO THE SWIVEL LOWERING OPERATION Once completed the entire process, in retrospect, two improvements arise to the design of the swivel lowering operation that would have made it more effective. Both are related to the balance breaking stage, clearly the most critical phase: – To slump the semi-arches over the river, instead of against the piers, at the initial position. – To reduce the number of strands in order to increase the working tension of the remaining strands. 605
Figure 12.
Position at the end of the second working day.
Figure 13.
Finished swivel lowering operation.
The first improvement would have prevented the movement of the center of gravity of the semi-arches over the theoretical vertical line of the pins, achieving the following: 1. To reduce the requirement for the total movement of the balance breaking hydraulic cylinders to reach the so called balance breaking angle. 2. To reduce the requirement for strength of the same hydraulic cylinders to withstand the reactions induced by an eventual wind trying to lift the semi-arch. 3. To avoid the reversal of the reaction of auxiliary devices. The criterion used to dimension the number of strands was 100 kN per strand, using nominal loads, without combination coefficients. It is equivalent to the 41% of the tensile strength reduced by the material coefficient. The second improvement would increase the safety against sliding of the guiding cables with respect to their respective anchor wedges. REFERENCE Ministerio de Fomento. Gobierno de España, 2010. Instrucción de Acciones a Considerar en Puentes de Ferrocarril.
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Deck forces of a cable-stayed bridge – “Analysis of the construction and the in-service phases” P. Almeida & R.C. Barros FEUP, Department of Civil Engineering, Structural Division, Porto, Portugal
ABSTRACT: This paper deals with the construction of cable-stayed bridges by the process of successive advances, with a concrete deck achieved by in-situ concreting. A case study of a cable-stayed bridge built in Brazil is presented, which analyzes the response of the structure under distinct evolutionary construction phases and in-service phase. In the analysis of the construction process it was considered the evolutionary character of the structure, verifying the structural system response in terms of generalized forces and strains. The analysis focused on variants of the method of successive advances to build the bridge deck. One of the analyzed variants is the application of OPS (Organic Pre-stressing System) by implementing an active bolting system that is an alternative to the construction processes currently used. From the analysis of the results it is possible to understand the influence of the selection of the construction process in the bridge deck design. 1 CASE STUDY 1.1 Cable-stayed Bridge Paulicéia The Paulicéia cable-stayed bridge is a bridge across the river Paraná between Paulicéia and Brazilândia in Brazil. The total length of the bridge including the accessing viaducts is 1705 m. The longitudinal configuration of the cable-stayed structure is that of a bridge with two towers, one central span and two side spans. The total length of the cable-stayed zone is 400 m, the main span length is 200 m, and the compensation side-spans have a length of 100 m each. The bridge deck in the cable-stayed zone has a width of 17.30 m. The main bridge structure consists of a deck integrally constructed in reinforced concrete, transversely pre-stressed and longitudinally cable-stayed with ties (Figure 1). The cable-stayed suspension system is lateral, achieved sideways on two levels with 56 pairs of tie-risers spread over two towers (Enescil, 2013). 1.2 Construction processes considered in the analysis The construction process of cable-stayed bridges by successive advances, under the conventional cantilevered system, introduces great generalized forces in the deck and in a few tie risers near
Figure 1.
Bridge Paulicéia –longitudinal configuration of the cable-stayed spans (m) (courtesy ENESCIL).
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Figure 2.
Feed-slide chariot with integrated OPS system (adapted from BERD).
the segment being built. The highest efforts are produced after concreting the cantilevered segment being built, with low stiffness and before placing the pair of tie-risers of the corresponding construction phase. With the conventional system during the execution of a segment occur high bending moments and high tensile stresses in the top deck fibers. Sometimes, in order to aid the construction phase, it is necessary to use temporary longitudinal pre-stressing, causing an oversizing of the structural elements of the bridge structure essentially of the bridge deck (Almeida, 2013). In the construction process by successive advances using organic pre-stressing system (OPS), it is incorporated into the feed-slide chariot (Figure 2) a system that allows tensioning the tension ties in which the hydraulic tensioning is applied, and through which the risers become active. The tensioning of ties is insured through the data associated with the deck displacements (Pacheco, 1999). The sensors receive information in the form of vertical movements, passing this information to controllers or drivers that activate the hydraulic jacks for tensioning of ties (reverse operation is also possible). 1.3 Construction phasing The modeling of the bridge structure for building the concrete deck by in-situ concreting using the cantilever constructive process (Figure 3a), with the conventional system (without the OPS), has the following construction sequence of repetitive stages for the symmetrical construction of 14 segments adjacent to each tower (Almeida, 2013): i. Advancement of the sliding chariots or construction cars from the preceding segments to the segments to be built; ii. Formwork, placing of reinforcement and concreting of the segments (A.04.L and A.04.C); iii. Placement and tensioning of the pair of tie-risers in the segments (T.04.L or T.04.C); iv. Advancement of the sliding chariots or construction cars for the subsequent segments (A.05.L and A.05.C). The modeling of the bridge structure for building the concrete deck by in-situ concreting using the constructive process of successive advances, with the organic pre-stressing system (OPS) as represented in Figure 3b, has the following construction sequence of repetitive stages for the symmetrical construction of 14 segments adjacent to each tower (Almeida, 2013): i. Advancement of the sliding chariots or construction cars from the preceding segments to the segments to be built; ii. Placement of the pair of tie-risers in the hydraulic tensioning system fixed to the sliding chariots (T.04.L or T.04.C). The construction of the segments is no longer cantilevered executed, and is now executed in a supported system, wherein the supports are the anchor points of the risers fixed to the hydraulic jacks of the advancement chariots or construction cars (Pacheco, 1999); iii. Formwork, placing of reinforcement and concreting of the segments (A.04.L and A.04.C); 608
Figure 3.
Construction phasing: a) conventional system; b) system OPS.
iv. Tensioning of the tie-risers for the values required at the end of the construction phasing, and release of the hydraulic jacks Advancement of the chariots or construction cars for the segments (A.05.L and A.05.C). The actions considered (permanent actions and actions during construction) are those associated with: Actions of self-weight of the deck; Actions of construction equipment (650 kN). 1.4 Bridge Paulicéia under service loading In modeling the bridge under service loading was used load model #1 (LM1), composed of concentrated and uniformly distributed loads, that covers most of the effects of lorry traffic (heavy traffic) and automobile vehicles (light traffic). This model is used for local and global checks, as regulated by the European norm EN 1991-2 (CEN, 2003). The deck of Paulicéia bridge does not have central separator, and the action of the standard vehicle type was applied in three reference circulation lanes. The action of the standard vehicle type consists in applying, along the longitudinal central axis of each reference lane, two concentrated loads that are spaced 1.20 m longitudinally. The partition of the concentrated and distributed loads by the reference circulation lanes, follows the numbering and the values of Figure 4 according to the European norm EN 1991-2 (CEN, 2003). 1.5 Selection of steel area of the tie rods For the case study of this analysis it was necessary to rationally design the section of the tie rods, admitting that each pair of tie rods balances the vertical load acting on a specific influence area. The self-weight of the deck for the permanent loads is 243.20 kN/m. Under permanent actions the tie rods are designed for a maximum tensile force of 0.30 (fptk ). 2 ANALYSIS OF RESULTS 2.1 Analysis of the construction phasing and of some serviceability characteristics The bridge software CSI Bridge (CSI Bridge, 2015) was used extensively in the analysis of the present case study. The modeling of the bridge structure for its construction either with the constructive process by successive advances using conventional system (without OPS) or using the organic pre-stressing system (OPS), started from the intended final position of the structure, vertically non-deformed, under the action of dead loads (of deck self-weight, of the pavement surface layer, of the edges and sidewalks, of the guardrails curbs and drainage system). The results presented from Figures 5 to 8, relate to the most unfavorable situation of modeling the construction phases of the bridge segments. The results presented correspond to those of the construction phase of the segments: A.06.L, A.10.L, A.14.L, A.06.C, A.10.C and A.14.C. 609
Figure 4. Transversal representation of distributed and concentrated vehicle loads on the deck lanes.
Figure 5.
Shear force distribution in the bridge deck during construction phasing.
The shear force distribution in the bridge deck (Figure 5) maintains a similar value in all the segments when using the OPS system of construction. Nevertheless when using the conventional cantilevered system, by successive advances (without OPS), occur significant higher values of shear forces in the segment just prior to the segment under construction. During the construction phasing with the conventional system (without OPS), the segments are subject to the loadings on the segment being constructed, without rigidity and with the load of the construction equipment applied on the cantilevered segmental portion. In the distribution of bending moments occur quite different values between the two construction solutions (Figure 6). The conventional cantilevered system by successive advances (without OPS) leads to higher values of bending moments; in the case of using the alternative construction method with the OPS system, the values of the bending moments are quite reduced. During construction of the 6th segment with the conventional system, the negative moments obtained the maximum value of 13135.31 kN.m; while with the OPS system the maximum value of bending moments was 2844.65 kN.m, therefore much lower. During construction of the 10th segment with the conventional system, the highest value is found in the segments A.07.L and A.07.C with 14875.68 kN.m; with the OPS construction system the maximum value of bending moments is much lower: 2892.30 kN.m, a value that was recorded in segments A.09.L and A.09.C. During construction of the 14th segment with the conventional system, the highest value of the negative bending moments (15599.72 kN.m) is recorded in segments A.11.L and A.11.C; with the 610
Figure 6.
Bending moment distribution in the bridge deck during construction phasing.
Figure 7. Tension developed in the tie risers during the bridge construction phasing.
OPS system the value is much lower (2495.51 kN.m) and was recorded in the segments A.13.L and A.13.C. The tension installed in the tie risers, mounted along the bridge deck, also show quite significant differences between the two building systems (Figure 7), although with the conventional system this difference is more pronounced in the last three pairs of rods installed prior to the segment being constructed. When using the cable stayed bridge construction phasing with OPS system, the tie risers keep their tension values with little variation from phase to phase; while when using the construction phasing with the conventional system (without OPS) the tie risers tension values undergo quite significant changes from phase to phase on the order of 30% to 40%. At the end of the construction phasing, both construction systems (conventional cantilevered system and system with OPS active rods) led to exactly the same forces in the tie risers; the only variations occur in the interim envelopes of the generalized forces during the construction phases. The differences of the interim envelopes are a direct result of the construction system used. The vertical displacements have significant differences between the two construction systems (Figure 8); with the conventional cantilevered system excessive displacements occur. With the conventional cantilevered system (without OPS) at the end of construction of the 6th bridge segment the vertical displacement is 0.1399 m below the reference value; when using OPS system, such vertical displacement at the end of the 6th segment is 0.0077 m above the reference value. That is, when using OPS construction system the vertical displacements not only decrease (in terms of absolute value) but also minimize construction phasing errors, so contributing to more efficiency of the construction process and objectives. The vertical displacements, evaluated with the conventional cantilevered system, increase as the deck construction advances. At the end of construction of the 10th segment using conventional 611
Figure 8. Vertical deck displacements, for both construction systems.
Figure 9. phase.
Envelope of deck bending moments, for the two bridge construction systems and for the in-service
system, the vertical displacement is 0.2343 m below the reference value; when using OPS system, such vertical displacement at the end of the 10th segment is 0.02 m above the reference value. With the conventional cantilevered system (without OPS) the deck displacements, already with the 14th segment built, are similar to those determined in other segments and have quite excessive displacements; while when using the OPS system the deck vertical displacements are perfectly controlled. With the conventional cantilevered system, the vertical displacement in the last built segment is 0308 m below the reference value; with the OPS system, this vertical displacement is 0.0565 m above the reference value (Almeida, 2013). 2.2 Envelope of bending moments during construction phasing and in service conditions The bending moment envelopes of the construction phases, either with conventional cantilevered system or with OPS system, are determined considering all the actions defined for these phases. The envelopes are produced taking into account the actions resulting from the construction of all the segments. Figure 9 shows the two pairs of bending moment envelopes of the construction phasing with both construction systems, and also the pair of in-service envelopes with the action of highway overload (Almeida, 2013). The analysis of the construction phasing with the conventional cantilevered system to carry out the segments A.14.L and A.14.C, led to the development of negative bending moments with the value of 15599.72 kN.m. With the bridge in service, the highest bending moments occur: (a) at segment A.11.L (lateral span) with a negative bending moment of 6673.71 kN.m; (b) at segment A.11.C (central span), with a negative bending moment of 6659.45 kN.m; (c) at segment A.01.C (tower support), with a value of 9484.19 kN.m. The maximum positive bending moments were obtained for the analysis of the bridge in service: (a) at segment A.12.L (lateral span), with the value of 16126.69 kN.m; (b) at segment A.12.C (central span), with the value of 13413.23 kN.m. 612
Figure 10. Tension installed in the tie risers, for the two bridge construction systems and for the in-service phase.
The tension installed in all the bridge tie risers, always attains the lowest value with the OPS system compared to the conventional cantilevered construction system (without OPS) and when the deck is submitted to the highway overload. The highest tension value in the tie risers occurs, for all the tie risers, during the conventional cantilevered construction system (Figure 10). The average registered tension in the risers, related to the maximum value registered in all the phases, varies significantly. When using the conventional cantilevered construction system (without OPS): during the construction phasing was obtained an average tension of 783.87 MPa; but if already in the in-service phase with the highway overload, the average tension was 712.22 MPa. When using the OPS construction system, during the construction phasing occurs a much lower value of the average tension of 540.37 MPa; the tension with the highway overload has the lowest value in the lateral compensation spans, where other generalized forces can be higher. 3 CONCLUSIONS The modeling of the construction phasing in the design of cable-stayed bridges is a step of paramount importance. During the construction phase are developed quite significant generalized forces that are potentially detriment for the design of the deck and tie risers, as was seen by the analysis of the results obtained for this case study. During the construction phasing, the bridge deck is subjected the great forces, mainly originated from bending behavior. When using the conventional cantilevered construction system, it is evident from the results that the generalized forces are much higher than those that occur with the implementation of the OPS system. The analysis of the results leads to the conclusion that the negative bending moments are the force constraint to verify the real advantage of the construction system OPS. As it was verified for all the construction phases, when using the conventional cantilevered construction system quite high negative bending moments occur when the segments are executed without the pair of tie risers of the corresponding phase; in fact these values are more than double of the corresponding values produced by the highway overload, and still are several times higher than the corresponding values produced for the OPS construction system. It is found that the bending moments in the in-service phase with the highway overload have more approximate values (presenting lower variations along the bridge deck) in the construction with OPS system, than with the conventional cantilevered construction system. ACKNOWLEDGEMENTS The first co-author deeply and publicly acknowledges and thanks the thematic of his integrated master of science thesis, developed during 2013 under the supervision of Prof Pedro Pacheco. 613
Both co-authors acknowledge and express their gratitude for the generosity of colleague Prof Pedro Pacheco, in the sequel of further studies and developments associated with the inscription of the first co-author in the doctoral program PRODEC at FEUP in July 2013. Nevertheless it is still thought that decoupling would not be neither necessary nor advisable, in view of several formal PRODEC institutional contacts during summer 2013 and of final formal PhD supervision attributions allocated in extremis on 11th November 2013, but personally accepted and actively initiated during December 2013. REFERENCES Almeida, P., 2013. Cálculo de Esforços em Tabuleiros de Pontes de Tirantes Durante a Fase Construtiva. Dissertação de Mestrado, FEUP, Porto. CEN – EN1991-2, 2003. Eurocode: Actions on Structures - Part 2: Traffic loads, Comité Europeu para a Normalização (CEN). CSI Bridge, 2015. Integrated 3-D Bridge Analysis Design and Rating, Computers & Structures Inc., Structural and Earthquake Engineering Software, California – USA. Enescil, 2013. Elementos sobre a ponte Paulicéia. Brazilândia, Brasil. Pacheco, P., 1999. Pré-esforço Orgânico - Um exemplo de sistema efector. Dissertação de Doutoramento, FEUP, Porto.
BIBLIOGRAPHY Bangash, M.Y., 1999. Prototype Bridge Structures: Analysis and Design, Thomas Telford, London. Benaim, R., 208. The Design of Prestressed Concrete Bridges: concepts and principles details. Taylor & Francis, London, 2008. Brien, E.J., 1999. Bridge Deck Analysis, E & FN Spon, London. Cruz, J.S., 1997. Controlo da Fase Construtiva de Pontes Atirantadas, Dissertação de Doutoramento, IST, Lisboa. CEN – EN1992-1, 1992. Eurocódigo: Projecto de Estruturas de Betão – Parte 1: Regras gerais e regras para edifícios. Comité Europeu para a Normalização (CEN). CEN – EN1991-6, 2005. Eurocode: Actions on structures, Part 1–6: General actions, Comité Europeu para a Normalização (CEN). Gimsing, N.J., 1983. Cable Supported Bridges – Concept and design. John Wiley & Sons Ltd, Chicheste. Gimsing, N.J., Georgankis, C.T., 2012. Cable Supported Bridges: Concept and Design. WileyBlackwell, Chichester. Manterola, J., 2006. Puentes II: Apuntes Para su Diseño, cálculo y construcción. Colegio de Ingenieros de caminios, canales y puertos, Madrid. Menn, C., 1990. Prestressed Concrete Bridges. Birkhauser Verlag, Basel. Pedro, J.J., 2007. Pontes Atirantadas Mistas – Estudo do comportamento estrutural. Dissertação de Doutoramento, IST, Lisboa.
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Building the decks of the world’s largest high speed train arch bridges with movable scaffolding systems António Albuquerque Póvoas Bridge Construction Systems, Lda., Portugal
ABSTRACT: High Speed Train (HST) viaducts have very heavy concrete decks due to the live loads required for such superstructures. In Spain, in order to perform the construction of such viaducts from 2004–2005 on, very strong MSS were designed and supplied to the Construction Industry. In this article we will present their utilization building the largest HST arch bridges in the World – Viaducto del Rio Tajo (central span 324 m) and – Viaducto de Almonte (central span 384 m) – where MSS were used for the construction of the multi-span access viaducts with spans of 60 m and 45 m repectively, and were also used to build the decks over the arches.
1 OVERHEAD MOVABLE SCAFFOLDING SYSTEMS FOR HST VIADUCTS IN SPAIN 1.1 First overhead MSS used in Spain for HST viaducts – 40 m span – 368 KN/m The first modern overhead MSS used in Spain for building HST viaducts was used in 2004–2005 in Cornella Viaduct near Barcelona. It was designed to cast a concrete deck with a self weight of 368 KN/m with some heavier zones of 512 KN/m, with spans of 40 meters between piers and 36 m between supports of the MSS. The construction joint was at 1/5th of the span. 1.2 Second overhead MSS – 50 m span – 334 KN/m Another project followed in 2005 – Llobregat Viaduct (Barcelona) – with a deck dead load of 334 KN/m, with some zones weighting 377 KN/m, with a span of 50 m between piers. Distance between MSS supports for pouring the concrete was 45 m. The section was an U type with a very complex and heavy formwork. The construction joint was made at 1/4th of the span. 1.3 Third overhead MSS – 50 m span – 281 KN/m A third overhead MSS was produced and built 3 HST viaducts – Rio Portos (Orense), La Robla (León) y Pk. 206,2 (Padrón). This machine achieved in the first viaduct a performance of pouring one span per week in the last seven spans. When building the viaducts that followed, this type of performance was maintained. Distance between MSS supports for pouring the concrete was 45 m. The construction joint was made at 1/5th of the span. 1.4 Fourth overhead MSS – 55 m span – 260 KN/m A stronger overhead MSS model was made to build 3 more HST viaducts – Toxa, Martixe and Sar Viaducts with 55 m span between piers. The optimization of the deck weight and of the MSS design enabled a large step for the distance between pouring supports −54 m – what opened the door for the 60 m span HST bridges. The construction joint was made at 1/5th of the span. This machine proved that it was also possible to build 55 m spans weekly. 615
Figure 1. Tagus Viaduct.
Figure 2. Tagus Viaduct – Cross sections of the deck and of the arch.
2 TAGUS VIADUCT 2.1 The Viaduct layout — Total length – 1488 m This Viaduct has a central zone of 6 spans of 54 m over the arch of 324 m. The North access viaduct of 642 m has a first span of 45 m, 9 spans of 60 m, and one of 57 m. The South access viaduct of 522 m has a first span of 45 m, 7 spans of 60 m, and one of 57 m. 2.2 Deck section – Arch section The deck of this viaduct is a box girder type, with an upper slab 14 meters wide as in most of the Spanish HST bridges, and with a height of 3.6 m on the centerline of the box girder. The same section is used all along the Viaduct. The deck self weight was optimized to 250 KN/m. The distance between pouring supports of the MSS is 54 m on the access viaducts and 45 m on the central viaduct decks. The construction joint of the deck is made at 1/4th of the span The arch is a single box girder with a variable section starting with a width of 12.0 m and a height of 4 m, finishing with a width of 6 m and a height of 3.5 m. 2.3 Piers The piers of the bridge have an octagonal cross section, with variable width of the section, and constant height of 3.0 m. The tallest pier on the access viaducts is P9 – 71.5 m high. The smallest pier is P1 – 9.6 m high. Over the arch 5 piers are erected, being P12 and P16 about 29 meters high and P13 and P15, about 7 meters high. The central pier P14 is just 0.565 m high, because a tangent look is wanted for the connection between the deck and the arch box girder. 616
Figure 3. Tagus Viaduct – Piers.
2.4 Construction of the access viaducts The piers of each access viaduct were built using a jump form system. The access viaducts were built with an MSS from the first to the last span just before the arch. The same MSS was used for both sides, achieving performances one span per week in most of the 60 m spans. 2.5 Construction of the arch The arch was built using form travelers. Pylons were placed on top of piers P11 and P17 to hold the cables that anchored the arch girder to the pylons and piers. 2.6 Construction of the decks over the arch The decks over the arch are built by 2 MSS working almost simultaneously. One of the MSS will dismantle the nose of the other in order to perform the 2 spans before the 2 central spans simultaneously. The north MSS will pour the 2 central spans. 3 ALMONTE VIADUCT 3.1 The Viaduct layout – Total length – 996 m This Viaduct has a central zone of 2 spans of 45 m and 7 spans of 42 m over the arch of 384 m. The North access viaduct of 261 m has a first span of 36 m and 5 spans of 45 m. The South access viaduct of 351 m has a first span of 36 m and 7 spans of 45 m. 617
Figure 4. Tagus Viaduct – MSS finishing North access Viaduct deck.
Figure 5. Tagus Viaduct – Construction of the arch.
3.2 Deck section The deck of this viaduct is a box girder type, with an upper slab 14 meters wide as in most of the Spanish HST bridges, and with a height of 3.1 m on the centerline of the box girder. The same section is used all along the Viaduct. The distance between pouring supports of the MSS is 45 m on the access viaducts and 40.5 m on the central viaduct decks. The construction joint of the deck is made at 1/5th of the span. The deck weight is 256 KN/m.
3.3 Arch sections The arch is bifid type starting with 2 legs closing up while climbing to reach the union of both boxes into a single one, finishing with a width of 6.0 m and a height of 4.2 m at section 33 at the top of the arch. 618
Figure 6. Tagus Viaduct – MSS construction of deck over the arch.
Figure 7. Almonte Viaduct.
Figure 8. Almonte Viaduct – Cross section of the deck.
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Figure 9a. Almonte Viaduct – Cross sections of the arch girder – double box.
Figure 9b. Almonte Viaduct – Cross sections of the arch girder – single box.
3.4 Piers The piers of the bridge have an octagonal cross section, with variable width and variable height of the section. The tallest pier on the access viaducts is P15 – 65.1 m high. The smallest pier is P22 – 10.75 m high. Over the arch 8 piers are erected, being the tallest P7 – 35,77 m high and P14 – 37.43 m high. The piers of the central span P10 and P11, are about 0.3 m high as a tangent look is wanted for the connection between the deck and the arch box girder. Piers 6 and 15 are fully solid in their upper part due to the heavy loads brought by the suspension cables of the arch that pass through the piers and the vertical reaction of the pylons. 3.5 Construction of the access viaducts The piers of each access viaduct were built by jump form system. The access viaducts were built with an MSS from the abutment to the span just before the arch. Two equal MSS were used. The formworks for the access viaducts were replaced by different formworks for the spans over the arches. 620
Figure 10. Almonte Viaduct – Piers.
3.6 Construction of the arch The arch was built by form travelers. Pylons were placed on top of piers P16 and P15 to hold the cables that anchored the arch girder to the pylons and piers. 3.7 Construction of the decks over the arch The decks over the arch are built by 2 MSS working almost simultaneously. Each MSS will travel to the beginning of the decks over the arch (P6 and P15). Then they are launched for the first span on the same day. The decks are poured separately, one in one side, another one on the other side, in order to compensate the horizontal forces transferred to the arch and the consequent deformations. One of the MSS will dismantle the nose of the other and also its own nose in order to perform the 2 spans before the central span simultaneously. That MSS will then pour the center span, reduced to 30 meters. 4 CONCLUSIONS – The construction of such high multi-span bridges shows the importance and capacities of the overhead Movable Scaffolding Systems (MSS) for high bridges. – These machines are bringing to the construction industry solutions to build multi-span bridges with great efficiency and remarkable construction speed. 621
Figure 11. Almonte Viaduct – MSS finishing South access Viaduct deck.
Figure 12. Almonte Viaduct – Construction of the arch.
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Figure 13. Almonte Viaduct – MSS construction of deck over the arch.
– Bridges for High Speed Trains in Spain are already being designed for spans larger than 60 meters what is an important step in order to reduce the cost of some of those viaducts with expensive foundations or very high piers. – The site is cleaner by using these machines that don’t spread or leave wood parts and small pieces all over the site, a very important issue in what concerns safety and environment protection when working in so well preserved areas and natural reserves. – When comparing this method with old fashioned shoring and scaffolding systems, it is relevant to see that the machines do not need any support apart from the piers and the deck, being therefore much safer than all systems that require supports over the arches. – The construction joint at 1/4th of the span is a strong contribution for the increase of the span that can be built by these machines in one stage only, pouring the entire deck in the same day. The existing machines can nowadays build HST bridges with 70 m spans.
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The Patani Bridge (Nigeria): Innovative construction methods P. Stellati & L. Marenzi TEAM Engineering SpA, Rome, Italy
ABSTRACT: During bridge design and construction, it is always a challenge to find technical solutions that can bring to the project significant savings in construction cost and time. This report describes the innovative construction methods adopted for this bridge, and applies in particular to the construction of the pier foundation bases in water. 1 INTRODUCTION “TEAM Engineering SpA” (herein after TEAM) has performed the preliminary, final and construction designs of the Patani Bridge over the river Forcados in the Delta State of Nigeria. TEAM has also carried out the engineering and technical assistance during all the construction phases of this 849m long bridge, on behalf of the Construction Contractor (Setraco Nigeria Ltd). Site mobilisation started in November 2010 and the bridge was structurally completed at the end of 2014. The Forcados and Nun Rivers are the two main effluents of the River Niger; the bridge in subject overpasses the River Forcados in Delta State in Nigeria (Fig. 1), A similar bridge over the River Nun at Kaiama is now under construction, both structures having been designed by TEAM. This new two-lane bridge over the River Forcados is part of the dualisation of the East-West Road from Port Harcourt to Warri through Rivers, Bayelsa and Delta States (Route F103). The bridge is formed by six central spans of 95 m and four outer spans of 49.2 m and 90.3 m, two on each side; the total bridge length is 849.0 m (Fig. 2).
Figure 1.
Geographical location map and bridge view from Patani.
Figure 2.
Schematic layout of this ten span bridge.
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Figure 3.
Figure 5.
Precast of deck segments.
Figure 4.
Deck erection.
Bridge geotechnical profile.
The seven central piers have double legs and are structurally continuous with the bridge deck, while the two outer piers, still with double legs, are connected to the deck via bearings. The bridge deck has a single concrete box section, a width of 11.0 m and height ranging between 3.0 m and 4.9 m (Fig. 3). The deck is formed by precast, pre-stressed, concrete segments erected by a launching girder with the balanced cantilever method (Fig. 4). The subsoil of the area crossed by the alignment of the road is made up of geologically recent Delta deposits, represented generally by clayey-silty sands and sandy-silty clays, inter-bedded with sands and gravel. The hydraulic situation and the characteristics of the subsoil, require pile foundations (Fig. 5). Each pier and abutment is founded over 16 (4 × 4) bored piles, with a diameter of 1.3 m and length of up to 50 m; permanent steel casings, with a length of up to 35 m, are necessary for the pile boring. The maximum calculated pile scour is equal to 8.0 m, for the piers in water. The river water level varies significantly at this location, from 1.9 to 11.4 m asl, due to the seasonal rains characteristic of these areas, and constrict the construction method and schedule of the foundations of the six bridge piers permanently in water.
2 FOUNDATIONS OF THE SIX PIERS IN WATER 2.1 Description of pier foundation base Here below the particular method adopted for the foundation of piers permanently in water (P3 to P8) is described. The foundation piles are built with the following nominal sequence: – insertion of the permanent steel casings, up to a length of 30–35 m (Fig. 6); – pile boring (bentonite is required for the soil stabilization of the pile bottom part not provided with steel casing); – erection of steel reinforcing cage and final concrete pour. The construction of the pier foundation base (13.2 × 13.2 × 2.0 m), for the six piers in water, is carried out using a bespoke method, now patented by TEAM. This innovative construction method 626
Figure 6.
Erection of pile steel casing.
Figure 7.
Precasting of formwork units.
Figure 8. Transportation of formwork units.
Figure 9.
Erection of formwork units.
Figure 10. All formwork units mounted onto piles.
falls within the field of reinforced concrete basements, on foundation piles, partially or totally submerged in water. It basically consists in the prefabrication of a reinforced concrete formwork of the foundation base, sub-divided into various units, mounted and connected onto the piles above water level and then lowered down into the water with a special jacking down system and pile clamping collars. 2.2 Construction sequence of one foundation base This procedure foresees the following main construction phases: a) The reinforced concrete units for the foundation base formwork are constructed off-shore with holes in positions that correspond to those of the piles already built in water (Fig. 7). b) The precast units are transported with cranes (Fig. 8) and mounted on the piles above water level, suspended on the piles with a support system (Figs. 9–10) and connected to each other with steel brackets (Fig. 11). 627
Figure 11. Temporary connection details of formwork units.
Figure 12.
Jacking down system.
Figure 13.
Figure 14.
Base before reinforcement erection.
Box filled with water at lower position.
c) The units are then permanently connected to each other with reinforced concrete cast in-situ joints. The jacking down system for the lowering of the entire formwork is then assembled in correspondence with the four corner piles (Fig. 12). The formwork lowering is carried out with a series of cycles of the jacking system. When the formwork enters the water, this will logically penetrate through the spaces between the piles and the formwork (Fig. 13). When the formwork has reached the desired level, the clamping collars that have previously been mounted round the piles, can be closed. d) The jacking lowering system can then be removed, as the support of the formwork is now guaranteed by the connection of the clamping collars. These guarantee a connection between the platform and the piles that is carried out very quickly and, above all, ensuring the feasibility of this connection in water. The joints at the piles are then made waterproof. The internal water can now be pumped out of the formwork. e) At this point, the pile casing can be cut and the concrete piles are lopped (Fig. 14). The reinforcement of the platform and the concrete casting (Fig. 15) inside the formwork is then carried out. All the six foundation bases (Fig. 16) were assembled and completed in four months, during the low water level period, from the end of January up to the end of May; thus providing a great saving in construction time and costs if compared with the traditional method of steel sheet piles. 2.3 Description of the design system In synthesis, the illustrated construction method consists of a disposable, prefabricated formwork in reinforced concrete, sub-divided into various units and connected to the foundation piles with a system of clamping collars when lowered to the design level. 628
Figure 15.
Figure 17. piles.
Concrete pour.
Support system at
Figure 16.
Figure 18.
Six foundation bases in water completed.
Jacking system during lowering down.
The procedure for the connection of the formwork to the piles foresees the phases described below. After having constructed the formwork units in the precast yard, they are mounted on piles, above water level, with the clamping collars assembled around the piles but not locked. The formwork units are then suspended on the piles with a support system shown in Figure 17. Once this operation is completed for all the units and they are connected to each other, the assembly of the jacking system and the system for lowering the formwork is carried out in correspondence with the selected operating points, the four corner piles in this instance. The lowering of the formwork is done using the jacking system for a series of cycles, as shown in Figure 18. When the formwork has reached the desired level, the clamping collars, that have previously been mounted round the piles, can be closed. The lowering jacking system can be removed after tightening the pile collars, as the support of the formwork is now guaranteed by the friction connection of the collars. The collars guarantee a connection between the platform and the piles that is carried out very quickly and, above all, ensures the feasibility of this connection in water. 629
Figure 19. Pier foundations with bases above water level.
Figure 20. Tie back system at the two outer piers, structurally connected to the deck.
After waterproofing the piles/formwork joints, the water can be pumped out of the formwork and, as a consequence, the clamps have to take the upward hydraulic pressure. At this point, the pile cut-off can be done in order to proceed with the installation of the reinforcing bars and concrete pour. The entire cycle described above can be applied to another special case: that in which the platform remains totally submerged in water after construction. This is possible thanks to an additional operation, consisting in the use and successive removal of a temporary formwork only after casting the concrete for the platform base and all the parts that, at the end of the operations, will remain under water. 2.4 Comments on this construction method The current technology for the construction of reinforced concrete platforms in water, mainly foresees the use of steel sheet piles of the “Larssen” type: they are driven around the design area and then the water is pumped out until the level is lower than that required to construct the platform. This method requires rather lengthy execution times and high construction costs. In comparison with the traditional method, this system is simple, practical and reliable to execute and construct, offering many advantages, both in economical terms and construction time, especially in those cases in which the platform must be partially or totally submerged in water (obviously not beyond a certain limit). The alternative of having foundation bases above water level (Fig. 19) is not a good design option for the following reasons: – Hydraulic (floating transported items remaining trapped in between the piles) – Aesthetic (bad visual impression) – Structural (reduced ship impact load capacity and corrosion in splash zone of top piles)
3 OTHER CONSTRUCTION PECULIARITIES OF THIS BRIDGE 3.1 Tie-back system at outer piers structurally connected to the deck During the design phase it was decided to apply the tie back system at the two outer piers of the seven central piers structurally connected to the deck; in other words, these two piers (P2 and P8) were pulled backwards at the start of the corresponding deck cantilever with a tie back horizontal force of approximately 2500 kN (Fig. 20), in order to achieve a pull-back displacement of approximately 40 mm. The tie back loads applied at these two piers have to be released after completion of the bridge deck between piers P2 and P8. The removal of the temporary tie back forces will generate a compression force in the deck between piers P2 and P8 and the given pull-back displacement shall 630
Figure 21. Removal of sand jacks after connection of 1st and 2nd deck cantilevers.
Figure 22. clamps.
FE model of foundation box and
remain locked in. The initial pull back displacement at these two piers is in the opposite direction to the deck shortening and partially compensates its shortening caused by concrete creep, shrinkage and deck prestressing forces. The tie-back system is only carried out on the two outer piers P2 & P8, that are subject to the greatest effect caused by deck shortening and has allowed to keep the scheme of the piers structurally connected to the deck, instead of deck onto bearings. 3.2 Sand jacks system at outer piers with permanent bearings Sand jacks are certainly not new construction devices, however these days Contractors prefer to use commercial hydraulic jacks which can be procured with less hassle. The two outer deck cantilevers are supported on permanent bearings, two for each pier; as a consequence temporary deck restraints are required during the balanced deck segments erection, in order to held the unbalanced moment generated during the most critical erection stages. In this case sand jacks and tie rods have been employed on the two sides of the permanent bearings (Fig. 21). The sand jacks are formed by a steel pipe (1000 mm dia.) stiffened by four annular rings and with a top steel plate, 30 mm thick, working as a piston. The sand jacks are removed after connection of the deck cantilevers; they have proved to be very efficient and cheaper than the alternative option with hydraulic jacks, especially in Nigeria, where the supply of special equipment is always complicated by the customs time and cost. 4 GENERAL ANALYSIS CONSIDERATIONS The structural check of the temporary systems has been confirmed, after preliminary hand calculations always recommended, with finite element analysis (SAP2000). The foundation box has been modeled with shell elements with variable support geometry and loads, in accordance with the different temporary erection conditions, as clarified below, in chronological construction order: a) b) c) d) e)
Box supported at the 16 pile supports. Box supported at the 4 corner pile supports. Installation of the pile clamps and removal of the 4 corner pile supports (jacking frames). Dewatering of the box and consequent hydraulic upward pressure. Pour of concrete inside the box with consequent downward gravity forces. 631
Figure 23.
Double leg pier structurally connected to the deck with foundation base permanently in water.
The above finite analysis is nonlinear for the activation and deactivation of elements and temporary supports at the above mentioned erection steps. The special element of the pile clamping collars have been also analyzed with a finite element model, with particular attention to the friction contact interaction between collar and pile (Fig. 22). 5 CONCLUSIONS There is always a challenge to seek innovations in construction methods and the challenge is greater as the scale of the projects increases. It is probably the acceptance of this challenge that motivates those in the business and contributes to the satisfaction that working on these projects brings. A key factor for a successful project is the mutual trust and honest collaboration between the Designer (TEAM) and the Construction Contractor (Setraco Nigeria Ltd.) and in this case, this synergy worked particularly well. The Authors of this article wish to give special thanks to Eng. Raed Saliba (Project Manager of Setraco). REFERENCES O. Belluzzi. Scienza delle Costruzioni. J.E. Bowles. Foundation Analysis and Design. C. Viggiani. Foundation. K. Terzaghi & R.B. Peck. Soil Mechanics in Engineering Practice. V.N.S. Murthy. Geotechnical Engineering. V.N.S. Murthy. Advanced Foundation Engineering. W.K Elson. Design of laterally-loaded piles. PhD CEng MICE, CIRIA publication n.103. R. Butterfield & R.A. Douglas. Flexibility coefficients for design of piles and pile groups. CIRIA publication n.108. M. Tomlinson & J. Woodward. Pile Design and Construction Practice. 5th Edition.
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Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
Innovative spliced girder method for multi span bridges I.Z. Stern Y.D.E. Engineers Ltd.
ABSTRACT: Spliced girder bridges are the most competitive solution for highway and railway bridges with spans ranging from 40 to 90 meters. In addition, the high span length to depth ratio leads to better looking bridges. The cost and the material quantity savings average about 60% when compared to the alternative of a segmental balanced cantilever bridge. The Histadrut Bridge runs over a wide intersection as well as a railway. The entire area is congested with underground utilities, leaving no room for either pier foundations or temporary towers. Although the Spliced Girder method usually requires temporary supporting towers, an innovative construction technology has been developed to avoid the need for temporary towers. This bridge is a continuous 400 meter long spliced girder bridge, composed of precast channel girders 2.5 meters deep with 72 meter spans. In order to avoid temporary shoring, the following static scheme was used for the connection between the superstructure and the pier: The pier girder segments were monolithically connected to the pier cap with vertical posttensioned bars. In addition, the transversal pier cap beam was pre-stressed for both bending and large torsional moments (36000 KN · m) created due to the lack of temporary shoring. This bridge was a Design Build project and saw a reduction of materials of 40–50% compared to the segmental balanced cantilever bridge proposed by the owner. This paper will also discuss other spliced-girder multi span bridges, a 2100 meter long railway bridge and a 320 meter long curved highway bridge.
1 INTRODUCTION Spliced Girder Bridges became the best solution for multi span bridges with spans ranging from 40 to 90 meters. The advantages of this method stem from the fact that spliced girders treat self-weight and live loads as continuous in exactly the same way that cast-in-place bridges do. In addition, this method allows for ease of construction and better quality control for bridges composed of precast elements. The result is maximum span to depth ratio, much better looking bridges and savings from the height of the approach embankment. Most importantly this method seriously reduces construction materials, and by extension the cost of the project, as can be seen in Table 1. 1.1 Construction method concept The basic idea behind the concept of spliced precast girders is to place a splice in the low moment zone as shown in Figure 1. This redistributes and ultimately reduces the maximum moment in the beams. The erection sequence shown in Figure 1 can be altered for various conditions including the bridge’s number of spans, span length, and other site limitations (see Figure 3). This will be discussed in greater detail later on. 633
Table 1. Cost comparison of the Histadrut Bridge design with a tender design of a cantilever segmental bridge. Construction Materials
Units
Tender design
Final design
Saving
Total volume of foundation concrete piles Total volume of above ground concrete Mild reinforcement Posttensioning tendons and bars
m3 m3 ton ton
9075 21200 4240 680
5335 12260 2460 312
41% 42% 42% 54%
Figure 1.
3 stages of construction for Spliced Girders.
Figure 2 shows the moment distributions for various construction method alternatives. As can be seen the moments due to self-weight for a spliced-girder are about half of those in other precast element methods. As a result the beam section itself can be reduced resulting in still lower selfweight moments. This effect is amplified as the span length increases. 2 HISTADRUT BRIDGE The Histadrut Bridge is located in an area congested with underground utilities. To meet these challenging site conditions an innovative and improved spliced girder method was developed. Due to the lack of room for the temporary shoring footing at the Histadrut Bridge, the cantilever pier girder segment was placed over the transversal pier cap beam with a moment connection 634
created by four 70 mm diameter, vertical Freyssinet HTSR bars each pretensioned to 350 tons. (See Figures 4 and 6). This system accomplishes several goals: 1. By placing an exterior drop-in-girder segment, the moment that results in the connection prevents overturning in the cantilever pier. This, in turn, makes temporary shoring superfluous. 2. The moment connection of the girders to the pier cap beam and the pier itself create a static frame system for horizontal loads. It reduces the elastic moment in the pier, and allows for the use of a reduction factor on the moment in the pier due to seismic loads. 3. A standard pier cap comprises of longitudinal girders sitting on top of bearings that are attached to the pier cap. A monolithic pier cap is one that integrates the longitudinal girders into its head by pouring concrete in between the girders and then putting posttensioned tendons through the pier cap perpendicular to the girders. The pier cap section that results is much shallower. This reduced height allows maintaining the bridge’s top elevation while also keeping the same clearance beneath the bridge.
2.1 Pier cap beam design Due to site conditions the pier and the footing had to be placed at the center of the 33 meter wide bridge, this required a unique and advanced design for the pier cap beam. The moment connection between the pier’s longitudinal segment and the pier cap beam is designed to transfer 18,000 kN·m. The pier cap beam with a cantilever length of 12 meters was designed for a combination of 135,000 kN·m bending moment, 36,000 kN·m torsion moment and 22,500 kN shear force. 2.2 The bending and torsional moments design The pier cap beam was built in two stages. The first stage includes the bottom part of the cap beam, below the longitudinal girders. This part is intended to carry the weight of the pier girders during construction. The second stage takes affect after the pier girders’ erection. It involves pouring cast-in-place concrete between the girders and the posttensioned transverse pier cap tendons. The post-tensioned tendons provide the longitudinal reinforcement required to resist bending and torsional moments. The vertical post-tensioned bars provided assist in resisting the moments developed in the connection between the longitudinal girders and pier cap beam, and also provide shear reinforcement. Together with the longitudinal post-tensioned tendons and bottom mild reinforcement they also create closed torsional reinforcement. (See Figures 5 and 6).
3 2100 METERS LONG HAEMEK RAILWAY BRIDGE SECTION A1 The typical span for this bridge is 25 meters. Section A1 consists of one additional span that runs over a highway and has a width of 52 meters. This project was designed and constructed within the framework of design-build. The tender design called for simply supported beams at each span, which translates into about 460 bearings, 80 expansion joints and three expensive and complicated rail expansion joints needed for flexible bridges in case a train must come to an emergency stop atop the bridge. The alternative design that we proposed, and that eventually went into construction, consists of a continuous section over 3 spans with monolithic connections at the pier, and an expansion joint with bearings to one side at each third pier. This design reduced the number of bearings to 50 and the expansion joints to 26. 635
Figure 2.
Continuous bridges composed of precast concrete elements self-weight moments.
Figure 3. Additional spliced girder configurations.
636
Figure 4a.
Histadrut Bridge longitudinal section.
Figure 4b.
Histadrut Bridge longitudinal section.
Figure 5.
Pier cap beam, elevation.
637
Figure 6.
Pier cap beam, transversal section (through longitudinal pier girder segment).
The bridge became more rigid, and an analysis of the rail/bridge interaction showed that there was no need for rail expansion joints. The long span 52 meter was solved using the spliced girder method.
4 HAEMEK RAILWAY SECTION B2 The same solution was provided for another segment of the same railway. The total bridge length for section B2 is 700 meters with typical 30 meter spans and a single span over a highway of 47 meters. The typical spans were also combined to 3 span continuous monolithic segments as was done in Section A1. Due to the rigidity of the bridge no rail expansion joint was required. The bridge’s 47 meters span was solved by spliced girder method. The typical precast girder depth of 2 meters needed to be increased by only 0.3 meters over the piers at each side of the long span. Notice in Figure 7 that regardless of the change in span length over the highway the section height stays constant throughout.
5 NAVOT BRIDGE This curved bridge is 320 meters long with a horizontal radius of 550 meters and was designed by the spliced girder method. The bridge crosses over a highway, railway and two streams at sharp angles. The bridge was constructed using 1.80 meter height U-shaped spliced girders. The bridge is located in a high earthquake risk zone. This fact facilitated the use of monolithic connections to the five interior piers. Monolithic connections are better at resisting earthquake loads and have the added benefits of cutting back on the use of bearings and increasing the durability of the connections. However, the use of monolithic bearings does require a very accurate analysis and design for the effects of creep, shrinkage, and, of course, for earthquake loads.
638
Figure 7.
Railway B2 Illustration.
Figure 8.
Railway B2 partial longitudinal section.
Figure 9.
Navot Bridge Illustration.
6 HIGHWAY 77 BRIDGE This project consists of two parallel 600 meter long bridges, currently in the design stage. The bridge consists of seven 82 m spans with varied depths of 4.95 m above the pier and 2.50 meters at the middle. The construction method is spliced girders. The girders are precast in site. The pier segment has a weight of 250 tons and the drop-in segments have a weight of 200 tons. 639
7 CONCLUSIONS The Spliced Girder Method is a very economical solution for bridges with spans of 40 to 90 meters. Furthermore, the system is very versatile and can be adjusted to an array of site conditions and limitations. Generally speaking, a long bridge consists of short spans, with longer spans being used to pass over rivers or highways. These longer spans require sections with increased depth. Utilizing the Spliced Girder Method allows for the use of the same cross section in longer spans as was used for shorter spans. Continuous and monolithic connections should be used as often as possible for long bridges with interior expansion joints for creep, shrinkage, and temperature effects.
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Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
Innovative formwork systems in bridge construction – Case studies A. Preuer, M. Kamleithner, M. Mihal & C. Beer Doka GmbH, Austria
More and more complex bridge structures are typical of the architecture of our times. All over the world. Doka technicians work closely with the client’s planners to put together the most suitable solutions, exactly tailored to each individual situation and to the requirements of the structure. Doka helps you decide which is the right system for you – and for your success on the site. Customtailored, versatible, and thus efficient. Doka’s modular system, based on its decades of experience of hugely diverse construction tasks in many different countries, means that it can deliver the right solution for the entire spectrum of architecture. Doka has the biggest and most flexible range of offerings on the market – anywhere in the world. What all Doka systems have in common is this: The very highest standard of safety, combined with extremely high cost effectiveness and easy, safe handling. Doka knows what demands are made, and with ist modular system it can always offer the ideal equipment.
1 TERCER PUENTE SOBRE EL RIO ORINOCO (BRIDGE NO. 3 OVER RIVER ORINOCO), CAICARA DEL ORINOCO, VENEZUELA: AUTOMATIC CLIMBING FORMWORK FOR THE 2 PYLONS 1.1 Diamonds for the Orinoco Taking a road and rail link across one of the biggest rivers in South America, together with its swamps and flood-plain, calls for a new crossing of truly superlative dimensions. Two 135.5 m pylons for the third bridge across the Orinoco River in Venezuela are taking shape with a formwork solution and automatic climbing technology from Doka. The Venezuelan government is investing in this showcase project at Caicara del Orinoco. The bridge will have an overall length of 11.125 km on completion, which is scheduled for 2016. The main bridge is 2.28 km long, and the roadway is 55 m above the water-level of the Orinoco. The two identical diamond-shaped pylons of the 360 m main span will be the centrepiece of this high-capacity road and rail link. The formwork solution for both the North and South Pylon comes from Doka. When the contractors Odebrecht first approached Doka Brasil, the forming operations for the South Pylon had already commenced. By this time, however, it had become clear that the local crane-jumped formwork originally fielded here was no longer up to the task. Following initial discussions with Odebrecht, Doka developed a solution based on the crane-independent Automatic climbing formwork SKE100. This had many advantages: “The formwork solution presented by Doka convinced the client in terms of both safety and cost-efficiency, as well as speedy construction progress”. 1.2 Flexible formwork solution needed The 135.5 m tall pylons are basically identical, but with certain minor design differences. They incline at an 18◦ angle below the cross-beam, and at 13◦ above it. The most difficult aspects to 641
Figure 1.
Regular section and Plan view.
Figure 2. Automatic climbing formwork.
deal with in the planning work were the changes in the cross-section, the catwalk between the legs of each pylon, and incorporating an extra suspended platform to provide access to the passenger hoist. With significant changes in cross-section between the foundations and the cross-beam, and a continued upward taper until the point where the pylon-legs meet, the structure geometry demands a great deal of adaptability from the formwork systems. Automatic climbing formwork SKE100 and the versatile Large-area formwork Top 50 system meet these requirements without needing time-consuming adaptation work. The load capacity of 10 t per bracket allows work to proceed simultaneously on several levels. An extra work-deck level has been provided above the pouring platform, for mounting the reinforcing cages. This allows the forming and reinforcing operations to run in parallel.
1.3 Safety writ large A protection screen has been deployed to ensure workplace safety. The working platform and Level +1 are enclosed with trapezoidal sheeting, and all other levels safeguarded by standard handrails. A catwalk is the main route for site-traffic between the legs of each pylon. It is attached to the automatic climbers and is raised along with them up to the next section. An extra – third – telescopic suspended platform provides access to the passenger hoist. 642
Figure 3.
South pylon completed end 2014.
1.4 On-site technical expertise After an approx. five-month planning stage, work on pre-assembling the SKE100 units began in September 2011. Doka took over the forming operations on the South Pylon from the 17th casting step onward; on the North Pylon, Doka automatic climbing technology was used right from the outset. Two sets of Automatic climbing formwork SKE100 with 54 units in total, a catwalk and Large-area formwork Top 50 are in use here. Doka supplied all the systems for on-site pre-assembly and installation. In the crucial assembly phase, Doka project manager and international Formwork Instructors were on the site. They gave professional introductory training to the site crew and made sure that the formwork systems were properly assembled, installed and handled. A Formwork Instructor has been on-site ever since, to assist the project team with his technical expertise. After seven cycles on one pylon and eight on the other, the Doka automatic climbers had climbed to heights of 65.7 and 33.8 m respectively by the beginning of 2013. The Odebrecht company, which has worked closely with Doka on projects in Brazil and other countries, is very satisfied with its choice of self-climbing solution. “Doka convinced us with a well thought-out formwork solution. Both the all-in package and the level of detail in the offer documents were truly impressive. The help the Doka Formwork Experts gave us with on-site assembly laid down an important basis for the project to progress smoothly”, notes Vicente Rodrígues from Odebrecht. 1.5 In brief Project: Third bridge over the Orinoco River (Tercer puente sobre el Río Orinoco) Location: Caicara del Orinoco, Venezuela Contractor: Odebrecht, Venezuela Start of construction: 2007, pylons: 2011 643
Figure 4. Two pairs of Cantilever forming travellers in operation.
Completion scheduled for: 2016 Start of planning work at Doka: March 2011 First formwork deliveries: September 2011 Type of structure: Diamond shape pylons Structure height: 135.5 m Total length of bridge: 11.125 km Length of cable stayed bridge: 2.28 km Number of spans: 17 Bridge width: 21 m Systems in use: Products: Automatic climbing formwork SKE100, Large-area formwork Top50 Services: Engineering, Formwork Instructors, on-site engineering support 2 RV. 80 LODING-VIKAN, TVERLANDSBRUA, NORWAY: DOKA CANTILEVER FORMING TRAVELLER FOR SUPERSTRUCTURE CONSTRUCTION 2.1 High seas in the Norwegian Arctic Tailor-made Doka formwork solutions are ensuring swift, safe progress during construction of the Tverlandsbrua bridge between Løding and Vikan in the Nordland region of northern Norway. Among the unusual challenges of this project are the high seas encountered out on the fjord, with waves up to 10 m high putting the technicians and the site team severely to the test. The Norwegian construction firm Reinertsen is erecting a new bridge between Løding and Vikan in northern Norway, for the Rv 80 trunk road. This Trondheim-based company decided to work with Doka because of the positive experience it had had on previous projects: In fact, the extensive advisory and planning services and high safety standard provided by Doka were crucial factors underlying the decision. The Rv 80 road-building project aims to link the Tverland peninsula better to the regional capital of Bodø – some 1200 km north of Oslo. 2.2 Doka Cantilever forming traveller The Tverlandsbrua is 670 m long, has a longitudinal gradient of as much as 4.5%, and is being built in the form of a spiral transition curve with varying radii. It crosses the Saltfjord between Løding and Vikan, in seven spans borne by six piers of up to 13 m in height – three of them are twins. Three main spans max. 180 m, double box girders with inclined webs, were built by the balanced cantilever method. Because of the complexity of the superstructure geometry (longitudinal and transversal gradient, spiral transition curve, double box with inclined webs), a strong interaction between bridge designer and supplier of forming traveller was required. The high safety demands need special procedures for launching and checking the main suspension tie roads (simulating the fresh concrete weight) before every casting. Doka has four of its Cantilever forming travellers 644
Figure 5.
Cross section of superstructure with CFT and formwork.
Figure 6.
Longitudinal section of superstructure with CFT and formwork, unbalanced start at pierhead.
(CFTs) in action at this site. Because of the large – 22.6 m – width of the superstructure deck, they are built with three longitudinal trusses, from rentable components. The slanted webs made it necessary to widen the bottom formwork; this, in turn, meant that the bottom grid – including the platforms – had to be slide-mounted. In this way, the site crew can adapt the bottom formwork to the next segment quickly and safely. The modular design concept of the CFTs also makes for efficient, cost-saving adaptability to the different cross-sections of the superstructure deck. 2.3 Pierhead construction Beams rested on horizontally mounted Doka supporting construction frames, enabling them to sustain the load from the widely cantilevering pier-heads. The pier-heads were poured in three casting steps: Large-area formwork Top50 was used to form the bottom slab, Large-area formwork Top50 and Wall formwork FF20 to form the webs, and the Bridge formwork ParaTop, Largearea formwork Top50 and Load-bearing tower Staxo systems to form the deck slab. Using ParaTop meant that no shoring was needed for the cantilever arm formwork – which also kept the equipment 645
Figure 7.
Supporting construction and formwork for pierhead.
commissioning-quantity small. A particular challenge here was that it had to be possible to lower the screwjack mechanisms after the first casting segment, so that the bottom slab could carry its own weight: this meant that when the webs were poured, it was already possible to transfer the load into the piers by way of the bottom slab. 2.4 In brief Project: RV. 80 Loding-Vikan, Tverlandsbrua, Norway Location: Norway Contractor: Reinertsen, Norway Start of construction: April 2011 Completion scheduled for End of 2013 Total length: 670 m Number of spans: 7 Bridge width: 22,6 m Systems fielded: Large-area formwork Top50, Wall formwork FF20, Bridge formwork ParaTop, Load-bearing tower Staxo 100, Staxo stair towers, ‘Universal’ supporting construction frames 3 VIADUCT NUTTLAR, NORDRHEIN-WESTFALEN, GERMANY: UNDERSLUNG FORMING CARRIAGES FOR DRIVE DECK CONSTRUCTION 3.1 Superstructure of the Nuttlar steel composite bridge: Formwork engineering at a high level Four identical underslung composite forming carriages are casting the superstructure of the Nuttlar viaduct in Germany’s hilly Sauerland region. The bridge is part of the new section of the A 46 autobahn between the Bestwig-Velmede and Nuttlar junctions. When finished, it will be the highest viaduct in North Rhine-Westphalia. 646
Figure 8.
Four units of underslung forming carriages in operation.
660 metres long and with a radius of 1000 m, the single-cell steel composite bridge spans the river valley of the Schlebornbach. Its superstructure is a closed steel box section 6 m high, with external angled struts for bracing and a steel-composite longitudinal girder. The roadway slab, 28.6 m wide and with a constant 4% sloping pitch, is being cast in place in a back-stepping process. The centre sections are cast first as the primaries, so that the load pattern always remains the same as on completion. The secondary steps above the bridge piers are then cast, closing the structure. Consequently, the formwork system has to be movable along the entire length of the bridge – preferably without the need for crane time. Another requirement is that the forming-carriage structure should have no more than a minimal effect on the bridge’s load-bearing capability. Over-heavy and rigid superstructure formwork would inevitably impose unnecessary constraints. 3.2 Uninterrupted progress The underslung composite forming carriages traverse at a height of 115 m above the valley floor. They are designed for the two slabs, each some 9 m wide and both cantilevering over the bridge edges and propped in thirds. The roadway slab is 40 cm thick in the infield, thinning down to 25 cm at the outer edges of the cantilever arms. The formwork is anchorless, so there are no holes to be filled in the roadway slab. Firstly, because the project owner included this specification in the design brief. Secondly, because this ensures uninterrupted work topsides for the crews placing the reinforcement, pouring and smoothing the concrete. The composite forming carriages introduce their deadweight and the live concreting loads into the steel-composite longitudinal girders of the cantilever slabs. Heavy-duty roller assemblies carry the carriages on the box girder. The steel-composite longitudinal girders are of limited load-bearing capacity, so the engineers were faced with the challenge of coming up with a high-strength, weightoptimised design. Concreting lengths vary from 19.2 m to 25 m, so the superstructure will be finished in 28 one-week cycles. 3.3 Site infills are of minor significance The engineering solution consists of a kinematic system involving the interaction of carriage level, formwork level, and lifting and lowering unit. Having the carriage defined as a level in its own right means that all bearing forces are introduced optimally into the bridge’s steel structure. Formwork panels adapted to the structure keep the set-up process moving ahead rapidly, because site infills are no more than minor. And the composite forming carriages, moreover, evince a unified and all-inclusive safety concept across all four platform levels, with trapdoors and integrated ladders for access. 647
Figure 9. (right).
Underslung forming carriages ready for casting (left), stripped formwork ready for launching
While the C35/45 concrete is being poured, the composite forming carriages are suspended in SL-1 beam clamps. Parts of the formwork are swung inboard hydraulically by a rocker beam, so that the carriage can advance past the angled struts of the bridge’s structure. Each carriage has six synchronised hydraulic cylinders, vastly simplifying and facilitating work throughout the setup and stripping procedures. An hydraulic stepper mechanism moves each composite forming carriage on its steel rollers. 3.4 Straightforward assembly on site Contractors Max Bögl deployed site personnel for assembly and disassembly of the composite forming carriages. A Doka site foreman assists with in-project advice. Lead engineer Ulrich Rödel has this to say: “To my mind the formwork concepts for the automatic climbers for the bridge piers and for the underslung composite forming carriages are totally convincing.” 3.5 In brief Project: Roadway slab, Nuttlar autobahn viaduct Location: Germany Contractor: Max Bögl Bauunternehmung GmbH & Co. KG, Germany Start of construction: 2010 Completion scheduled for 2014 Height above valley floor: 115 m Total length: 660 m Number of spans: 7 Structural height: 6 m Roadway slab width: 28.6 m Cantilever arm width: approx. 9 m Concreting lengths: 19.2 m to 25 m Number of one-week cycles: 28 Radius: 1000 m Constant slope pitch: 4% Formwork on site: 4 underslung Doka composite forming carriages
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Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
Prestressed I-beams made of ultra-high performance concrete for construction of railway bridges ˇ P. Tej, J. Kolísko, P. Bouška, M. Vokáˇc & J. Cech Klokner Institute, Czech Technical University in Prague
ABSTRACT: This paper focuses on research into prestressed I-beams made of ultra-high performance concrete, which are designed to be structural elements in small and medium span railway bridges. Prestressed concrete I-beams are designed with ten prestressing cables in the bottom flange. The prestressed beams are laid close together in the actual structure with panels inserted between them. The entire structure will subsequently become monolithic. At the present time, I-beams made of rolled steel are commonly used as structural elements in this type of structure. The advantage of these types of structures lies in their having a low construction height. This paper presents a computer and experimental analysis of the loading of UHPC prestressed I-beams. For the purpose of the experiments, three specimens of 7 m span and two specimens of 12 m span were made. The specimens were subsequently tested in the laboratory in four-point bending tests. The paper presents the process and results of the experiments and a comparison of failure modes of the beams of different spans. Simultaneously with the experiments, computer analyses were created in which optimization of the material and geometric parameters of the beams were carried out. The paper demonstrates the correspondence of the experimental and computer-simulated load test results.
1 EXPERIMENT The subjects of the testing were prestressed concrete I-beams of class C110/130 UHPC with steel fibres. The span of the beams is a) 7 m and b) 12 m, with a cross-section height of 400 mm. Prestressing is implemented using straight ten cables in the lower flange. Five of them are at a distance of 1.3 m from the supports stored in the protectors in order to prevent excess of the stress
Figure 1. Scheme of the loading tests of a) 7 m span, b) 12 m span UHPC prestressed I-beams and c) their cross section.
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Figure 2, 3. The development of shear cracks during the loading tests of 7 m span UHPC prestressed I-beams.
650
Figure 4, 5.
Setup of the experiment, deflection and bending failure of 12 m span UHPC prestressed I-beams.
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Figure 6. Computer model – Isosurfaces of crack width (load bearing capacity to a value of the force 2 × 151 kN, deflection 41 mm).
Figure 7. Computer model – Isosurfaces of crack width (load bearing capacity to the value of the force 2 × 150 kN, deflection 173 mm).
at the ends of the beam. One straight cable leads in the top flange. The cables are tensioned by the force of 206.25 kN, ie. stress in each cable is 1375 MPa. The strength of the cables is 1570/1770 MPa, the diameter is 15.7 mm. Prestressed concrete I-beams were tested in 4-point bending tests (refer with: Fig. 1). 2 COMPUTER ANALYSIS A computer model of an UHPC prestressed I-beam was created in a 3D environment using GiD 11 software, whereby the calculation was prepared. The calculation itself was carried out with ATENA Win Statics software. The computer model of the specimen was modelled as a half of a symmetric structure by means of one macro-element. The prestressing was modelled using lines elements with assigned profiles of 15.7 mm according to the actual structure. The surface of the symmetry of the model is prevented from rotating about a horizontal axis (the axis perpendicular to the span) and the 652
Figure 8.
Comparison of L-D diagrams of laboratory test and FEM analysis.
Figure 9.
Comparison of L-D diagrams of laboratory test and FEM analysis.
horizontal displacement perpendicular to it. Steel spreading elements were added for transferring the load to the model. The loading was performed by displacement per 0.1 mm. The displacement and the corresponding force were monitored. In addition to this monitoring the volume of concrete was monitored for checking the crack width. The model was meshed by hexahedra elements of a size of 0.05 m.
3 CONCLUSIONS The UHPC prestressed I-beam of a span of 7 m failed during the experiment in shear at value of the force 2 × 161/212/158 kN with a deflection of 50/60/56 mm in the middle of the span (refer with: Fig. 2 and 3). The numerical analysis showed a very similar load bearing capacity to a value of the force 2 × 151 kN with a 41 mm deflection. Isosurfaces of the cracks width at the maximum 653
force are shown in the following figure (refer with: Fig. 6). Figure 8 shows a comparison of the L-D curves of the experiment and the FEM analysis. It shows a very good congruence between laboratory tests and the FEM analysis. The UHPC prestressed I-beam of a span of 12 m failed during the experiment in bending at value of the force 2 × 150/138 kN with a deflection of 173/167 mm in the middle of the span (refer with: Fig. 4 and 5). The numerical analysis showed a very similar load bearing capacity to the value of the force 2 × 129 kN with a 200 mm deflection. Isosurfaces of the cracks width at the maximum force are shown in the following figure (with refer: Fig. 7). Figure 9 shows a comparison of the L-D curves of the experiment and the FEM analysis. It shows a very good congruence between the laboratory tests and the FEM analysis. ACKNOWLEDGEMENT This research is supported by grant GACR 104/15-22670S Experimental and numerical analysis of bond behaviour between steel reinforcement and ultra-high performance concrete. REFERENCES AFGC/SETRA, 2013. Bétons fibrés à ultra-hautes performances, Recommandations. Documents scientifiques et techniques, Association Française de Génie Civil, Setra. J. Kolísko, M. Rydval, P. Huˇnka, 2013. UHPC – Assessing the Distribution of the Steel Fibre and Homogeneity of the Matrix, Tel Aviv, Israel, fib Symposium, Tel Aviv. ˇ ˇ V. Cervenka, J. Cervenka and R. Pukl, 2002. ATENA – A tool for engineering analysis of fracture in concrete, Sadhana 27/4.
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Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
Preliminary assessment of wind actions in large span MSS A. Resende & H. Coelho BERD SA, Oporto, Portugal
P. Pacheco BERD SA, Oporto, Portugal & FEUP, Oporto, Portugal
ABSTRACT: Historically the use of Movable Scaffolding Systems (MSS) in bridge deck construction was restricted to the range of spans approximately between 30 m and 60 m. In the recent years, the use of MSS expanded to increasingly larger spans. These applications were made possible not only by the progress of structural analysis techniques and computational power but also the use of innovative structural solutions, such as an actively controlled prestressing system (Organic Prestressing System – OPS). However, as the span of the bridge deck increases, the span of the MSS increases as well. The MSS structure results much more slender and thus more vulnerable to Wind Action. As the design approach for wind actions in MSS is not included in the Wind Action Codes, it is necessary to create a specific set of rules. Even if basic specifications can be adapted from the rules for general metallic structures, some of the procedures have to be explicitly developed for this type of structure with very particular characteristics: changing support conditions, changing span distribution, changing in mass and mass distribution, changing location. Due to this variability, the design of MSS is divided in several operations and it is attributed a wind velocity for each of the operations. This paper gives an overview of the probability of occurrence of wind velocities chosen for each operation. 1 INTRODUCTION The assessment of Wind Actions and the reliability of structures subjected to them has received growing interest during the last few decades mainly due to the increasing boldness and complexity of the structures. The designers must, as always, assure that the Wind Action evaluation is accurate and adequate because boldness in structural design is often legitimated by a better evaluation of the loads. This increased knowledge on loads is clear even in the standards evolution. It has been observed that new standards are increasingly not only more detailed and complex but also more accurate and closer to reality. Although some load types acting on structures are almost invariable considering the geographical location, there are a few actions that are highly dependent on environmental loads (wind loads and snow loads) and also seismic loads. The construction of prestressed concrete bridge decks with Movable Scaffolding Systems (MSS) (Fig. 1), a tridimensional lattice steel structure that supports the formwork used to construct one entire span of the bridge deck that additionally has the ability to self-launch between adjacent spans, is normally used for a 40–60 m span range. Until the last few years, bridges with 70–90 m were normally constructed with precast solutions, metallic solutions or cantilever method. However, over the last few years, experiences have been made and new solutions have been developed for the 70–90 m span range (LMSS – Large Movable Scaffolding Systems) (Pacheco, 2011). BERD is currently developing a 90 m span LMSS (designated as M1-90-S). The M1-90-S is programmed for the partial construction of 4 viaducts in Turkey with a maximum span of 90 m. The 655
Figure 1.
MSS for the construction of a 70 m span bridge in Slovakia.
estimated time for execution is around 3 years including all assemblies and disassemblies. During this project the M1-90-S will operate at around 75 m above the ground level. Unlike a permanent structure, a LMSS has to undergo several operations with very different characteristics, namely: – Static phase (usually called Concreting): suspension of the deck concrete weight while the deck is not self-supporting. The concrete weight is substantially bigger than the LMSS weight (in this case a ratio around 2.45). During this phase it is very important that the LMSS deformation is kept small in order to achieve a bridge deck with the desired geometry; – Movable phase (usually called Launching): movement of the LMSS between adjacent spans facing an evolving structural system (the LMSS moves above supports that run through the entire main girder). During the movable phase the LMSS is more vulnerable to wind because fixation is limited especially in the transversal direction. The disparity between operations that the LMSS is subjected usually leads to the consideration of different velocities for each stage (SEOPAN, 2007). In this paper the Wind Action considered in the M1-90-S design is analyzed and is calculated the probability of occurrence of wind velocities based on statistical data available for Ankara, Turkey.
2 EXTREME VALUE DISTRIBUTION The Extreme Value analysis of wind velocity and other geophysical variables, like floods or even seismic acceleration, is based on the application of at least one of the 3 Extreme Value Distributions identified by Fisher and Tippet. These 3 distributions are usually called Type I or Gumbel distribution, Type II or Frechet distribution and Type III or Weibull distribution (Bastos, 2008). This study uses the cumulative distribution function for the Type I distribution of the largest values FI (also referred as the Type I Extreme Value distribution, or the Gumbel distribution) (Simiu & Scalan, 1996):
In Equation 1, U is the extreme wind velocity for a specific period (in this paper it is considered 1 year or 1 day), µ and σ are referred to the location and the scale parameter, respectively. It can 656
be shown that the mean value of U , E(U ) and the standard deviation of U , SD(U ) are:
Equation 1 may be inverted to yield the percent point function, that is, the value U of the random variable wind velocity that corresponds to any given value of the cumulative distribution function. In the case of the Type I distribution:
A classical method approach to the problem of estimation is the method of moments (Simiu & Scalan, 1996). In this method it is assumed that the distribution parameters can be obtained by replacing the expectation and the mean square value of the random variable U by the corresponding statistics of the Sample, using Equation 2 and Equation 3. In this case:
Where σˆ and µ ˆ are estimations for σ and µ in the Equation 1–4. The values U and s are the Sample mean and the Sample standard deviation of a Sample with dimension n. From the Sample data it is possible the calculate U and s:
3 MEAN RECURRENCE INTERVAL The Mean Recurrence Interval – MRI(U), corresponding to the generic value U of a random variable is defined as the mean number of events that it is necessary to register to obtain a value bigger than U (Bastos, 2008).
If all the extreme values of wind velocity are considered independent, the probability pL that all extreme values of wind velocity do not exceed the value U during a period of L events (equally spaced in time) for the MRI(U) is:
4 WIND VELOCITY IN TURKEY (ANKARA) Since there is very few information available about the wind in Turkey, it was necessary to conduct a specific study to characterize adequately the wind velocity for the region of Ankara, the place where the LMSS will be used. 657
Table 1. Annual maximum wind gust velocity at 10 m for Ankara [m/s] (Firat, 2007). Year
Umax [m/s]
Year
Umax [m/s]
Year
Umax [m/s]
1960 1961 1962 1963 1964 1965 1966 1967 1968 1969 1970 1971
22.7 24.8 25.4 28.3 19.1 29.6 23.0 23.2 32.1 24.4 28.5 29.2
1972 1973 1974 1975 1976 1977 1978 1979 1980 1981 1982 1983
25.1 23.2 23.3 17.8 22.0 16.9 17.6 16.5 14.7 17.7 17.0 17.2
1984 1985 1986 1987 1988 1989 1990 1991 1992 1993 1994 1995
24.1 22.2 19.2 18.2 16.7 15.8 14.4 15.3 13.5 13.5 14.5 18.1
Year
Umax [m/s]
1996 1997 1998 1999 2000 2001 2002 2003 2004 2005
17.4 17.7 16.3 20.2 17.7 19.4 19.4 19.3 16.8 17.0
Figure 2. Cumulative Distributions for the Sample of Annual Maximum and for the adjusted Gumbel Distribution.
This statistical study of wind velocities with very small probability of occurrence – extreme wind velocities – was based on the annual maximum wind velocity for Ankara found on (Firat, 2007) and collected by the Turkish Meteorological Department. The data for the annual maximum wind gust velocity at 10 m above the ground for Ankara is presented in the Table 1. For the wind velocities with infrequent probability of occurrence – strong wind velocities – it is considered the daily maximum wind gust velocity for the year 2004 collected by the Turkish Meteorological Department and also available also on (Firat, 2007). It was noticed some incongruity in the cross reference between the annual maximum for the year 2004 in the Table 1 and the data from the daily maximum since there were several values of daily maximum greater than the value stated for annual maximum. However, this incompatibility is considered to do not significantly change the conclusions intended for this study. Therefore, the data considered is exactly the information by (Firat, 2007). For lack of space, the data from daily maximum it is not reproduced.
5 FITTING OF GUMBEL DISTRIBUTION TO THE SAMPLE OF ANNUAL MAXIMUM Based on the Sample of Table 1 and using Equation 5 and Equation 6:
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Figure 3. Cumulative Distributions for the Sample of 2004 Daily Maximum and for the adjusted Gumbel Distribution.
From the estimates above it is drawn in Figure 2 the Cumulative Distribution FI (U ) based on Equation 4. For comparison it is also included the Cumulative Distribution for the Sample. Comparing the Cumulative Distributions from the Sample and from the Gumbel Distribution, apparently the Gumbel Distribution is well suited to represent the Sample. Inherently from the definition of the method of moments the mean value and standard deviation from the Sample and the Gumbel Distribution are the same. The Gumbel Distribution with the annual maximum wind velocity is used to characterize the probability of extreme wind velocities, typically from the magnitude of expected to be an annual maximum. 6 FITTING OF GUMBEL DISTRIBUTION TO THE SAMPLE OF DAILY MAXIMUM Based on the Sample of the daily maximum of wind velocity for the year 2004 found on (Firat, 2007) and using Equation 5 and Equation 6:
Repeating the procedure from the previous chapter, it is drawn in Figure 3 the Cumulative Distribution FI (U ) based on Equation 4 considering the estimates above. For comparison it is also included the Cumulative Distribution for the Sample. Comparing the Cumulative Distributions from the Sample and from the Gumbel Distribution, it is clear that the Gumbel Distribution is adapted to the Sample. 7 EN 1991-1-4 RULES The Standard EN 1991-1-4 (CEN, 2005) defines the velocity profile as a function of the fundamental value of the basic wind velocity, the height above the ground and the terrain roughness (considering 5 types of normalized terrain). The fundamental value of the basic wind velocity, vb,0 , is the characteristic 10 minutes mean wind velocity, irrespective of wind direction and time of year, at 10 m above ground level in open country terrain with low vegetation such as grass and isolated obstacles with separations of at least 20 obstacle heights. Since the LMSS will operate around 75 m above the ground it is necessary to include the height effect when estimating the wind velocity acting on LMSS. The EN 1991-1-4 provides relations to 659
Table 2. EN1991-1-4 relations between peak wind velocity for 75 m and 10 m. Terrain Category
z0 [m]
Iv (75)
vp (75)/vp (10)
II
0.05
0.1367
1.2675
where z0 is the terrain roughness length, Iv is the turbulence intensity and vp (z) is the peak wind velocity.
estimate proportion between the peak wind velocities at 75 m and 10 m. Considering the terrain category II from EN1991-1-4 (CEN, 2005) in Table 2. Being a structure with particular characteristics with narrow application worldwide, the LMSS are not directly covered by traditional standards and calculation rules are usually defined by the designer from its own experience. However, as it is clear from this chapter, there are some rules in the traditional standards that provide useful information and can be used. The use of traditional standards must be cautious and should be complemented with additional studies whenever it is possible.
8 WIND ANALYSIS FOR LMSS As the LMSS is not attached to a specific place (it is intended to be used in the construction of several bridges possibly in different countries), it is difficult to establish a fundamental value for the basic wind velocity, because this value is specific to each place. For this reason, the wind velocity to be considered in the LMSS analysis is normally defined for the LMSS and not for an exact place. Furthermore, as the LMSS is not a static structure, being subjected to several structural systems and very different loads, the wind velocity considered in each operation should be adjusted. For example, in the construction of these bridges in Turkey, 3 different wind velocities are considered (60 km/h, 120 km/h and 180 km/h for 75 m above the ground). Using the Equation 1 it is possible to estimate the probability for each wind velocity to be exceeded in the reference period, 1 day for the Gumbel Distribution adjusted to the values of daily maximum of wind velocity and 1 year for the Gumbel Distribution for the annual maximum. Additionally it is possible to establish also the MRI for each one of the 3 wind velocities using the Equation 10. The probability wind velocity of 60km/h is better estimated using the data from the daily maximum since it is almost certain that the wind velocity will be greater than 60 km/h at least once each year. In this case it is important also to determine how many occurrences of wind velocities greater than 60 km/h it is expected. For the velocities of 120 km/h and 180 km/h, it is used the data from the annual maximum since the probability of one exceedance in one year is relatively small. If these wind velocities are surpassed it is very probable that it is only in one occasion. For all wind velocities it is considered that each occurrence of exceedance has the duration of one day. Considering the rules in the EN1991-1-4, it is possible to estimate the wind velocity for a height of 75 m for a terrain category II by multiplying the peak wind velocity for 10 m by 1.2675. The strong wind velocity – 60 km/h – is infrequent and is apparently a well suited for operational limit. The extreme wind velocities, 120 km/h and 180 km/h, have very low probability of occurrence. In order to establish a probability of each position of the LMSS it is estimated the amount of time that LMSS will stay in each specific position in Table 4. The difference between operations O.4, O.6 and O.5, O.7 is that in the operations O.4 and O.6 it is considered that concrete is still fluid while in operations O.5 and O.7 it is considered that concrete is not fluid. The fact that concrete is not fluid allows assuming that the wind in the formwork is not transmitted to the metallic structure of the LMSS. 660
Table 3. Probability of exceedance at 75m of the velocities 60km/h, 120km/h and 180 km/h for 3 years. 3 YEARS vp (75) [km/h]
vp (75) [m/s]
vp (10) [m/s]
MRI
Number occurrences
Probability [%]
60 120 180
16.7 33.3 50.0
13.1 26.3 39.4
15.1 days 10.2 years 364.4 years
72.3 0.293 0.0082
6.61 0.027 0.00075
Table 4. Time and probability of each MSS position. Operation
Description
Time [months]
Probability [%]
O.1 O.2 O.3 O.4 O.5 O.6 O.7
Assemblies and disassemblies Without fixations – ready for launching With fixations, without concrete With fixations, concrete 1st stage, concrete fluid With fixations, concrete 1st stage, concrete not fluid With fixations, concrete 2nd stage, concrete fluid With fixations, concrete 2nd stage, concrete not fluid
11.3 2.2 8.4 2.1 5.2 2.1 4.7
31.4 6.1 23.3 5.8 14.4 5.8 13.1
The analysis of operation O.1 is not included in the scope of this paper because it is an accessory operation (corresponds to the preparation of the equipment for bridge deck construction) and also because it comprises very different operations. In some of the operations, the more sensible ones, a limitation is usually imposed to the wind velocity in the Technical Manual of the equipment in order to guarantee the adequate conditions to accomplish the operation. This limitation must be followed by the Responsible for the LMSS operation who must decide before initiating the operation if there are adequate conditions to complete the respective operation comprising the limitation imposed in the Technical Manual at all times. In order to decide to start, the Responsible is also obliged to fill and sign a check list to engage his responsibility to start the operation. In this check list, the Responsible has to: 1) check the forecast for the period of the operation and some additional time (often 0.5–1.0 the duration of the operation to account for unpredictable situations) and 2) check the wind velocity in the anemometer of the LMSS. If any of these conditions is not verified, the Responsible must not start the sensible operation. Because these conditions are normally clear to the Responsible of the LMSS and other operators and also because the non-accomplishment of these conditions may imply major injuries or even death, the probability of non-accomplishment is normally very small. About the wind velocity forecast, as the sensible operations are generally of short duration (maximum around 1 day), the probability that the wind velocity forecast matches the wind velocity verified is relatively high. For these sensible operations it is assumed that the typical probability of forecast mismatch or Responsible bad decision to start the sensible operation without adequate conditions is in the interval [1%;5%]. Conservatively, in this paper it is adopted a value of 5% for all sensible operations where the wind velocity is limited in the Technical Manual. Assuming that the events A (the MSS is in the operation O.n), B (the wind velocity surpasses U.n, being U.n the wind velocity considered in the calculation of the operation O.n) and C (forecast mismatch or operation responsible bad decision) are independent, the probability of all events are verified: P (A ∩ B ∩ C) = P (A) × P (B) × P (C) This probability value can be seen as the probability of wind velocity considered in the LMSS calculation being surpassed. 661
Table 5. Probability of wind velocity surpasses the wind velocity used in LMSS calculation. Event A
Event B
Event C
Operation O.n
P(A) [%]
Wind velocity U [km/h]
P(B) [%]
Restriction in Technical Manual
P(C) [%]
P(A)∩P(B)∩P(C)
O.1 O.2 O.3 O.4 O.5 O.6 O.7
31.4 6.1 23.3 5.8 14.4 5.8 13.1
N/A 60 180 120 180 60 180
N/A 6.61 0.00075 0.027 0.00075 6.61 0.00075
N/A Yes No Yes No Yes No
N/A 5 100 5 100 5 100
N/A 2.0E-4 1.7E-6 7.8E-7 1.1E-6 1.9E-4 9.8E-7
As it can be seen in the Table 5, the total probability of a wind velocity not considered in the LMSS calculation is small for all operations, being relatively homogeneous for 2 groups of operations. The bigger probabilities are associated to sensible operations with wind velocity limitations in the LMSS Technical Manual. 9 CONCLUSIONS It is important to consider the indications prescribed in the standards. However, in some particular cases like the LMSS, the applicability of the standards is often difficult or even inadvisable. An additional aspect to be taken into account is that the LMSS market is worldwide and sometimes it is not possible to verify simultaneously a wide variety of regulations. In both cases, the judgment of the LMSS designer is very important to define the particular set of rules to be used in LMSS calculation. Through the analysis of the wind actions in the LMSS that is being developed for Ankara, one can verify that the probability of exceeding design wind velocities is small enough to consider that operation safety is assured and wind velocities are adequate. REFERENCES Bastos, F.J. 2008. Comportamento Aerodinâmico de Estruturas Esbeltas: Análise de Efeitos de Rajada. MSc Thesis. Porto: Faculdade de Engenharia da Universidade do Porto. CEN (Comité Européen de Normalisation). 2005. EN 1991-1-4 Eurocode 1: Actions on Structures, General Actions Part 1-4: Wind Actions. Brussels: CEN. Firat, F. K. 2007. Development of Load and Resistance Factors for Reinforced Concrete Structures in Turkey. PhD Thesis. Ankara: Middle East Technical University. Pacheco, P., Coelho H., Borges P., Guerra A. 2011. Technical Challenges of Large Movable Scaffolding Systems, Structural Engineering International (IABSE), Vol. 21, Number. 4. pp. 450–455. Roldsgaard, J. H., Kiremidjian, A., Georgakis, C. T., Faber, M. H. 2013. Preliminary probabilistic prediction of ice/snow accrestion on stay cables based on meteorological variables. 11th International conference on Structural Safety & Reliability. New York. Schneider, J. 2006. Introduction to Safety and Reliability of Structures 2ed. Zürich. International Association for Bridge and Structural Engineering (IABSE). SEOPAN – Comisión Tecnológica. 2007. Manual de diseño y uso de cimbras autolanzables. Confederacion Nacional de la Construccion (CNC). Simiu, E. & Scalan R. H. 1996. Wind Effects on Structures: Fundamentals and Applications to Design. New York: John Wiley & Sons, Inc.
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Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
Cabriel River Viaduct in Cofrentes (Valencia, Spain) bypass at N-330. Construction design J.F.M. Soriano, J.I.C. Vázquez & B.D. Santana VALTER valenciana de estructuras S.L., Valencia, Spain
ABSTRACT: The Cabriel River viaduct has an overall length of 520 m, distributed by 8 spans, 2 × 50 m spans at either end and 6 central spans each 70 m in length, conceived as a continuous tapered curved beam formed by a post-tensioned box-girder deck. Located in a natural environment characterized by abrupt orography requires piles which reach around 50 m in height. The contractor was invited by orographical, hydraulic and works schedule to change the conventional self-bearing falsework system originally defined by a large movable scaffolding system (LMSS) developed “ad-hoc” and provided with an innovative OPS system. This enabled the development of a 14-day cycle to complete pouring, curing, hardening and scaffolding self-launching to the next span stage. The present article presents and discusses the problems to be solved using this “state of the art” scaffolding system, and keywords to be considered during the design process. 1 INTRODUCTION: STRUCTURAL SYSTEM FEATURES The Cabriel River Viaduct has an overall length of 520 m, distributed by 8 spans, 2 × 50 m spans at either end and 6 central spans each 70 m in length, conceived as a continuous parabolic varying curved beam formed by a post-tensioned box-girder deck. Located in a natural environment characterized by an abrupt orography requires piles which reach around 50 m in height. The contractor was invited by determinant factors such as orographical, hydraulic and works schedule to change the falsework system originally defined. Hence, the conventional self-bearing scaffolding system which needed to be dismantled and moved span by span, was substituted by a large movable scaffolding system (LMSS) developed “ad-hoc” and provided with an innovative OPS system developed by a specialized bridge engineering company. The use of the LMSS enabled the development of a 14-day cycle to complete pouring, curing, hardening and scaffold self-launching to the next span stage. On the contrary, using of this construction systems implied some deck design and slender pile troubles to be solved, inducing some unfavorable load state compared to deck ultimate state design considered in the original project, also laying out new dynamical analysis scenarios and affecting the design carried out of some deck principal resistant elements. 1.1 Deck The box-girder post-tensioned deck of the viaduct has varying parabolic depth between 2.50 m at mid span and 3.47 m at supports, and an 11.00 m width. Section lateral walls are inclined to facilitate demoulding. Lateral cantilevers have 3.00 m length. Pre-stressed active steel has been arranged in deck section webs, through 12 tendons of 31 wires each, tensioned to 75% of its maximum load (12 × 6061 kN). 1.2 Piles & abutments The structure piles are made of reinforced concrete with a rectangular hollow tapered section of 35 cm thickness, varying linearly in both, longitudinal direction from 2.20 m at top with 1:50 gradient, and transversal direction from 4.00 m at top with 1:70 gradient. 663
Figure 1.
Cabriel River Viaduct general view.
The lateral piles reach heights of 17.00 m whilst the central ones reach 47.00 m. Foundations are resolved by direct footings in the former case and by deeper foundations (1.50 m diameter ground piles) in the latter Both deck endings bearing over closed reinforced concrete abutments with lateral walls, with superficial foundations. 1.3 Bearing system In order to avoid the use of deck intermediate expansion joints, a bearing system was defined, formed by sliding confined devices at end and lateral piles, unidirectional POT type, and standard neoprene rubber devices in the rest. This solution allowed to practically mitigate important horizontal reaction transmission to the more rigid substructure located at either end.
2 CONSTRUCTION PROCESS ASSOCIATED TO LARGE MOVABLE SCAFFOLDING SYSTEM The contractor was inclined towards utilisation of Large Movable scaffolding System in order to reduce and simplify construction process. It implied some differences to the one originally defined in viaduct project: – Self-bearing falsework formed by an upper bow-string arch, with active pre-stressed tie (Pacheco; da Fonseca, 2002) & (Pacheco, 2004), which allowed self-winding deformation correction. – The scaffolding system was supported on one side by the deck cantilever of the prior constructive phase of 17.50 m in length from the previous pile and supported on the other side by the next pile. – Necessity of anchorage to deck to permit the scaffolding system shift over previously built deck, and to guarantee its stability. The large movable scaffolding system had an active pre-tensioned tie registered system OPS (Pacheco; da Fonseca, 2002), and was “ad-hoc” developed by a company specialized in large span viaduct construction. It was 146 m in overall length and has a weight of 730 t, excluding formwork and poured concrete (da Fonseca; Pacheco, 2009). Its use enabled the completion of 70 m-long deck spans including concrete moulding, pouring, curing and strengthening, together with shifting to the next span construction stage position in 14-day cycles. 664
Figure 2.
Large movable scaffolding system view in 2nd span construction stage position.
Figure 3. View of the large movable scaffolding system in 7th span construction stage.
The deformation of the tensioned chord of the arch was controlled by OPS ® system which consists in a non-adherent active pre-stressing system that allowed to adapt to any load requirements, reaching 22 t/m in deck weight, controlling its maximum deformations to a threshold of 40 mm during concrete curing, hardening and strengthening (Pavasal, 2010). In general, structurally speaking, the constructive procedure was in concordance with that originally defined, but an important exception arose from new significant load cases applied in deck and piles. This load state was transmitted in each construction stage to deck and piles through advance, concreting and pile portico of the auto-scaffolding system as outlined in the next figure. 3 CONSTRUCTION DESIGN The construction re-design of the viaduct was marked by new load cases and analysis scenarios of the originally defined bridge resisting elements derived from the self-bearing movable scaffolding system use. They can be classified into 6 types: 1. Introduction of provisional deck anchorage to fixed points in order to guarantee the movements and operating safety of the self-bearing falsework. 665
Figure 4.
Load porticos of scaffolding system outline (source: (da Fonseca; Pacheco, 2009)).
Figure 5. Geometry, supports, constraints and discretising of one-dimensional finite elements of global structural model.
2. Instability buckling analysis derived from falsework – pile interaction. 3. Aerodynamic analysis due to falsework – pile interaction, detecting inadmissible top pile displacements induced by aeroelastic phenomenon. 4. New load states in deck construction stages to take into account a complete review of the global and local deck sizing, particularly in the load introduction points, together with the new definition of deck camber evolution. Moreover, a beam stresses revision should be carried out, in each case adopting the adequate layouts to guarantee its limitations fulfilment. 5. Design of reinforcement detailing and strengthening to allow the portico cross through pile deck diaphragms, remaining the twisting circuit continuity in the critical points of load transmission interfaces. 6. Transversal bending of box-girder deck section analysis due to modifications in maximum cantilever lengths, due to slight geometrical modifications in the plan of the beam axis induced by falsework geometry demands. Viaduct behaviour was analysed by a global evolutionary finite element model, which replicates real geometry, elements transversal sections and supporting conditions in each construction stage and service final state. Load cases associated to shifting, parking and concreting of falsework was introduced in each stage acting according to outline of Figure 4. Likewise, derived from global model results, partial models were implemented to analyse certain static and dynamic structural behaviour, such as: second order analysis and aeroelastic analysis of piles, stress analysis and transversal bending of box-girder deck. 3.1 Piles From partial models of one-dimensional finite elements a non-linear geometric and mechanic analysis of piles was carried out taking into account the maximum falsework reaction associated to concreting operation, analysing both, buckling instability and reinforcement piles optimization. 666
Figure 6. (Left) Vortex-shedding frequency vs. Wind velocity over an elastic structure (source: (Simiu; Scanlan, 1996)); (Right) Stream lines of flow around an object inmersed for Reynolds number, Re = 250 (source: (Simiu; Scanlan, 1996)).
Despite not having an excessively high slenderness in bridge longitudinal direction (Btrans /hmax,pila ≈ 1/10), the presence of the mass associated to 70 m of poured deck stretch, 1450 t, implied an excessive sensitivity of the higher piles to aeroelastic phenomenon during construction stages (Dominguez; Company, 2009a), associated to longitudinal vibrations induced by vortexshedding in the pile wake, dynamically amplified by resonant effects. Since, considering a simple cantilever static system restrained at top by a longitudinal spring representing falsework system longitudinal stiffness as Hooke constant (since deck concrete was still hardening and strengthening), the natural structural frequency was close to that of vortex-shedding. 3.1.1 Vortex – shedding These kind of periodic oscillations are produced by alternating impulsive periodic forces induced by vortex-shedding starting inside the boundary layer, and translated as a difference in dynamic pressure resultants acting in both “long” faces of the pile section. The main problem of this phenomenon is the low wind velocities threshold above it appears, with a high presentation probability, in which resonance takes place between the vortex-shedding frequency and natural frequency of the dynamic system. The evolution of the vortex-shedding frequency with wind velocity over an elastic structure can be seen in next figure, which shows that the “lock-in region” corresponds to a wind-velocity range in which frequencies coupling takes place. In this region, the structure what controls system oscillations, being dynamically amplified, as may be noted in Figure 6. One of the parameters that best describes the vortex-shedding vibrations is the Strouhal Number, which connects the object section shape with fluid velocity. This dependence on the fluid velocity makes the study of vortex-shedding on physical models very sensitive to scale effects. 3.1.2 Problem particularization to viaduct piles Modal dynamical analysis was carried out over partial one-dimensional finite elements models of the most slender pile, considering different constraints-support conditions on top of the pile, obtaining the natural frequency of the structure in each analysed case, associated to longitudinal displacements vibration modes. The shape and geometry of the pile section presented a low Strouhal Number, a parameter for which a low vortex-shedding frequency was obtained for moderate wind velocities with a high probability of occurrence. These similar low values obtained for natural frequencies of piles in longitudinal direction and for vortex-shedding frequency associated to moderate wind velocity induced the coupling and consequently the resonance which derived in dynamic magnification of top piles displacements. These motions were theoretically evaluated with the rough method proposed by EC.1 – P2.4 for the case of no restraint at the top of the pile, reaching values of approximately 1 m (Dominguez; Company, 2009a) for lower wind velocities than the wind design velocity of the environment. 667
Figure 7. Impulsive response time series induced by vortex-shedding in displacements and spectral density, registered in boundary layer wind tunnel tests on a large scale box-girder deck model of the “StoreBaelt Bridge” (Denmark), (a) before “lock-in”; (b) during “lock-in”; (c) after “lock-in” (source: (Dominguez, 2006)).
Figure 8. Vibration mode and natural frequency (0.47 hz) associated to displacements in bridge longitudinal direction; concentrated top mass = 1450 t and top restraint free.
In spite of the roughness and conservatism of the hypothesis and simplifications adopted in the restraints of the pile’s top modelization, the results obtained revealed the necessity to adopt measures to mitigate or reduce the piles aeroelastic induced vibrations during concreting stage, so that: – It would be able to represent a hazard to adequate concrete hardening and strengthening of the viaduct deck. – It would be able to imply dangerous oscillations and displacements for worker and formwork and falsework systems. – It would be able to generate overstresses in piles. – It would be able to damage falsework system sensors and mechanism. Thus, different solutions were defined to reduce vibrations. They were set out in order to obtain higher wind critical velocity (Vcri ) than the threshold condition established in EC.1 – P2.4 to avoid 668
Figure 9. Bending moment outline due to deadload, pre-stressing and concreting portico load case during 2nd deck construction stage (Dominguez; Company, 2009f).
resonance between natural frequency of piles during deck construction stage and vortex-shedding frequency (excitation force). The condition is summarized in next equations:
where: vcalc = wind design velocity evaluated at a height of effective correlation length where shedding take place, establish by EC.1 – P2.4 hypothesis and equations.
where: b = reference object immersed section width in the vibration direction; ni,y = natural frequency of ith vibration mode of the system; St = Strouhal Number, function of shape’s section and flow velocity. Basically, the solutions might be shortened by the subsequent options: – Improve aerodynamic response of piles by using aerodynamic smooth and light profiles attached to pile faces throughout the superior effective length where the aeroelastic phenomenon took place, increasing vortex-shedding frequency and consequently critical wind velocity. – Enlarge the longitudinal stiffness of the pile’s top restraint during the construction stage, modifying natural frequency through an increase in the longitudinal stiffness of the falsework system. Due to economic and simple availability reasons, the latter option was chosen and materialized by increasing the post-tensioned high yield tension bars area arranged (da Fonseca; Pacheco, 2009) anchorage to bridge deck, to the one which guaranteed a higher enough critical wind velocity in any operating phase of the system. Thus, a safe operational wind critical velocity was established at 220 km/h, which had a low probability of apparition at the viaduct location in view of its wind velocity register enclosed by the contractor. 3.2 Deck 3.2.1 Ultimate limit & service stresses verifications The utilisation of the self-bearing falsework induced the inversion of bending moments in the critical sections of mid span, and also an appreciable increase of negative moments at supports section in comparison to ultimate state design carried out in original project. Hence, the pre-stressed effects were enlarged during deck execution (Dominguez; Company, 2009f). 669
Figure 10. View of ballast used for counterbalance deck, formed by deadload concrete blocks and active steel spools.
Figure 11. View of detail of deck anchorage to abutment during construction stages.
Despite critical deck sections having enough resistance in ultimate limit state verifications, a high risk of compression cracks in the lower slab of the box-girder was detected, setting out some alternative options based on the maintenance of concrete and bar steel volumetric quantities. 3.2.1.1 Concrete quality upgrade Some concrete additives were used in order to reach the concrete design strength in 14 days (execution cycle period). Moreover, an upgrade of concrete quality (higher compressive resistance) was analysed in order to solve the high compressive stress problems of the lower plate of box-girder deck. 3.2.1.2 Counterbalance of (n-1) prior spans In order to reduce excess of negative bending moment in mid-span of the former executed span during the 2nd one construction stage, and in general in mid-span of (n-1) span during the construction stage of the n span, the necessary counterbalance bending moment were defined as a function of concrete quality. Finally, the arrangement of some kind of ballast in particular positions, mainly at mid-span, was the option adopted to achieve goal effects (Dominguez; Company, 2009f) in each construction stage. 3.2.2 Provisional deck anchorage In order to avoid a remain deck displacement that affects its positioning over bearing devices, together with the necessity of guaranteeing work and operational safety, it was decided to anchor of deck to former executed abutment during construction process by use of high yield strength pre-stressed bars (Dominguez; Company, 2009d). 670
This kind of anchorage but in vertical direction was needed to solve the deck raising on exterior bearing of the fixed abutment which arose during 2nd span construction stage due to unbalanced reaction detected. 3.2.3 Deck camber control during construction New beam camber analysis was carried out in order to define the required execution precamber to avoid arranging pavement overthicknesses. The total precamber for each (n-2) construction phase was defined by the sum of the next load cases (Dominguez; Company, 2009c): (n-2) phase Deadload and Pre-stressing; (n-1) phase concreting portico load case; (n-1) phase release of concreting portico; (n-1) phase deadload and pre-stress load cases; (n) phase concreting portico load case; (n) phase release of concreting portico load case; (n) phase deadload and pre-stressing load cases. Finally, after precamber analysis carried out, it was decided to not precamber any of front contruction phase section or mid-spans section, due to low permanent deformation account as previously depicted sum of different construction stages. 3.2.4 Pile portico of falsework interference with pile diaphragms The presence of the falsework system pile portico during concrete hardening and strengthening generated some interference over pile diaphragms, vital elements in load transmission from deck to piles and large strength and stiffness demanded. It implied the cutting of the torsional circuit at these points, which was replaced by the use of double thread couplings arranged after pile portico removal, together with the reinforcement resizing of diaphragms. 4 CONCLUSIONS In principal, as has been seen in previous sections, the elements that needed a detailed analysis and practical solutions given primarily under economic and ease factors are summarized in the following list: – Ultimate limit state resistant deck verification & Service limit state stresses analysis of viaduct box-girder post-tensioned concrete deck. – Deck counterbalance during construction process in order to avoid compressive cracks. – Horizontal and vertical displacement control during the construction process. – Provisional anchorage deck definition to fixed point during construction stages. – Pile portico of falsework system interference over pile diaphragms. – Precamber deck analysis. – Dynamical pile analysis in order to avoid inadmissible displacements in pile’s top induced by aeroelastic phenomenon. – Wind velocity establishment compatible with falsework system functionality. The adequate analysis and control of the previous list factors are the keywords to take into account to form the basis of the Construction Design of long-span box-girder posttensioned concrete decks, guaranteeing a safe, economic and quality execution of such a kind of bridge as Cabriel River Viaduct. REFERENCES da Fonseca, A.; Pacheco, P. 2002, Organic Prestressing, Journal of Structural Engineering, ASCE 2002, pp. 400–405. da Fonseca A.; Pacheco, P. (BERD). 2009. Anejo 2, Interacción de la Autocimbra con la Estructura, Flasework system MS-70 Project. Domínguez Santana, B. 2006. MScThesis: Vortex-induced motions in box-girder decks of Long Span Suspended Bridges. BYG – DTU, Denmark Technical University, FORCE Technology, Copenhagen.
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Domínguez Santana, B; Company Vazquez, I.J. (Valter). 2009a. Viaducto Río Cabriel – Análisis dinámico pilas, IN-005.2, Valencia. Domínguez Santana, B; Company Vazquez, I.J. (VALTER). 2009b. Viaducto Río Cabriel – Propuesta de soluciones para mejora del comportamiento dinámico de pilas en fase de construcción, IN-008, Valencia. Domínguez Santana, B; Company Vazquez, I.J. (VALTER). 2009c. Viaducto Río Cabriel – Estimación Contraflechas tablero, IN-006, Valencia. Domínguez Santana, B; Company Vazquez, I.J. (VALTER). 2009d. Viaducto Río Cabriel – Materialización del punto fijo en estribo E2 en fase de construcción, IN-011, Valencia. Domínguez Santana, B; Company Vazquez, I.J. (VALTER). 2009e. Viaducto Río Cabriel – Comprobación tablero con cargas actualizadas de la autocimbra, IN-012, Valencia. Domínguez Santana, B; Company Vazquez, I.J. (VALTER). 2009f. Viaducto Río Cabriel – Definición de lastres de compensación en fase de construcción, IN-016, Valencia. Pacheco, P. 2004. Auto-adjustable prestressing, PCT Patent, pct/pt2004/011, WO2004/109018, Gazette OMPI. Pacheco, P.; Coelho, H.; Borges, P.; Guerra, A. 2011. Technical Challenges of Large Movable Scaffolding Systems, in Structural Engineering International, Journal of the International Association for Bridge and Structural Engineering, Vol. 21, Number 4, pp. 450-455. Pavasal. 2010. CN-330 Variante de Cofrentes. Autocimbra sobre el Río Cabriel, Valencia. Simiu, E.; Scanlan R.H. 1996. Wind Effects on Structures: Fundamental and Applications to Design. Wiley – Interscience, 3a edition. John Wiley & Sons. New York.
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Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
Segmental precast technology for multi-span bridges (production, transportation and launching) V.N. Heggade Board of Management, Gammon India Limited
ABSTRACT: In case of bridge building by precast segmental construction method, the precast elements that are cast in the casting yard through specially designed molds are transported to the location and post tensioned together. The small segments are made transversely which is in contrast to precast girder bridges where the overall lengths of the precast girders are almost same as that of spans. Generally the joints which are either glued by specially designed epoxy formulation or in some cases dry are orthogonal to the longitudinal axis of the bridge. Normally the reinforcement is not continuous through joints unless it is a must for from structural reasons. The decisions on the permanent structure designs are dictated by shape of segments, weight of precast elements, arrangement of pre-stressing cables and methods of casting, transportation and erection. Generally there are two principle methods of construction which are span by span construction and cantilever method of construction. For both these method of construction the precast segments have to be realized in casting yard either by long line casting method or short line method of casting. In span by span construction the erection or launching of the segments can either be realized by under slung erection method or overhead launching system. Balanced and even free cantilever method of construction can be realized both by form travelers or overhead launching system, while the progressive cantilever construction a variation of free cantilever construction may require temporary towers. In the detailed paper below, the above nuances and subtleties of pre casting, transportation and erection methods are discussed by way of illustrations and evolution of precast segmental bridge engineering in India. 1 INTRODUCTION Long multi-span bridges have always been designed taking into account certain aspects such as geographical location, available materials, known and practical construction methods imbibing durability, speedy delivery and economic and aesthetic considerations. The design approach to long bridges consisting of multi-span has an inherent distinction from short bridges involving less number of spans as the repetitive production characteristic in multi-span bridge provides with an opportunity of mass industrial production. In addition, concrete by its very nature renders to be moulded to produce exciting shapes.it is a versatile material that can reproduce the texture from steel forms. This quality of the material can be put to use in obtaining exciting shapes and textures that enhance the appearance of the built environment. It is the material most suited to express the ideas of form, which is true to its function, both efficiently and economically. Precast segmental construction grew out of mass production concept to prefabricate off site as much of the bridge deck as possible which has many advantages that include: – Enhanced quality control of casting in factory conditions. – Ability to adjust the rate of casting to suit the rate of erection by increasing the number of moulds thereby flexibility to control the construction programme. – Reduction of disruption to the existing users of the site. 673
Figure 1. First pre stressed concrete bridges in India, Assam railway link bridges (1948).
Figure 2. First cast-in-situ cantilever construction in India, Barak bridge at Silchar (1961).
Figure 3. Top: span-by-span construction method. Bottom: cantilever method of construction. Source: Precast segmental bridges, Guide to good practice for structural and site engineers. Prepared by fib Task Group 10.3.
– The flexibility to run the casting of the deck in parallel with the construction of foundations and sub-structures. – Overall reduction in construction programme due to possible parallel activities and hence cost savings. The technological evolution of segmental construction has never been unremitting. The progression is hall marked all over the world by heaps, leaps and also significant plateaus. The economic and competitive necessities had given rise to innovations through incremental improvisations by virtue of new structural forms, new methods of analysis and designs, new material properties and materials itself, new pre stressing ways, new construction methodologies and value engineering enabled contracting environments. The history of segmental construction in India is not very different. The technology was pioneered for bridges (Fig. 1) way back in 1948 for Assam railway link bridge load testing, which incidentally holds record for first time application of pre stressed concrete technology for bridges in the country. The technology was extended first times to cast-in-situ cantilever construction for Barak bridge at Silchar (Fig. 2) in 1961. In precast segmental construction method, the precast elements that are cast in the casting yard through specially designed moulds are transported to the location and post tensioned together. The small segments are made transversely which is in contrast to precast girder bridges where the overall lengths of the precast girders are almost same as that of spans. Generally the joints which are either glued by specially designed epoxy formulation or in some cases dry are orthogonal to the longitudinal axis of the bridge. Normally the reinforcement is not continuous through joints unless it is a must for from structural reasons. The decisions on the permanent structure designs are dictated by shape of segments, weight of precast elements, arrangement of pre-stressing cables and methods of casting, transportation and erection. Generally there are two principle methods of construction (Fig. 3) which are span by span construction and cantilever method of construction. For both these method of construction the 674
Figure 4. Top left: Krishna bridge at Deodurg (1971–72); Top right: Long Key bridge in US (1981); Bottom left: Ganga bridge at Buxer (1971–75); Bottom right: Ramp I bridge Florida (1985).
Figure 5. Top left: Fort Lauderdale-Hollywood International Airport expansion required the removal of three precast segmental concrete bridges in 2002. Top right: dismantling of one span of Ganga bridge at Patna in 2014.
precast segments have to be realized in casting yard either by long line casting method or short line method of casting. In span by span construction the erection or launching of the segments can either be realized by under slung erection method or overhead launching system. Balanced and even free cantilever method of construction can be realized both by form travelers or overhead launching system, while the progressive cantilever construction a variation of free cantilever construction may require temporary towers. It is very interesting to note that first application of span by span precast segmental construction was realized in India for Krishna bridge at Deodurg in 1971 while the technology was used in United States first time for Long Key bridge in 1981. Precast segmental for balanced cantilever method (BCM) of construction was adopted first time for Ganga bridge at Buxer in 1971–75 while in United States technology using BCM adopted first time for Ramp I bridge Florida in 1985. On the other hand the dismantling of pre cast segmental bridges started in United states as early as 2002 for Fort Lauderdale-Hollywood International Airport expansion while in India the same has begun in 2014 for Ganga bridge at Patna. It is bit surprising that precast segmental construction technology for both span by span and Balanced cantilever method of construction started in India almost a decade earlier and dismantling began almost a decade later. In the recent years, segmental construction technology has been extensively adopted for multispan fly-over, bridges and elevated transportation structures, though there was a significant hiatus from 1980 to 1995 in India, in exploitation of this technology. 2 METHODS OF CONSTRUCTION Pre cast methods of construction have the following process of value addition: – Pre casting of segments – Transportation & Handling of pre cast segments – Erection of pre cast Segments 675
Principally there are two methods of construction in precast segmental bridging, VIZ, span by span method of construction (SSMC) and cantilever method of construction (CMC). Span by span construction have similarities to that of cast in situ simply supported method of construction, the broad distinction being instead of whole span being cast at the location, the segments are cast in factory environment somewhere else, transported to location and segments are assembled and post tensioned at location to make a span or spans. 2.1 Span by span method of construction (SSMC) In SSMC span length is determined by the gantry which depends upon the deck width. In India a maximum span of approximately 50 m is achieved. Typically two spans per week is possible in case of simply supported schemes while cast in situ span of 50m is achieved in six weeks. Erection speed could be as fast as 3-4 spans per week for 35 to 40 m spans. When large deck elements working at height supported on slender piers, extra stability precaution is warranted as overhead gantries are potentially susceptible to instability also foundation design is governed by this construction state particularly under high winds. This method of construction can be realized by a larger range of types of available erection equipment, such as overhead, underslung or falsework systems. This gives flexibility in the reuse of already available equipment. Depending upon the equipment chosen, the delivery of each segment can take place over or below the deck which needs to be taken into account during the equipment selection process. Invariably, overhead gantries are self-launching which involves more complexity and the consequence cost. Overhead gantries are designed and fabricated by some contractors who have separate enabling structures design department complimented by plant engineering and the contractors who don’t have in house design facilities procure the same from specialized suppliers. The underslung trusses are launched from one span to the next with the support of cranes or where access is restricted the trusses can be self-launching whose design being simple are normally carried out by the contractors themselves. The false work or ground supported trestles are used in urban environment where access is easily available or where the radius of curvature of span is too sharp (lesser than 300 m) as the same introduces much complexities of geometric control during precasting and transverse adjustments in overhead gantries during erection of segments. The heaviest pier segments are stiffened with diaphragms as the same are used for anchorage, deviation of longitudinal external post-tensioning cables and to transfer the loads to the bearings. In addition the large concentrated forces are acting on this segment during erection, particularly when overhead launching system is adopted as the front leg is supported on this segment in one variation of SSCM. Obviously, reduction in weight is very important for precast segmental bridges than cast-in-situ SSCM bridges. The same is the derivative of aspects like transportation of the segment, capacity of cranes and winches of the launcher, etc. The weight reduction of cross-sections are generally achieved by: – – – –
Reducing weight of webs (use of steel/concrete trusses or corrugated steel plates), reducing thicknesses of bottom and top slabs, two stage construction for some parts of the segment in transverse direction, transverse pre-stressing to reduce number of cells in the box and increase the lengths of cantilevers, – provision of ribs (steel or precast) with in the box as well as outside to support cantilevers, – reducing the length of segments, – use of high strength concrete. The variety of cross-sections (Fig. 6) are used for precast segmental bridges. Among the number of aspects enumerated below, the choice principally depends on span and deck width. In addition to the torsional stiffness of a box section which provides some tolerance in the construction, the reasons why box sections are preferred in precast segmental construction are: – during construction, temporary out of balance forces, lateral torsion deformation and buckling effects are taken care of by torsional stiffness, – for curvilinear alignments closed box sections are required to cater for permanent loads, 676
Figure 6. Various structural cross sections used precast segment bridges.
– In segmental construction using overhead launching,hogging capacity of section near the pier is required to allow for additional hogging moment effects from a launching girder. In box girder it is possible to lower neutral axis due to wide bottom slab than an open section which allows an efficient use of prestressing and high moment capacity in hogging at the pier. – the external prestressing can be effectively used in box section, there by reduction in weight Thus the governing segment dimensions are width of the top slab, depth of the box girder, width of the bottom slab, spacing of the webs and length of the segments. Normally, the width of the top slab is considered to be equivalent to the width of bridge in transverse directions and up to 12.0 m two webs are considered to be adequate from the premise that the girder behaves like a beam up to width to depth ratio of 6.0, exceeding which cross section starts behaving like a slab. In the case where the cross-section has to be analysed on the principles of beam theory for the width to depth ratio exceeding 6.0, multiple celled box or single box with multiple webs is adopted. The depth of the box section could be anywhere between 1/18 to 1/30 of the span for uniformly deep girders, which are normally used for span-by-span construction. For less than 40m span in some cases, voided slab with transversal cantilevers are also used. Where the roadway width is greater than the achievable maximum for single cell cross-sections also for reducing the weight of segments, the basic box sections are over the period value engineered with number of variations such as box with ribs, box with webs in steel, box with struts, box with transverse pre-stressing. Fish belly closed sections with defined and undefined webs and the spine with precast cantilever ribs and strut are also tried out in the recent past from aesthetic and economic considerations. External post-tensioning has many advantages for precast bridges constructed by SSMC. The external cables can be installed very quickly and a near ideal vertical PT arrangement can be realized without affecting the required web thickness. They also offer benefits with regard to durability, as the corrosion protection is pre-applied under a controlled environment and can be inspected during the operational period of the structure. The disadvantages if any could be additional anchor blocks and deviators as well as the reduced available internal lever arm and inefficient resistance to ULS as the tension in the cables increase due to rotation of section. Despite the obvious advantages, the external prestressing for SSMC is not caught up in India. The first externally pre-stressed bridge using this technology in India is Noida toll bridge at Delhi (Fig. 7). Continuity of pre-stressing especially for continuous bridges in SSMC method of precast segmental bridges can be achieved in many ways: – – – – –
long tendons over multiple spans after spans constructed cables terminating at or near 1/4 span crossing cables close to the pier segment cable Couplers straight cables adjacent to the pier
The bridge across River Krishna at Deodurg, constructed in 1971–72 was the first bridge where span by span method of construction adopting precast segments was used in India. This 540 m long 677
Figure 7.
First pre cast segmental bridge in India with external prestressing. Noida toll bidge (2000).
Figure 8.
Krishna bridge, many firsts in SSMC technology in India.
submersible bridge across River Krishna at Deodurg in Karnataka, having 18 spans of 30 m each, was considered to be the most ideal bridge for adoption of this technique. The bridge was also the first submersible bridge in pre stressed concrete. The cross section of the decking is a three-celled box, in the shape of an aerofoil required from design considerations of 5.5 m submergence and a maximum mean velocity of 5.5 m/s under peak flood condition. The decking is semi-continuous, with six span continuity for superimposed loads and live load. This bridge should be considered turning point in precast segmental bridge technology in India as many firsts of precast segmental technology was adopted. 2.2 Cantilever method of construction (CMC) Cantilever method by precast segmental construction is similar to that of cast-in-situ method of construction, the difference being that the segments are precast in casting yard in the former while concreting is done in place for the later. In CMC, the pier head or the pier table is constructed first and the segmental construction progresses from the pier table or pier head on either side or on one side. There are three variations in CMC depending upon whether the segmental construction from pier head progresses on one side or on either side simultaneously. – Balanced cantilever method of construction (BCM) – Free cantilever method of construction (FCM) – Progressive placement method (PPM). In this method, it is not necessary to support the segments of a whole span like in SSM as such the spans up to 120 m are achieved in India though longer spans are possible. Though the launching and assembling of segments can be faster than SSM, other operations like pre cambering adjustments, 678
continuity prestressing for closing segments, etc. are time taking as such the method is suitable for long spans. The stability overturning loads controlled during construction can be governing factor in the design of foundation, piers and pier heads in this method. As launchers or form travellers are more slender than in SSM and spans are larger, stability of launcher during erection needs to be given special attention. In CMC the variety of equipment options are available like erection gantries resting on piers, form travellers, lifting frames, cranes and bed or barge mounted gantries depending upon the delivery of segments like whether from below the deck or above the deck. Similar to SSM, temporary prestress is used with bars for gluing the joints while permanent prestress is carried out for self-launching of the gantry considering the local loads and deflections. The launcher capacity depends upon the weight of the segment and heavier the weight of the segment, the design for advancing stage of launcher may govern. For the longer spans, geometric control during erection becomes critical as such the practice is to construct the cantilever over jacks enabling three-dimensional geometric control at the completion of each segment. This needs to be correlated during match casting of segments at casting yard. The radius of curvature in plan, longitudinal and transversal slopes. etc. are limited by geometric control requirement and span lengths. Normally for 100 m span, the minimum radius of curvature is limited to 700 while it may go up to 750 for 120 m span. As CMC are of larger span lengths, from economic considerations variable depth segments, i.e. deeper segments at and near the piers and less deep segments at mid-span are preferred. Though from the precasting point of view, variable depth segments are more complicated, once the complexities are addressed at casting yard, the production efficiency can be more or less the same as that of constant depth segments. As far as possible, combination of inclined webs with variable depth shall be avoided as the same will lead to a more complex formwork solution. Normally, second or third grade parabolic variation akin to moment diagram is adopted for variable depth spans from the structural and aesthetical considerations. For variable depth spans, the ratio over the pier is 1/17 to 1/20 with a minimal depth at the mid-span of 2.5 m, to enable movement inside the box (minimum 2.2 m). In CMC invariably, the superstructure have to be continuous over the piers either having the deck monolithic with the pier or being supported on bearings. The choice depends on many structural factors, such as the span length, total bridge length, pier height and flexibility requirements. In this method for the structural optimum, the end spans at either side of these intermediate piers are approximately 0.65 x the intermediate span length. By virtue of construction method large negative moments due to cantilever dead load and erection load are to be resisted that warrants internal pre stressing tendons to be arranged in the top slab by being usually anchored at the intersection of web and top slab, also in some cases blisters inside the box can also be an option. The mid-span continuity by closure pour (both top and bottom) is typically realized using internal tendons. In some cases where additional pre stressing is required for service conditions over the construction condition, external tendons could also be used for the continuity post-tensioning with a high point over the piers and low points at mid span. Ganga bridge at Buxer (Fig. 4) was the first bridge in India realized by CMC using precast segmental construction technology. This balanced cantilever bridge is situated in Bihar on an important highway link is still in an excellent condition. Total length of bridge is 1122.26 m comprising of 10 intermediate spans of 101.22 m and two end spans of 55.03 m. The superstructure consists of post-tensioned single cell balanced cantilever box girder of depth varying from 6.1 m near pier to 2.13 m at the tip of the cantilever. The cantilever arms are connected with central hinges. Precast as well as in-situ segmental construction was adopted for construction of box decking. Out of eleven T arms central nine T arms were constructed using precast segments and one each on either bank was cast in-situ. In case of precast segmental construction, pier head segment was also precast (Fig. 9). The weight of precast segment was restricted to 65 t. With this the length of segment was varying between 2.13 m near pier to 3 m for the balance portion. For pre casting of segments, long line method was used. Casting bed, extending for about 100 m, for all the segments of cantilever arm including pier head segment was set on one of the banks. 679
Figure 9. For the first CMC in India (Ganga bridge at Buxer), pier head was precast.
Figure 10.
First precast segmental Bridge over the Marne River at Luzancy, France.
3 PRE CASTING TECHNOLOGY Match casting is the essentially used technology for precast segmental bridges. In match casting each segment is individually cast using already cast adjacent segments as face shuttering form. Thus the system ensures a perfect fitting between adjacent segments in their final position. So a span in SSMC or a cantilever in CCM is incrementally cast segment by segment. The first precast segments were designed by Eugène Freyssinet, for the bridge built in 1944–1946 over the Marne River in France (Fig. 10). The bridge was idealized as a two-hinged portal frame with multiple hollow box cross-section and adjustable hinged bearings. The joints were filled with mortar to equalize for the tolerances between adjacent segments. The segments were post tensioned in transverse, longitudinal and vertical direction. Though, this bridge was precast segmental, the technology used was not match casting as such the gaps between the segments were to filled up with mortar. It is interesting to note here that for the first time, match casting technology was used in the world for a railway bridge in Assam (Fig. 1) of India in 1949 by J C Gammon. After Independence in1947, Assam Rail Link was taken on priority which included many bridges. All railway bridges used to be of steel. Due to non-availability of structural steel five bridges were constructed with PSC beams. Gammon founder Mr. J.C. Gammon offered to construct based on alternative design. Design was to be validated by load test to failure using heaviest available locomotive in India. Load test was conducted in Kalyan yard to verify behaviour under dynamic loads. Two 13.4 m long beams were match cast in three segments each, assembled and pre stressed (Fig. 11). Decking was load tested using heaviest B.G.locomotive.Engine was made to derail on beams. Construction was taken up only after satisfactory test results. Incidentally this is the first match cast segmental beam. 680
Figure 11.
First match casting technology in 1949 for Assam rail link in India.
There are two methods of match casting. VIZ. Long line match casting (LLMC) and Short line match casting (SLMC). In LLMC, the segments stay on the casting bed after concreting and the formworks are moved forward to cast the next segment. In SLMC, the formworks are at a fixed position and the concreted segment is moved from the casting bed in order to make room for the next one. Though it is not psychologically palatable, it is more efficient in terms of time and labour to move the segments than the forms in spite of segment being heavier than forms. 3.1 Long line match casting (LLMC) In LLMC the entire span is cast in segments for SSMC. Similarly in CMC match casting is done for entire double cantilever length over a pier (or “T” shape), including camber, which caters for deflections during the erection. Though for a long line system in balanced cantilever, the pier segment is often preferred to be cast- in-situ to enable control of the erection geometry, precast pier segments are also resorted to in some cases. However, in case of precast pier head, the same has to be trapezoidal (Fig. 9) to enable stripping as the pier segments in cantilever construction are flanked by match cast cantilever segment inducing pinching effect. The method does not warrant constant geometrical control adjustments but there is a large space requirement for casting bed. 3.1.1 LLMC for SSMC The LLMC for SSMC was used for the first time in 1952 for a single span bridge in the New York State near Shelton. The bridge was designed by Jean Muller who was pioneer of the precast segmental bridge technology. The superstructure was divided longitudinally into three precast RCC beams which were cast end to end (match-casting). The elements were stressed together with post-tensioning tendons. Though short line match casting had begun way back in 70s in India for SSMC, the long line match casting for SSMC for the first time was adopted as late as in 1996 for Sirsi circle fly over (Fig. 12) in Bengaluru. In this method, casting moulds (Fig. 13) consist of external shuttering, internal shuttering, soffit formwork and bulkhead (end shuttering). Special steel moulds are fabricated for producing the segments within tolerances permitted. Provision of turnbuckles and four screw jacks at corner enables the casting mould to be rotated to suit the geometric profile to maintain horizontal and vertical curve of span. The bulkheads are fabricated in such a way that they are capable of connecting the sheathing without intrusion of grout and producing shear keys of required shape. Normally the end diaphragm segment is cast on separate casting (S1) bed independently and same is placed on long line bed to act as match cast segment for further casting of intermediate segments. In the 1st stage (Fig. 14) for casting intermediate segments soffit formwork for one span 681
Figure 12. Sirsi Circle Viaduct, Bangalore, India 1996–1999.
Figure 13. Typical long line casting machine.
Figure 14.
Process of LLMC in stages.
is aligned as per required curvature. End diaphragm segments S1F and S1R are placed on casting bed at required locations with proper alignment by means of adequate capacity gantry or crane. In the 2nd stage (Fig. 14) external shuttering is fixed at S2R and S2F location, bulkhead (end shuttering) are placed at S2F and S2R joints i.e. joints 2 & 8 in the Figure. Reinforcement cage is placed at S2R and S2F locations either with the help of a small capacity crane or bed gantry which is used for handling segments. Internal shuttering is placed at position with the help of carriage platform. Inserts are left at lifting position before the segments S2F and S2R are concreted. 682
Figure 15.
First time application of LLMC for CMC in 1962 in France.
Figure 16. Casting machine for LLMC for CMC at Ganga bridge Buxer.
In the 3rd stage (Fig. 14) after concrete has achieved required strength shuttering are removed. First external shutter is removed and shifted to S3 position, then bulkhead is removed and finally internal shuttering is shifted to S3 position. Similar procedure is followed for casting remaining segments. After segment has achieved required strength segments are lifted with the help of gantry and shifted to stacking yard. 3.1.2 LLMC for CMC The bridge over the Seine River near Choisy-le-Roi, France (Fig. 15) built in 1962 was the first application of LLMC for CMC. The superstructure (37.5 – 55.0 – 37.5 m) consisted of one or two standard hollow boxes assembled by free cantilever method of construction joined later by link slabs. The segments had three shear keys per face and epoxy glued joints were used. Segments were produced by the long-line match casting method to achieve a perfect fitting of adjacent segments. Ganga bridge at Buxer (55 + 10 × 101 + 55) constructed during 1971–75 was the first bridge (Fig. 4) in India to use LLMC for CMC. This balanced cantilever bridge is situated in Bihar on an important highway link. Total length of bridge is 1122.26 m comprising of 10 intermediate spans of 101.22m and two end spans of 55.03m. The superstructure consists of post-tensioned single cell balanced cantilever box girder of depth varying from 6.1 m near pier to 2.13 m at the tip of the cantilever. The cantilever arms are connected with central hinges. Precast as well as in situ segmental construction was adopted for construction of box decking. Out of eleven T arms central nine T arms were constructed using precast segments and one each on either bank was cast in situ. In case of precast segmental construction, pier head segment was also precast. The weight of precast segment was restricted to 65 t. With this the length of segment was varying between 2.13 m near pier to 3 m for the balance portion. Casting bed, extending for about 100 m, for all the segments of cantilever arm including pier head segment was set on one of the banks. The casting bed (Fig. 16) consisted of brick walls below the webs, profiled to the shape of the soffit. The soffit shutter of the bottom slab was supported by independent tubular staging. The formwork for sides and the soffit of box, both inside and outside were supported from a mobile transverse girder which could be moved on longitudinal rails to various positions to enable casting 683
Figure 17. Typical casting machine for SLMC.
of precast segments from pier head up to the tip of cantilever. Three such transverse girders were provided for each half of ‘T’ arm to speed up the casting progress. Initially field tests were carried out to examine the smooth working of the system and easy removal of precast segments without damage. Eighty ton capacity Goliath gantry, traversing over entire length of casting bed and stacking bed was used for lifting, stacking and loading of precast segments. It was noticed that removal of earlier cast segments near pier head caused breakages in shear keys of webs and deck and the keys were therefore, reshaped. For convenience of removal of the precast segments (to avoid pinching effect), pier head segment was kept trapezoidal in shape (Fig. 9). The shuttering for central pier head segment was supported on a combination of sand jacks and hydraulic jacks so that the segment could be raised, lowered or tilted as required, to facilitate longitudinal gradient and transition in the decking. Pre casting was started from pier head segment. Following the casting of this segment, subsequent segments were match cast. Ordinary soap solution was used to prevent development of bond between the joining surfaces during match casting. For removal of precast segments, pier head segment was first lowered using sand jacks. Subsequent segments were shifted longitudinally and then lifted and transferred to stacking yard. The segments were stacked horizontally up to 3 segments piled one over the other, but not exceeding 8 m height. Structural adequacy of lower segments was checked and the support points were selected carefully so as to avoid distortion of stacked segment. The segments of each T arm were marked and numbered to ensure perfect matching during erection. 3.2 Short line match casting (SLMC) In SLMC, all the segments are cast in the same place, using stationary form (Fig. 17) against the previously cast segment in order to obtain a match cast joint. Hence space requirement is less as compared to long line method of casting wherein all the segments in the span are cast in their correct relative position on a casting bed that exactly reproduces the profile of the span. The geometric control during casting in SLMC is of paramount importance. The casting machine allows for movement of the previous segment in order to establish a relative position. This unique positioning of each segment, throughout casting, is critical in ensuring that the final geometrical shape of the installed segments over the pier is correct. To do this, geometrical modelling 684
Figure 18.
Unique vertical short line method of casting for Krishna bridge.
Figure 19.
SLMC for SSMC for various projects.
calculations of the segments for final positions are performed. Then the segments are to be positioned accordingly in the mechanism. 3.2.1 SLMC for SSMC While the bridge over the river Rhône at Pierre-Bénite (1962–1965) was the first structure where the precast segments were produced by the short-line match casting method, in India for the first time a unique SLMC was adopted for Krishna bridge at Deodurg (1971–75). Keeping in view easy handling of segments with available equipment, the weight of segment was restricted to 10 t, with segment length of 1.2 m. The box segments were precast in a casting yard and transported to erection location. To facilitate proper access for shuttering, concreting,etc. the 1.2 m long segments were precast one above the other with 1.2 m length kept vertical. Geometric control was achieved by sensitive levelling frame while alignment for match casting were enabled by insertion of rods in 4 number of holes left in the sectional plan (Fig. 18). Since 90s many elevated structures (Fig. 19) are built by SSMC where SLMC is adopted using modern horizontal casting machine. This modern SLMC machine (Fig. 17) consists of of XYZ geometric control adjustment carriage which is a frame supporting the previously cast segment on its bottom form and is provided with horizontal and vertical jacks allowing the accurate positioning of the segment. It is used also to transfer the completed segment on its bottom form to the match cast position avoiding lifting at early age. The machine also consists of internal core form, which allows shuttering for the inner faces of the webs, the bottom face of the upper slab and the vertical faces of the blisters. Translation gantry or internal form carriage of the machine supports the internal core form.All the above 3 mechanized and specialized units of casting machine are operated by hydraulic jacks. 685
Figure 20. Typical casting sequence of SLMC for SSMC.
In the sequence of casting (Fig. 20) soffit formwork is aligned as per required geometry which is preceded by adjusting the fixing of the external form. After this reinforcement cage is kept in the mould and the internal shuttering is fixed, the mould is closed with stop end form. After casting the segment, it is shifted to match cast platform with help of hydraulically operated trolley. Then this segment will be used as match cast segment for casting next segment. After casting next segment, this segment is shifted to the stacking yard. 3.2.2 SLMC for CMC SLMC for BCM (a variation of CMC) was very successfully adopted for Narmada Bridge at Zadeshwar (Fig. 21) in Gujarat in 1971 for the first time in India. Casting yards were setup on both the banks of River. The segment length was varying between 1.5 m to 3 m. The later this technique was extensively used for 5.5 km Ganga bridge at Patna. A meticulous planning of pre casting process including stacking and handling of the segments by SLMC were prepared as shown in the Figure 22. A goliath gantry was used in the casting yard to handle the segments. Segments were transported using specially designed trailer. Short line system demands high accuracy and exact leveling of the segment to achieve the desired deck profile and transverse camber. Systems were not so sophisticated as available now, for accurate positioning during the match casting of segments. The segment had a maximum weight of 90 t. The pre casting beds were located at one end of the bridge and precast segments were stacked on the river bed itself. At one stage for Ganga bridge at Patna, more than 1200 segments were in the stacking yard 3.3 Casting yard organisation and stacking of segments Casting yard (Fig. 23) for pre casting of the segments generally consists of: – – – – –
Casting area including the formworks and survey stations, Segment attacking area, Reinforcement cutting and bending area, Concrete batching plant, if concrete is produced on the yard; Additional areas for workshops, stores and so on.
The kern of the casting yard is the casting area. The other areas are designed in such a way that the handling and transportation within the yard is optimised. 686
Figure 21.
First time application SLMC for CMC in 1971 in India.
Figure 22.
Casting process design of SLMC for CMC of first time application in India.
The design of the lay out should consider the following: – – – – – –
The area required depending upon whether LLMC or SLMC for SSMC or CMC. Casting machine supporting foundations Gantry span and capacities to handle formwork, reinforcement cage and pre-cast segments. Batching plant capacity for providing concrete during casting the segments. The logistics of reinforcement bay for cutting steel, storage space for pre-cast segments, fabrication yard and site office. The segment storage area has to be designed to cater for two major objectives which are:
– The segments shall be kept on the storage area until concrete reaches the required strength for erection. – Segment casting and segment erection activities have different paces and shall be synchronised to arrive at quantity of segments to be stored. 687
Figure 23. Typical casting yard.
Figure 24.
Single, double and triple decker stacking.
Figure 25.
Lifting device privisions & low bedded trailer for transportation.
Generally segments are stored in casting yard itself for limiting transportation and double handling costs. Typically, the span segments are usually stacked (Fig. 24) on two levels. However, stacking on 3 levels or more could be considered for segment’s having higher load bearing capacity. The precast segments are generally provided with lifting holes or hooks (Fig. 25) to fit the lifting devices to enable handling of the segments. The segments are normally transported by specially designed multi-axle trailers whose design has to cater for the limitations of transportation route. 688
4 METHODS AND ERECTION The terminologies for methods of construction i.e. whether SSMC or CMC have been derived from methods of erection adopted. There have been a number of developments in the erection methods for precast segmental bridges. In the early days it was common practice to erect the segments in balanced cantilever, with one segment either side of the pier stressed onto the previous segments using bars, followed by the permanent pre stress. The segments were erected either using land or water based cranes or shear legs. Segment delivery mechanisms are also important in achieving rapid construction methods. Very seldom the permanent works layout can be adjusted to suit the preferred delivery method. In balanced cantilever construction, it is necessary to ensure the stability of the balanced cantilevers during erection. Typically this was done with a symmetric pair of false work towers either side of the pier. However, on a number of recent projects a combination of a prop and vertical pre stress ties have been adopted which has the advantage of reducing the amount of steelwork, and minimising the foundations and work at ground level required during construction. In this solution, one can use ties only on the centreline of the prop to control overturning away from the prop, or where the prop is kept close to the pier, one can also use ties on the pier centreline to control overturning towards the prop. On bridges which are tightly curved in plan, it is possible to use the prop/tie solution to assist in ensuring the stability transversely as well as longitudinally. As the curved balanced cantilever grows one can increase the tie forces at the pier on the outer edge of the curve thereby ensuring the transverse stability of the curved cantilever. Although these methods of erection are still appropriate for many bridges, gantry erection methods are often used for larger projects. Gantries usually allow more rapid construction and they can allow the segments to be delivered at deck level to minimise ground level working. There are a number of types of gantry, which can erect segments either in balanced cantilever or span by span. Whilst developing the design, it is necessary to consider in detail the interaction between the permanent works, the construction method and the construction programme. For example there is no point in detailing the design for rapid erection, if the supply of the segments is slow. Similarly the relationship between the permanent prestress design and the temporary pre stress bars used to hold the segments together during construction can be modified to suit different construction cycles. Whereas it used to be common to use bars to stress the segments onto the previous segment followed by permanent prestressing, installing the permanent prestress is a relatively slow process and it is now more common to fix 2 or 3 segments on temporary bars before applying the permanent prestress. For shorter spans it is even possible to erect all of the segments on bars and then follow up with the permanent prestress. However, with such long segments one needs a lot of temporary prestress, both inside the box and outside it and this creates its own difficulties. 4.1 Span by span Method of erection (SSME) In SSM of erection, an entire span (pier to pier) is erected with special equipment. Depending upon the accessibility and delivery of segments, the equipment could be (Fig. 26): – False work or trestles from ground – Overhead girder – Under slung girder These erected segments become self supporting only after application of posttension as such during non self supporting stage of the span, segments are to be individually supported either by: – Hangers connected between deck top slab and overhead truss for overhead girder, – Point support below the segment wings (part of deck cantilevering out from the web) for under slung girder, – Point supports below segment soffit for under slung girder and false work. 689
Figure 26.
Structural fit & SSME variations.
Figure 27.
Unique combination for SSME used for Krishna bridge in 1971.
There could be 3 structural fit depending upon the determinacy of structural configuration and also provision of cast in situ stiches for SSME, which are: – Isostatic span structure which are simply supported superstructure – Continuous pan structure made of in-fill-span between the pier table with two cast in situ stitches – Continuous span made of isostatic span having cast in situ stitch right on top of the Centre line of pier. 4.1.1 SSME using false work or ground supported trestles For the first time SSME (Fig. 27) was adopted in India for Krshna bridge at Deodurg in 1970. The method used was unique in the sense, the segments were supported on false work and the erection of segments were done using specially designed travelling gantry moving on top of already placed segments. Since the decking was designed as six span semi continuous structure, it was necessary to erect each span on temporary staging. The precast segments were assembled over pipe staging erected in the span. A specially designed gantry moving over already placed segments was used for erection of segments. The segments were checked for proper alignment prior to gluing together. Then the cables were stressed. After establishing semi-continuity through gap concreting, the decking was lowered over permanent bearings. The gluing material was selected by conducting field tests on various materials like epoxy, mortar etc. Ordinary Max Pulls were used to impart temporary pre stress while gluing the segments together. In late ninties, India embarked upon the mammoth infrastructure development especially in urban areas where elevated corridors for both roads and metro became popular. Especially where ground accessibility is there, this method of erection (Fig. 28) became very handy due to its simplicity and also as it could accommodate sharp curvatures. 690
Figure 28.
Classical SSM underslung erection using scaffolding/Trestles.
Figure 29.
Modern erection system using Trestles.
The modern Erection machine (Fig. 29) consists of trestle resting on screw jack assembly, knuckle bearings, main girder and sliding trolleys. Footings are placed on the ground and trestles are erected on the footings supported with screw jack assembly at bottom. Screw jacks enable the lowering and rising of main girder for final alignment vertically. Main girder is placed on knuckle bearings provided over trestles. Knuckle bearings enable main girder to rotate horizontally to facilitate as per horizontal curvature. Trolleys are placed over the main girder to support the segments. Two trolleys are used to support each segment. Arrangements for raising, lowering the segments as well as positioning the segment in longitudinal and transverse direction are required to be provided. The segments are transported (Fig. 30) on special trailers having required capacity (60 to 100 ton) from casting yard to site. The required capacity crane or bed gantry is used to place the segments, dry matching needs to be carried out to ensure that segments are aligned as per required geometry of the structure. After dry matching the segments are to be glued with epoxy from one end to another. PT bars are normally used for temporary pre-stressing. Normally pressure of 3 kg/cm2 is specified to be applied between faces of two segments. After permanent pre-stressing cables are threaded through the ducts and stressed, PT bars or any other temporary external pre stressing are distressed and removed. On an average, erection of 5 spans of around 50 m can be achieved in a month with one set of under slung erection machine in spite of traffic constraint and other constraints. 4.1.2 SSME using special underslung girder and camel gantry This modern SSME is in a way similar to that of erection method adopted for Krishna bridge at Deodurg except that in place tubular staging underslung gantry is used and in place of travelling 691
Figure 30. Transportation & handling of segments for SSME in underslung.
Figure 31. Top right: first time application of special underslung for SSME inYamuna bridge at Delhi (2000).
gantry, sophisticated camel gantry/segment handler is used. This sophisticated and improvised version over Krishna bridge at Deodurg was adopted first in India for SSME for Yamuna bridge at Delhi (Fig. 31) in 2000. In this method of SSME (Fig. 32), the precast segments are erected by means of a 3 span long underslung type erection girder. The erection girder consists of two structural steel box girders, which are supported over RCC/steel bracket one on either side of bridge decking projecting from pier. These erection girders may be supported on foundations also if sufficient place is available. Specially designed erection gantry (Camel Gantry) is used to feed the segments to Launching girder. Sufficient capacity trailers are used which travel over already erected deck to bring precast segments below Camel Gantry. Segment carriages with hydraulic jacks (for final alignment purpose) are placed on the Launching girder over which Camel Gantry places the Precast segment. After placement of all the segments and aligning them to proper geometry epoxy gluing followed by temporary prestress is carried out. If there are any in situ joints between pier segment and penultimate segments, the same are then to be concreted. After assembly of all the segments, permanent prestressing is carried out to make the span self supporting. The Launching girder is then auto launched to the next span by means of winches and Camel gantry is brought to the next feeding location. The main components of the launching system are: – Launching Girder: For placement of precast segments till permanent stressing – Camel Gantry: For feeding the segment to Launching Girder – Required Capacity Hoist ( connected to Camel Gantry): For lifting the segment from Trailer and then placing it on Launching Girder – Wheel Block Assembly: For forward movement of Launching Girder 692
Figure 32.
SSME using underslung girder & over head segment launcher.
– Outrigger: To support the Launching girder during shifting of wheel block assembly – Required capacity Hydraulic Jacks mounted in Wheel Block: Supports Launching Girder during erection of Segments – Segment Carriage – Supports individual segment during erection and also used for geometry adjustment during dry matching – Trolleys: For shifting the wheel block assembly – Winches: For pulling the launching girder during auto launching operation and for longitudinal movement of segment carriages – Front and rear boggies of camel Gantry: For forward movement of Camel Gantry – Required Capacity Hydraulic Jacks (connected to front leg of Camel Gantry) : For transferring reaction on Diaphragm segment during placement of segment on launching girder and also to transfer camel gantry load on wheel boggies before commencing forward movement of Gantry – Trailer of required capacity: For transportation of Precast segment from casting yard to erection site. 4.1.3 SSME using overhead gantry Overhead gantry is a steel girder specially designed to hang the assembly of segments of the entire span and spanning over minimum 2 consecutive piers. This temporary girder has self-launching capability with mechanized launching system to be able to move longitudinally over the successive piers. Overhead head gantry for SSME was perhaps first time introduced to Delhi metro viaduct constructions and there after the technology was picked up extensively even for road elevated structure and bridges later (Fig. 33). The main components of Overhead gantry (Fig. 34) are: – Main girder could be single or two. – Carriage trolley able to pick up and handle segment. Sometimes used also for launching the main girder. – Launching system like hydraulic jacks, chain pully blocks, etc. – Main supports like Front and Rear legs. – Auxiliary supports like Rear leg, Front leg and intermediate legs if required. – The bracket to support the front leg if there is no space over the pier cap. – The assembly of beam and Hangers to handle the precast segment with the carriage trolley – Suspenders to support the segments. 693
Figure 33. SSME using over head gantries.
Figure 34. Components of overhead gantry for SSME.
Figure 35. Typical erection of overhead launching girder.
Typically overhead launching girder itself) has to be erected first to enable the segments erection subsequently. Typical erection sequence of launching girder could be (Fig. 35): – – – –
Temporary trestles placed on footings for erecting boxes of launching girder. Then front support has to be erected on Front pier (Pn + 2). Middle support has to be erected on built up stool provided on Centre pier (Pn + 1). One by one box of launching girder are erected on temporary trestles and joined to each other by splice joint. – After joining all the boxes slider Beam are to be erected, – Temporary trestles are to be removed.
Once the launching girder itself is erected, SSMC using overhead launching is done typically in 5 stages. In the 1st stage, the segment is brought on trailer and lifted with help of crab hoist provided in launching girder.Then segment is hung to the slider beam and shifted to its position 694
Figure 36.
1st stage, the segment is brought on trailer lifted to over head girder with crab.
Figure 37.
Segment are positioned in 2nd stage.
with help of slider beam.In the next stage, all the segment are positioned by procedure given in stage-1. In the 3rd stage, the following activities are carried out: – – – – – –
Dry matching is carried out between segments. S1 segment is positioned accurately as per required level and alignment. Remaining segments are shifted back by 200 mm for applying epoxy. Then epoxy is applied to the surface S1 and S2 segment. Segment S1 and S2 are joined by applying force with help of attachment frame and tension bar. Similarly all the segments are joined with epoxy by applying force with help of attachment frame and stressing bar. – Permanent pre stressing cables are stressed. – Load is transferred from slider beam to pier cap with help of span releasing jack. – Segments are released from the suspenders of slider beam. In stages 4 and 5, principally auto launching of launching girder itself is implemented for erection of next span in various steps as below: – All the sliders beam are moved back at the rear end of the launching girder. – Rear support is shifted and activated on S1 segment. – Middle support is deactivated so that load is transferred on rear support and middle support is free to move. – Middle support is shifted and positioned near front support. Middle support is activated at that position. 695
Figure 38.
3rd stage, dry matching to release of suspenders.
Figure 39.
Finally auto launching of the girder.
– Telescopic leg of front support is closed so that front support is deactivated and load is transferred on middle support. – Launching girder is pushed forward with help of longitudinal jack (having stroke of 1000 mm) provided on middle support, until it reached near next pier location. – Front support is activated by opening the it’s telescopic leg. – Sliders are brought to the front for erecting the segments of next span. There could be minor variations in SSME using overhead launching gantries, but the principle of erection is the same as explained above. The minor variations could be that there could be single or two girder in the cross section, the girders could be plate girders or truss girders, the front and intermediate legs could be supported on the piercap itself or the temporary brackets attached to pier cap or on the diaphragm segment of the superstructure depending upon the availability of space on the piercap. 4.2 Cantilever Method of erection (CME) CME has principally three variations VIZ Balanced cantilever method of erection (BCME), Free cantilever method of erection (FCME) and Progressive placement method of erection (PPME). Further in BCME three types of launching equipment can be used VIZ, form travelers, under slung launching girder and overhead launching girder. In BCME (Fig. 40), the segments are erected by pair, symmetrically on either side from the pier, each of them being cantilevered from the preceding one. The erection of a pair of segments can be 696
Figure 40.
BCME where cantilevers are erected in pairs progressively.
Figure 41.
First precast segmental bridges in India by BCME.
Figure 42. Use of bed gantry & floating crane for BCME.
done either simultaneously or one after the other. Equipment used for handling and delivering the segments could be mobile crawler or barge mounted cranes, Lifters sitting on each of the last pair of segments already erected, various type of launching girders, barges, bed gantries. 4.2.1 Balanced Cantilever Method of erection (BCME) In India, for the first time BCME (Fig. 41) was extensively adopted for Ganga bridge at Buxer and Narmada bridge at Zadeswar in 1971 and later around 1977, the same technology was adopted for Ganga bridge at Patna too. The erection equipment were bed gantry on land and barge mounted crane in water (Fig. 42). The segments were hoisted with the main gantry and placed over power driven trolleys, running on the completed deck. The segments were picked up by floating crane in the water portion, and transported to their erection location. Before its final erection, the segment was checked for its perfect matching and alignment with the previously erected segment. The segment was held with the gantry and cables were threaded. A specially developed epoxy formulation, successfully tested in R&D wing of the company provided an interfacial bonding surface with the match cast segments. The segment was held in position with the help of a system of hydraulic jacks placed over deck slab and soffit slab, till the cables were stressed and the segment was attached with the previous segment. 24/7 HT cables were used and the pre stressing was carried out after the symmetrical segments of T arm were erected. The complete system for match casting, transporting and erecting the segments in position was developed in-house by a team of engineers and was applied successfully in 1971 itself. In Ganga bridge at Patna the segments for land spans were erected using a bed gantry (Fig. 42) where as water spans were erected using a floating crane as explained earlier. Precast segments, after matching with epoxy interface, were stressed at the deck and soffit levels with high tensile bolts. This is required for supporting the segment temporarily, till the main pre stressing cables terminating at the section, were stressed. The entire temporary stressing system was developed to ensure uniform compression on the epoxy joint. Another variation of BCME where lifting beams are used as depicted in Figure 43, has been perhaps tried out first time in India for 2nd Vivekanada extradosed bridge in Kolkotta. Lifting 697
Figure 43.
BCME using lifters. Right: 2nd Vivekananda bridge, Kolkotta.
Figure 44.
Use of underslung girders for BCME in 2nd Narmada bridge at Zadeswar.
Figure 45.
Use of over head girders for BCME.
beams is a simple way to erect balanced cantilevers. The system is composed of a steel structure sitting and fixed to the segments already erected. The segments are fed under the cantilever by trucks, barge or other equipment and lifted directly to the right position. The lifting operation can be carried out either by winches or strands. BCME using under slung launching girder (Fig. 44) was perhaps first time successfully executed for 2nd Narmada bridge at Zadeswar (1997–2000). It consisted of 13 main spans of 96.2 m with 4.63 km of approach roads and was built in a record period of 32 months The two-lane deck consists of a 7.5 m wide carriageway with 1.5 m wide footpaths on either side and is located 29 metres downstream adjacent to the existing bridge. The self launching overhead gantry was used for the first time for the Viaduc d’Oléron, built in 1966. With a total length of 2862 m it was at that time the longest prestressed bridge in France. The hollow box cross-section had a width of 10.62 m and a height of 2.50 to 4.50 m. The balanced cantilever method and epoxy joints were used for gluing. The method basically consists in installing a steel girder with two legs (one in the centre and one at the back end), the length of which is somewhat greater than the maximum bridge span. The centre leg is on the first segment of the new symmetrical deck cantilever to be constructed and the back leg is on the end of the deck that has already been built (Fig. 45). The girder is equipped with a trolley, which runs on the lower chords of the girder, enabling the successive segments to be installed. The steel truss girder is built-in two tunnel legs which provide the opening for the segments to pass through. The central leg and rear leg are supported on steel beams, which enable the necessary adjustments to be made according to the curve and inclination of the deck. A temporary front leg enables the first segments of each 698
Figure 46.
First bridge to be constructed by FCME in India for PSB.
Figure 47. Sequence of construction for FCME in Pragathi maindan bridge.
double cantilever to be installed. BCME using overhead launching girder has to be tried out yet in India. 4.2.2 Free Cantilever Method of erection (FCME) The first bridge to be built by free cantilever method of erection (FCME) using precast segments is Choisy-le-Roi Bridge in France (1962). This was constructed using a large capacity floating crane which carried and installed the segments symmetrically on either side of the piers. FCME was first time adopted in India for Pragathi maidan extradosed bridge at Delhi. This extradosed bridge had a total length of 196 m with the central span of 93 m having sharp radius of curvature of 300 m. The lateral spans consisting of U trough cross section were erected by SSME using overhead launching system. The segments for cenral free cantilevering spans were brought on top of already erected side spans and the specially designed segment launchers which has the arrangement of rotating the segments, lifting and lowering segments in addition to the facility of telescoping and holding the segment for free cantilever erection. The sequence of free cantilevering and their illustrations are depicted in Figure 47. Progressive placement method of erection (PPME) is similar to FCME, the difference being that the structure is continuously built from one abutment to another by placing precast segments regularly which are transported on already erected deck using trailors. The specially designed segment launcher as used in Pragathi maidan extradosed bridge above is used for free cantilever progression. In this method, as the entire span has to be cantilevered during erection, a temporary staying mast (Fig. 48) is used for erection purpose. The temporary staying mast which lets the segments through consists of two vertical transversally braced steel legs, on which regularly spaced 699
Figure 48.
PPME using temporary mast in CMC.
hydraulic jacks allow the tensioning of the cable-stays by moving back their anchor plates. The mast is transferred from one pier to another on a trailer drawn by the multi-wheeled trailer used for segment transportation. 5 CONCLUSIONS Precast prestressed segmental is tailor made for long multi spans and fast track bridges. As the entire superstructure is cast away from actual site location, minimizes the hindrance to traffic & inconvenience to public in urban environment.Prefabrication being done In factory like environment of casting yard better control on quality and dimensional tolerances can be achieved. Prefabrication of superstructure segments are done independently as work on foundations progresses that reduces overall completion time. Cost effectiveness is inherent in terms of total construction period, less labour and repetitive use of formwork etc. Future trend could be the combined use of concrete and steel to reduce weight consisting, concrete core segment, external strut and rib of steel and Precast panels as deck slab on ribs and struts. REFERENCES Precast Segmental Bridges. Guide to Good Practice for structural and site engineers, Prepared by fib Task Group 10.3, Commission 10. Combault, J. 2004. Precast Concrete Segments for Bridges Fabrication and Assembly Fundamental Details, fib Symposium 2004 on Segmental Construction in Concrete, November 26–29, 2004, New Delhi, India. Raiss, M. 2004. Developments in precast segmental and incrementally launched bridge Construction methods, fib Symposium 2004 on Segmental Construction in Concrete, November 26–29, 2004, New Delhi, India. Heggade, V.N. 2004. Segmental Construction-some issues, fib Symposium 2004 on Segmental Construction in Concrete, November 26–29, New Delhi, India. Dharap, V.M. & Joshi, G.P. 2004. Evolution of pre cast segmental construction for bridges and fly overs in India, fib Symposium 2004 on Segmental Construction in Concrete, November 26–29, New Delhi, India Heggade, V.N., 2010. Bridge aesthetics-Some issues, 3rd International fib Congress 2010, Think Globally & Build Locally, May29–June02, 2010, Washington.D.C. Jatkar, M.V. 2011. Extra-dosed Bridges for Delhi Metro Projects, FIB days-2010, An International Conference, Jan 14–15, 2011, New Delhi. Heggade, V.N. 2012. Construction Technologies for concrete Bridges, fib- days 2012, International conference, 10–11 January 2012, Chennai. Heggade, V.N. Aesthetics & Creative aspects of bridges, Invited lecture by ICI-Chennai, IIT Chennai. Heggade, V.N., 2014. A construction case study of Kolkota Metro EWE-01 Corridor, ING-IABSE Seminar on Elevated Transport Corridors, Mysore, 27–28 June 2014, pp 236–251. Heggade, V.N. 2014. Bridge on river Kosi by shortline precast segments and special under slung erection technique, International seminar on recent trends in segmental construction and retrofitting of bridges & flyovers, IIBE in collaboration with BRO, 5–6 September 2014, New Delhi.
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Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
Construction of Panipat Elevated Expressway on NH-1 on BOT basis P.N.S.S. Sastry Larsen and Toubro Infrastructure Engineering Limited, Chennai, India
ABSTRACT: Time is the essence of “Build Operate and Transfer (BOT)” type of contracts. Construction of long elevated corridors in urban busy traffic environment is challenging one in BOT projects especially in meeting the construction time schedule. Panipat Elevated Corridor on National Highways (NH)-1, India is one such project. The total time frame for completion of elevated corridor and complete service and main carriage At-Grade was 36 months. The elevated corridor falls in the road project having a six-lane facility, covering a length of about 10 km from km 86.00 to km 96.00 through Panipat city on NH 1, India. The elevated corridor is 3.4 km long with dual carriage way. The over-all deck width of each carriage way is 11.75 m. From the economics and construction point of view Pre-Tensioned I Girder with cast-In-place deck slab was adopted. Typical five span continuous modules of 29.5 m + 31.0 m × 3 + 29.5 m are adopted for the project. All continuous spans are supported with two POT-PTFE bearings beneath each diaphragm. Hammer head piers with Pile foundations is adopted for substructure and foundations respectively. Among various structural component cast-in-place deck slab takes long duration with minimal staging arrangement. In this situation, generally contractors will opt for permanent shuttering, which is an additional cost and increases dead weight and quantities in substructure and foundation. To achieve minimum cost solution and increase in speed of construction, for the first time in India, partial precast deck slab is used for such long corridor. The paper presents “Design and Construction Methodology” adopted
1 INTRODUCTION National Highways Authority of India (NHAI) awarded Construction of Six Lane access controlled highway from Km 86.00 to Km 95.700 and 2 lane side roads of NH-1 including elevated structure covering Gohana road junction to Skylark tourist complex at Panipat in the State of Haryana on Built Operate and Transfer Basis to Concessionaire M/s Larsen and Toubro Limited (L&T). The challenge before L&T was to meet the scheduled time of construction, while maintaining the dense traffic movement at grade during construction, where traffic snarls, congestion bottlenecks and crowded market places is order of the day. Once the project is thrown open, expressway will provide comfortable, hindrance free smooth sail-through at Panipat city.
1.1 Salient feature of the project Major portion of the scope of work of Concessionaire are – Access controlled 6-lane highway in the reach km 86.000 to km 96.000 – Separate 2 lane peripheral road with paved shoulder on either side of access controlled 6-lane highway for the local traffic. – Three underpasses to accommodate 4-lane divided carriageway. – Construction of 2 new minor bridges and road furniture including toll plaza. 701
Figure 1.
Deck cross section.
2 ELEVATED PORTION 2.1 Choice of basic spans Overall cost of superstructure, substructure and foundation including construction methodology per meter run governs the span configuration. Since deep pile foundations are required, after studying several alternatives a typical module of five span continuous unit of 29.5 m + 3 × 32 + 29.5 m appeared to be the best choice. This configuration also meets obligatory span requirements at important road crossings and also provides maximum repetitions. 2.2 Deck cross section The elevated two lane carriage way for each direction shall be provided on separate substructure and foundations as per concession agreement. Deck cross section is shown in below Figure 1. 2.3 Choice of basic structural forms for superstructure In BOT projects, early completion of project drives all the engineering decisions. Therefore, precast superstructure is the most preferred option. Two options were studied namely Precast I girder cum cast in-place deck slab and Segmental Precast Box Girder. From Transportation and Erection point of view Pre-cast I girder was preferred over Precast Segmental Box Girder. For speedy production of girder, Pre-tensioned I girder was preferred over Post-Tension girders, moreover it is economical as 1200 girders were to be cast in short time. 2.4 Feature of the typical span configuration The deck cross section is divided into two independent carriageways. Each carriageway consists of six girder of 2.0 m overall depth. The girders get connected at the top by deck slab and at the ends by diaphragms. The deck slab and diaphragms are cast-in-place simultaneously with the permanent bearing is position. The behavior of the superstructure is simply supported on temporary bearing up to stage of casting of deck slab and end diaphragms. For all subsequent loads such as superimposed dead loads and live loads, the behavior is changes to continuous deck of five spans. In simply supported condition only precast girders are operative and full composite section is effective for subsequent loads. In conventional method of construction, in-situ deck slab is cast over either on sacrificial shutters or removable shutters which will either add additional dead weight on girders or add considerably to the time taken to cast the deck slab, hence the idea of partial pre-casting of deck slab was adopted. 702
Figure 2.
Isometric view of precast slab with truss.
Figure 3. Temporary bearing for deck panel.
For the first time in India for such a long flyover Precast-Deck slab panels have been used as load bearing members in combination with cast-in-situ slab. Partial precast against full precast of deck has two advantages namely lesser weight from handling and erection. Thus longer panels can be used and minimize the number of in-situ stitch especially over piers. The in-situ concrete topping fills the gap over beams and shear connectors on the precast girder. The deck is designed as full composite section between pre-cast panel, in-situ topping and the supporting beams (longitudinal girders).The in-situ concrete topping provides flexibility to adjust roadway profile and for differences in beam elevations. 2.5 Design of deck slab The non-composite panel which is a steel truss with concrete slab of 75mm thick at bottom is designed for stresses occurring only in the precast units resulting from lifting, transportation, wet concrete and live load due to construction working personnel. The space truss arrangement with re-bars as shown below in Figure 2. Spacing of the trusses is governed by construction loads and permissible tensile stresses in the concrete. When top in-situ concrete is poured, the truss elements of the precast panel act as shear connectors to ensure monolithic slab action. To ensure performance one composite panel was tested for service loads. Truss web members acted as shear transfer reinforcement. Precast panels are supported on temporary bearing pads to achieve the cross slope. Refer below Figure 3. The in-situ portion of deck accommodates reinforcement for negative moment over supports for both transverse direction as well as longitudinal direction. 703
Figure 4.
Erection of Precast Panel is in progress.
Table 1. Construction schedule. Age (days)
Activity
0 day 3–5 days 40–45 days 60–70 days >90 days
Casting of Girders in Yard Transfer of Pre-stress to Girder Launching of Beams and Casting of deck Shifting of Supports (temporary Supports to Permanent Bearings) Laying of SIDL and allow live load
Figure 5.
Casting of girders in yard.
2.6 Design of girders The pre-tensioned girders are designed for stage by stage constructions and continuous five span unit. The superstructure is analyzed as grillage model for loading coming after composite action. The portion of the superstructure between the diaphragms is designed as fully pre-stressed a member with zero tensile stresses during service conditions. The part portion of deck over pier is designed as reinforced concrete member. The longitudinal reinforcement in the deck slab is designed for superimposed dead load, live loads and time depended effects such as differential shrinkage and creep. The superstructure is also checked for differential settlement of 6mm and temperature gradient as per IRC:6-2000. The pre-tension girders are monolithic with Reinforced concrete cast-in-situ diaphragms, at this junction prestress is absent due to transmission length of strands and continuity is achieved through 704
Figure 6. Erection of girders with gantry.
Figure 7. Tying of reinforcement over erected precast panels and girders.
reinforcement. At interface of girder and cast-in-situ diaphragm positive moments develop due to creep as well as live load effects on remote spans. For positive moments adequate reinforcement is provided to avoid opening of the joints during service.
3 CONSTRUCTION PROGRAM For composite construction of the type in which the girders and deck slab/diaphragms are cast at different time, the construction program has a profound influence in the design of superstructure. 3.1 Casting of girder All girders are cast at yard by using long line bed. At any given time five girders are produced on a single bed. To maximize the production and early transfer of prestress stream curing was adopted. At the time of transfer of prestress minimum 40 Mpa strength was achieved. Totally 1200 numbers of I girders precast in a nearby casting yard were transported to site through specifically modified trailers and erected using gantries and heavy lifting cranes. 705
4 CONCLUSIONS By modifying the construction scheme from conventional method of construction of deck slab to partial pre-casting there is substantial saving of time and cost of construction. This innovative method helped L&T to execute the project most efficiently ahead of the construction schedule. From the date of appointment M/s Larsen & Toubro completed the elevated corridor of 3.6 km within a time frame of 18 months which includes design, design approval from third party and construction activities.
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Mold for full span method Moonkyo Kye BTU Korea, South Korea
1 INTRODUCTION 1.1 Overview and current conditions of full span method Full span method is an engineering technique to create a casting yard in a field site where girders are produced in full span without separation, and then move them onto a bridge pier with the use of such equipment as launching girder. Therefore, in the case of full span method, mold and equipment for installing girder (e.g., straddle carrier, tire trolley, launching girder, and barge crane) are critical. It is fair to say that selecting and operating such equipment determines failure or success of construction. Full span method has strong and weak points as follows: Strong points:
Weak points:
Girders are produced in a casting yard so that it is easy and safe to control quality. Simple and repeated working leads into duration shortening (in the case of pre-tension, it takes 2 days to produce one girder). The longer a bridge is and the higher a bridge is, the better the method is. Thanks to high-cost mold and equipment, it requires a lot of initial investment in purchasing the equipment. If bridge linearity changes a lot, it may be impossible to apply the method.
In Korea, since full span method was applied to Incheon grand bridge completed in 2009, the features and strong and weak points of the method has widely been known. Up to now, three projects and nine field sites used the method and successfully completed construction. The method is applied or will be applied to the three foreign sites in which Korean engineering & construction companies participate. – Kuwait Causeway project: 60 m & 40 m span – Kuwait Doha link project: 40 m span – Brunei Temburong project CC2: 50 m span 1.2 Components of full span method mold The basic components and quantities of full span mold are presented as follows: – – – – –
Outer mold: 1 set Bottom mold: 2 sets Inner mold: 2 sets Bulkhead mold: 2 sets Thrust head: 1 set (only pre-tension)
Although there are basic components, the operating method and quantities of mold can be changed differently depending on field conditions, the circumstances of a casting yard and a storage area, and duration and the lifting and transporting method of completed girders. Based on the completely constructed Honam high speed railway site and a foreign construction site under construction, the case of the full span method which is applied to a multi span large bridge will be introduced to provide more understanding of full span mold. 707
Figure 1.
PSC box girder section of Honam high speed railway .
Figure 2.
Basic mold sections.
2 INTRODUCE OF FULL SPAN MOLD FOR HONAM HIGH SPEED RAILWAY SECTION 2-2 2.1 Main specification of box girder • • • •
Height = 3.5 m (center) Width = 12.6 m Span Length = 35 m, 30 m, 25 m Maximum box girder weight = 960 ton
2.2 Components of mold The components and quantities of full span mold are presented below, and the total weight of Mold is around 484 ton. 2.3 Typical Mold operations (a) Outer Mold Folding System Hydraulic system and folding brace were used to implement the system which makes an 35 m outer mold open and close by one button. Fig. 4 illustrates Outer Mold Folding System. When the two hydraulic cylinders installed in the ends of 35 m Outer Mold start to work, the braces supporting the mold operates at the same time and thereby the whole Mold with 35 m in length is opened and closed. 708
Figure 3. Weight table of mold.
Figure 4.
Figure 6.
Figure 5.
Outer mold folding system.
Inner mold folding & pulling.
Launching system.
(b) Inner Mold Folding System The easy and convenient way of drawing Inner Mold from the inside of Box Girder which completed casting and curing processes is a key to duration shortening. As shown in Fig. 5, with some button operation, the Inner Mold with hydraulic system can be folded small enough to be passed freely through the diaphragm’s narrow halls. (c) Launching System It is a system to draw the completely produced Box Girder and Bottom Mould weighting 960 ton out of a casting yard in order to lift them with construction equipment. In the system, the pipes and hydraulic cylinders buried in the floor at the time of making a casting yard are used to transport heavy objects at a low cost. 709
2.4 Production procedure
Figure 8.
Figure 7.
710
Figure 9.
Figure 10.
Figure 11.
Figure 12.
Figure 13.
Figure 14.
Figure 15.
Figure 16.
711
3 OTHER FULL SPAN METHOD PROJECT 3.1 Kuwait Doha link project
3.2 Brunei Temburong project CC2
4 INTRODUCE CONCLUSION Many matters need to be taken into account in order to design, manufacture, and operate an efficient full span mold. Among them, information sharing and work cooperation between a bridge designer and a mold manufacturer is especially important. Let’s take an example of Brunei Temburong project. A mold designer participated in the bridge design step and thus applied sensitive elements of a mold, such as a diaphragm hole size and an inner slab hunch size, in the design. The participation, though limited, helped to design full span method a superstructure suitable for full span method. Such a way will be able to reduce trials and errors in future construction and make duration shortening effect. REFERENCES Bridges for High speed railways. Edited by Rui Calcada, Raimundo Delgado, Antonio Campos e Matos. K.D. Kang & S D. Suh. Experience with the precast span method on the Korea high-speed rail project.
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Special foundations and geotechnical site investigations
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Offshore pile driving foundations monitored by PDA® Test at Puente Nigale M. Rojas, I. Miquilena & A. Souza Odebrecht Venezuela S.A., Venezuela
ABSTRACT: This document presents PDA Tests performed as part of monitoring and control of Pile Driving at Offshore Foundations in the construction of Puente Nigale. The project consists of a 12 km cable stayed link for both road and rail traffic located in Maracaibo, eastern Venezuela and will support the existing 8.6 km link, Puente General Rafael Urdaneta by releasing traffic load. The design proposes a curved alignment with two separated but parallel superstructures, carrying road and rail respectively. The foundations and pile caps will be common for both superstructures. At the present time, 9 Piles with an inner diameter of 1.8 m were monitored as part of control of driving activities and verification of capacities using Impact Hydraulic hammers obtaining a set-up of 1.14 for all piles after PDA Test. Driving equipment was verified by reaching bearing capacity and minimum tip level established by designers.
1 INTRODUCTION 1.1 Project description The urban and industrial growth in the area, the need to establish new routes and the idea of releasing traffic load from the original bridge engender the construction of a second overcrossing on the Maracaibo strait, Puente Nigale. The project consists of a 12 km link for both road and rail traffic connecting Santa Cruz de Mara on the east coast to Punta de Palmas on the west coast, subdivided in two viaducts, a low level bridge, two elevated bridges and cable-stayed bridge over the Maracaibo lake navigation channel. In case of the Foundations, piers with 18, 21 and more piles will be built to support road and rail superstructures. In total, 128 piers will be executed in offshore conditions, including two pylons at both sides of navigation channel. It was planned the execution of the foundations located approximately at center of alignment, P54 to P64, at the Low Level Bridge section. Those piers require driving 21 steel pipe piles with 1.8 meters of inner diameter. The target depth fixed as tip level ranging from −43 to −65 meters below Medium Level of Maracaibo’s Lake (MLMN). The minimum Pile Capacity in compression for these piles is 27 MN according to the geotechnical design report submitted by designer. 1.2 Geotechnical conditions At the pier locations (P54 and P55) were executed 4 boreholes (two per Pier) and two CPTu Tests. Information available at submission of Geotechnical report including all data from basic lab tests and SPT Tests executed in the geotechnical boreholes. SPT Test were executed each 1.5 m and disturbed or undisturbed (according to the type of soil founded) samples were recovered between SPT’s for laboratory tests. Design Profiles at P54 consist of approximately of 4.0 meters of silt with sand (ML) just below the seabed level followed by 21 meters of normally consolidated lean clay (CL1). This is underlain by a 6 meter thick layer of fat clay (CH). From approx. 29 below the seabed level decomposed and soft siltstone (CL3/CL4) has been encountered. The siltstone has 715
Figure 1.
Idealized Geotechnical Profile from P54 to P64.
been assumed continued to lower depth. At P55 were found approx. 4 meters of silt with sand (ML) just below the seabed level, followed by 18 meters of normally consolidated lean clay (CL1). This is underlain by a 5.5 meter thick layer of fat clay (CH) and 4.5 meter thick layer of dense to very dense silt sand (SM3). From approx. 32 below the seabed level decomposed and soft siltstone/claystone (CL3/CL4) has been encountered. The siltstone/claystone has been assumed to continue to great depth. Idealized Geotechnical Profile at these Piles is show in figure below. Calculations of Axial Bearing Capacity were carried out with CPTu data by direct CPT methods (AASHTO LRFD Bridge Design Specifications, 5th Edition 2010) and, as part of construction specifications, PDA Test was determined as verification technique of End Bearing Capacity, including CAPWAP analysis after collection of data on site. 1.3 Piles design Piles for the section P54-P64 consist in driven ID 1.8 m steel piles up to 65 m long and a wall thickness varying between 25.4 mm and 31.75 mm. In all piles, the uppermost 9.3 m are filled with a reinforced concrete. In order to verify the design, the uppermost 5 m of the piles are assumed as a concrete cross section alone, and in this area all load is taken by the concrete and reinforcement. A transition zone with shear rings is placed from level −3.5 m below pile cap and 4 m further down. In the transition zone either the reinforced concrete or the steel pile is assumed to provide the necessary structural capacity. The full steel capacity is verified from −5 m and down. At level −9.0 m the pile is assumed as a steel cylinder cross section alone and therefore all loads in this area are taken by the construction steel. Design states that, in general, the highest utilizations are found in the corner piles and their nearest neighboring piles, as indicated in the figure below. 1.4 PDA Test Pile Dynamic Analysis is a well-known technique for bearing capacity verification of a wide range of piles (concrete, steel, wood) and the guidelines for its execution are established in ASTM D 4945 – 12 Standard Test Method for High-Strain Dynamic Testing of Deep Foundations. In Nigale fixed link were implemented the methodology established by ASTM D 4945 – 12 with execution of dynamic test with equipment provided by PDI, Inc: Pile Driving Analyzer or PDA. The devices 716
Figure 2.
Layout of design approach used by the designer.
Figure 3. lowers.
Plan view of a pile cap. Darker piles have the highest utilizations, meanwhile lighter ones have
consists in a PDA model PAX-8 which is capable of carried out 4 and 8 channels tests with End Of Driving (EOD), End Of Restrike (EOR) and Beginning Of Restrike (BOR) monitoring. In total, 19 EOD monitoring were executed and 18 tests as EOR/BOR as well. 2 INITIAL DRIVING RESPONSE – EOD TEST As result of a detailed analysis of geotechnical conditions on site, a 3 phases of installation were identified: 1) a self-weight installation of piles (because until 20 to 28 meters of depth SPT N60 values reported by geotechnical boreholes were less than 10) retaining a 12 meters section per time on the template installed on the barge; 2) a vibratory driving with a PTC 100 unit, generally this driving phase reached 32 to 33 meters of depth and; 3) Impact hammer with a single hydraulic unit (CG300 - CGL 590) until target depth (43 to 45 meters into the ground). At beginning of activities, a hydraulic hammer BSP CG300 with 20 Tons of drop weight was used. It was recommended by designer the use of a unit with at least 500 KN-m as minimum capacity for driving at this site. In case of BSP CG300 a total of 294 KN-m could be applied as maximum capacity noticed by provider BSP in their technical brochures. Motivation for using that 717
Figure 4. Comparison of signals from EOD monitoring with both driving equipment (CG300 upper and CGL5900 lower) at Pile 17 (at north) and Pile 19 (at south).
Figure 5. Comparison of signals from BOR monitoring with both driving equipment (CG300 upper and CGL590 lower) at Pile 10 (at north) and Pile 11 (at south).
unit was due to unavailability of a bigger one (CGL 590, ordered by Odebrecht) as recommended by design reports and technical notes. As result of this issue, in the first stage of offshore driving works, an ‘activated capacity’ were reported after CAPWAP analysis reported by GRL Engineers (Hussein et al, 2002). This condition is reached when a hard driving with an underestimated unit (hydraulic or other) is used as driving equipment. In those CAPWAP reports a superimposed capacity of 22.5 MN (as minimum on Pile 3 at P54) to 26.4 MN (as maximum at Pile 10 on Pier 55) were reported, recommended to use a bigger unit (i.e. CGL590) for final driving and further verification of end bearing capacity after a BOR monitoring of piles. 718
Figure 6.
ETR – Energy Transfer Ratio for unit CG300.
Figure 7.
ETR – Energy Transfer Ratio for unit CGL590.
3 PERFORMANCE OF DRIVING EQUIPMENT After completion of driving with a major unit, CGL 590 from BSP with a 590 kN-m as maximum energy could be applied per stroke; capacities higher than 27 MN were reached. In general, according to the maximum energy rated by each unit (CG300 and CGL 590) more than 50% of time the energy applied exceeded 70% of maximum energy for both hammers. In figures 6 and 7 is shown frequency distribution analysis from all monitoring recorded by PDA model PAX-8. 4 ANALYSIS OF PILE SET UP After complete installation, a setup phenomenon was analyzed. In case of piles at P54 and P55, an average value 1.14 after 8 days in average was observed. In case of piles 3 and 10 on P54 setup analysis was made with superimposed capacity as reported by CAPWAP report. 719
Figure 8. Set Up and Days after EOD monitoring for Piles at P54 and P55 (in case of Piles P54 10, P54 3 and P54 17 setup was calculated with the ‘superimposed’ capacity).
In figure 8 is shown the comparison for this values reported at piles P54 and P55. 5 CONCLUSIONS From Pile Dynamics Test at piers 54 and 55 in offshore activities for Puente Nigale project was found that unit CGL590 has enough capacity for mobilized total resistance established by designer. The Setup factor for those piles was 1.14 after 8 days in average; those values could be adopted at least for Piers 56, 57, 58 and 59 as references values for planning activities and considerations for PDA Tests which will be performed in the piles of those piers. It is highly reccommended to perform EOD and BOR monitoring at least on opposite corners of piles for each pier as indicated by construction specifications. In the same way, is recommended increase capacity for vibratory driving, reducing the use of unit CGL590 consequently. REFERENCES Axelsson, Gary. 2000. Long-Term Set-Up of Driven Piles in Sand. August. Doctoral Thesis. Royal Institute of Technology, Sweden. Frederiksen, A. 2013. CW-GN-RE-9-01062 Puente Nigale, Design Report – Low Level Bridge, P54-P64. Denmark. Hussein, M.H., Sharp, M. & Knight, W.F. 2002. The Use of Superposition for Evaluating Pile Capacity. Deep Foundations 2002, An International Perspective on Theory, Design, Construction, and Performance, Geotechnical Special Publication No. 116, O’Neill M. W., and Townsend, F. C. Eds., American Society of Civil Engineers: Orlando, FL; 6–21. Komurka, Van E.; Alan B. Wagner and Tuncer B. Edil. February 2003. A Review of Pile Set-Up, Proceedings of the 51st Annual Geotechnical Engineering Conference, University of Minnesota, St. Paul, Minnesota, pp. 105–130. Likins, G. E., Rausche, F., September 2008. What Constitutes a Good PDA Test? Proceedings of the Eighth International Conference on theApplication of StressWaveTheory to Piles 2008: Lisbon, Portugal; 403–407. Lutful Khan, PhD, PE Kirk Decapite. February 2011. Predicting of Pile Set-up for Ohio Soils. PDA-W Manual of Operation. February, 2009. Pile Dynamics, Inc. Ohio. Rausche, F., Nagy, M., Likins, G. E., 2008. Mastering the Art of Pile Testing. Proceedings of the Eighth International Conference on the Application of Stress Wave Theory to Piles 2008: Lisbon, Portugal; 19–32, September. Weng Ng, Kam. 2011. Pile Setup, Dynamic Construction Control, and Load and Resistance Factor Design of Vertically-Loaded Steel H-Piles. Iowa State University. Ames, Iowa.
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Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
Ceira bridge foundations: Combined Micropile and Footing Foundations (CMFF). Static load tests J.M.S. Cruz & M.S. Neves LCW Consult, Algés, Portugal
S. Gil Mota-Engil, Linda-a-velha, Portugal
ABSTRACT: The Ceira Bridge, a stretch of the IC3/A13 highway build under the contract awarded to Ascendi within the Pinhal Interior subconcession, was built over the river bearing the same name. Due to the local geological and geotechnical constrains related with a heterogeneous foundation therefore with varying resistance, a Combined Micropile and Footing Foundation (CMFF) was designed for the main piers P2 and P3. The foundation of each pier is composed by two spread footings each one reinforced with 120 micropiles. The design of the micropile foundation was achieved by a tridimensional finite element analysis. “In situ” tests were also performed in order to validate the applied design methods, mainly static micropile load tests and static load tests on footings reinforced with a single deep element. These tests were important to estimate several mechanical parameters such as: foundation stiffness, lateral friction distribution, load distribution over the spread foundations and micropiles and to confirm the geological and geotechnical conditions previously accessed by boreholes. This paper presents both the design phase with a tridimensional finite element analysis and the subsequent load tests phase.
1 INTRODUCTION The work presented in this article deals with the design of the Ceira bridge foundations. The bridge is located south of the city of Coimbra and spans over the river bearing the same name. The project lies within the Pinhal Interior sub-concession and it is a stretch of the IC3/A13 highway, which opened partially to the public in April 2014. The Ceira Bridge is composed by the main structure and by an access viaduct bearing a total length of 930 m; it was constructed in precast reinforced concrete. Due to the geological-geotechnical constrains related with a heterogeneous foundation therefore with varying resistance and the importance of the permanent loads to be transmitted to the foundations of the piers P2 and P3, there was the need to ensure transmission of part of the loads at depth, which resulted in an increase of the foundation bearing capacity and stiffness and overall uniformity of loads. The adopted solution consists of spread footings combined with arrays of micropiles. 2 CEIRA BRIDGE DESCRIPTION 2.1 General design The bridge is defined geometrically in plan with an “S” shape. It is composed by two distinct types of structures: the main bridge, of the segmental type and with a box girder cross section that allows the Ceira river crossing and the access viaduct (Fig. 1 and Fig. 2). The structural distinction between these two structures is mainly their span dimension and constructive method used in deck construction. 721
Figure 1.
Plan view of Ceira bridge.
Figure 2.
Longitudinal cross section of Ceira bridge.
In the main bridge (582 m length), the river crossing is done 140 m high above the riverbed by a large-span multiple framed structure with a continuous prestressed concrete deck between abutment E1 and pier P4; the main span is 250 m long. The segmental single-cell box girder deck is defined by an upper 26.40 m wide slab and inclined webs with heights ranging from 5.5 m to 14.5 m. Its cast-in-place construction was done with the balanced cantilever method. The access viaduct comprises two prestressed concrete decks, with a double beam cross-section of 345.5 m long. It was constructed span-by-span with an auxiliary steel launching girder incorporating concrete joints at 1/5 of the span length in each movement direction. The main piers (P2 and P3) are about 100 m high and are fitted with a twin blade box girder section spaced 20 m between axes. Continuity is assured between the blades and the deck diaphragms. This geometry ensures adequate stability during deck work construction and safeguard unobstructed views of the valley. 2.2 Bridge foundations The abutments and subordinate bridge piers are supported on shallow foundations due to nearsurface presence of the phyllite bedrock with competent geotechnical characteristics. In the case of the main piers P2 and P3, they are located on the river banks in areas with low to moderately inclined slopes, and exposed to high transverse bending moments resulting from the deck torsion. The geological device at the pier P2 consists of phyllites and phyllites – greywackes with interspersed graphitic greywackes. These rocks exhibit lateral differential weathering degrees and the weathering progression in depth does not follow a typical sequence, i.e. is frequent to observe layers with lower weathering degree (W4 to W3-4 – ISRM 1981) on top of residual soils (W5) or heavily weathered rock (W4-5). The rock with better geotechnical quality (W3 to W2-3/F4) was found at a depth of about 30 m. On the other hand at pier P3 it was observed a completely to highly weathered (W5 toW4-5) phyllite-greywacke rock from surface level to a depth of about 12 m. Below this depth, the rock was found highly to moderately weathered (W4 toW3-4) and with the mechanical characteristics of a soft rock (ISRM 1981: can be peeled with a pocket knife and shows indentation by firm blow with geological hammer); it was also observed that the mechanical characteristics of the rock increases at depth. 722
Figure 3.
Piers P2 and P3: combined spread footing and micropile geometry.
Considering these geological-geotechnical constraints and the high permanent loads to be applied to the foundation, two findings were reached: the need for implementation of eccentric footings from the piers axis in order to uniform the stress diagrams at the foundation; and to design spread footing foundations combined with micropiles capable of guaranteeing the required load distribution at surface and at depth. Each one of the foundations (Fig. 3) is defined by a rectangular spread footing, with dimensions of 21.5 m × 14.0 m and a variable thickness between 2.0 m and 4.0 m; the footing is then connected to a set of 120 micropiles 12 m long. These deep foundation elements were constructed with N80 (API 5A) tubular steel tubes with Ø177.8 mm/9.0 mm and set in an array of alignments spaced 1.25 m and 0.88 m in the most loaded area of the spread footing and 2.50 m in the remaining areas. The drilling diameter was 250 mm and the steel tubes were sealed from the inside and outside. The grouting system used was IGU (Injection Global Unique). 2.3 Structural modeling The design of the structure in order to satisfy the criteria of ultimate limit state and serviceability state was based on the Portuguese legislation and complemented with the Eurocodes. The analysis of the different elements that comprises the main bridge and access viaduct were done using several differentiated models according with the desired objectives, such as: – linear-elastic and nonlinear three-dimensional frame models for the analysis of constructive method and the global bridge behavior; – the shell finite element models for the study of the slabs and deck girder webs behavior; – nonlinear three-dimensional models of piers P2 and P3, to simulate excavation phase, the construction of the micropiles and spread footings and the incremental foundation loading. These calculations allowed the design of the combined foundations (Combined Micropiles and Footing Foundation – CMFF), the evaluation of the stress distribution and the quantification of settlements and rotations. 3 CMFF FOUNDATION DESIGN ANALYSIS 3.1 Design concept CMFF-type foundation is a complex geotechnical system (Fig. 4a) where one can pinpoint several interactions that manages its global behavior (e.g. pile/pile or pile/footing…). The combination of two types of traditional foundations (shallow and deep foundations) is a quick and effective way to resolve constraints associated with the bearing capacity and/or with settlement control in heterogeneous bedrocks with lateral and at depth bearing capacity asymmetry. Overview of the existing State-of-the-Art shows few references to the design of micropiles combined with footings. However, if the element of deep foundation is a pile, then one can easily find several study cases where this combined behavior was studied and well complemented with various 723
Figure 4. Combined Pile Raft Foundation (CPRF): a) Geotechnical interactions; b) CPRF coefficient (adapted from Katzenbach et al. 2005).
Figure 5.
Flow chart for combined foundations design. Adapted from El-Mossallamy (2009).
experimental tests. Katzenbach (2005) also recently presented a methodology for the analysis of Combined Pile-Raft Foundation (CPRF) and the concept of shared coefficients (Fig. 4b). In a simplified manner, this type of foundation can be designed by using the superposition of effects and considering for the shallow and deep foundation elements, resistance values near to their ultimate bearing capacity. However, the main criterion that rules the concept of combined foundation is based on the knowledge of the load sharing distribution between elements and the effect of an additional one in the settlement control. These two parameters are difficult to quantify, since they depend not only on the type of terrain but also on the foundation geometry and the type of applied load. Notwithstanding this aspect, if we consider as a first approximation that the interaction between elements is done in parallel, then is possible to estimate the load percentage shared with the micropiles (αM ) by using a parallel stiffness system – Equation 1. As a reference value and for preliminary studies, is usual to adopt a load sharing distribution factor αM of 50%.
where F = total load applied to the system; F2 = partial load shared with the micropiles; K1 = soilfooting spring stiffness; and K2 = soil-micropile spring stiffness. The simplified methodology presented above shows itself as an essential tool for a preliminary assessment, but a detailed design will only be possible if there is a knowledge of the geological and geotechnical conditions and if there is the possibility to have access to numerical models that characterize the nonlinear and three-dimensional foundation behavior. The Figure 5 summarizes 724
Figure 6. Three-dimensional finite element models of the P3 CMFF foundations.
the design sequence of a combined foundation; one can observe that the geotechnical investigations phase plays a significant role in the design method. 3.2 Numerical models The foundations of the main piers P2 and P3 was designed using numerical models, an essential key for the characterization of their nonlinear behavior. The three-dimensional simulations were performed on the software Plaxis 3D Foundation using volumetric finite elements of 15 nodes and rheological models Hardening Soil and Mohr Coulomb. Regarding micropile, their behavior was modelled by frame embedded pile-type elements and the average skin friction was considered within a range of 300 kPa to 500 kPa, depending on the quality of the rock at depth. As is common in such type of analysis, model complexity is proportional to data processing time; as such there was the need to make some geometric simplifications (Fig. 6). 3.3 Static load tests As previously stated, load tests always play an important role in the design of a combined foundation, as they can provide so much more information to the designer the more complex they are. In the case of the Ceira Bridge a total of three (3) load tests were performed: – one (1) simple load test, located near to the foundations of pier P2; the purpose of this test was to analyze the bearing capacity of the tested micropile and to obtain the load-displacement curves to be used on back analysis modelling; – two (2) CMFF load tests, each one located near to the foundations of piers P2 and P3, in this case with the purpose of characterizing the combined foundation behavior and aimed at: • finding the load sharing distribution between the shallow and deep foundations; • the analysis of the axial load variation the micropiles; • the analysis of the variation of the grout-terrain bond adhesion (qs;m ) along micropiles; • to check the foundation load/deformation behavior when subjected load/unload cycles; • to estimate the foundation stiffness and the bedrock mechanical parameters; • back-analysis modelling. In the simple load test (Fig. 7a) a sacrificial micropile was installed with the same characteristics of production micropiles of the bridge foundations. The micropile and the reaction apparatus were instrumented with deal indicators of manual reading in order to control the settlements and the hydraulic jack load applied to the system was measured with a pressure cell. The maximum compressive load was N = 2400 kN and it was applied with incremental load steps. Two load/unload cycles, one for 50% (0.5 N = 1200 kN) and another for 100% were performed. The results of this load test allowed one to obtain micropile stiffness and update the design foundation mechanical parameters. It was concluded that the rock foundation was softer than initially expected, which resulted in an increase of the sharing load with the micropiles. Consequently, additional measures were considered to reinforce the foundation, including the installation of four (4) micropiles in each one 725
Figure 7.
Set-up of the load tests. a) Simple micropile load test; b) CMFF load test.
Figure 8.
CMFF load test at pier P3. a) Set-up of the load test; b) Automatic data acquisition system.
of the footings corners; negatives were also foreseen if an additional reinforcement was needed. This proved not necessary after the knowledge of the results obtained in the CMFF tests. Each one of the CMFF load tests was performed with a sacrificial micropile that bared the same characteristics of the one used in the simple load test. Its head was connected monolithically to a circular concrete footing with a diameter of 1.20 m and 1.0 m height where the hydraulic jack wielded the load test against the reaction structure (Fig. 7b and Fig. 8a). Given the complexity of the CMFF load test, several redundant devices were installed in order read the desired parameters and to control any anomalies that might arise during the test. The desired parameters were measured by: – the applied load: one (1) load cell and one (1) pressure gauge installed on the hydraulic jack; – the vertical displacements of the test structure: four (4) displacement transducers (LVDT’s) arranged orthogonally at the top of the concrete footing, fixed to independent metallic profiles and firmly keyed far away from the test structure pressure bulb; – the vertical displacements of the reaction structure: controlled by three (3) deal indicators of manual reading; – the calculus of the shared load with the micropile and the estimation of the axial load variation along its shaft was done with the measures of seven (7) levels of three (3) strain-gauges distributed in the interior of each section of the micropile steel tube; – the load sharing estimation with the shallow foundation was measured with four (4) load cells installed orthogonally at the base of the concrete footing. 726
Figure 9. Back analysis of P2 CMFF load test. a) Axisymmetric numerical model; b) Load-Displacement curves; c) Micropile axial load distribution.
The load cells, the transducers and the strain gauges were connected to an automatic data acquisition system that allowed for the real-time control of the results during the load test (Fig. 8b). The test procedure of both CMFF tests was done with few differences between them in terms of load cycles. At pier P3, the load test was carried out in 16 (sixteen) steps, distributed by two (2) load/unload cycles and with a total duration of 23 hours. It was reached 50% of the load test (0.5 N = 1950 kN) in the first cycle and 100% (N = 3900 kN) in the second one. At pier P2 a third cycle was added with a maximum load corresponding to 150% (1.5 N = 5850 kN) of the load test. The total duration of this second CMFF test was 25 hours. In the phase of post processing of the obtained data, the quantification of load distribution between the shallow and deep foundation was carried out according to three (3) distinct calculation approaches: – only taking into account the load cells; – only taking into account the strain gauges; – taking into account both load cells and strain gauges. On the other hand, the method to post process the data depends on the measuring equipment. For the load cells, once known the calibration constants, it was possible to obtain by a simply and expeditious procedure the load applied to the cells. In the case of the strain gauges it was necessary to convert the obtained shortenings in compression forces. For that, it was assumed that the micropile section has a planar behaviour, without sliding between the steel tube and the grout. To accommodate both the discrepancies of values obtained in these three calculation approaches as well as the loading level dependency, a range of values was attributed to the load distribution. The results of the CMFF load tests showed a micropile sharing load range factor of 50% to 80% and 25% to 50% of the total load at piers P2 and P3, respectively. This difference on the obtained range of values could be mainly related with the stiffness of the geotechnical arrangement of the foundation. In the case of the load test performed at pier P2 it was also shown an interaction footing-soil-micropile with a load transfer at depth to the micropile. Regarding grout-terrain bond adhesion strength, the obtained values ranged with the loading level and with the geotechnical arrangement at depth. The maximum average values (qs;m ) were 500 kPa and 1000 kPa at pier P2 and P3, respectively. The CMFF load tests were later simulated numerically by axisymmetric models (Fig. 9). The results of this back-analysis modelling ensured good correlation with the in situ load tests, as well as the rheological behaviour of the combined foundation and the load distribution along the micropile; thus it was possible to feasible calibrate the models used for the design of CMFF-type foundations. 727
4 CONCLUSIONS The constructive solution designed for the Ceira Bridge safeguards unobstructed views of the Ceira river valley through non-current solutions, in particular the adoption of: an exceptionally long span (250 m long); the 26.40 m wide segmental single-cell box girder deck with an upper slab; the nonprovisional devices needed for the cast-in-place construction of the balanced cantilever method and finally the adoption of Combined Micropile Footing Foundations (CMFF) at piers P2 and P3 of the main bridge. Concerning the foundations, the CMFF proved to be an excellent choice with very interesting technical-economic advantages. This type of solution requires, however, a high understanding of the geotechnical behaviour of the soil-structure interaction and because of that, simple and innovative load tests that allowed the direct evaluation of the combined behaviour were performed. The obtained results showed a micropile sharing load range factor of 50% to 80% and 25% to 50% at piers P2 and P3, respectively. These differences should be mainly related with the general stiffness of the geotechnical arrangement of the foundation. CMFF load tests were later simulated numerically also. It was also concluded that the complexity of the load tests instrumentation constrains the reliability of the results and that the geometry of the circular footing limits the effect of the shallow foundation pressure bulb mobilization at depth. For these reasons and given the relative low cost-benefit ratio of the CMFF load tests, one must recommended undoubtedly that in future projects simple load tests are performed to calculate the micropile stiffness and its bearing capacity; One must stress however, the importance of complement these load tests with back-analysis numerical models. In the construction phase and depending on the degree of importance of the structure, it is recommended the installation of monitoring devices on the micropiles in different locations at the combined foundation and the use of load cells in contact with the footing in order to complement the results of the load tests and thus validate/monitor the true combined foundation behaviour during construction phase and throughout the structure serviceability. As a criterion of preliminary design and for this type of foundations it should be adopted a micropile sharing load factor of 50%. However when deformations are a factor of major importance in the design, a higher degree of sharing should be adopted: around 75%. REFERENCES El-Mossallamy, Y.M. & Lutz, B. 2009. Special aspects related to the behavior of piled raft foundation. 17th International Conference on Soil Mechanics & Geotechnical Engineering, Alexandria. IOS Press. Katzenbach, R., Bachmann, G., Boled-Mekasha, G. & Ramm, H.. 2005. Combined pile raft foundations (CPRF): An appropriate solution for the foundations of high-rise buildings. Slovak Journal of Civil Engineering.
BIBLIOGRAPHY ISMR, 1981. Basic Geotechnical Description of Rock Masses. Int. Jour. Rock Mech. Min. Sci. & Geomech. Abstr, Vol. 18, pp. 85-110. Pergamon Press Ltd. Ballouz, M. 2005. Micropiling in Kartsic Rock: New CMFF foundation solution applied at Santa Factory. ASCE Special Publication: 10th Multidisciplinary Conference on Sinkholes & the Engineering & Environmental Impact of Karst TM, San Antonio. LCW Consult. 2012a. Adenda à Nota de Cálculo 1. Ajustamento das Fundações do Pilar P2, Subconcessão do Pinhal Interior. IC3 – Condeixa/Coimbra (IP3/IC2). Ponte sobre o rio Ceira, Projecto de Execução, Lisbon. LCW Consult. 2012b. Nota Técnica 1. Ensaio de Carga da Fundação Mista do Pilar P3. IC3 – Condeixa/Coimbra (IP3/IC2). Ponte sobre o rio Ceira, Projecto de Execução, Lisbon. LCW Consult. 2013. Estudo Geológico-Geotécnico. Revisão A. Ponte sobre o rio Ceira, Projecto de Execução, Lisbon. Mahmoud, Mahomed A. Baqi & Elarabi, Hussein. 2003. Verification of Simplified Approach With Numerical Approach in Analyzing Piled – Raft Foundations. Proceedings of 5th International Young Geotechnical Engineer’s Conference – 5 thiYGEC, Amsterdam. IOS Press.
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Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
Tresfjord Bridge – Foundation of main span on 40 m caisson on soil seabed K.B. Dahl, L. Toverud & D.E. Brekke Multiconsult AS, Oslo, Norway
ABSTRACT: The main span of the Tresfjord Bridge currently under construction in Norway is founded on 2 caissons. The largest is built on soil seabed, and is the first big caisson in Norway not founded on bedrock. Moreover, it is the biggest caisson ever installed in Norway. The bottom slab consists of posttensioned lightweight aggregate concrete and has a diameter of approximately 24 m, whereas the shaft diameter ranges from 9–15 m. The fender slab has a diameter of 12 m and a thickness of 3 m, constituting a monolithic base for the bridge pier. The caisson is filled with water saturated gravel. It was designed with the in-house code MultiCon, which is a postprocessor specifically developed for large concrete offshore structures.
1 INTRODUCTION 1.1 Tresfjord bridge The Tresfjord Bridge currently (2012–2015) being built by Bilfinger Construction at the Norwegian west coast is a concrete bridge with a total length of 1290 m. It consists of 19 spans, 60 m each, and a FCM (free cantilever method) main span of 160 m, see Figure 1. The bridge has a box cross section with a maximum height of 9.2 m and a total width of 12.5 m, with 2 traffic lanes and 1 pedestrian lane. The navigational channel has a width of 60 m and a height of 32 m. The bridge is part of highway E136, and will shorten the distance between the local towns Åndalsnes and Ålesund by approximately 15 km. The bridge contract cost is approximately €100 mill making it one of the biggest concrete bridges ever built in Norway. The engineering is done by Multiconsult AS in Oslo for the Norwegian National Road Authority. 1.2 Foundation system The maximum water depth in the fjord is approximately 40 m. 13 axes are founded on pile groups based on Ø1200 mm steel pipe piles reinforced and concreted, with a free length in water up to 43 m. The main span is founded on two caissons, one on bedrock and the other on soil seabed, consisting of dense ground moraine.
Figure 1. Tresfjord bridge with total length 1290 m, built 2012–2015 at the west coast of Norway.
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Figure 2.
Caisson geometry, with bottom slab and fender/top slab shown to the right.
1.3 Objective This paper addresses the deepest foundation (axis 8) in the FCM main span, with a water depth of approximately 37 m. Here, the seabed consists of soil, i.e. foundation on bedrock was not possible. In Norway, this is the first large caisson ever founded directly on soil seabed. Thus, the caisson resembles a GBS (gravity based structure) similar to offshore structures. This, in conjunction with its size and general complexity required a comprehensive design procedure, which is the basis for this paper. 2 CAISSON GEOMETRY 2.1 Bottom slab The loads from the bridge pier include considerable bending moments, especially from the FCM construction stages. This is taken up by eccentric ground pressure, requiring a considerable bottom slab in order to ensure stability and acceptable ground pressure values. A circular slab was chosen, see Figure 2, with a diameter of 23.7 m and a thickness of 1.2 m at the edge, linearly increasing to 1.4 m in the middle. Lightweight aggregate concrete LC60/66 is used in the slab, as it had to be constructed on a barge with limited weight capacity. It is posttensioned with 20 cables 30ø0.6” in both directions. Successive tensioning of cables oriented in each direction was prescribed in order to avoid tension fields developing perpendicularly. The ordinary reinforcement was arranged orthogonally in the middle, and else in a circular-radial system. 2.2 Shaft In order to keep bending moments in the slab at a reasonable level, a circular shaft was found advantageous. The inner shaft diameter is 15 m at the bottom and 9 m at the top, resulting in a 730
conical shape vertically. The shaft walls are 0.5 m thick; at the bottom the thickness is increased by 50% in order to handle enhanced forces. Concrete C55/67 and ordinary reinforcement is used in the shaft; the bottom and top is designed with shear reinforcement. The caisson is filled with water saturated gravel (2.17 ton/m3 ), giving a considerable contribution to stability. 2.3 Fender slab The fender slab at the top of the shaft has a diameter of 12.0 m including the outer wall. The thickness is 3.0 m, constituting a monolithic base for the bridge pier, which has a height of approximately 35 m in this axis. Concrete C45/55 and ordinary reinforcement is used, i.e. no posttension cables. A concrete cover of 120 mm is specified in the fender slab due to the tough environment in the splash zone. In total, the height of the caisson itself is 41.1 m, reaching 2.0 m above mean sea level. 3 CONSTRUCTION 3.1 Construction The bottom slab and the lower shaft were constructed on a submersible barge at a quay. The caisson was launched at a shaft height of 8.5 m by submerging the barge, with further construction based on climbing about 10 stages, each typically 3.5 m high. The caisson was successively immersed for each stage by adjusting the water level inside so that a reasonable working height could be managed from barges positioned at each side, keeping the caisson in position at the quay. The last stage comprised the outer wall around the fender slab, constituting a formwork for casting the slab after ballasting was completed. 3.2 Buoyancy and hydrostatic stability The hydrostatic stability was controlled at each stage by calculating the metacentric height GM. Similarly, the buoyancy was ensured by hand calculations, with a minimum freeboard of 2.5 m, keeping the scaffolding dry. The size of the bottom slab, and the fact that water saturated gravel was used for ballasting instead of water only, gave relatively good stability for all stages (min GM = 1.3 m). 3.3 Installation Before installation, the seabed was excavated some 1–2 m and levelled with 0.5 m underwater (UW) concrete. The caisson was towed from the quay by a tug-boat, and immersed by further ballasting into position by means of a steel arrangement at the seabed. The caisson rested temporarily on 3 concrete “buttons” 6 m2 each, until 0.4–0.6 m of UW-casting could be completed between the levelling UW-concrete and the tapered bottom of the slab. Steel tubes down along the shaft walls and through the slab were used for the casting, being inspected by divers. Subsequently, further ballasting could be finished. In order to ensure stability, a weather window was defined until this work was completed. Finally, the fender slab was cast, followed by construction of the pier. 4 DESIGN 4.1 The software MULTICON Multiconsult has a long tradition in designing large offshore structures in Norway and abroad, and has over the last 25 years developed a code for design of concrete shells, MULTICON (Multiconsult, 2014). This is a post-processor, for which the FEM-code ANSYS 12.0 (www.ansys.com) 731
Figure 3.
Caisson geometry shown in MULTICON, with vertical membrane forces in the shaft from Fz.
was used for the analyses. One of the advantageous features of MULTICON is code check for solid/3D-elements against i.a. EC2 (CEN, 2004). This allows a geometrically accurate structure being modelled with solid elements, with a fairly realistic application of loads and soil behavior, making use of multiple point constraints (MPC), surface load elements, and concentrated and distributed springs. MULTICON also comprises load combination, plotting routines of forces, utilization ratios etc., hence offering a good design tool for this type of bridge foundation. A more comprehensive presentation of MULTICON is given by Brekke et al. 1994. 4.2 Loads and design combinations Unit forces (1000 kN, 1000 kNm) representing the loads from the clamped bridge pier (Fx, Fy, Fz, Mx, My, Mz) was subjected at the top of the FEM-model by MPCs, see Figure 3, for which the results were combined in the postprocessor MULTICON according to the results from the global analysis. The combinations were limited to maximum and minimum bridge forces within each limit state for each of the 6 load components, respectively, with corresponding values for the 5 other, resulting in some 300 combinations of loads from the bridge pier alone. Buoyancy was modelled by applying water pressures inside and outside the structure. The accurate geometry model proved to be beneficial for handling the water pressures in such a structure. Similarly were wave forces in both directions modelled as varying compressive and tensile surface loads at each side of the shaft, on the basis of Morison’s equation (Morison et al. 1950). Pressure from gravel ballast was likewise included inside the shaft. It can be noted that a vertical component from possible gravel friction along the shaft walls was included in the load combinations. Concrete shrinkage and creep was neglected due to the submerged condition. Likewise were effects from ice and dynamics considered negligible. Ship collision was included in the loads from the bridge pier. In total, this resulted in some 2000 design combinations. 732
Figure 4.
Buckling analysis of caisson at touch-down with inner and outer water pressures.
4.3 Stages Various stages were included in the design combinations: – – – – – –
Posttensioning of the bottom slab, at shaft height 5m Immersed, at different shaft heights Immersed at touch- down, with a maximum difference in water pressures inside and outside Caisson supported temporarily on 3 “buttons” (concentrated supports) Caisson with pier and FCM beams before superstructure keystones Completed bridge
4.4 Shaft walls – geometric imperfection and 2nd order effects The shaft wall appears quite slender, and is during immersing subjected to relatively large hoop compression due to different water pressures inside and outside. In the completed phase the shaft is subjected to hoop tension, however, vertically the shaft is subjected to compressive forces. EC2 (CEN, 2004) chapter 5 was used to assess whether the corresponding 2nd order effects should be included in the design. The buckling load NB was calculated for the wall both analytically (Timoshenko & Gere, 1963) and numerically inANSYS, as shown in Figure 4, resulting in realtively good correspondence. Based on NB , the effective buckling length l0 and thereby the imperfection ei and the 2nd order deflection e2 and could be calculated. Here the method based on nominal curvature according to part 5.8.8 in EC2 (CEN, 2004) was used for e2 . For the construction stage, water pressures at touch down was used as load vector, resulting in:
For the permanent stages, the load combination giving maximum vertical membrane force was used. This resulted in a very low e2 and hence a deflection equal to the minimum imperfection value emin according to part 6.1(4):
The eccentricity etot may be given as a direct parameter in MULTICON, increasing the bending moments Mi from the first order analysis, given that the membrane force Ni is compressive:
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Figure 5.
Sectional forces and reinforcement for a typical shell design section (Multiconsult, 2014).
Conservatively, the same value of etot was used in the whole shaft for the different stages, respectively. This resulted generally in somewhat higher utilization ratios. 4.5 Design The FEM-analysis is performed on a linear basis. Subsequently, sectional forces according to Figure 5 are established by MULTICON based on imported gauss point stresses from each solid element in ANSYS. These are integrated and transformed into a grid and coordinate system in accordance with the prescribed reinforcement system. The load cases from ANSYS are combined into numerous design combinations. For each combination and element, a local design is performed, taking into account the non-linear behavior of concrete and reinforcement. The program steps reinforcement until design criteria are met for ordinary and shear reinforcement. Moreover, control of crack widths and concrete compression stresses are carried through. It can also be noted that MULTICON includes the effect from extra water pressure in cracks in submerged concrete, a feature that was utilized here. Based on this, the capacities in each element for each stage and design combination were checked. An example of design plot is shown in Figure 6. Decompression was checked by fictitious reinforcement. 5 GEOTECHNICAL ASPECTS 5.1 Geotechnical springs The vertical soil reaction was modelled in ANSYS as distributed linear springs under the bottom slab. Because of uncertainty in the soil quality, the analyses were conducted with both an upper and a lower geotechnical spring value, 3000 and 23 000 kN/(m m2 ). Both sets of results were included in the design combination database. Concentrated linear springs connected to MPCs under the bottom slab were used to constrain the caisson horizontally and against torsion. 5.2 Ground pressure Independent analyses using geotechnical software (PLAXIS 3D, 2014) gave comparable ground pressures to ANSYS, underpinning the adequacy of the design procedure. The maximum pressure was found to be about 700 kPa, see Figure 7. Conventional hand calculations resulted in 500 kPa, which as expected is lower since ANSYS includes peak effects, being particularly pronounced at the edge and under the shaft wall. Sufficient bearing capacity was verified by geotechnical analyses. The effect from cyclic loading (wind) on the bearing capacity was also controlled by geotechnical analyses. Moreover, the global caisson eccentricity (e = M/N) was checked against stability criteria for ULS and SLS in codes given by the Norwegian Road Authority. 734
Figure 6.
Lower reinforcement in bottom slab with max UR = 84% in ULS (plots from MULTICON).
Figure 7.
Ground pressure (ANSYS) for the load combination resulting in maximum base moment Mx.
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5.3 Nonlinear analyses Tensile spring values indicated uplift of a smaller part of the bottom slab for some load combinations. Hence, a separate nonlinear model with contact elements between the concrete and a 1 m thick elastic material was modelled, allowing physical openings without tension to occur. Complete load situations were subjected to the model, as the principle of superposition no longer was valid. Highest astronomical tide (HAT = +1.29 m) was used in order to obtain a condition of maximum eccentricity. The maximum ground pressure was found to be 730 kPa, i.e. somewhat higher than 700 kPa from the linear analyses, as expected. Only minor effects on the concrete utilization ratios were observed. 5.4 Settlements Settlements were controlled in geotechnical analyses. In total, they can be expected to be in the range of approximately 100 mm. However, a considerable part (ca. 70%) occurs before the pier is finished, i.e. much is compensated during construction. Some skew settlement may occur due to dominating bending moments from the superstructure and somewhat varying soil depth to bedrock. 6 CONCLUSIONS The caisson described is a relatively complex concrete structure subjected to large membrane and plate bending forces resulting from water pressures, posttensioning, ballast, construction stages and soil interaction. Moreover, the fact that the structure was immersed in water during construction and installation, requiring sufficient buoyancy and hydrostatic stability, made extra demands on the design process. For this purpose, the software MULTICON in conjunction with ANSYS proved to be a useful design tool, giving good overview of global load responses as well as local design in each element. The ability to check solid/3D FEM-elements against EC2 (CEN, 2004) allowed accurate modelling, which is beneficial for such a complex structure. Moreover, the possibility to easily rerun analyses numerous times in a complete manner, even for small changes in configuration or loads, was deemed very valuable from the viewpoint of practical engineering. The maximum support reaction is about 200 000 kN vertically, with an extreme base moment of approximately 500 000 kNm. In Norway, the caisson is the biggest ever installed, and furthermore the first caisson founded directly on soil seabed. REFERENCES ANSYS, Inc. 2014. Ansys . Canonsburg, PA, USA: www.ansys.com. Brekke, D.-E., Åldstedt, E., & Grosch, H. 1994. Design of Offshore Concrete Structure Based on Postprocessing of Results from Finite Element Analysis: Methods, Limitations and Accuracy. International Offshore and Polar Engineering Conference (pp. 318–328). Osaka, Japan: The International Society of Offshore and Polar Engineers. CEN. 2004. Eurocode 2: Design of concrete structures – part 1–1: General rules and rules for buildings. Comité Européen de Normalisation. Morison, J., O’Brien, M., Johnson, J., & Schaaf, S. 1950. The forces exerted by surface waves on piles. Pet. Transactions 189, 149–154. Multiconsult. 2014. MultiCon software manual. Oslo, Norway: Multiconsult. PLAXIS 3D. 2014. Plaxis. Delft, The Netherlands: www.plaxis.nl. Timoshenko, S. P., & Gere, J. M. 1963. Theory of elastic stability. Singapore: McGraw-Hill.
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Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
Chiapas Bridge G.R. Argüelles Grupo Triada, Mexico City, Mexico
ABSTRACT: Chiapas Bridge is located on Las Choapas – Ocozocuautla Highway, crossing the second largest dam in Mexico. To develop the constructive project, the site’s topographic-hydraulic, geotechnical, seismic risk, wind incidence, geological and geophysical studies were carried on. The substructure consists of 8 jackets formed each by 4 steel pipe legs, floated and sunk, through which big diameter concrete piles were drilled and cast in place. Superstructure is a steel box-girder deck, launched from one side with 168 m cantilevers.
1 INTRODUCTION 1.1 Type selection Chiapas Bridge is located at km 961 + 731 of the Raudales – Ocozocoautla stretch of Las Choapas – Ocozocoautla highway, in the state of the same name. This segment is 202 km long, and was constructed to communicate the state capital Tuxtla Gutiérrez more directly and efficiently with Mexico City and the center of the country, generating significant savings in the operating costs for road transportation of people and loadings. This location was mainly due to ecological reasons, to avoid passing the road by the ecological reserves of Los Chimalapas and Los Montes Azules, so it was necessary to project the bridge crossing the reservoir of the Nezahualcoyotl or Malpaso dam, with adverse engineering conditions. Thus, prior to the design of the bridge structure that was finally developed, it was necessary to make a comparative analysis of alternative solutions summarized below, calling a) to c) as “fixed” structures, and the one designated d) as “semi-fixed”: a) Two incremental launched bridges, with orthotropic steel box-girder decks for two and four traffic lanes, respectively. b) A ribbed double cantilever bridge of prestressed concrete, built for four traffic lanes. c) Two cable-stayed bridges for four traffic lanes, one with a mixed steel-concrete deck, and one with an orthotropic steel box-girder. d) Two alternatives of floating bridges with steel deck supported by pontoons of traditional prestressed concrete, and of steel anchored to the bottom of the reservoir. As a result of the performed analysis, the alternative a) was selected as the most suitable, by aspects of cost, time, construction procedure and operation, combining a foundation, substructure and superstructure design for two traffic lanes, but prepared for an extension to 4 lanes in a second stage. 2 SITE FEATURES 2.1 Horizontal and vertical alignment In order to reduce as far as possible the length of the bridge, the crossing line was selected through a significant narrowing of the reservoir, located about 5 km south of the curtain and the hydroelectric 737
central of the dam, and also about 6 km south of the town of Raudales, Chis. Thus, the bridge is in a horizontal straight tangent between km 961 + 127 and km 962 + 335, so the total length is 1,208 m. To define the vertical alignment of the bridge, operational data were provided by the Comisión Federal de Electricidad (CFE), about the project of Nezahualcoyotl Dam and the historical behavior of the water levels in the reservoir, since its first partial filling in 1964. Also, the CFE informed that “in times of scarce water inputs, it has been necessary to lower the levels to meet the electricity demand, having arrived even to the Minimum Water Level Operation (MWLO) in 1980”. Based on the above data, a grade line of the horizontal bridge was defined at Elev. 184.75 m, 3.55 m above the crown of the curtain of the dam (Elev 181.20 m), where the vertical free space is 7.50 m with respect to OHWL (Ordinary High Water Level, Elev. 171.70 m), and 2.00 m compared to EHWL (Extraordinary High Water Level, Elev. 177.20 m). The OHWL is the maximum water level in the reservoir, that is usually reached almost every year since 1982, and EHWL had never been achieved in the 35 years of historical record since the construction of the dam, so that the probability of occurrence in the bridge life is minimal. 2.2 Subsoil conditions The results of the geological – geophysical survey, and geotechnical borings made, indicated the following: • The bottom of the reservoir is formed by interstratifications of sedimentary rocks of Oligocene to Paleocene age, of shales, basically sandstones and fine grain conglomerates type, with variable degrees of fracturing, superficially altered to more or less resistant residual clay or sandy soils, in turn covered by a sediment layer of low consistency. • The main channel of the Grijalva River, flooded by the dam, followed the master fault known as Chiapa – La Venta or Malpaso – Muñiz. • In the Las Choapas (Raudales) side of crossing, corresponding to the right bank of the Grijalva River, some minor faults were detected, which favoured the formation of a smooth hillocks relief, that when being flooded defined a series of small islands and peninsulas. • In the Ocozocoautla side, that is the left bank of the Grijalva River, topography corresponds to a steeper hill with a monotonously increasing elevation. • The seismic-acoustic subsoil profile, determined with dual frequency profilers, allowed the detection of 5 different types of materials: – Sediments deposited in the main river channel after filling of the dam (siltation). – Sediments (alluvium) deposited in the main river channel before the reservoir. – Natural soil and talus deposits, undifferentiated (unconsolidated material). – Decompressed rock. – Compact rock. • The results of exploratory drilling confirmed that in both sides (Piers 1 and Abutment 9; Piers 2, 7 and 8), the thickness of residual soils grows up to 20 m and the underlying rock is more altered and has lower quality, while in the central part with deeper water (Piers 3 to 6), the residual soils thickness is less than 5 m and the rock has better characteristics. 3 SUBSTRUCTURE AND FOUNDATIONS As a first step, the length of the bridge was defined according to the position of the extreme abutments. In the Ocozocuautla side, the location of Abutment No. 9 was set up by the EHWL that the water in the reservoir reaches, so that it was fixed at kilometer 962 + 335; on the margin of Raudales side, Pier No. 1 was located at km 961 + 127, to give common support to another adjacent structure of less importance, called Chiapas Bridge II. In consequence, the total length of Chiapas Bridge is 1208 m, divided into 8 spans. 738
The main problem in the bridge design crossing the dam, was the construction of the substructure and its foundation with water depths up to 90 m, in an area of high seismicity. To resolve this issue, it was considered important to take advantage of the great international experience achieved in the design and construction of offshore oil platforms. Thus, for the 7 central piers, numbered from 2 to 8, a substructure of the type known as “Jacket” was selected, each made up of four main vertical pipes of steel A-36, separated “center to center” 10.00 m in the longitudinal direction and 18.00 m in the transverse direction (except Piers N◦ 2 and 8 having 10.00 × 10.00 m), functioning as “legs” of the piers. These pipes, with 2.78 m (109 ) of external diameter and 2.54 cm (1 ) thick, were joined together by horizontal and vertical bracing systems, also formed by welded smaller diameter steel pipes. The Jackets were fully assembled in a manufacturing park, conveniently located on the edge of the reservoir, and floated to the site of each support, where they were gradually turned up to the vertical position, allowing controlled access of water inside the pipes and using counterweights. To keep the Jackets fixed in position, auxiliary piles were installed around each, also made of steel pipes with diameter of 0.90 m, penetrating as needed in firm ground of the bottom of the dam. According to its location, the piers had heights of 27 to 89 m, and were installed in water depths ranging from about 5 m to 88 m, depending on the year’s season. Once in position, big reinforced concrete caps were casted over each of them. The foundation of the Jackets consisted of reinforced concrete piles of 2.5 m diameter, drilled and casted through the main pipes, penetrating into the bottom of the reservoir from 25 to 40 m, until materials of adequate resistance were found, making the transference of charge to the steel pipes by means of bulbs. In order to drill the piles, modern equipment and techniques proven in similar works were used. One important condition was the fact that drilling of the piles casted in place had to be done vertically, according to requirements of the manufacturers of the equipment. This meant that the main pipes of the “Jacket” type piers, should remain in vertical position during the works, with minimal deviation tolerances. By the other hand, Abutment N◦ 9 consisted in a reinforced concrete box, housed in an excavation in the hill of Ocozocoautla side margin; the superstructure was anchored to it, to restrict its longitudinal movement, so it will receive the total horizontal seismic forces acting in the same direction. Support in Pier N◦ 1, at the Raudales Side, allows free displacement of the superstructure.
4 SUPERSTRUCTURE The superstructure consisted of an orthotropic A-50 steel box-girder deck, with 5.5 m of constant camber in all the length of the bridge, supported at 8 spans of 124, 168, 168, 168, 168, 168, 152 and 92 m, respectively. The relationship between the extreme and adjacent clearings is equal or greater than 0.605, and the less favorable camber/clear ratio is 1/30.5, values both that are suitable for launched bridges with constant height camber. The fundamental oscillation period of the structure was 3.39 seconds, corresponding to a transverse vibration mode. At the first stage of construction the deck has a width of 10.00 m for two traffic lanes, but is prepared so that, in the future, when the traffic demand requires, it will be extended to 16.00 m to allow circulation in four lanes. Enlargement will be performed by welding steel plates on both sides of the box, to extend the cantilever outside; these extensions will be supported by inclined steel shoring, longitudinally spaced every 4 meters, in turn supported at the bottom of the box-girder. The superstructure was launched from a manufacturing park, built in the access on Ocozocoautla side, by excavating below the road grade level, to house a concrete “U” form box, of 375 m length, 6 m deep and 10 m wide. Attacched to the foundation slab two prestressed concrete beams were built, which served to anchor a device consisting of clamps and push-jacks with 50 cm stroke, used to displace the steel box-girder. 739
Due to the great length of the clearings between supports, it was necessary to use a light steel nose and a steel mast to set temporary stay cables. The length of the maximum launched cantilever was 168 m, which constituted a worldwide reference for this installation procedure. Once the superstructure was located in its final position, the concrete box was filled up to the grade level indicated in the earthworks project, and proceeded to place the asphalt layer of 5 cm thick, parapets and finishings. 5 TERMINATION Construction of Chiapas Bridge began on April 1999, to be finished 4 years and 8 months later, on December 2003, when the bridge was opened to traffic. The Chiapas Bridge Project was awarded in Spain with a Special Diploma for it’s participation in the International Alcantara Bridge Prize 2002–2004. 6 PHOTOGRAPHS AND FIGURES
Figure 1.
Photography of Chiapas Bridge.
Figure 2.
Elevation Chiapas Bridge.
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Life cycle
Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
Life-cycle costs of bridge bearings – Key considerations for bridge designers and owners T. Spuler & N. Meng Mageba SA, Bulach, Switzerland
G. Moor Mageba USA LLC, New York, USA
ABSTRACT: This paper presents an overview of the considerations affecting life-cycle costs for bridge bearings, with particular reference to selection of bearing type, design, fabrication, installation, inspection, maintenance and replacement. With an understanding of all these issues, bridge designers and owners will be well equipped to properly address the challenge of minimising bearing life-cycle costs. 1 INTRODUCTION That life-cycle costs should be considered when specifying and selecting bridge components such as bearings, as in any modern construction project, is today widely accepted. It is particularly true in the case of long and multi-span bridges due to the increased complexity of the demands and challenges they generally present, such as higher loading, greater multi-directional movements and more significant vibrations. This paper presents an overview of the considerations affecting life-cycle costs for bridge bearings, with reference to factors right throughout the life-cycle of the bridge. It is important to consider the life-cycle of the bridge as opposed to that of its bearings, in order to include the costs of bearing replacements during the bridge’s life. The costs of bearing replacements include not only supply costs but also installation costs (generally requiring lifting of the bridge deck), traffic management costs and the bridge user costs associated with delays, and are thus far higher than the initial bearing supply and installation costs at the time of the bridge’s construction. It is therefore critical, in managing and minimising life-cycle costs, to ensure that the number of times the bearings need to be replaced during the bridge’s life is kept as low as possible. 2 THE CHALLENGES FACED BY BRIDGE BEARINGS The bearings that support a bridge deck are not only (or always) required to transmit vertical loads from the deck to substructures. They must also, in many cases, resist horizontal forces (longitudinal and/or transverse), while accommodating deck movements and multi-axial rotations as required by the bridge’s design. They are thus critically important, relatively complex structural components. They are also much less robust than the main bridge structure, and are thus likely to require replacement, perhaps several times, during the bridge’s long service life. As a result, it is important to carefully consider, during selection and design of the bridge’s bearings, the complete life-cycle of the main structure and of the bearings themselves. 3 LIFE-CYCLE COST ANALYSIS A great deal has been written to assist engineers and owners in the assessment of life-cycle issues, and the field of bridges is no exception – for example, with the 2003 report, “Bridge life-cycle cost 743
analysis” (Hawk et al. 2003), published by the Transportation Research Board of the American National Research Council as Report 483 of the National Cooperative Highway Research Program (NCHRP). This report notes, in relation to Life-Cycle Cost Analysis (LCCA): “LCCA is essentially a technique for considering the economic efficiency of expenditures”. It goes on to define LifeCycle Cost (LCC) for a bridge in terms of its constituent parts, as follows:
where DC = design cost, CC = construction cost, MC = maintenance cost, RC = rehabilitation cost, UC = user cost, and SV = salvage value. Life-cycle cost analysis thus represents a great improvement on the “traditional” approach often used in the construction of infrastructure, which considers only the initial direct costs of design and construction (i.e., the terms DC and CC in the equation above). This is explored in more detail in the next section, for the specific case of a bridge’s bearings. 4 THE IMPORTANCE OF CONSIDERATION OF LIFE-CYCLE COSTS IN THE SELECTION, DESIGN, INSTALLATION AND MAINTENANCE OF BRIDGE BEARINGS It can be inferred from the foregoing statements that it is important that the complete life-cycle of a bridge’s bearings be carefully considered when selecting, designing, fabricating, installing and maintaining them. This can be confirmed by closer analysis of the life-cycle costs, starting with a definition of what they include. Equation 1 above, formulated to define the life-cycle costs of a bridge as a whole, can reasonably be considered generally applicable also to the bearings within the bridge. Adapting Equation 1 slightly for use in relation to bearings (with salvage value neglected):
where LCC = life-cycle cost, ISC = supply cost, IIC = initial installation cost (at time of bridge construction), IMC = inspection and maintenance cost, DRC = direct replacement cost, and UC = user cost. It is important that the life-cycle to which reference is made is that of the bridge structure, and not of particular bearings which are installed in the bridge at a particular point in time (e.g. at the time of the bridge’s construction). This is an important distinction, because only consideration of the bridge’s life-cycle will take account of the most significant costs associated with its bearings: the cost to the owner of periodic replacement works, and the user costs that accompany those works. The first 4 of the 5 costs on the right of Equation 2 above (ISC, IIC, IMC and DRC) are classified as Agency Costs, which are carried by the responsible agency or bridge owner, as opposed to the User Costs (UC) which are carried by the bridge’s users (which include motorists and others who cross the bridge, and possibly the businesses and residents of nearby areas that rely on the bridge for access). The significance of each of these cost groups is discussed below. 4.1 Initial supply and initial installation costs The costs of supply and installation in a new bridge of its bearings depends on many project-specific factors, such as (perhaps most significantly) the forces the bearings must carry/resist and the deck movements they must facilitate. But the costs are also somewhat related to the construction costs 744
Figure 1.
Representation (sequential) of the life-cycle costs of a bridge’s bearings.
of the bridge as a whole: the forces the bearings must carry/resist depend on the size/weight of the bridge, and the deck movements to be facilitated by the bearings increase with the bridge’s length and width. 4.2 Inspection and maintenance costs There are two general approaches to infrastructure management, proactive and reactive, and, in general, only the proactive approach can be recommended in the case of a bridge’s bearings. Inspection and maintenance work is an essential part of the proper management of any bridge, and even more so in the case of its bearings, which as noted previously are less robust but subjected to greater demands than the bridge as a whole. Unfortunately, the reactive approach is applied far too often, with bearing issues only being addressed when a safety hazard has developed or the deck is being rehabilitated or replaced. A change of mind-set is therefore required of many of those who are responsible for inspection and maintenance activities; it should be recognised that the costs of a sensibly planned inspection and maintenance regime are well invested, and will likely result in much greater long-term savings by reducing the need for expensive reactive repairs and by delaying or avoiding the need for bearing replacement work. 4.3 Direct replacement costs As noted above, a bearing of any type has a shorter life expectancy than the main structure on which it is installed. The direct cost to the owner of the replacement works that become necessary at the end of the service life of a particular bearing can be very significant. At any rate, due to the costs of site mobilisation, provision of access to the bearings, lifting of the deck, traffic management, etc., the costs are likely to be much higher than the initial supply and installation works that were carried out when the bridge was under construction (Fig. 2). Therefore, in order to minimise the life-cycle costs of a bridge’s bearings, during the life of the bridge, it is clearly necessary to minimise the number of bearing replacements – by the use of bearings of suitable quality and durability, and proper attention to inspection and maintenance activities. 4.4 User costs The user costs associated with a bridge’s bearings result primarily from the disruption to traffic that is caused by bearing maintenance or replacement works. The assessment of these costs requires the estimation of such factors as the number of vehicles and occupants which will suffer delays, the average length of delays, the cost per hour per vehicle or occupant, and increased fuel consumption. User costs will therefore vary greatly from one structure to another. This emphasises once again the importance of minimising the frequency at which the bearings of a bridge will have to be replaced; by the use of bearings of high quality and high durability, and proper attention to inspection and maintenance, life-cycle costs can be minimised. 745
Figure 2.
Representation (proportional) of the life-cycle costs of a bridge’s bearings (typical).
5 HOW THE LIFE-CYCLE COSTS OF A BRIDGE’S BEARINGS CAN BE MINIMISED The results of the above analysis of the life-cycle costs of a bridge’s bearings, as presented in Equation 2, might be summarised as follows: In order to minimise the overall life-cycle costs of a bridge’s bearings, it must be ensured that adequate resources are devoted to maximising the suitability, durability and quality of the bearings selected for use, and enabling them to perform as well as can be expected, for as long as can be expected, by ensuring the quality of their installation and the adequacy of subsequent inspection and maintenance activities. Special attention must also be paid to the bearing replacement process.
This overall strategy is broken down and discussed in the following sections. 5.1 Maximising the suitability, durability and quality of the bearings selected for use The bridge designer can play an important role in optimising the overall costs of a structure’s bearings, by paying due attention to the issues described below. Many such issues require proper understanding of the capabilities of particular bearing types and awareness of the capabilities of the supplier that will design and manufacture them, so support from an experienced supplier can be of great value, even in the early stages of a bridge design and construction project. 5.1.1 Specification of demands which must be satisfied by the bearings It is, of course, important to define the demands to which the bearings will be subjected, and to ensure at an early stage that this will not present major difficulties for supply. This assessment should not be limited to just the six degrees of freedom and the deck forces to be carried and resisted, but should consider all other relevant factors such as the cumulative movements during the lifetime of the bearings. “Slow” thermal movements are relatively straightforward to calculate, and are much more significant for steel than for concrete bridges. “Fast” movements related to live loading are a phenomenon typically observed only in the case of suspension bridges, where they can lead to massively increased wear of components such as bearings and expansion joints. When these factors have been assessed, the selection of bearing type may be optimised. Such issues as choice of sliding material (e.g. UHMWPE instead of PTFE) can then also be properly considered, and it can be confirmed whether the best materials available can be expected to satisfy durability requirements. If they cannot, then this should be recognised and suitable allowance made for maintenance and replacement, or another solution sought. 5.1.2 Selection of the optimal bearing type The selection of bearing type (pot, spherical, disc, elastomeric, rocker, roller, etc.) will be largely defined by the principal functions the bearing is required to fulfil for the bridge – in particular, in terms of resisting forces and accommodating movements. Once a short-list of suitable bearing types has been created, based on these primary requirements, a secondary analysis should be performed to optimise the final selection, with consideration, in particular, of long-term performance. Maintenance effort should ideally be minimised, as should the frequency of bearing replacement works and the required effort when replacement becomes unavoidable. In some exceptional cases, such as that of the horizontal force (shear key) bearings of the north approach viaduct of the new NEAT Ceneri Alpine railway tunnel with their great specific challenges (Spuler et al. 2013), it 746
may even be possible to design and produce the bearings for the structure’s full service life, thus avoiding the need for any bearing replacement works at all. Durability may be maximised, minimising bearing maintenance and replacement effort, in a number of ways. For example, bearing designs that avoid the use of deformable materials such as elastomer will generally offer better long-term performance, as will those that minimise the risk and effects of corrosion. And the use of sliding surfaces, which are particularly prone to damage and deterioration, may be avoidable if any structure movements can be accommodated by the (limited) deformations of elastomeric bearings. Further benefits, resulting in significant life-cycle cost reductions, may also be realised by the selection of bearings (e.g. pot bearings) that can be designed with lifting and/or force-measuring capabilities. The elastomeric pad of an injection pot bearing can be raised at any time after installation without the need for hydraulic jacks, making such bearings ideal for use wherever substructure settlements are anticipated.. And measuring bearings, which enable the forces they carry to be easily read on a connected device, facilitate easier inspection and more reliable identification of problems that may arise. 5.1.3 Verification of bearing performance It is most important that the ability of the selected bearing, as designed and fabricated by its manufacturer, to withstand the loads and facilitate the movements to which it will be subjected during a long life on a structure, should be verified in advance of its use. The best verification of this is a strong track record on the part of the bearing supplier, with evidence of satisfactory performance of similar bearings over many years on comparable structures which place similar demands on the bearings. Laboratory testing also serves a useful purpose, but it must be recognised that the degree to which it can replicate actual service conditions is limited by the need to make testing practical and affordable. This dictates that any particular test can only assess certain defined performance characteristics, and that such assessments will be based on various simplifications and assumptions. Nonetheless, laboratory testing is often necessary, even if only to give new or improving suppliers an opportunity to demonstrate the quality of their products. The specification of standard testing requirements by road authorities/agencies also ensures that a certain level of quality, which is necessary in minimising life-cycle costs, will be demanded in the procurement of the bearings that will be installed on their structures. 5.1.4 Evaluation of the needs of the preferred bearing type Once the type of bearing which can optimally satisfy the structure’s needs has been identified, it is important, at an early stage, to ensure that the bridge deck is designed to receive the selected bearings, with proper access, adequate clearance between substructure and superstructure, correctly sized and reinforced connecting surfaces and allowance for suitable anchorage. Inadequate access to the bearings may cause difficulties with inspection and maintenance at a later stage, and incorrectly dimensioned main structures may necessitate changes to approved plans, or even adaptations to the constructed bridge structure on site. 5.1.5 Designing to maximise durability and extend service life A key factor in maximising durability is the proper application of a suitable corrosion protection system. The corrosion protection applied to any exposed steel on the bearings should be appropriate to the bridge’s environment – meaning that an appropriate system (e.g. painted, galvanised etc.) and appropriate level of protection must be specified, and properly applied, with adequate verification of quality and particularly layer thickness and adhesion. It must also be considered that such bearings will require re-application of corrosion protection (generally by painting) some time after the bearings have been installed, so access to the susceptible parts of the bearings should allow this work to be done well and without great difficulty. Another important factor is the design and orientation of bearings to protect sensitive sliding surfaces. As a rule, all sliding bearings should be fitted with some form of sliding surface protection, designed to prevent the ingress of dust and other contaminants. The sliding interface of a sliding 747
bearing, generally consisting of a suitably greased PTFE or UHMWPE sliding material at one side (that of the bearing’s core) and a larger stainless steel sheet on a steel sliding plate at the other, is generally its most vulnerable part. This is primarily because, with time, the sliding material wears away and becomes scratched and damaged as it moves along the sliding plate, especially when the interface has become contaminated by dirt and debris. Therefore, the bearing, and its position and orientation in the structure, must be designed to minimise the risk of damage and deterioration to the sliding interface’s materials. In particular, it is generally recommended to place the sliding plate, which is larger than the PTFE/UHMWPE disc, at the top of the bearing – unless it must be placed at the bottom, for example, for more efficient load transfer to a steel bridge deck. If placed at the bottom, the relatively large stainless steel sliding sheet would be liable to collect any dirt and debris that comes to rest on its surface, contaminating the sliding interface. Such an “upside down” bearing orientation would be of particular concern in the case of a bridge in a marine environment, as the collection of seawater on the stainless steel sliding sheet would result in accelerated corrosion. While it may be possible to prevent the ingress of wind-blown seawater, to a large extent, by the provision of a waterproof shroud or similar protective cover, it should be ensured that excess humidity under the cover is avoided. Such humidity, which may develop inside a cover that is waterproof but not airtight, is also likely to accelerate corrosion and deterioration, and the presence of the cover prevents the dispersion of the moisture in the air at the bearing by wind – the normal mechanism. 5.1.6 Bridge design measures which can protect bearings and extend their lifespan The bridge designer may consider measures, separate from the bearings, which have the effect of protecting the bearings and enabling them to serve a longer life. For example: – the fitting of hydraulic dampers to a bridge deck which would otherwise experience fast, erratic movements at its bearings, could reduce the movements and their detrimental effect on the bearings; or – the use of an automated structural health monitoring (SHM) system may provide ongoing confidence in the performance of the structure’s bearings, especially where the magnitude, frequency or nature of the deck’s movements, or other influences, cannot be predicted with confidence (see Section 5.3.2 below). 5.1.7 Quality of design and manufacture A comprehensive QA/QC system, for example in accordance with ISO 9001, and approvals of design and manufacturing processes in association with the issuing to suppliers of general approvals to supply the product in certain countries without further evaluation, can also provide confidence in the ability of a particular supplier to provide a product of the required quality. 5.2 Ensuring the quality of installation The importance of proper installation to the correct functioning and durability of a bridge’s bearings should be fully appreciated. For example, a bearing should be installed in such a way that all its parts are properly supported and will not be subjected to any unnecessary forces. In the case of a sliding bearing that accommodates deck movements, it is vital that the sliding surfaces be parallel to each other and to the direction of deck movement. The pre-setting of a sliding bearing (i.e. the position of its sliding plate relative to the rest of the bearing) must be appropriate for the location of the connecting point on the deck at the time of installation, considering the prevailing structure temperature, with allowance for the future deck movements in all directions that the bearing must accommodate. And sliding bearings must have their transport fittings, which hold them together until fully installed, cut at the correct time to avoid damaging constraint forces. The probability of a bearing’s being installed properly can be enhanced by the provision of proper access for the installation crew, and by limiting the size of each bearing for improved constructability – e.g. by the use of spherical bearings with UHMWPE sliding surfaces, which are typically roughly twice as strong as other steel bearings containing PTFE or elastomer. Many other factors must also be considered and checked during installation. But all too often, bearings are installed with insufficient care or expertise. It is thus important that the installation of 748
a bridge’s bearings is supervised by a competent person who is familiar with the design and needs of the particular bearing type. Supervision provided by the bearing manufacturer may be the best solution and is generally to be recommended. 5.3 Ensuring the adequacy of inspection and maintenance activities As noted in Section 4.2 above, proper inspection and maintenance are essential for the long-term functioning of a bridge’s bearings. As also noted above, these activities often do not get the attention or resources they deserve, resulting in durability and other problems. The main factors inhibiting good inspection and maintenance practice are likely to include – required effort/lack of sufficient inspection personnel – access limitations and – lack of technical understanding among engineers and maintenance personnel. In commentary on each of these points, it should be noted: – Avoiding inspection and maintenance work to save money is generally counter-productive, as this is likely to lead to far higher maintenance costs and earlier bearing replacement; – Access limitations should certainly be avoided during the bridge design and bearing selection phase, as noted in Section 5.1.4 above; and – Technical understanding is certainly important, so that the potential consequences of any issues observed can be investigated, and so that the correct course of action to address such issues can be selected. An enhanced recognition of these facts among those who are responsible for funding and arranging bearing maintenance will play an important role in improving the performance of the bearings and thus minimising their life-cycle costs. In undertaking inspection and maintenance work, reference should be made to all applicable regulations and guidelines – such as, in Europe, the European standard EN 1337-10, “Structural bearings – Part 10: Inspection and maintenance” (CEN, 2003). 5.3.1 Measures that support future inspection and maintenance efforts Bearing inspections can be made considerably easier by the provision, at the time of the bearing’s supply, of a type plate on each bearing and a movement scale on each sliding bearing. Type plates should identify the bearing type and its unique location, stating the year of installation and summarising its key parameters and characteristics, while the movement scale should enable the bearing’s correct sliding functionality to be confirmed. Of course, type plates and movement scales should be located in such a place that enables them to be read during inspections with a minimum of difficulty – ideally without the need for special access equipment. Inspection and maintenance efforts can also be supported by the keeping and consultation of proper records of previous installation, inspection and maintenance work on the same bearings. 5.3.2 The role of modern structural health monitoring Of course, inspections, and the investigations associated with remedial works, are no longer limited to purely manual efforts; great benefits, in terms of both ability and expense, may be offered by automated SHM systems (Spuler et al. 2011). The formidable data measuring power of today’s SHM technology can be used to measure and report on any chosen variables in the bridge’s condition (saving manual inspection visits), or to precisely define a maintenance or repair challenge and thus refine and optimise the selected solution. Such systems can even be programmed to send alarm messages to the bridge engineer, by email or SMS, should any measured variable exceed a predefined boundary value – thus ensuring that any potential problem can be dealt with before it develops into a costly repair project. The use of such systems can thus, in many cases, enable the life-cycle costs of a bridge’s bearings to be optimised. 5.4 Designing for exchangeability When the time comes to replace a bridge bearing that is anchored to a concrete structure, the effort required will be greatly reduced if the existing bearing, when installed, was equipped with anchor 749
plates. The bearing itself, bolted or keyed to each anchor plate, can then be removed without any breaking out of concrete and with only minimal lifting of the bridge deck. The design of the bridge structure should also consider future bearing replacement needs, with proper access allowed and adequate space for bearing removal and insertion. When guided bearings are replaced, horizontal loads must typically be temporarily transferred to neighbouring axes. Ideally, the bridge’s design should include consideration of this situation. Alternatively, non-guided bearings may be designed to serve temporarily as guided bearings. Such “hybrid” bearings may be the preferable solution for large single span or cable-supported bridges. 5.5 Planning of bearing replacement works Planning of replacement works should consider the complete history of the bearings and the structure, with reference to all available records from installation and inspections, and should confirm the demands (forces, movements, rotations etc.) that must be fulfilled by the new bearings. It should be noted that the demands on new, replacement bearings may differ from those applied to the design of the original bearings – for example, with higher loads arising due to heavier traffic, or smaller sliding movements due to the dissipation of concrete shrinkage and creep. The use of an SHM system may prove particularly beneficial for such purposes. 6 CONCLUSIONS Bridge bearings are critically important, relatively complex structural components. They are also much less robust than the main bridge structure, and thus require proper periodic maintenance and, from time to time during the longer life of the bridge, replacement. Consideration of the long-term costs associated with the bearings – including for supply, installation, maintenance, and replacement of the bearings, and other costs such as those resulting from the traffic disruption caused by replacement works – demonstrates the importance of devoting adequate attention and expenditure to the procurement, installation and maintenance of high-quality bearings. Recognition of this, and consideration of the measures which are proposed to assist in implementing a long-term strategy, can help minimise the life-cycle costs of a bridge’s bearings – for the benefit of owners, users and society at large. Although much lower than replacement and user costs, supply costs unfortunately still often play a dominant role in the selection process, because a lower-cost, low end product may fulfil short-term needs. It is therefore important that bridge construction contracts are so devised, that the party that selects the bearings has a real incentive to ensure their long-term performance. Of course, life cycle considerations should not consider only financial and user costs, but also the wider impacts on society and the environment. It seems likely that the future generations who will be responsible for the management of structures built today will be yet more concerned with environmental protection and conservation of natural resources than we are today, and the provision of bearings which achieve longer life with less maintenance and repair effort, and which can be replaced with less effort and waste, will be highly appreciated. REFERENCES CEN. 2003. EN 1337-10: Structural bearings – Part 10 Inspection and maintenance. European Committee for Standardization. Brussels, Belgium. Hawk, H. et al. 2003. Bridge life-cycle cost analysis (NCHRP Report 483 by the Transportation Research Board, National Research Council). Washington, USA. Spuler, T., Moor, G. & Berger, R. 2011. Automated monitoring systems for bridge bearings and expansion joints. Proc. 7th World Congress on Joints, Bearings and Seismic Systems for Concrete Structures. Las Vegas, USA. Spuler, T., Moor, G. & Savioz, P. 2013. Exceptional bearings for landmark structures – examples of challenges posed by selected bridges. Proc. 8th International Cable Supported Bridge Operators Conference. Edinburgh, UK.
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Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
Application of the Monte-Carlo method to calculate the life-cycle costs of bridges C. Hofstadler & M. Kummer Graz University of Technology, Institute of Construction Management and Economics, Graz, Austria
ABSTRACT: The application of the net present value method for the calculation of life-cycle costs provides a deterministic calculation where each computation step results in a single net present value that is associated with many uncertainties whilst allowing no conclusion as to the chances seized or risks taken. In fact, this single value represents only one of many possible scenarios that might arise from the combination of the input variables. The use of Monte-Carlo simulations can systematically account for uncertainties that input variables are associated with. This method allows to consider varying anticipated service lives of structural components, and distribution functions can be applied to costs, rehabilitation intervals and – last but not least – the computed interest rate. This approach delivers a larger amount of information whilst also increasing the level of decision confidence. This paper outlines a possible approach to the chance- and risk-based life-cycle costing of bridges. 1 INTRODUCTION Life-cycle costs serve as a benchmark, provide important arguments in favour of choosing a particular construction method and form the basis for any investment decision. In the construction sector, life-cycle costs are essentially determined by the type of building or structure, the characteristics of discretionary and elementary production factors, the quality, combination and processing of these factors, the type of use, maintenance and care, and by the applied calculation methods and their input variables. Investment decisions for construction projects are made not only on the basis of environmental and societal considerations but also on the basis of economic aspects. In this respect, the focus is not only on construction costs but is also increasingly shifting to the costs generated over the entire life cycle of the building or structure. These factors will become even more important not only in building construction but also for engineering structures, such as bridges. To be able to plan and assess repair and maintenance work to be performed on infrastructural facilities, it is necessary to ensure a life-cycle cost estimate for these structures that is as accurate as possible. Irrespective of the applied method, the quality and plausibility of results will depend on the reliability and accuracy of input variables. In most cases, a deterministic method is used. Singular input variables are applied to arrive at an output that is again just a single figure. To improve plausibility and reliability, individual results are analysed as a function of time, and their curves are represented for the considered period. Such a calculation is not suitable to systematically account for input parameter ranges; nor does it provide an appropriate basis to determine probabilities of occurrence. This paper refers to an example (based on Jodl 2010a and 2010b) to demonstrate how a probabilistic calculation of life-cycle costs can be performed for bridges. Any calculation of life-cycle costs includes elements that stem from costing and investment calculations. This paper outlines the fundamentals of the net present value and terminal value methods. What is crucial to arrive at sound comparisons of various options is the difference in the number and type of included categories of costs of use, as well as the quality and completeness of the data on which the calculation is based. This paper demonstrates the advantages of a probabilistic method compared to the deterministic approach, which has previously been the method of choice. 751
2 FUNDAMENTALS The results of life-cycle costing exercises deliver conclusions as to the economic viability of the project. In requirements planning, for example, they reveal the maximum range of management options within a given project. User requirements, planning or refurbishment scenarios are used as a basis to develop and analyse life-cycle models, which serve as a basis for decision making and/or certification. The basic formula shown below (see Equation 1) is at the very heart of life-cycle costing. These costs include the costs of a property (such as a production line or a building) and its components that are incurred over the entire life cycle during which they fulfil relevant use(r) requirements. Life-cycle costs (LCC) are composed of construction costs (C), operating costs (O), maintenance costs (M) and demolition costs (D).
2.1 Methods of investment calculus Specific investment calculation methods have been developed because investment decisions are one of the most important decision categories in the economic field, which is why they should not be made intuitively. The diverse range of economic criteria for investment alternatives is usually reduced to a single metric that should be as plausible as possible. Methods of investment calculus can be distinguished depending on their micro- or macroeconomic focus (Veit 2012). Static investment calculus is based on costing and profit and loss accounting; it is also referred to as the “classical” method. The basic principle of this approach is to relate the economic criteria that apply to the entire service life to an average period, which is usually an average year of use for which annual averages are used. This method neglects the temporal aspect of payments, i.e. the exact times or dates of payments (Bauer 2012). For the purposes of the calculation, it is thus irrelevant if scheduled major maintenance works are performed in the fourth or fifth year of use, for example, as long as total maintenance costs incurred over the entire service life are identical in both cases (and thus the average maintenance value per each year of use). This method has been strongly criticised not least because of its neglect of exact payment times. Dynamic methods avoid averaging; instead, exact payment times are considered over the entire service life. Accordingly, all cash inflows and outflows that occur during the life-cycle in the form of revenues and expenses must be known, or related assumptions must be made. These flows of funds are compounded or discounted using the compounded interest method known from financial mathematics and the imputed interest rate q (Bauer 2012). Dynamic methods are also often specified in the various sets of criteria of certification systems (in the field of building construction) if a life-cycle costing exercise is required. Furthermore, a distinction can be made between different sets of expectations. The use of deterministic methods requires known, robust input variables. In the case of uncertainties, however, it is recommended to at least consider ranges of values or to apply probabilistic calculation methods. On the most basic level, any life-cycle costing analysis is to capture costs and to evaluate cash flows in terms of their amounts and points in time when they occur. Dynamic methods are predominantly applied to comparatively long periods, which is why they are particularly suitable for infrastructural projects. Discounting makes these cash flows comparable to the interest rate, which enables a phase-specific identification and analysis of life-cycle costs. The following sections outline the principles of the net present value and terminal value methods. 2.2 Net present value method The net present value method starts from the assumption that the net present value of any and all future expenses and revenues plus construction cost permits a conclusion as to whether an investment is profitable or not. The net present value is the total of the cash flows discounted 752
by q, adjusted for the investment amount at a given point in time. It represents the amount that the investor generates over and above the capital employed. For projects that cannot generate directly attributable revenues (such as bridges constructed by public institutions that are not covered by a road toll system that would generate revenue), the net present value is used to compare the costs of various project options. The following sections of this paper relate net present values and terminal values exclusively to expenses (costs) whereas revenues are not considered. Accordingly, the related amounts are shown as positive amounts. The profitability of an investment will thus increase the lower the costs and thus the net present value or terminal value are. Accordingly, the net present value (“goodwill”) of an investment property can be linked to the following parameters to be determined for this purpose: – – – – –
the accuracy of construction costs cash flows – (revenues and) expenses – during use the anticipated residual value (revenue or cost) the selected or determined imputed interest rate the period under consideration The following equation is used to calculate the net present value:
where AInvest is the total of cash flows pertaining to land purchase, planning, design and construction; AOperation is the total of periodic and non-periodic cash flows in relation to use, operation, management, maintenance and repair and replacement; ERevenue represents periodic revenue (for example from road tolls); EResidual is the terminal value at the end of the period under consideration; q is the discount rate; µI is the inflation index, which depends on the cost group (taken as 0 in the worked example); t is the considered period; and tb is the point of reference. (Girmscheid 2006). Jodl states that the interest rate has a major influence and may result in changes to the entire maintenance approach, alter a potential ranking of bidders or pose a challenge to the profitability of a project. However, he uses a constant interest rate of 4.00% in all his calculations (Jodl 2010a). There is thus a strong discrepancy between the attribution of a major influence on net present value on the one hand and the use of a specific interest rate on the other. Since it is the very essence of life-cycle costing exercises to look very far into the future, it is useful to apply ranges of input variables to the calculations, rather than using deterministic values, to make the effects of uncertain parameters on the overall result transparent and to ensure that these effects can be assessed in an appropriate fashion. Jodl adds that the service lives of the individual structural components do not have any significant influence on the maintenance philosophy as long as they vary within realistic ranges. He considers annual repair and maintenance costs to be crucial for individual projects. 2.3 Terminal value method Compounding makes it possible to directly convert the net present value to a terminal value. The latter represents the value of cash inflows and outflows at the end of the period under consideration. Figure 1 shows the qualitative relationship between net present value and terminal value for the deterministic calculation. The upper portion of the diagram includes bars that symbolically represent individual expenses. At the time t0 , an investment is made (construction costs), which is then followed by annual expenses in the same nominal amounts (t1 ; t2 ; t3 ; and t4 ). At the time t3 , an additional expense is incurred due to the repair of specific structural components. All expenses incurred 753
Figure 1.
Relationship between net present value and terminal value.
after the time t0 must be discounted using Equation 2. Together with the investment amount, they represent the net present value in relation to the time t0 . Accordingly, the terminal value of expenses is calculated by compounding this net present value up to the time t4 (i.e. the end of the period under consideration). Net present values and terminal values can be calculated for any point in time by compounding and discounting; their graphical representation is a curve over time. 3 APPLICATION IN THE CONSTRUCTION INDUSTRY The last few years have seen a growing demand for and heightened interest in life-cycle costing exercises in the construction industry. This is due to increasingly demanding requirements in the context of investment decisions, but also to a greater awareness of options to manage opportunities and risks. Life-cycle analyses are also gaining in importance when it comes to engineering structures such as bridges or tunnels with exceedingly long service lives. 3.1 System of targets for life-cycle analyses Quality, time, costs, quantity, susceptibility to disruptions and process quality are the key decision and management variables that need to be analysed more thoroughly in order to adequately assess a given construction project both in quantitative and qualitative terms. Project quality is a major driver of follow-up costs because a higher quality standard usually implies longer maintenance intervals (for instance, due to the robustness and ease of maintenance of surfaces) and thus enhances operational quality. Consequently, shorter product life cycles cause an increase in life-cycle costs. The growing demand for a holistic approach in the construction industry requires an extension of the conventional “magic triangle” of costs, deadlines and quality that goes beyond project targets and encompasses operational targets as a new criterion (Hofstadler 2014). Operational targets focus on life-cycle costs, sustainable quality standards and long-term, timerelated targets such as warranties, service agreements and contracts. Furthermore, it is necessary to overcome the previously applied rigid set of targets and to move towards a strategic sourcing strategy that includes long-term goals, as shown in Figure 2. In addition, process quality and 754
Figure 2.
System of targets for the life cycle of construction projects (Hofstadler 2014).
susceptibility to disruptions should be assessed both in the construction project phase and in the operational phase (Bichler 2014). Prior to performing life-cycle cost calculations, the system boundaries relevant to the considerations need to be defined, and the existing cost base must be taken into account. When determining alternatives, the options to be compared need to be defined. This step should ascertain whether a comparison is both possible and useful (functional equivalent). The following phase involves the gathering of information to determine all required parameters. Using established methods of investment calculus, options can be analysed and related decisions be made. Calculation of the net present value or terminal value is a key tool in this process. Probabilistic methods, such as Monte-Carlo simulations, can be applied to include uncertain parameters in the calculation of life-cycle costs. 4 MONTE-CARLO SIMULATION Monte-Carlo simulations make it possible to include uncertainties of input variables in the calculation process. For this purpose, distribution functions are allocated to the calculation parameters of a deterministic computation model. For each of the iterative steps of the simulation, random values within the ranges of the specified distributions are selected (Latin Hypercube Sampling is used in the case discussed in this paper). This process is repeated several thousand times, and the results of the individual iterative steps are represented in histograms. These histograms provide information on statistical variables (such as ranges, mean values, modal values, spreads or skewness). Furthermore, selecting a certainty level represented by a deterministic value provides direct visibility of the risk taken, or the existing chance. There is thus a shift in the risk/chance ratio depending on the selected certainty level. If the certainty level is exactly identical to the median, there are identical risks and chances of exceeding or failing to achieve this value. (Hofstadler & Kummer 2014) 755
Table 1. Expected input values for the life-cycle costing exercise. Component
Constr. cost [€]
Service life [Years]
Maintenance cost [% of constr. cost]
Demolition and admin. [% of constr. cost]
Substructure Skeleton Equipment 1 Equipment 2
3,100,000 4,900,000 600,000 1,400,000
129 77 20 30
0.41 0.72 1.50 1.20
22 22 22 22
Selection of the probability distributions for the input variables is crucial because it defines the “character” of the input parameters: Are open-ended (infinitely negative or positive) distributions permissible and plausible; are distributions continuous or discrete, or can a higher proability be assigned to any value within the range? For instance, an expert survey conducted at Graz University of Technology found that the characteristics of the log-logistic distribution best corresponded to the information provided by the respondents with respect to labour consumption rates of shuttering works. Further studies are currently being conducted at Graz University of Technology to determine the shape of certain distribution functions for input parameters. For reasons of simplicity, the example outlined below uses triangular distributions that are defined by minimum, expected and maximum values. 5 WORKED EXAMPLE This worked example should demonstrate the application of the Monte-Carlo method to calculate the life-cycle costs of a bridge. Baseline data is taken from an example cited by Jodl (Jodl 2010a and Jodl 2010b) and refers to the “Wientalbrücke” project. The substructure of the bridge was built from concrete; the skeleton (superstructure) consists of reinforced concrete. Table 1 shows the expected values for the bridge’s construction cost and service life. Jodl used a constant interest rate of 4.00% to compound and discount expenses over the years. To calculate total construction costs, construction costs in the stricter sense are multiplied by 1.10 to take administration costs into account. Life-cycle costs are calculated for a service life of 70 years. During this period, pieces of equipment are repeatedly removed and replaced. The costs of demolition/removal, including administration costs, amount to 22% of relevant construction costs. Furthermore, varying maintenance costs are incurred in each year for the individual structural component groups, which are shown as percentages of the relevant construction costs in Table 1. This deterministic data is fraught with major uncertainties; it does not provide any information on the probability the calculated life-cycle costs are associated with. To incorporate these uncertainties in the calculation process, symmetrical triangular distributions are applied to the input parameters instead of deterministic values. Any available historical data or specific distributions that are known for certain parameters (such as from analyses of expert surveys) can be used for the simulations. Tables 2 & 3 state the minimum, expected and maximum values that define the triangular distributions of the calculation parameters for construction costs, expected service lives and maintenance cost percentages. Costs of demolition and administration associated with the individual components amount to a constant share of 22% of construction costs. A triangular distribution is also used for the imputed interest rate (3.00%, 4.00% and 5.00%). For the purpose of calculating life-cycle costs, maintenance costs need to be considered that are incurred each year; these are determined as a percentage of construction costs. For each iterative step, the use of distribution functions on the input side results in random maintenance costs that lie within the defined ranges. For each iteration, a service life within the defined ranges is randomly selected for the maintenance and repair (demolition and replacement) of equipment. Stated costs of demolition, administration and replacement are then incurred in this randomly determined year. Equation 2 can be applied to calculate a net present value for each year. Net present value totals can 756
Table 2. Selected range of construction costs and expected service lives. Construction cost [€]
Expected service lives [years]
Component
MIN
ERW
MAX
MIN
ERW
MAX
Substructure Skeleton Equipment 1 Equipment 2
2,900,000 4,500,000 550,000 1,200,000
3,100,000 4,900,000 600,000 1,400,000
3,300,000 5,300,000 650,000 1,600,000
110 70 18 28
129 77 20 30
148 84 22 32
Table 3. Selected range of maintenance costs. Maintenance cost [% of constr. cost] Component
MIN
ERW
MAX
Substructure Skeleton Equipment 1 Equipment 2
0.35 0.68 1.00 0.80
0.41 0.72 1.50 1.20
0.47 0.76 2.00 1.60
Figure 3. Terminal values for the entire project for a 70-year period, including a terminal value histogram at the end of this period.
be converted to terminal values in a compounding step. Revenues are not included in this example, but can be integrated in the calculation very easily by stating annual revenues, for example (note the inverse prefixes: expenses are positive, revenues are negative). Nor are inflation effects (inflation proper as well as price increases, such as for energy or services) taken into account for simplicity reasons. Figure 3 shows the terminal value curves during the considered period of 70 years. The middle curve represents the mean terminal value. The other two curves define the 5% and 95% quantiles of 757
the terminal value determined in a probabilistic calculation applying the Monte-Carlo simulation method. Despite the fact that symmetrical distributions were used for all input variables, terminal value distributions become increasingly skewed over the length of the considered period, resulting in the distribution shown in Figure 3. This is due to the fact that the distribution for the interest factor also shows an increased skewness for longer periods due to the mathematical relationship applied [= (1 + q)(t−tb) ]. Spreads of simulation results, mean values and other statistical variables can be compared to each other when considering other project options to arrive at a conclusion regarding the opportunities and risks associated with the individual options. 6 SUMMARY AND OUTLOOK When performing life-cycle calculations, it is crucial to consider the system boundaries in terms of underlying costs and related cost groups, as well as the system boundaries defined to evaluate various options in order to ensure their comparability. To analyse various investment decisions, the defined system boundaries and selected methods need to be thoroughly evaluated to enable this comparability in the first place. This paper demonstrated the probabilistic calculation of life-cycle costs by means of a MonteCarlo simulation (Latin Hypercube Sampling) using the net present value and terminal value methods. This procedure is associated with the benefit that it not only calculates a deterministic value for life-cycle costs but also enables histograms to be shown that permit an estimate of probabilities of occurrence of certain deterministic values. The use of Monte-Carlo simulations can systematically account for uncertainties that calculation parameters (particularly service lives, costs and interest rates, but also revenues, such as from road toll schemes) are associated with on the basis of distribution functions. Only a probabilistic calculation will permit conclusions as to how uncertain life-cycle costs stated as single figures are for exceedingly long periods. The spread of probabilistically determined terminal values increases continuously in line with the length of the considered period. Application of probabilistic methods thus improves results whilst increasing the level of decision confidence in the case of several project options with varying maintenance costs and service lives. REFERENCES Bauer, U. 2012. BWL Bau/BWL Enzyklopädie/Grundlagen der BWL. Lecture Notes, Graz University of Technology. Graz. Bichler, D. 2014. Lebenszykluskostenrechnung in der Bauwirtschaft. Master thesis at the Institute of Construction Management and Economics. Graz University of Technology. Graz. Girmscheid, G. 2006. Risikobasiertes probabilistisches LC-NPV-Modell – Bewertung alternativer bau-licher Lösungen. Bauingenieur 81, September 2006: 394–405. Düsseldorf: Springer-VDI-Verlag Hofstadler, C. & Kummer M. 2014. Systematischer Umgang mit Produktivitätsrisiken in der Auftragskalkulation. In Heck, D.; Mauerhofer, G.; Hofstadler, C. (ed.), Risiken im Bauvertrag – Bau-betriebliche, bauwirtschaftliche und rechtliche Aspekte; Proc. 12. Grazer Baubetriebs- und Bauwirtschaftssymposium, 11 April 2014. Graz: Verlag der Technischen Universität Graz. Hofstadler, C. 2014. Methoden zur Ermittlung von Lebenszykluskosten. In Nguyen, V.T.; Maydl, P.; Freytag, B. (ed.), Nachhaltig Bauen mit Beton; Proc. 2. Grazer Betonkolloquium, 25–26 September 2014. Graz. Jodl, H.G. 2010a. Lebenszykluskosten von Brücken – Teil 1 – Berechnungsmodell LZKB. Bauingenieur 85, May 2010: 221–230. Düsseldorf: Springer-VDI-Verlag Jodl, H.G. 2010b. Lebenszykluskosten von Brücken – Teil 2 – Software LZKB. Bauingenieur 85, Mai 2010: 231–240. Düsseldorf: Springer-VDI-Verlag Veit, P. 2012. Betriebswirtschaftslehre – Bau – Kostenrechnung, Investition. Lecture Notes, Graz University of Technology. Graz.
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Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
Selective use of non-corrosive rebar to increase concrete durability Alexis E.C. Borderon Valbruna, Vicenza, Italy
ABSTRACT: Concrete life is submitted to the influence of climate and time. Much is done to design mixtures that ensure concrete durability, as well as preventing its collapse. Many solutions are today available to increase service life; coatings of the rebar or the concrete surface, composite materials as reinforcements, galvanized steel, cathodic protection. This paper is a pragmatic guideline based on leading engineers methods to achieve durability and sustainability of the structures, through the use of non-corrosive concrete reinforcements where needed only.
1 INTRODUCTION Service life of concrete constructions has been increasing since the economical crisis started, due to obvious financial issues as maintenance is costly and difficult to predict. The demand from the builders has been either monitoring of the structure in order to be able to plan and finance in time maintenance or repair, or specify the project with a longer durability from the start, as the savings in time exceed by far the design and construction costs.This paper examines the different options to access durability and their benefits, and the technical solutions to provide it. 2 RESEARCH SIGNIFICANCE Statistics on concrete repair show clearly that corrosion and concrete disintegration happen differently on structures, defying hereby the notion of exposed zones. There are several solutions to increase durability precisely where needed, using non corrosive reinforcements. However, all reinforcing materials do not perform equally in time, given exposure, concrete mix and cover. Furthermore, we will study the relevance of predictive service life models, designed to predict concrete durability. 3 CONCRETE DETERIORATION Most of the deteriorating concrete in infrastructures is due to the corrosion of the carbon steel rebar. The cost is recognized to be 1% of the project price per year. (cf. U.S. Federal Highway Administration) For the public administration or the private builder, it means build 1 and pay for 2 for the expected 100 years lifetime. If costs are important, structural integrity is crucial. Still, concrete decay may lead to collapse, and 70% of structural failures happen unloaded, due to only 5% of the reinforcement been corroded. 4 CONCRETE COVER High performance concrete (HPC) is nowadays preferably used in civil engineering projects, to protect reinforcement, to resist deterioration and to provide adequately high strength to fulfil 759
Figure 1.
Service life in years as actual value on-site is 0.6 in (15 mm) lower than the nominal design value.
the structural requirements. As it strength usually remains within the range of 7300–11600 Psi (50–80 MPa) or higher, microcracking and cracking is inevitable due to the high filler content of the concrete. As higher concrete cover causes cracks, and low concrete cover does not protect the rebar from corrosion in time, because highest chloride content is 1–1.2 in. (25–30 mm) in the concrete. Corrosion of reinforcing steel and concrete cover are therefore closely related, in a exponential relation according to the Dura Crete probabilistic design methodology. (i.e. [2]). As service life is exponentially related to concrete cover, little variation on-site means therefore extensive service life reduction. In this example, the structure’s design life of 100 years is reduced to 15 years, as concrete cover had been measured to be just 0.6 in. (15 mm) lower. (See figure 1). Differences in concrete cover might result in severe durability decrease, depending on the exposure. 5 CORROSION RESISTANT REINFORCEMENT Durability is also highly dependent on the rebar type as there are different materials available with different properties. Mild steel, black steel or carbon steel reinforcement – is efficiently protected against corrosion when cast into a good quality alkaline and chloride free concrete. Only when chlorides in sufficient quantity reach the surface of the reinforcement will the passivating effect be eliminated and corrosion may start. The following products with different degrees of resistance against corrosion are available: • Stainless steel reinforcement (SSR). • Epoxy coated reinforcement. • Hot dip galvanised reinforcement (zinc coating). This application is very limited. In general zinc coating is not considered adequate – or cost-effective – for structures exposed to chlorides. • High-Chromium steel rebar, mostly used in North America. low carbon, chromium, microcomposite steel with high tensile strength. • Non-metallic reinforcing bars such as reinforcing bars from glass fibres, aramid fibres or carbon fibres. The non-metallic reinforcing bars have limited applicability due to the major differences needed when constructing; they can’t be bent. 760
5.1 Stainless steel reinforcements (SSR) The use of SSR in zones being exposed to high chloride concentrations is considered a highly reliable solution. It can ensure a very long problem-free service life in that part of the structure, provided the concrete itself is made sufficiently resistant to avoid other types of deterioration such as alkali-aggregate reactions, sulphate attack or salt scaling. In addition, there are regions in the world where the chloride contamination is so widespread that all aggregates and mixing water are more or less chloride contaminated. Sometimes a 10–20 year service life has become the accepted norm in such regions, or continued repair works have been the accepted solution. Using SSR may often solve this problem completely. Used selectively in the most exposed zones of the structure, the increased costs per kg of SSR compared to the costs of normal steel will most often have only marginal or negligible effect on the overall initial construction costs. In addition the service life costs will be reduced considerably due to savings in future repair and maintenance. From a practical point of view this technology is particularly interesting because it “only” solves the corrosion problem. All other techniques and technologies within design, production and execution of reinforced concrete structures remain practically unchanged. Of particular importance is the often-overlooked fact that SSR can be coupled with normal mild steel reinforcement (carbon steel) without causing galvanic corrosion. The reason is that the two types of steels reach nearly the same electro-chemical potentials when cast into concrete. This leads to the possibility to use SSR only in those parts of the structure where it is considered necessary, and then reinforce the remaining parts with ordinary mild steel reinforcement. A very convincing documentation of the performance of stainless steel reinforcement in highly chloride contaminated concrete is presented by the 77 year old 1.36 mi. (2.2 km) long concrete pier out into the Mexican Gulf at Progreso in Mexico reinforced with stainless steel. No corrosion has taken place within the structure, despite the harsh environment and poor quality materials used in the construction. The chloride levels, at the surface of the reinforcement were more than 20 times the traditionally assumed corrosion threshold level. A newer, only 35 years old parallel pier has already perished due to reinforcement corrosion of the ordinary carbon steel reinforcement used in this structure. (i.e. [3]).Finally, an additional benefit of using SSR is the fact that stainless steel is a poorer cathode than carbon steel. Therefore, SSR can be beneficial in those repair cases where ordinary carbon steel has corroded to such an extent that local replacement or added reinforcement is needed as part of a repair. In this way, the traditional problem of new corrosion developing on the reinforcement adjacent to the repaired area – and initiated in part by the repair – can be reduced or fully avoided. As it is recognised that the most serious durability problem for concrete structures is reinforcement corrosion it becomes evident that, the reliable and readily availability of stainless steel reinforcement at reasonable and foreseeable prices may change – or revolutionise – major parts of the building sector in aggressive environments, simply by solving the corrosion problem. 5.2 Epoxy coated reinforcement Epoxy coating of reinforcement has been used in North America since the mid 70’ies. The nearly unavoidable fine cracking occurring during bending, the pinholes occurring in the coating, and the inferior protective ability of the patch repaired zones and cut ends have led epoxy coated reinforcement to be a non reliable solution. The key source of uncertainty, namely the individual execution phases. North American experience has thrown serious doubts on this approach, when following the traditional procedure of coating straight bars individually, then cutting them to length and bending them to the required shape, see Figure 2 from the Florida Keys. In several states epoxy coating is not allowed by the local Departments of Transport and in Ontario, Canada, SSR is taking over within bridge construction and bridge repairs. The first public report on failing performance of epoxy coated bars was from January 10th 1992, which concluded that the “Epoxy coated rebar technology is flawed”. The technology is now slowly being phased 761
Figure 2.
Close-up of corroded epoxy coated reinforcement.
Figure 3. Details of the corroded epoxy coated reinforcement shown on Figure 2. The loose layer of epoxy on the corroding bar is clearly visible.
out, also in the Middle East and Gulf Countries, (see Figure 3–4, for corroded epoxy coated reinforcement.) As each type of reinforcement has different capabilities, their performance depends on corrosion resistance, measured by corrosion threshold.
6 CORROSION THRESHOLD 6.1 Concrete durability calculation Rebar corrosion depends on many and various factors; exposure, concrete mix design, concrete cover, cracks, and time. The resulting contaminated concrete due to chloride penetration will deteriorate over time depending on the corrosion resistance of the rebar. There are several models to calculate concrete durability, or corrosion initiation time. Notable ones are the RILEM report 14, the ACI committee publication 365. 1-R 42, the FIB Bulletin 34. Lab tests of accelerated concrete decay have shown significant variation between test results and estimated durability. Even minor changes in the original assumptions of the models can possibly result in major variations over a longer period of time. The longer the period, the more insecure concrete durability calculation is. (i.e. [4]). This has undoubtedly an influence on the safety factor one must base calculation on, when variations reach 25%. As the chloride threshold level for stainless steel bars is much higher than that carbon steel, the accuracy of the calculation is therefore increased. (i.e. [5]). 762
Table 1. Chloride threshold types for various reinforcing types.
CI-/OH% of CI- by weight of cement Kg of CI-/m3 concrete
Carbon steel
Stainless steel 316/1.4404*
Stainless steel lean duplex 318/1.4362*
24 >10
2,8
0,71
10,8
Epoxy coated rebar
Galvanized rebar
1,96
1,52
6.2 Chloride threshold Initiation of corrosion on reinforcing bars occurs when a critical chloride concentration is achieved. The corrosion resistance ability of the reinforcing material is therefore depending on the chloride threshold of the material and the surface condition altogether. There are several chloride threshold methods and units; CI-/OH- molar ratio, kg of CI-/m3 of concrete or % of CI- by weight of cement. Many tests have been done considering one or more steel qualities, however, as methods differ, results can’t be compared. The emerging chloride threshold seems to be % of CI- by weight of cement, representative of concrete corrosion resistance based on cement and chloride content only. It is also the only method performed for most kinds of reinforcing products, carbon steel, stainless steel, epoxy coated carbon steel and galvanized carbon steel. Table 1 shows the different chloride threshold values obtained through different sources tests so far. It is clear then that stainless steel can stand much higher chloride content before corrosion initiation than any other product. However, chloride content also depends on other factors, which can be calculated. 6.3 Chloride threshold to calculate corrosion initiation time Diffusion coefficient of chloride ions in concrete and remaining time before rebar start to corrode are estimated by Fick’s second law of pure diffusion (Equation 1):
where, C(x,t): chloride ion content at certain depth (x) and time (t) Ci: chloride ion content introduced as materials of concrete Co: chloride ion content at concrete surface x: distance from concrete surface D: diffusion coefficient of chloride ions in concrete t: time after construction The accuracy of this calculation is limited due to lack of an accurate D, which varies with CIand concrete age, quality and external conditions. However for the same D value, it is possible to compare the performance of the different types of reinforcing materials under equal conditions. Table 2 shows test performance of the different reinforcing products against carbon steel rebar, at an average depth of 2 in. (50 mm) in the concrete. It shows how effective each product compared to carbon steel before initiation of corrosion. Whatever part of the concrete structure, the performance of the reinforcing products is therefore higher than carbon steel, but they’re not needed everywhere in the structure. 763
Table 2. Test performance of reinforcing products against corrosion initiation time. Rebar type
Performance (× times better than carbon steel)
Carbon steel Epoxy coated carbon steel Galvanized carbon steel Stainless steel
1 2 2 40 to 50
7 SELECTIVE USE OF STAINLESS STEEL REBAR 7.1 Exposed zones As stainless steel is obviously an expensive but reliable solution, it should only be used where necessary to optimize project costs. Corrosion damage is linked to concrete cover (cc) and exposure, but rebar is assumed to be protected regardless of the exposure at cc > 4 in. (100 mm). It implies non-corrosive reinforcement only to be relevant in the outer layer of the structure. (i.e. [6]). Outer layer of the concrete is therefore regarded as exposed zone, with various positions:
7.1.1 Edges Unless isolated from water, winds and salts, edges are very much exposed to concrete deterioration and spalling or cracking particularly due to rebar corrosion. There is no evidence horizontal edges show more damage over time than vertical edges, though it seems likely.
7.1.2 Tidal zones The area between minimum and maximum height of water tide defines the tidal zone. This wet/dry environment of the concrete favours constant chloride/hydroxide presence, leading to rapid rebar corrosion. Surface chloride content is not as much an issue as diffusion (D) of chlorides in the concrete, which is the highest in the structure precisely in the tidal zone. It means the chloride content at rebar depth increases to reduce corrosion initiation time. It automatically limits the durability of the concrete if non-corrosive reinforcing material is used.
7.1.3 Splash zones The splash zone is where seawater particles wash the concrete surface, and is situated above the tidal zone. The surface chloride content is higher there than the tidal zone, but the diffusion is lower. Corrosion will therefore begin later than in the tidal zone if concrete cover is greater than 1.4 in. (35 mm) for carbon rebar. Chloride concentration is notably highest 1–1.4 in. (25–35 mm) in the concrete.
7.1.4 Joints As joints mostly transmit loads in the structure, they are crucial to its stability and durability. Whatever solution is adopted to avoid corrosion of steel in the joints, it will demand maintenance over time as leakage occurs even when sealed; dilatation hampers durable joint closure. The very nature of a joint also makes it difficult to inspect as well, so the best alternative is to use a non-corrosive material there. 764
Figure 4. Section view of stainless steel mesh and carbon rebar in concrete. Concrete is micro cracking up to the carbon rebar, but stainless steel mesh limits crack width to 0.01 in. (0,3 mm) in this case.
Figure 5.
Section view of stainless steel rebar in concrete. Concrete Cover is min. 1.2 in. (30 mm).
8 METHODS 8.1 Solution 1- edges, splash and tidal zones When concrete cover is over 2 in. (50 mm), carbon steel rebar is well enough protected according to the standards. To avoid cracks leading to earlier corrosion, using stainless steel mesh on the outer layer of the concrete is a known method. Stainless steel mesh is placed within 1.4 in. (35 mm) of concrete, mask size according to the crack width calculated against Euro Codes 2, Design of concrete structures.The obvious advantage of this solution is that reinforcing bar design is done as usual with carbon steel, but service life is prolonged as cracking is prevented through use of stainless steel mesh, that will not corrode even when concrete cover is low (0.4–1.4 in. (10–35 mm)). 8.2 Solution 2- edges, splash and tidal zones Most leading consulting engineers do adopt this solution with concrete cover of min. 1.2 in. (30 mm) to stainless steel rebar. It gives the necessary service life of at least 100 years according to the standards, and an actual high reliability when executed. The rebar is also structurally active, and entirely part of the reinforcing. Carbon steel rebar remains inside the structure with cc > 4 in. (100 mm).This solution is also practically easy to perform, as carbon steel rebar does not physically interfere with stainless steel rebar. 8.3 Solution 3 – day and construction joints Structural or not, joints are almost always exposed to chloride penetration and should therefore be reinforced with high corrosion resistance stainless steels. The following figure shows the joint 765
Figure 6.
Section view of a bridge deck, pile and beam. The rebar is stainless steel.
Figure 7. Carbon rebar solution on the left to 50 years service life, same structure on the right with stainless steel rebar on exposed zones to 150 years service life.
section of a bridge, and the stainless steel rebar designed to couple the three precast concrete elements; deck, pile and edge beam. 9 COSTS Proper use of stainless steel represents a small part of the reinforcement, as figure 7 summarizes. The figure represents the very same bridge with respectively carbon steel only on the left, and carbon steel plus stainless steel rebar on the right. Using stainless steel on exposed zones extends the service lifetime of the structure to 150 years instead of 50 years.The case of this Danish bridge, stainless steel rebar was only 1,7% of the total reinforcing steel, or 5,8 % price increase compared to carbon steel rebar only. 10 CONCLUSIONS Although there are alternative reinforcing solutions to carbon steel, they differ in local performance due to factors not yet accurate to calculate. It leaves the owner with a risk in time, even when 766
execution on site is flawless, as variations in service life keep high.With extensive service life, the reliable solution is to chose the product that limits the risk most, where needed only. Doing so, the costs of using stainless steel rebar remains low compared to the mainteance costs over time. REFERENCES Proceedings of the ICE – Forensic Engineering, issue 1, 2011. Concrete technology and durability design, 1999, Dr.-Ing.Carola Edvardsen, Section Manager concrete technology and durability Design, Cowi. Durability design of concrete structures in severe environments, second edition 2014, by Odd E. Gjørv. Durability of Reinforced Concrete Structures, Theory vs Practice. Albert K.H. Kwan and Henry H.C. Wong,Department of Civil Engineering, The University of Hong Kong. 04/2008 Chloride Threshold Levels in Clad 316L and Solid 316LN Stainless Steel Rebar, Michael F. Hurley, John R. Scully Center for Electrochemical Science and Engineering Department of Materials Science and Engineering University of Virginia. Guide for the use of stainless steel reinforcement in concrete structures, Gro Markeset, Steen Rostam and Oskar Klinghoffer – Nordic innovation centre project 04118, Norwegian Building Research Institute 2006. Concrete Diffusion Coefficients and Existing Chloride Exposure in North Carolina, Janos Gergely, Principal Investigator; Joshua E. Bledsoe, Graduate Research Assistant; Brett Q., Tempest Graduate Research Assistant and Iosif F. Szabo, Post-Doctoral Fellow. North Carolina Department of Transportation Research Project No. HWY-2004-12, Department of Civil Engineering, University of North Carolina at Charlotte – 9201 University City Boulevard Charlotte. 07/06/2006 Les armatures passives d’acier inox, Y.Tardivel – SETRA and C.Tessier – IFSTTAR. Les Plénières Journées Techniques Ouvrages d’Art 2011.
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Monitoring, maintenance and management
Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
Dynamic characterization and continuous dynamic monitoring of long span bridges E. Caetano, A. Cunha, C. Moutinho & F. Magalhães ViBest, Faculty of Engineering of the University of Porto (FEUP), Portugal
ABSTRACT: This paper characterizes the extensive research activity developed by the Laboratory of Vibrations and Structural Monitoring (ViBest) of FEUP in the fields of dynamic testing and continuous dynamic monitoring of long span bridges, briefly presenting results of some relevant applications as, for instance, the dynamic tests of Vasco da Gama Bridge, in Portugal, of the Millau Viaduct, in France, of the Humber suspension bridge, in UK, or the continuous dynamic monitoring of various Portuguese bridges, like the Pedro e Inês footbridge and the Infante D. Henrique roadway bridge.
1 INTRODUCTION During the last twenty years, the Laboratory of Vibrations and Structural Monitoring (ViBest, www.fe.up.pt/vibest) of the Faculty of Engineering of the University of Porto (FEUP) has developed a remarkable research effort in the field of Dynamic Characterization and Continuous Dynamic Monitoring of Long Span Bridges. This extensive work begun, in a first instance, with the development of very efficient and robust tools (hardware and software) for ambient, free and forced vibration tests on large bridges, with a special emphasis on the automatic implementation and application of the most sophisticated and powerful output-only modal identification techniques. In a second stage, special attention has been devoted to the development and implementation of long-term dynamic monitoring systems (active since 2007) in several kinds of large structures, with different typologies and materials, and targeting four levels of objectives: – The implementation of alert systems used for the safety checking of vibration serviceability limit states and the verification of the efficiency of vibration control devices; – The implementation of Structural Health Monitoring (SHM) systems enabling the vibration based damage detection; – The investigation of wind engineering problems based on continuous measurements on prototypes; – The experimental assessment of local vibration fatigue in old metallic railway bridges. This paper attempts to make a brief characterization of this extensive research activity, briefly presenting results of some relevant applications as, for instance, the dynamic tests of Vasco da Gama Bridge, in Portugal, of the Millau Viaduct, in France, of the Humber suspension Bridge, in UK, or the continuous dynamic monitoring of various Portuguese bridges, like the Pedro e Inês footbridge and the Infante D. Henrique roadway bridge. The mentioned applications involve different aspects, namely: the huge amount of high quality experimental data collected, the portability and wireless nature of measurement systems used in ambient vibration testing, the exhaustive application of the most recent and powerful output-only modal identification methods and the development of efficient software for continuous dynamic monitoring, removal of the effects of environmental, and operational factors and damage detection. 771
The implementation since 2007 of about ten continuous dynamic monitoring systems in large Civil structures with different typologies, working as demonstrators for owners of infrastructures, construction companies and designers, has led to a very high volume of experimental data, which have been conveniently organized in a digital data repositorium, enabling the future scientific cooperation and benchmarking with other international research groups focused on Structural Health Monitoring. 2 BASIS OF RESEARCH About thirty years ago, the experimental identification of relevant dynamic properties of large bridges was done based on conventional modal testing procedures previously developed in the fields of Mechanical and Aeronautical Engineering. Such tests involved the estimation of a set of frequency response functions (FRFs) relating the applied force and corresponding response at several pairs of points along the structure with enough high spatial and frequency resolution, and required the use of an instrumentation chain for structural excitation, data acquisition and signal processing. Though Forced Vibration Tests may lead to very accurate modal estimates, they present a strong drawback when dealing with large structures, which is the difficulty in exciting the most significant modes of vibration in a low range of frequencies with sufficient energy and in a controlled manner. In order to overcome this difficulty, some pioneering studies have been developed (e.g. Carder 1936, Vincent 1958, McLamore et al 1971, Trifunac 1972) in which only the response to natural excitation was measured and assumptions were made on the characteristics of this excitation, due to wind, traffic and environmental actions. The idealization of this ambient excitation as a zero-mean Gaussian white noise allows the construction of “pseudo” frequency response functions relating the responses at two points of the structure, which are the basis for identification of natural frequencies, damping ratios and the so-called operational modes. Despite the limitations of the former studies, which essentially provided an estimate of the fundamental natural frequency and the characteristics of the corresponding mode shape, the new ambient vibration test technique fostered intensive research, both in terms of hardware and of system identification algorithms. In fact, ambient excitation typically generates vibrations of low amplitudes, which measurement requires sensitive sensors. At the same time, the stochastic nature of excitation demands a statistic signal processing, which requires the recording of long time series. This implies important storage capacity, as well as more complex identification algorithms, able to separate noise from structural response and to identify closely spaced vibration modes excited in a wide frequency range (Felber, 1993; Van Overschee and DeMoor 1996; Peeters & De Roeck, 1999; Brincker et al 2001; Peeters et al 2005). The studies of Brownjohn et al (1989) conducted on the Humber Bridge, in UK, using a very limited number of sensors and recording the structural response in magnetic tape demonstrated the power of ambient vibration testing (AVT) in the dynamic characterization of very long span bridges. Further technological developments in transducer and digitizer technology combined with an increased computational power and storage capacity have permitted the improvement of the accuracy of measurements and of the spatial characterization of the dynamic behavior. More recently, the use of sensors conjugated with high level digitizers with local storage and time synchronization capabilities made feasible the accurate modal testing of very large structures in a short time and in a rather comfortable way. The tests at the Vasco da Gama Bridge (Cunha et al 2001) clearly mark a new era for the so-called operational modal analysis, which has further been emphasized with the application at the Millau Viaduct (Flamand and Grillaud 2006) and the re-testing of the Humber Bridge (Brownjohn et al 2010). Ambient vibration tests have therefore become a testing technique of great importance in the field of civil engineering, allowing accurate identification of modal properties of large structures at construction, commissioning or rehabilitation stages without interruption of normal traffic (Cunha et al 2007, Cunha et al 2013). In some applications, for instance related with the aerodynamic behavior of slender structures, the accurate identification of modal damping factors is a major problem in the identification process 772
due to the considerably larger scatter of modal damping estimates with regard to the corresponding natural frequencies and modal shapes counterparts. This is also true because the viscous damping assumption does not correspond exactly to real damping characteristics. Modal damping ratios increase gradually with levels of oscillation. This concern has led frequently to the complementary performance of free vibration tests, based on the measurement of the free vibration response of the bridge after release of a weight. In this way, a considerable increase of the level of vibration with regard to ambient excitation can be achieved, without the need to use an exciter, and a damping ratio estimate can be obtained from the fitting of an exponential curve to the envelope of the free decay response of the bridge, eventually conjugated with some filtering, in the case of closely spaced mode contribution evidence (Bietry and Jan, 1995; Cunha et al., 2001; Flamand and Grillaud, 2006). Extending dynamic testing campaigns to the continuous monitoring of dynamic properties is another direction allowed by the developments in the wake of AVT. In the past, monitoring of large structures was conducted with the main purpose of observing the evolution of the structural behavior by means of the measurement of deflections and inclinations, as well as other quantities related with the materials’ degradation. The introduction of acquisition systems based on data acquisition cards enabled the progressive incorporation of sensors for observation of the dynamic behavior. In an initial stage, minimal observation systems were installed, formed by two or three tri-axial accelerometers, placed for instance at the mid-span section or at the top of one or two towers, in the case of a cable-stayed or suspension bridge, and including a trigger system to record the dynamic response under specific dynamic conditions, such as earthquakes or wind storms. The reduction of the cost of computers and sensors allowed a progressive increase of the dimension of the monitoring systems, and in some countries the installation of dynamic monitoring systems with several tenths of channels became a normal practice. Taking into account the amount of acquired data, and the impossibility of permanent processing and management of collected data, many of the dynamic measurement systems rapidly became totally inefficient. More recently, a better definition of the necessities for continuous characterization of the dynamic behavior has taken place and information transmission through the Internet made feasible the continuous dynamic monitoring of the structural behavior (Wong 2004, Ko and Ni 2005), which may complement other components of structural monitoring, such as the monitoring of loads (e.g. wind, traffic loads), static behavior and durability (FIB 2003). These systems can presently play a very important role in the understanding of the structural behavior either during the bridge construction, or during the service lifetime, as will be shown with the examples presented in this paper. 3 DYNAMIC CHARACTERISATION OF LONG SPAN BRIDGES 3.1 General aspects Long-span bridges are typically characterized by very low natural frequencies and damping ratios. Depending on the structural type, a significant number of vibration modes may exhibit natural frequencies lower than 1 Hz. These modes will be excited by the wind and, if closely spaced in frequency, the structural response will exhibit a significant interaction. The characteristics above described result in important demands whenever their identification is aimed by means of ambient vibration tests. The accurate identification of low and close natural frequencies demands a high frequency resolution, which implies the recording of long time series. At the same time, sensors need to be sensitive in the very low frequencies, ie., from almost DC, and preferably less sensitive in the high frequencies, say above 50 Hz, in order to avoid the use of analogue filters or the requirement for a very high sampling frequency. Then, the identification of modes exhibiting high interaction or coupling implies the use of techniques for separation of modal contributes in the response, or the use of sophisticated identification algorithms. Finally, the correct identification of the characteristics of vibration modes demands a good spatial coverage, meaning the need for measurement of the ambient response in a large number of points along the structure. To accomplish the above mentioned purposes, the use of force balance accelerometers has been a normal practice, actually followed in the former tests (Carder, 1936; Vincent, 1958; MacLamore 773
Figure 1. bridge.
Recorders, sensors and GPS antennae during cross-calibration measurement at Humber suspension
et al., 1971). However, the evolution of electronics has led to a progressive improvement of the sensitivity and signal-to-noise ratio of these devices, so that accelerometers 165 dB in dynamic range and 2 micro-g peak-to-peak noise can presently be found in the market. A second key point in the testing of very long span bridges refers to the characteristics of data acquisition systems. Traditional systems used in dynamic testing have the so-called “star” configuration, which comprehends a computer and a data acquisition modulus located in a central part of the bridge, receiving data from the sensors distributed along the bridge by means of electrical cable. In the case of the Humber Bridge, the former tests conducted in 1985 (Brownjohn et al., 1989) were based on 3 accelerometers roving a number of points along the bridge, and 2 km of electrical cables were laid along the deck. For individual measurement campaigns, this strategy is very limiting, given the time and effort required to lay the cables along the bridge deck. This difficulty motivated the development of more comfortable and portable alternative decentralized systems, minimizing or even totally avoiding the use of electrical cables. In this perspective, a pioneering work has been developed at LNEC and FEUP (Cunha et al 2007), applying a system composed by several independent tri-axial seismographs conveniently synchronized on the basis of GPS sensors (Figure 1). Each of these devices incorporates a triaxial accelerometer, a digitizer, a memory disk and a battery with autonomy for eight to ten hours of continuous operation. The connection to a GPS antenna is further used for a time synchronization of the devices, which are initially programmed to sample at specific times and are roved along the bridge deck and the towers, in a completely independent form regardless of any cables, and stored in a convenient way in a rock bag for transportation in more complex sites. The measurement system above described was used at the Vasco da Gama Bridge (Cunha et al., 2001), as well as at the Millau Viaduct (Flamand and Grillaud, 2006) and the Humber Bridge (Brownjohn et al, 2010). A final aspect to mention refers to the required length of the collected time series for data processing. An empirical estimation of the adequate duration of the records collected at each measurement point has been considered in the past, using non-parametric identification methods, as 1000–2000 T, where T is the longest period present in the measurement signals (Cantieni, 2005). This implies that, in the case of long span bridges, with very low natural frequencies, very long time series need to be sampled. To exemplify, 16-minute records were collected at each measurement point for the tests at the Vasco da Gama Bridge and the Millau Viaduct, and the sampling started every 20 minutes, leaving 4 minutes to move the sensors into a new measurement position. For the Humber Bridge, almost 1-hour records were collected, and about 10-minutes were left to move the sensors, considering the longer distances. So, while for the Vasco da Gama Bridge, an ambient vibration test covering 29 pairs of measurement points took 2,5 days, for the Humber Bridge, the measurement at 70 pairs of points took 5 days. Fortunately, since measurements are conducted using natural excitation, it is not necessary, except for accurate damping tests, to interrupt the traffic on the bridge. At the same time, since hardly can stationarity of the recorded signals be ensured for very long time durations, the estimates of damping extracted from ambient vibration tests need to be regarded with significant care. 774
Table 1. Ambient vibration modal parameters of the Vasco da Gama Bridge obtained with peak picking (PP) and the MACEC implementation of SSI method. Type of mode
AVT PP f (Hz)
BT1 BV1 BV2 T1
0.298 0.341 0.437 0.471 0.572-0.624 0.651 0.693-0.755 0.817 0.895 0.985
BV3 T2 BV4 T3 BV5
AVT SSI (MACEC) f (Hz)
σf (%) (*)
ξ (%)
σξ (%) (*)
0.302 0.339 0.458 0.468
1.66 0.29 0.22 0.21
1.47 0.52 0.44 0.43
61 39 31 22
0.649 0.711 0.817 0.917 0.987
0.46 0.56 0.37 0.00 0.51
0.72 1.09 0.44
45 50 17
0.74
23
(∗) – standard deviations based on the setups used.
Figure 2.
Example of 2nd bending mode and 1st torsional modes identified at the Vasco da Gama Bridge.
3.2 Ambient vibration test results The analysis of the large databases collected on a bridge for extraction of natural frequencies and vibration modes can be made using traditional signal processing techniques and the basic frequency-domain method (peak-picking) (Felber 1993), or else the more recently developed system identification algorithms, which yield also estimates of damping, as the enhanced frequency domain decomposition (Brincker et al 2000), the covariance-driven stochastic subspace identification (SSICOV) (Peeters 2000; Peeters & De Roeck, 1999), the data-driven stochastic subspace identification (SSI-DATA) (Van Overschee and DeMoor 1996) or, more recently, the Polymax method (Peeters et al 2005). The former methods can provide excellent estimates, requiring however a certain amount of manual processing and personal judgement, while the later can be applied in a more systematic form and, in principle, can be applied to separate coupled modes. To illustrate the capabilities of the cited methodologies and tests, some results are shown. For the Vasco da Gama Bridge tests, conducted in 1997, the peak-picking method was applied, after separation of vibration components, namely the bending and torsion modes, with almost coincident frequencies of 0.44 Hz and 0.47 Hz (see Table 1 and Figure 2), by simple sum and subtraction of the signals collected simultaneously in the upstream and downstream points of the same section. The latter re-analysis of data using the most recent methods confirmed the modal parameters previously identified and yielded estimates of damping ratios that are systematized in Table 1. For the Millau Viaduct, the eight spans defined in an extension of 2460 m were covered in one-day test. Figure 3, representing average singular values of the spectra matrices for the vertical and transversal accelerations, and colour maps of the 1st singular values during the test, evidences the high density of modes in the range 0–1 Hz. The processing of data using the PolyMAX and the SSI-COV methods (Cunha et al 2007) enabled the identification of 20 vibration modes in this 775
Figure 3. Average singular values of the spectra matrices and colour maps with the variation of the 1st singular values during the ambient vibration test.
Figure 4. Example of (a) vertical and (b) transversal bending modes identified in the range 0- 1 Hz for the Millau viaduct: identified and calculated.
interval (10 vertical and 10 transversal bending modes), some of which are presented in Figure 4. The superimposition of these estimates to the corresponding configurations calculated at the design stage shows the excellent agreement of experimental data and evidences the quality of the numerical model. For the Humber Bridge, the various techniques above described were applied at different stages, which provided estimates of 30 vibration modes in the interval 0 to 1 Hz, some of which are presented in Figure 5, illustrating the extremely low significant natural frequencies and the closeness between parameters of subsequent modes. In fact, the lowest frequency identified of 0.056 Hz for the first transversal mode is well in the limit of the sensitivity of the force-balance accelerometers, 776
Figure 5. Example of (a) vertical; (b) transversal; and (c) torsion modes identified in the range 0–1 Hz for the Humber Bridge.
which are state-of-the-art devices. This fact could explain some irregularity of the identified modal configuration. Another interesting aspect in the observation of identified parameters in Figure 5 respects the relatively high identified damping ratios of the vertical and transversal bending modes, although not the torsion modes. The fact that tests were conducted in uncontrolled traffic and wind conditions may have contributed to the presence of aerodynamic damping and of some additional effect due to traffic, as the bridge is intensively used along the day. In this respect, it is noteworthy that whenever accurate damping ratios are required, ambient vibration tests are complemented with free vibration tests, based on the sudden application of an impulsive load which produces a free decay response with an initial amplitude significantly higher than ambient vibration. These tests are conducted with no traffic under controlled wind, say with a wind speed no greater than 5 m/s. As a reference, it is mentioned that the suspension of a 60 ton barge from a point close to 1/3rd the central span at the Vasco da Gama Bridge produced a maximum acceleration of about 30 mili-g, while the root mean square value of vertical acceleration at the reference section at Vasco da Gama bridge during the ambient vibration test was 1.7 mili-g. At the Millau Viaduct, the release of a cable tensioned to 1000 kN produced a maximum vertical acceleration of about 1 m/s2 , and a peak displacement of about 50 mm. 4 CONTINUOUS DYNAMIC MONITORING OF BRIDGES 4.1 General aspects Extending the use of modern data acquisition systems to the permanent observation of the dynamic response has been a natural consequence of the improvement of the hardware reliability, taking profit of the increased storage and improved communication capacities. In terms of the offer to the owner or operator of the structure, the on-line assessment of the dynamic behavior of a relevant bridge can provide information on the comfort, serviceability, the behavior under extreme conditions and even allow the detection of damage at early stages, as will be illustrated in the 777
Figure 6. Time evolution of the first natural frequency (lateral bending mode) (13/9/2007 to 12/9/2012).
following section. However one major aspect requires consideration: the permanent observation of the dynamic response leads to huge amounts of stored data, which have no value if not processed automatically and reduced in such a way information on the structural behavior can be extracted. Clearly this issue has been overcome in recent years, with the introduction of fully automated identification algorithms (Magalhães et al, 2009). Some examples will be described in the next section, showing applications developed both in terms of consultancy and research projects. 4.2 Case studies One of the simplest purposes in the continuous dynamic monitoring relates to the need to characterize vibration levels, either induced by traffic or explosions associated with construction works. This can be exemplified with the implementation of a monitoring system at the Pedro e Inês footbridge, in Portugal (Caetano et al., 2010, Hu et al., 2012), which was observed from design to be prone to lateral and vertical vibrations induced by pedestrians. A total of 33 tons of tuned mass dampers (TMDs) were installed on the bridge. However, the reduced efficiency of the lateral TMD with regard to expected motivated the preoccupation of the owner of the structure to ensure the comfort of the users and therefore the temporary installation of a dynamic monitoring system was contracted with the authors for a period of five years. The system comprised a total of 6 accelerometers and provided estimates of peak response every 30 minutes and was prepared to issue an alert in case of excess of vibrations, which was never activated due to the good behavior of the structure. A second and more complex application relates to the purpose of detecting damage on a structure at an early stage. In principle, modal parameters, in particular natural frequencies and vibration modes, should be constant in time. In case of structural degradation, a modification of natural frequencies and mode shapes might occur. The question that has been raised and discussed along the years, and considering that overall the natural frequency varies with the square of the structure stiffness, is that the level of degradation required to be detected by frequency variation would be enormous. It is necessary however to understand the amplitude of frequency variations that can actually be detected on the basis of experiments. To this purpose, a very relevant research project is mentioned, consisting in the installation of a dynamic monitoring system on the Infante D. Henrique roadway Bridge (Magalhães et al., 2012). This system was based on 12 force-balance accelerometers distributed along the structure, connected to local digitizers and transferring acceleration records sampled at 50 Hz to a computer at the University using an ADSL connection every 30 minute. Besides the most conventional and simple statistical data of the records, the continuous and automatic identification of natural frequencies, mode components and damping ratios for a period between 2007 and 2015) provided an enormous understanding of the capabilities of continuous monitoring data. The natural frequencies shown in Figures 6 and 7 for two of the twelve tracked modes in the range 0.5–5 Hz evidence a cyclic variation of modal parameters, which has been correlated with temperature (Figure 8) and traffic intensity. These environmental and operational seasonal and daily effects on the modal variability can be however removed (see Figure 7) using suitable input-output or output-only statistical techniques (e.g. multivariate linear regression, principal components analysis, etc), enabling the construction of appropriate control charts that can flag the occurrence of early damage based on modal frequency changes (Magalhães et al, 2012, Hu et al 2012). A third type of application of continuous dynamic monitoring, using a set of electrical strain gages, has been developed on an old riveted railway bridge (Trezói bridge), not only with the purpose of creating, updating and experimentally validating a sophisticated finite element modeling of the 778
Figure 7. Comparison between estimated and corrected natural frequencies of 4th vertical bending mode using a regression model just on the frequencies to remove operational and environmental effects.
Figure 8. Time evolution of bridge temperature measured at the deck (13/09/2007 to 12/09/2012) (top slab – solid black line, bottom slab – dashed blue line).
bridge, but mainly to accurately characterize real traffic loads and assess fatigue problems in critical elements affected by resonant behavior in higher order modes induced by the railway traffic (Marques et al, 2014). At last, it’s still worth noting that continuous dynamic monitoring has also been used with success in the context of aerodynamic studies involving in-situ measurements and not on physical models, as is the case of the Grande Ravine bridge (Bastos et al, 2014) or the Braga Stadium suspension roof (Martins et al, 2014), enabling the accurate characterization of the wind excitation and the structural response. 5 CONCLUSIONS The several case studies presented in this paper clearly evidence the high potential and usefulness of dynamic testing of large bridges, supporting the development of finite element model correlation, updating and validation, as well as the characterisation of the effects induced by different kinds of dynamic excitations. Moreover, such tests can be also used to identify the initial dynamic properties of the healthy structure in the context of the application of vibration based damage detection techniques. On the other hand, the implementation of long-term dynamic monitoring systems, duly combined with automated Operational Modal Analysis techniques and statistical removal of environmental and operational effects, made feasible the reliable detection of structural damages. Those systems can however have other types of applications, such as the vibration serviceability assessment and verification of efficiency of vibration control devices, checking if some vibration comfort limit is exceeded, the development of studies of Wind Engineering based on in-situ measurements or the experimental assessment of fatigue problems. REFERENCES Bastos F., Caetano E., Cunha A., Cespedez X., Flamand O. (2014). “Continuous aerodynamic monitoring of the Grande Ravine Viaduct”, 9th International Conference on Structural Dynamics”, EURODYN 2014, Porto, Portugal. Bietry, J. & Jan, P. 1995. Essais Dynamiques du Pont de Normandie. Report of “Mission du Pont de Normandie”. EN-D 95.5 C. Brincker R., Zhang L., and Andersen P. (2001). “Modal Identification from Ambient Responses using Frequency Domain Decomposition,” Proc. 18th Int. Modal Analysis Conference, Kissimmee, FL.
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Brownjohn J.M., Dumanoglu A.A., Severn R.T. and Blakeborough A. (1989). “Ambient Vibration Survey of the Bosporous Suspension Bridge” Earthquake Engineering and Structural Dynamics, Vol. 18, pp. 263–283. Brownjohn J.M.W., Magalhães F., Caetano E. and Cunha A. (2010). “Ambient vibration re-testing and operational modal analysis of the Humber Bridge”, Engineering Structures, Vol.32, Issue 8, pp. 2003–2018. Caetano E., Cunha A., Magalhães, F. and Moutinho C. (2010). “Studies for controlling human-induced vibration of the Pedro e Inês footbridge, Portugal. Part 1: Assessment of dynamic behaviour”, Engineering Structures, Engineering Structures, 32, pp. 1069–1081. Caetano E., Cunha A., Magalhães, F. and Moutinho C. (2010). “Studies for controlling human-induced vibration of the Pedro e Inês footbridge, Portugal. Part 2: Implementation of tuned mass dampers”, Engineering Structures, 32, pp. 1082–1091. Cantieni R. (2005). “Experimental Methods used in System Identification of Civil Engineering Structures”, International Operational Modal Analysis Conference, IOMAC, Copenhagen, Denmark. Carder D. (1936). “Observed vibrations of buildings”, Bull. Seismological Society of America, Vol. 26, No. 4, pp. 245–277. Cunha A., Caetano E. and Delgado R. (2001). “Dynamic Tests on a Large Cable-Stayed Bridge. An Efficient Approach”, Journal of Bridge Engineering, ASCE, Vol.6, No.1, pp. 54–62. Cunha A., Caetano E. and Magalhães F. (2007). “Output-only Dynamic Testing of Bridges and Special Structures”, Structural Concrete, Journal of FIB, 8, No.2, pp. 67–85. Cunha A., Caetano E., Magalhães F. and Moutinho C. (2013). “Recent perspectives in dynamic testing and monitoring of bridges”, Journal of Structural Control and Health Monitoring, Vol. 20, Issue 6, pp. 853–877. Felber A. (1993). Development of a Hybrid Bridge Evaluation System, Ph. D. Thesis, University of British Columbia, Canada. FIB (2003). FIB Bulletin 22: Monitoring and safety evaluation of existing concrete structures. Flamand O. and Grillaud G. (2006). “Dynamic testing of the Millau Viaduct”, In Proc. Third Int. Conf. on Bridge Maintenance, Safety and Management, Porto, Portugal. Hu, W.-H., Moutinho, C., Caetano, E., Magalhães, F. & Cunha, A. (2012) – “Continuous dynamic monitoring of a lively footbridge for serviceability assessment and damage detection”, Mechanical Systems and Signal Processing, Mechanical Systems and Signal Processing, Vol.33, pp. 38–55. Ko J.M. and NiY. Q. (2005). “Technology developments in structural health monitoring of large-scale bridges”, Engineering Structures, Vol. 27, Issue 12, pp. 1715–1725. Magalhães F., Cunha Á. and Caetano E. (2009). “ Online automatic identification of the modal parameters of a long span arch bridge”. Mechanical Systems and Signal Processing, 23(2), 316–329. Magalhães F., Cunha Á. and Caetano E. (2012). “Vibration based structural health monitoring of an arch bridge: from automated OMA to damage detection”. Mechanical Systems and Signal Processing, Special issue on interdisciplinary and integration aspects in structural health monitoring, Vol. 28, pp. 212–228. Marques F., Cunha Á., Caetano E., Moutinho C. and Magalhães F. (2014). “ Online automatic identification of the modal parameters of a long span arch bridge”. Analysis of dynamic and fatigue effects in a old metallic riveted bridge”, Journal of Constructional Steel Research, 99C, pp. 85–101. Martins N., Caetano E., Diord S., Magalhães F. and Cunha A. (2014). “Dynamic monitoring of a stadium suspension roof: wind and temperature influence on modal parameters and structural response”, Engineering Structures, Vol. 59C, pp. 80–94. McLamore V., Hart G. and Stubbs I. (1971). “Ambient vibration of two suspension bridges”, ASCE Journal of the Structural Division, Vol. 97, No. ST10, 2567–2582. Peeters B. (2000). “System Identification and Damage Detection in Civil Engineering,” Ph.D. Thesis, K. U. Leuven, Belgium. Peeters B. and De Roeck, G.. (1999). Reference-based stochastic subspace identification for output-only modal analysis. Mechanical Systems and Signal Processing, 13(6):855–878. Peeters B., Vanhollebeke F., and Van der Auweraer H. (2005). “Operational PolyMAX for Estimating the Dynamic Properties of a Stadium Structure During a Football Game,” Proc. 23rd Int. Modal Analysis Conference, Orlando, FL, USA. Trifunac M. (1972). “Comparisons between ambient and forced vibration experiments”, Earth. Eng. Struct. Dyn., Vol. 1, 133–150. Vincent G. (1958). “Golden Gate Bridge Vibration Study”, ASCE J. Struct. Div., Vol, 4, ST6. Van Overschee, P., and DeMoor, B. (1996). Subspace Identification for Linear Systems – Theory, Implementation, Applications, Kluwer Academic Publishers, The Netherlands. Wong, K.Y. (2004). “Instrumentation and health monitoring of cable-supported bridges”, Structural Control and Health Monitoring, 11, pp. 91–124.
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Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
Investigation and countermeasures for fatigue cracks that emerged on the finger joint of the cable-stayed bridge T. Kosugi Third Management Division, Structural Management Department, Shutoko Engineering Co. Ltd.
M. Takahashi Kanagawa Operation Bureau, Metropolitan Expressway Co. Ltd.
Y. Nakamura First Management Division, Structural Management Department, Shutoko Engineering Co. Ltd.
H. Dobashi Shutoko Engineering Co. Ltd.
ABSTRACT: Upon implementing follow-up investigations on the finger joint of the Tsurumi Tsubasa Bridge along the Bay Shore Route of the Metropolitan Expressway, as fatigue cracks have become apparent for the past 10 years, it was confirmed that there was an increasing tendency of damages. Analyzing the conditions of the damages, it turned out that weld detail at the time of fabrication was one of the causes, therefore countermeasures by improving details upon weld repair was implemented. In response to the occurrence of events on the Tsurumi Tsubasa Bridge, investigations of finger joint of the Yokohama Bay Bridge which also locates on the Bay Shore Route of the Metropolitan Expressway was carried out, however crack damages were undetected. Comparing structures of the finger joint of the Tsurumi Tsubasa Bridge and the Yokohama Bay Bridge, there were differences in the number of intermediate supporting beams and the cross sectional area of the faceplate, which resulted in difference of durability. 1 INTRODUCTION Fatigue crack damages of finger joints on the Metropolitan Expressway were detected for the first time in 2004 on the Tsurumi Tsubasa Bridge. The Tsurumi Tsubasa Bridge is a 3-span continuous steel cable-stayed bridge approximately 1km long in length, which was opened to traffic in 1994. The average volume of daily traffic is approximately 100,000 vehicles, and the percentage of over-sized vehicle traffic is more than 25%. Since then, cracks and fractures have been found intermittently, and exchange and/or replacement of faceplate blocks have been conducted as an emergency repair so far. A lateral-view of the Tsurumi Tsubasa Bridge is shown in Figure 1.
Figure 1.
Lateral-view of the Tsurumi Tsubasa Bridge.
781
Figure 2.
Structure of finger joint.
Figure 3. Weld structure of the faceplate block.
In the follow-up investigation of the finger joint of the Tsurumi Tsubasa Bridge carried out in 2012, which is intended for follow-up observation, it was affirmed that damages tended to increase. Therefore countermeasures were to be implemented by analyzing damage causes from tendencies for crack causing and weld detail at the time of fabrication. Also, crack investigation of the finger joint of the Yokohama Bay Bridge, which locates near the Tsurumi Tsubasa Bridge, was carried out from 2010 through 2013. In this report, contents of crack investigation and countermeasures taken for the damages of the two cable-stayed bridges will be described. 2 THE FINGER JOINT OF THE TSURUMI TSUBASA BRIDGE As for the finger joint of the Tsurumi Tsubasa Bridge, joint clearance is 3220 mm, the length of design movement is ±320 mm on a steady basis, and ±700 mm in times of an earthquake. The length of the faceplate is approximately 2000 mm, and as the cantilevered length is long, approximately 1700 mm, it is designed as a simple beam with an intermediate supporting beam. Structure of the finger joint is shown in Figure 2. The faceplate block of the finger joint is composed of faceplates cut out from thick steel plates and spacing plates that are welded mutually. Weld structure of the faceplate block is shown in Figure 3. As for a the blocks of the faceplate, welding is done around the spacing plate by fillet weld, but welding was not fully done in narrow spaces sandwiched between the faceplates, because it was physically difficult to weld at the time of fabrication. All crack damages found during the follow-up investigation in 2012 emerged from such parts. 782
Figure 4.
Cracks from the root of the welded part of the faceplate found in 2004.
Figure 5.
Fracture of the faceplate that occurred in 2010.
3 SEQUENCE OF EVENTS SO FAR AS FOR THE FINGER JOINT OF THE TSURUMI TSUBASA BRIDGE Fatigue cracks were found on the finger joint of Tsurumi Tsubasa Bridge for the first time during the inspection in 2004 (Figure 4). The cracks which were observed emerged from the root of the welded part of the top surface of the faceplate and had penetrated to the base metal. At this point, approximately 80 cracks were found in an emergency inspection, and as a countermeasure, removal of the cracks by machining and replacement of the faceplate blocks were implemented. Since then, follow-up investigations and support for damages have been carried out repeatedly, but a fracture of the faceplate occurred in 2010 (Figure 5). Fortunately, it did not affect the traffic, because it was only one faceplate that fractured. The cause of this fracture was a new type of crack that emerged from the corrosion part of the bottom surface of the faceplate. 4 CRACK INVESTIGATIONS ON THE TSURUMI TSUBASA BRIDGE 4.1 Contents of the investigation A follow-up investigation was implemented in 2012 for the finger joint of the Tsurumi Tsubasa Bridge. In this investigation, a magnetic particle examination was carried out for the corroded parts of the root of the top surface and the bottom surface of the faceplate where damages were found. Conditions of the follow-up investigation are shown in Figure 6 and Figure 7. As for the bottom surface, an investigation of the corroded parts was carried out in advance to the magnetic particle examination, however corrosion of the bottom surface had not progressed compared to two years before in 2010. This is assumed to be due to the countermeasure of detaching the cover plate for the bottom surface of the faceplate, where water does not gather. 783
Figure 6. (Left) Research status of investigation for the top surface of the faceplate. Figure 7. (Right) Research status of investigation for the bottom surface of the faceplate.
Figure 8.
Genesis location of cracks.
Table 1. Breakdown of discovered number of cracks. Discovered number of cracks (Points) Place
Direction
P1
Westbound Eastbound Westbound Eastbound
P4 Total
Cracks penetrated to the base metal
Cracks not penetrated to the base metal
Total
11 1 – – 12
45 72 3 8 128
56 73 3 8 140
4.2 Results of the investigation As a result of implementing a magnetic particle examination for 1100 faceplates, 140 cracks were detected at the root of the welded part of the top surface. Genesis location of the cracks is shown in Figure 8, and the breakdown of the number of cracks is shown in Table 1. Cracks on the bottom surface of the faceplate were undetected during this investigation. Most cracks which were found during this investigation emerged at the finger joint on the P1 pier. Of all 140 points, cracks penetrated to the base metal were found at 12 points. Points of the greatest damage is indicated in Figure 9. In the follow-up investigation implemented in 2012, 140 points and even more cracks were detected and it was confirmed that there was an increasing tendency in damages. In particular, although substantial replacement of blocks had been carried out from 2008 through 2010, more 784
Figure 9.
Cracks penetrated to base metal.
Figure 10. Weld detail of the root of the faceplate.
than 50 cracks were found on the lanes bound for west on the P1 pier. Therefore, it is considered that cracks detected at those points emerged 2–4 years after the replacement. 4.3 Considerations Out of the140 cracks that were detected during the follow-up investigation in 2012, 129 were cracks that emerged at the finger joint of the P1 pier. When the structure of the P1 pier and the P4 pier was compared, a difference in weld detail could be confirmed, which may be attributed to the difference of fabricating companies. The conditions are shown in Figure 10. Unwelded spots were left on faceplates on both piers, but for the finger joint of the P4 pier where there were less cracks, welding was done vertically within approximately 50 mm from the corner of the spacing plate. However, at the finger joint on the P1 pier where there were many cracks, welding was done only at the corner of the spacing-plate. Therefore, causing factors for cracks generating are considered as the following: – A breakpoint was installed in the weld structure with an overlapping joint that has low fatigue durability. – A start/end position of welding which is subject to damage was installed at the corner of the spacing plate, a point of stress concentration where stiffness changes. 5 COUNTERMEASURES IMPLEMENTED FOR DAMAGES ON THE TSURUMI TSUBASA BRIDGE Of the 140 cracks detected during the investigation in 2012, for the 12 points (7 blocks) where there was penetration to the base metal, blocks were replaced by robust ones installed under the road shoulders as an emergency measure, and thereafter were replaced by newly fabricated ones. As for points without penetration to the base metal, there were up to 37 blocks intended for repair, therefore it was impractical to newly fabricate all blocks. Blocks were temporarily removed and replaced one after another, and weld repair was carried out within the factory, not targeting restitution but bearing in mind increase of fatigue durability by detail improvement. 785
Figure 11.
Status of processing the weld toe (final status).
Figure 12.
Inspection robot.
5.1 Welding of narrowed sections Welding repair for crack damages was restored by re-weld, after weld bead of the cracks were completely removed by machining, and at the same time, insufficient welded places were restored by welding ensuring enough leg length of the bead. As for start/end position of the restored weld bead, a new bead was welded onto the existing one and then was removed by a grinder. 5.2 Process of weld toe finishing As for the processing of the weld toe, welding start/end position in narrow spaces between the faceplates were finished by a grinder, and the scope within 50 mm from the edge of the top surface of the spacing plate was processed furthermore by peening. The status of processing the weld toe is shown in Figure 11. 6 CRACK INVESTIGATIONS ON THE YOKOHAMA BAY BRIDGE 6.1 Contents of the investigation Following the series of events of the Tsurumi Tsubasa Bridge, investigation was implemented on the finger joint of the Yokohama Bay Bridge, which also locates on the Bay Shore Route of the Metropolitan Expressway. As for the top surface of the faceplate, a magnetic particle examination was carried out in 2010, and it was confirmed that there were no cracks. The bottom surface was unconfirmed because of the structure where an inspector cannot enter, therefore an inspection robot was developed as shown in Figure 12, and an inspection using the robot was implemented in 2013. 6.2 Results of the investigation From the pictures obtained from the inspection robot, it was confirmed that the bottom surface of the faceplate of the Yokohama Bay Bridge is under a stable condition, and corrosion that may be a point of crack origination was undetected. The status of robot inspection is shown in Figure 13, and pictures obtained during the robot inspection are shown in Figure 14. 786
Figure 13. Figure 14.
(Left) Status of robot inspection. (Right) Pictures obtained during robot inspection.
Figure 15.
Structure of the finger joint of the Tsurumi Tsubasa Bridge.
7 STRUCTURAL COMPARISON OF THE FINGER JOINT BETWEEN THE TSURUMI TSUBASA BRIDGE AND THE YOKOHAMA BAY BRIDGE Although many cracks emerged on the faceplate of the Tsurumi Tsubasa Bridge, there were no cracks confirmed on the Yokohama Bay Bridge. Although both bridges are cable-stayed bridges on the same route, and their finger joints have a similar weld structure, there was a significant difference between the investigation results. When the structures of both bridges were compared, differences shown in Figure 15, Figure 16 and Table 2 were confirmed. It can be considered that the finger joint of the Yokohama Bay Bridge is under a stable condition for the following reasons: – The cross sectional area of the finger joint of the Yokohama Bay Bridge is larger than that of the Tsurumi Tsubasa Bridge. – There is only one intermediate support on the Tsurumi Tsubasa Bridge, but there are two on the Yokohama Bay Bridge. Therefore, although the faceplate on the Yokohama Bay Bridge is longer than that of the Tsurumi Tsubasa Bridge, the maximum cantilevered length is less than one half. – One block of the faceplate of the Tsurumi Tsubasa Bridge is built up of 12, and 4 for the Yokohama Bay Bridge. Because the blocks for the Yokohama Bay Bridge are smaller than those of the Tsurumi Tsubasa Bridge, it may be easier to correspond to load and impact of running vehicles. 787
Figure 16.
Structure of the finger joint of the Yokohama Bay Bridge.
Table 2. Structural comparison of the finger joint between the Tsurumi Tsubasa Bridge and the Yokohama Bay Bridge.
Face PL material Face PL length Face PL height Face PL thickness Intermediate supporting beam Face PL numbers/block
Tsurumi Tsubasa Bridge
Yokohama Bay Bridge
SM400 2033 mm 210 mm 35 mm 1 point (max. span length: 1945 mm) (max. cantilevered length: 820 mm) 12 per block
SM400 3045 mm 233 mm 40 mm 2 points (max. span length: 1934 mm) (max. cantilevered length: 334 mm) 4 per block
8 SUMMARY The cracks detected on the finger joint of the Tsurumi Tsubasa Bridge emerged from the root of the welded part of the top surface of the faceplate. Many cracks were detected on the P1 pier, however there was a difference in the length of weld of the root of the welded part on the P1 pier and the P4 pier, and it was considered that the cause was the difference in weld detail at the time of fabrication. As for the crack damages emerged on the finger joint of the Tsurumi Tsubasa Bridge, countermeasures were determined by the level of damage. For points where crack damages had penetrated to the base metal, the whole faceplate block was replaced by a newly fabricated one, but for points where crack damages did not penetrate to the base metal, weld repair was carried out at the factory considering economic efficiency. Although the Yokohama Bay Bridge has a similar weld structure as the Tsurumi Tsubasa Bridge, crack damages were undetected. Upon comparing the two bridges, there were differences found in the cross sectional area of the faceplate, the number of intermediate supporting beams, and the size of the faceplate block, which were considered to be factors that caused difference in fatigue durability. 9 CONCLUSION Countermeasures for the finger joint of the Tsurumi Tsubasa Bridge using weld repair was completed in 2014. A follow-up investigation is scheduled in 2015, for follow-up observations. A magnetic particle examination is in practice since 2014, for the root of the welded part of the top surface of the faceplate of the Yokohama Bay Bridge, however cracks damages have not been detected up until now. 788
Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
Management of the Severn Bridge Suspension Bridge C.R. Hendy, C. Mundell & D. Bishop Atkins, Epsom, UK
ABSTRACT: The M48 Severn Bridge is a 988 m span suspension bridge carrying the M48 motorway across the River Severn in the UK. The bridge is substandard to current UK assessment codes but is permitted to continue in operation on the basis that application of the Highways Agency document, BD79, which allows a risk-based management approach to bridge operation, indicates the risk presented is acceptable given the mitigation and monitoring in place. A key aspect of the work involved in verifying that the bridge can remain in operation is undertaking continuous structural assessment, taking into account the results of previous cable intrusive investigations and data from the various continuous monitoring systems. This paper discusses the determination of the main cable strength allowing for its deterioration to date. It also discusses determination of the Bridge Specific Assessment Live Load model, used for the bridge assessment, from Weigh in Motion (WIM) data.
1 INTRODUCTION The Severn Bridge is a 988 m suspension bridge which, since opening in 1966, has carried over 300,000,000 vehicles. Due to increases in traffic load and deterioration, the bridge has received several strengthening measures since construction, including replacement of the inclined hangers, fatigue strengthening of the deck and bearing replacements. In 2006, a 40 year inspection of the cables was carried out to determine the condition of the cables. It was found that a number of the wires were broken or corroded, sparking concern over the bridge’s overall capacity. As a consequence, Heavy Goods Vehicles (HGVs) were restricted to the outside lanes, with weight restrictions in place elsewhere. Additional measures to protect and monitor the cables were installed, including a corrosion inhibitor within a dehumidification system, and acoustic monitoring to detect further wire breaks. In conjunction with the preventative measures in place, the M48 Severn Bridge is subject to continuous structural assessment, in particular for the main cables. The assessment involves developing the predictive future deterioration model for the cable; deterioration has now largely stabilised with respect to corrosion with the dehumidification system in place but wire breaks may continue due to overload events and fatigue. Weigh in Motion data for the traffic crossing the bridge is used to continuously adjust the Bridge Specific Assessment Live Load model used for the bridge assessment. This is used to justify the continued usage of the bridge where the traffic loading remains below a critical value and to design interventions where this critical traffic loading is exceeded, either for extreme events or due to a steady increase in traffic volume with time. This paper summarises the strength evaluation performed by Atkins and the methods used to monitor and calculate bridge specific live loads derived by the authors.
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2 BRIDGE SPECIFIC ASSESSMENT LIVE LOADING 2.1 Purpose of BSALL In the UK, the HighwaysAgency Design Manual for Roads and Bridges (DMRB) includes extensive guidance for the assessment loading criteria for bridges spanning up to 50 m. For the assessment of long-span bridges such as the M48 Severn Crossing, the use of such assessment loading becomes excessive, and not representative of real traffic loading conditions. Instead, a bridge-specific assessment live load (BSALL) can be determined, giving an assessment loading derived – in this case – from a probabilistic assessment of the actual traffic flows. With this approach comes the necessity to continuously monitor the traffic, as any changes in traffic trends will also affect the assessment loading. 2.2 Traffic monitoring There are two methods that are generally used to provide the necessary traffic data to calculate a BSALL. Traffic flows can be simulated using probabilistic traffic flows and vehicle weights (e.g. using Monte Carlo simulations), or they can be measured on site using live monitoring from Weigh-in-Motion (WiM) stations as is the case on Severn Bridge. Historically, the latter method has not been viable due to the computing power necessary to deal with such large volumes of traffic data; however modern processors have overcome such issues. The M48 Severn Crossing has four WiM sensors. These are located at the west side of the Wye Bridge under each lane. The lanes are numbered from 1 to 4, with 1 and 2 being Eastbound traffic, and 3 and 4 being the Westbound traffic. Lanes 1 and 4 are the offside lanes, with 2 and 3 the nearside lanes. The data acquisition equipment is recalibrated on a regular basis and captures the lane of traffic, date and time, vehicle type and length, number of axles, axle spacings and axle loadings. 2.3 WiM Trigger levels The WiM data must be routinely monitored to ensure that no changes in traffic trending are occurring. To achieve this, rather than continuously re-deriving the BSALL, three trigger criteria are considered for traffic flow in each direction. These are the number of HGVs as a percentage of the total traffic, the standard deviation of the HGV weights and the mean of the HGV weights. The trigger levels are considered for four scenarios during any given day: i. Early morning, when traffic volumes are low, but HGV percentages are respectively high; ii. Morning rush hour, when traffic volumes peak, but relative percentages of HGV traffic are lower than scenario i; iii. Evening rush hour, when traffic volumes again peak, but relative percentages of HGV traffic are lower than scenario i; iv. A five hour period between scenarios ii and iii. It has been determined that scenario iv) is the critical case, when the traffic volumes and percentages of HGV traffic combine to give the most onerous condition. It has also been found that these triggers are very sensitive to the sampling period, and large variations can occur on a daily basis. This highlights the necessity for caution when extrapolating trends and BSALL values from small samples of data, which may be subject to similar fluctuations. The WiM data is monitored on a monthly basis, with averages calculated for the critical scenario iv) time period. The exceedance of any one trigger criterion (% HGV, mean or standard deviation) is not indicative of an increase in the BSALL. Even the exceedence of two or more triggers in any given monitoring period does not give certainty of an increased BSALL; this acts only as a prompt that the BSALL should be checked and compared against previously calculated levels. Generally when trigger levels have been exceeded to date there has been either no or minimal increase in derived BSALL. 790
2.4 Derivation of BSALL A BSALL is a probabilistic traffic load based on a given set of traffic data. Live loading due to traffic flows will generally conform to a Gaussian distribution. However, to determine BSALLs for the purpose of assessment, these ‘normal’ traffic conditions must be extrapolated to provide safe but realistic load effects. In the UK, the characteristic highway traffic load model is defined as the load effect that has a 5% probability of being exceeded within a 120 year period. The BSALL methodology discussed in this section has been carried out independently for each lane of traffic. Initially, the WiM data is examined for a given time period. In this instance, five consecutive weekdays are considered in a month. If the BSALL is to be used for assessing annual trends, this will be the same month as for any previous assessments to reduce the impact of seasonal variations. The traffic from each lane is first “squashed” into queues formed of the vehicles crossing the bridge in each hour (the simulations presented in this paper have assumed that the traffic remains in the lanes as recorded by the WiM sensors). Therefore for each 24 hour period of data assessed, 24 No. hourly traffic queues are formed for each lane, giving 96 separate queues in total. These queues are formed with the clear gap between each vehicle of 5 m, measured as the distance between the rearmost axle of a vehicle and the front axle of the following vehicle. Each lane is considered independently of the others, with the hourly traffic queues used to determine the worst load effects for that train of vehicles. The worst case midspan bending moment, uniformly distributed load (UDL) and near-support shear forces are calculated for each load train. The assessed bridge length is taken to be 1298 m, equating to the M48 Severn Bridge main span and one side span (988 m + 310 m). The maximum load effects for each hour are tabulated and formed into a cumulative frequency distribution. These probabilities are then normalised to provide the Gumbel probabilities using the following formula: pGumbel = −Ln{−Ln(cumulative frequency)} The data, once normalised, can be presented as shown in Figure 1. This graph shows the normalised Gumbel probabilies from five days of WiM traffic queues. The Gumbel probabilities are given on the X-axis, with the associated load effects given on the Y-axis. The load effects for these data can also be shown instead as a ratio of either BD 21/01 or BD 37/01 values as required. Figure 1 also shows two trend lines; the whole-data trend line is generated by considering the entire dataset, whilst the red dashed trend line considers only the tail of the data where more linear behaviour is observed. If the entire dataset is considered, the extrapolation to the extreme 120 year event could be greatly over-estimated, as would be the case in Figure 1. By tailoring the trend
Figure 1. Typical Lane 1 Gumbel Distribution.
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line of each graph to suit the linear section of the data, the 120 year probability can be more accurately determined. In instances where the dataset is completely linear this tailoring may not be required. In all cases, increasing the volume of data considered will improve the accuracy of the BSALLs derived, as any linear trending will be made more apparent. The methodology presented in this paper uses five consecutive days of traffic, with 24 traffic queues per day, giving 120 data points for each lane. It has been found that in most cases this provides sufficient clarity to clearly determine a linear trend line. Once the maximum load effects for each traffic queue have been obtained, the nominalised value of the BSALL probabilities as derived from the hourly maximums as exemplified in Figure 1 is obtained as follows: The 5% probability of exceedence in 120 years equates (for small probabilities) to an hourly probability of exceedence of:
The probability of a queue forming, f , may be derived as a function of the hourly traffic, as discussed in the TRL report CR016 on long span bridge loading (1986), where:
Therefore, the probability of a queue forming will vary depending on the time of day and the traffic flow. The average peak hourly traffic for the WiM data obtained to date is 1984 vehicles per hour, giving a frequency f of 3.3. However, a frequency of 5% has been conservatively assumed, thus:
(for any given hour). This allows the 120 year Gumbel probability to be determined from the normalised distribution as: pGumbel = −Ln{−Ln(1 − p120 )} In the case where p120 is 9.513 × 10−7 , the pGumbel is 13.86. For the example dataset shown in Figure 1, for a Gumbel frequency of 13.86 and using a manually determined trend-line, the characteristic BSALL = 8.43 kNm−1 . As per BD 50/92, the nominal value is characteristic value/1.2 and therefore becomes 7.03 kNm−1 for Lane 1 in this instance. Should the queue frequency of 3.3 be used, this value would reduce to 6.90 kNm−1 ; a reduction of only 2%. If the derived load was used with Eurocodes, the characteristic value would be used directly to form combinations. 2.5 Calculated BSALL values The methodology presented in section 2.4 was carried out for eight months between June 2012 and January 2013 to monitor seasonal variations. However, the BSALL would only normally be calculated annually, with the WiM triggers used to identify any abnormal occurrences. Table 1 below presents the nominal BSALL UDLs for each lane, with the associated BD21/01 and BD37/01 co-existent values shown for comparison. The BSALL loading for each lane was
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Table 1. Nominal BSALL UDLs. Nominal Uniform Distributed Loads
June 2012 July 2012 Aug 2012 Sept 2012 Oct 2012 Nov 2012 Dec 2012 Jan 2013 Average BD 21/0 1 BD 37/01
Lane 1 (kNm−1 )
Lane 2 (kNm−1 )
Lane 3 (kNm−1 )
Lane 4 (kNm−1 )
Total (kNm−1 )
7.12 7.88 8.89 8.51 11.56 10.56 9.12 7.68 8.92 17.58 17.58
1.85 2.37 3.09 2.67 2.74 2.24 3.36 2.75 2.63 8.79 10.55
3.17 2.11 2.08 2.12 2.28 2.28 3.12 2.93 2.51 7.03 10.55
10.81 11.18 10.17 7.88 12.33 10.75 8.03 8.98 10.06 17.58 11.78
22.9 23.5 24.2 21.2 28.9 25.8 23.6 22.3 24.1 50.95 50.46
conservatively taken to be co-existent; a short study of the traffic showed this to be reasonable for Severn. 3 CABLE STRENGTH EVALUTION 3.1 Inspection Inspection of the main cables of the Severn Bridge was carried out in 2006 and subsequently in 2010, following which a dehumidification system was commissioned in 2008 (Cocksedge et al, 2010). The main cables have been subject to a yearly evaluation and this section of the paper describes the most recent evaluation performed. The inspections were carried out in accordance with the NCHRP Report 534, Guidelines for Inspection and Strength Evaluation of Suspension Bridge Parallel-Wire Cables, Mayrbaurl & Camo (2004). Initially tensile tests were carried out to determine yield strength (0.25% offset method), tensile strength, elongation in 254 mm gauge length, reduction of area and modulus of elasticity. The detailed visual inspection records the number of broken wires and categories each wire into corrosion stages. This inspection compliments the testing regime and enables wire properties derived through testing to be extrapolated to the full cable cross section. 3.2 Strength evaluation Initially the cable strength was evaluated using the brittle wire model described by Mayrbaurl & Camo (2004) to determine nominal cable strength capacity. However, for assessment purposes, a subsequent re-evaluation was carried out to determine the cable strength at a 5% probability of exceedence. This enabled the authors to apply nationally determined partial factors of safety to the derived capacity and calculated cable force, thus allowing an Ultimate Limit State assessment to be performed rather than a factor of safety calculation as described below. 3.3 NCHRP Report 534 strength evaluation The strength is calculated at a specific inspected location between two cable hangers which is called the evaluated panel. The strength of the cable in a specific panel is dependent not only on the deteriorated condition of the wires in that panel but also on the deteriorated condition of the wires in adjacent panels. A broken wire re-develops force away from the break due to friction from
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wrapping wire and clamping force at the cable bands. The strength of a cable at the evaluated panel is the sum of the strengths of wires in three categories: i. wires in the evaluated panel minus already broken wires in that panel and nearby panels; ii. wires that are already broken in nearby panels. These wires partially redevelop a proportion of their strength for the evaluation panel. iii. wires in the nearby panels that are assumed to become broken i.e. that are predicted to break by the brittle-wire model. These wires partially redevelop a proportion of their strength for the evaluation panel; 3.4 Evaluating inspection data Each wedge allows visual inspection of a number of wires within the main cable, but only a small percentage of internal wires can be visually inspected within each wedge. Therefore it is assumed that a wire observed at the surface represents all wires at the same radius in that half-sector between adjacent wedges. For illustration, Figure 2 below shows the location of the observed broken wires along the wedge lines within one typical panel for Severn Bridge. 3.5 Grouping of wires Each wire is assigned a stage of corrosion as described by Mayrbaurl & Camo (2004) as follows: Stage 1: spots of zinc oxidation on the wires; Stage 2: zinc oxidation on the entire wire surface; Stage 3: spots of brown rust covering up to 30% of the surface of a 75 mm to 150 mm wire length; and Stage 4: brown rust covering more than 30% of the surface of a 75 mm to 150 mm wire length. The stage of corrosion assigned to each observed wire is based on the maximum corrosion observed throughout a panel length. All wires in a specific stage of corrosion are assumed to have the same mechanical properties and distribution of these properties wherever they occur. Following tensile testing, samples wires are grouped as follows: i. ii. iii. iv.
Corrosion stage 1 and 2 wires are combined into a single group, called Group 2; Corrosion stage 3 wires that are not cracked are Group 3; Corrosion stage 4 wires that are not cracked are Group 4; Cracked wires are Group 5, irrespective of corrosion stage; they are identified by a brittle fracture type failure in tensile testing; v. Broken wires in the effective development length are treated as a separate group, irrespective of corrosion stage.
The tensile characteristics of each group are determined. The number of discrete cracked wires in the effective development length is subtracted from the appropriate group and assigned to a separate group for cracked wires. It was determined that for Corrosion Stage 3 sample wires 14.3% and 25.9% of Corrosion Stage 4 contain cracks. This is the assumed percentage of cracked wires within one evaluated panel length. The survivor function is used to calculate the percentage of discrete cracked wires (percentage of cracked wires in the entire development length) for each wire group. For the Severn Bridge this results in 72% of Stage 3 wires and 89% of Stage 4 wires being treated as discrete cracked wires in Group 5. The number of Stage 3 and Stage 4 wires visually observed during the inspection therefore had a significant impact on the overall strength of the main cables. 3.6 Broken wires and re-development length As discussed in section 3.3, wires broken in adjacent panels are assumed to partially redevelop their strength. The amount of redevelopment depends on the redevelopment coefficient. 794
Figure 2.
Location of observed broken wires.
Figure 3. (1968).
Cable Band and Hanger Detail Roberts
The force in a wire that is redeveloped by clamping friction at a cable band was estimated from site measurements of the gap between the ends of broken or cut wires. This essentially measures the amount a wire slips under service load. Note that although the inspected panel is unwrapped the adjacent panels are still wrapped and therefore the measured gap may include the effects of the adjacent panel wrapping in resisting slip. It is assumed in the evaluation panel however that the redevelopment force is applied only at cable bands and no intermediate friction occurs due to the wire wrapping. This is prudent as the potential friction from wrapping wire in the evaluation panel is lost during intrusive inspection. The cable bands orientation and connection detail for the M48 Severn Bridge is unique due to the inclined nature of the hangers (Fig. 3). Live load applied to hangers will cause a reduction in the clamping force. This reduction in clamping force was taken into consideration during the calculation of the re-development coefficient and re-development length by assessing the maximum reduction in clamping force applied to the cables. 3.7 Brittle-wire model The Brittle-Wire Model described by Mayrbaurl & Camo (2004) was used during the assessment to evaluate the strength of the cables. It assumes that all of the wires in each group follow the same stress-strain diagram, so that the stress in all of the intact wires is the same at any specific value of strain. An individual wire will share in carrying the tension in the cable until the stress in that wire exceeds its minimum tensile strength, whereupon all its strength is lost at the break location. Determining the cable strength requires increasing the cable stress in steps and calculating the number of wires that fail at each increment as they reach their tensile strength. The number of newly failed wires is subtracted from the number of previously intact wires to determine the number of unbroken wires. The cable force is calculated as the area of unbroken wires multiplied by the wire stress at that increment. When wires break, the stress in adjacent wires must increase to compensate. At some level of stress any further increase in stress will cause too many wires to fail and the overall strength will not increase. The stress at this optimal level determines the maximum force attained in the cable. The tensile stress which causes the highest cable capacity varies for each evaluation panel depending on the number of wires in each group for the given panel. Discrete cracked wires which fail in adjacent panels using the Brittle-Wire model are partially redeveloped by intermittent cable bands in the same way as wires which were already found to be broken. 795
Figure 4. Tensile Strength (Weibull Compound Cumulative Distribution Curve).
3.8 Evaluation of tensile test data The results of the tensile testing for each wire group on Severn Bridge was used to determine the compound cumulative distribution curves shown in Figure 4 below. The tensile data was assumed (and subsequently shown) to follow the Weibull distribution for all wire groups. A sensitivity check undertaken showed that the cable strength increased by only approximately 1% if a normal distribution was adopted so the additional effort associated with using a Weibull distribution was questionable. The minimum probable strength of a single wire which comprises groups 2, 3 and 4 were determined assuming a normal distributed one-sided confidence level of 97.5%. The minimum probable strength of group 5 cracked wires however was not determined at a minimum confidence level. It is assumed in accordance with Mayrbaurl & Camo (2004) that the minimum tensile strength of cracked wires between hangers for a given sample was recorded. This is because insufficient cracked specimen data points are available for each sample wire. This assumption could be optimistic as a relatively small length of each wire is subjected to tensile testing. For each sample wire 10 tensile test specimens are carried out for specimen lengths of 254 mm. This equates to 2.54 m out a possible 18.1 m wire length being tested.
3.9 Embrittlement of corroded wires and effect of reduced ductility Reduced ductility of corroded bridge cable wires have been recorded by Nakamura & Suzumura (2011). An evaluation was therefore carried out to assess the effects of any reduced ductility and it was noted that for Group 5 cracked wires the ultimate strain was much less than the average ultimate strain for all Groups. To assess the distribution of ultimate strain for each wire group the authors calculated the cumulative distribution curve at the 97.5% confidence level. The method adopted was similar to the method used to calculate the minimum tensile strength of each wire between cable bands assuming the normal distribution. Figure 5 below shows the results of this assessment and the distribution of ultimate strains for the Severn Bridge; the ductility of wires which contain cracks is significantly less than for those without cracks. Although the brittle-wire model takes different strengths into consideration for different wire groups it does not take into account varying stress-strain behaviour of each wire. This behaviour may be more different than the assumption of curtailing the plateau at the peak stress since each curve has a different shape and yield point. There is a risk that if the convergence stress calculated using the brittle-wire model was high some cracked wires would not reach the strain needed to mobilise the required level of stress. For the M48 Severn Bridge, however, convergence stress for the critical panel is reached at 1380 N/mm2 where the behavior is still elastic so the effect of reduced ductility was low and verified the use of the brittle-wire model in this case. 796
Figure 5.
Ultimate Strain (Weibull Compound Cumulative Distribution Curve).
4 ULTIMATE LIMIT STATE ASSESSMENT Further to the above evaluation, a method was adopted to calculate the cable strength at the 95% confidence level to allow an ultimate limit state assessment to be undertaken. The cable strength method adopted included allowance for the following likely errors in: i. ii. iii. iv. v.
calculating the minimum probable tensile strength of each wire calculating the cable strength due to the error in estimation of the no. of broken wires calculating the cable strength due to the approximation of the fraction of cracked wires calculating the cable strength due to errors in assignment of corrosion stage. extrapolation of the cable strength over the entire length of the cable.
To calculate the error caused by extrapolation of the cable strength over the entire cable length Mayrbaurl & Camo (2004) suggests that an exponential distribution is used. This type of distribution however does not fit in with the data returned for cable strength at each panel location. A relatively uniform reduction in cable strength has been calculated as the cable approaches the centre of the bridge. This relatively uniform change in cable strength has allowed the authors to allocate a number of cable zones for which the cable strength has been grouped. The zones were decided upon following review of the characteristics of derived strength, location and internal environmental conditions. It has been found during this evaluation that the normal distribution fits the strength of the cables within each zone. The error calculated between the mean nominal strength and the 95% confidence level of the datasets was determined. The error in strength calculated was then applied to the mean strength of each zone calculated from the 95% confidence level of each panel. This derives the ULS characteristic strength for all panels which were not investigated. Although the 95% confidence level is deemed sufficient to carry out an ULS assessment it is noted however that there is limited information available on tensile strength distributions of corroded wires. Therefore, a larger confidence level of 97.5% was adopted to calculate the strengths.
5 CONCLUSIONS This paper has briefly described the methods and techniques adopted to determine the bridge specific assessment live loading (BSALL) and deteriorated strength of the main cables to enable an Ultimate Limit State assessment to be carried out for the Severn Bridge. This has shown that the bridge has adequate reliability provided that traffic loading does not increase and the main cable 797
does not deteriorate further. Current monitoring suggests that both live load and cable condition have stabilized. REFERENCES Cocksedge C, Hudson T, Urbans B, Baron S., 2010. M48 Severn Bridge – Main Cable inspection and rehabilitation. Proceedings of the ICE – Bridge Engineering, Volume 163, Issue 4, 01 December 2010. Mayrbaurl & Camo, 2004. National Cooperative Highway Research Program (NCHRP), Report 534, Guidelines for Inspection and Strength Evaluation of Suspension Bridge Parallel Wire Cables, Transportation Research Board of the National Academies Nakamura & Suzumura, 2011. Corrosion of Bridge Cables and the Protection Methods, Journal of IABSE 0042, 20–23, 2011. TRL Contractor Report 16, 1986. Interim Design Standard: Long Span Bridges
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Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
Surveillance of continuous precast concrete bridge decks supported by monitoring-based techniques H. Sousa University of Surrey, Guildford, Surrey, UK
C. Sousa, A. Serra Neves & J. Figueiras CONSTRUCT, Faculdade de Engenharia, Universidade do Porto, Porto, Portugal
J. Bento Efacec Capital SA, S. Mamede de Infesta, Portugal
ABSTRACT: This paper presents the approach proposed for the surveillance of the structural behaviour of the South Approach Viaduct of the Lezíria Bridge, in Portugal, supported by monitoring-based techniques. The work focuses on one of the structure sub-viaducts, which is composed by 13 continuous spans with a span length of 36 m, and is supported by monitoring data recorded since the start of the construction and during the following three years. The main features of the surveillance system are a calibrated non-linear finite element model and a statistical model that establishes the normal correlation between environmental characteristics and material parameters with concrete strains and movements of expansion joints.
1 INTRODUCTION Prefabrication has been one of the most employed techniques for construction of multi-span large viaducts, for spans up to ∼35 m. For this span range, the decks are usually built with girders whose length is approximately equal to the span (Lounis et al. 1997, Miller et al. 2004). The girders are transversally connected by thin cast-in-place slabs whereas, in the longitudinal direction, continuous connections between spans are also cast-in-place (Valdés 1997). In these structures, the stresses due to permanent loads undergo important variations along time, owing to the construction procedure and the concrete creep and shrinkage deformations (Ma et al. 1998, Mirmiran et al. 2001, Ghali et al. 2002, Sousa et al. 2013a). Experimental results collected from real structures of this type are scarce (Cruz & Wisniewski 2004, Barr et al. 2008, Sousa et al. 2012). In this context, this paper presents the approach proposed for the surveillance of the structural behaviour of the South Approach Viaduct of the Lezíria Bridge, in Portugal, supported by monitoring-based techniques. This viaduct has a total length of 9160 m and was built with precast pretensioned “U”-shaped concrete girders with a typical span of 36 m. It is divided into 22 sub-viaducts, separated by expansion joints. The present work focuses on the sub-viaduct V14S, which is composed by 13 continuous spans, and is supported by monitoring data recorded since the start of the construction, i.e. July 2006, and during the following three years, i.e. until October 2009. In the paper: (i) the monitoring campaign is described and justified; (ii) illustrative experimental results are shown and discussed; (iii) the experimental results are used for calibration of a numerical procedure for long-term surveillance of this type of structures, i.e. finite element non-linear, phased, time-dependent analysis; (iv) finally, a statistical algorithm, developed by the authors (Sousa et al. 2013b), for long-term assessment of the structural behaviour is shown and discussed, i.e. a prediction model that establishes the normal correlation between environmental characteristics and material parameters with concrete strains and movements of expansion joints. 799
Figure 1.
Precast viaduct. (a) General view during construction, (b) precast beam detail.
Figure 2.
Mid-span cross section: geometry and transducer locations.
Finally, it is worth of mentioning that this work resulted from a collaborative work under the scope of the PhD of the first two authors. 2 STRUCTURE AND MONITORING CAMPAIGN The structure under analysis is composed by 13 continuous spans, with a span length of 36 m, and was constructed with U-shaped, pretensioned, precast concrete girders, as represented in Figure 1. The girder axes are separated 7.5 m apart and the girder depth is equal to 1.75 m. The girders are transversally connected by a 0.25 m thick cast-in-place slab (see Figure 2). In the continuity supports, the girders are monolithically connected to circular piles, visible in Figures 1 and 3, with a diameter of 1.5 m. This indirect foundation crosses alluviums with variable constitution and reaches a maximum depth of 47 m. In the region above the supports, the slab is posttensioned by straight tendons, which are extended approximately 6 m to each side of the pier axis. The monitoring system is composed by strain (STR), temperature (TEMP), rotation (ROT), joint displacement (JD) and relative humidity (RH) transducers. Four different cross sections were instrumented with embedded sensors: two mid-span cross sections and two cross sections above the pier axes. Figure 2 shows one of those mid-span cross-sections, with eight vibrating wire strain gauges (STR) and four temperature detectors (TEMP). Figure 3 displays the transducers installed in one of the support cross sections, which include also eight embedded strain transducers and two thermistors to measure the temperature in the girder concrete. The ambient conditions were characterized by measuring the temperature and the relative humidity both in the interior 800
Figure 3.
Cross-section above the pier axis: geometry and transducer locations.
and in the exterior of the box girder. Besides that, the concrete shrinkage deformations were measured by using four concrete prisms with dimensions 15 × 15 × 55 cm3 . Two of these prisms measure the concrete deformations in the interior of the box-girder (one for the girder concrete and other for the slab concrete). The remaining two prisms were kept above the transverse beam (at the top of the piers) in order to measure the shrinkage deformations in the exterior environment sheltered from rain (as shown in Figure 3). Inclinometers and linear variable differential transformers (LVDTs) were employed to measure girder rotations above the pier axes and joint displacements. The current quality control procedures followed during construction also provided important data for calibration of numerical models: the concrete compressive strength at different ages, for every concrete batch; the elongation of prestressing tendons; the relevant construction events and the dates in which they occurred.
3 CALIBRATED FINITE-ELEMENT ANALYSIS A calibrated finite element model provides an essential support for interpretation of the measured structural response and assistance in surveillance operations. In the present work, a time-dependent, phased, finite-element (FE) analysis was carried out to follow the viaduct response during the construction phase and in service. Beam finite-elements were used to model the global structure behavior, and the ordinary and prestressing reinforcements were simulated by embedded reinforcements, because a smeared crack approach was used to consider concrete cracking. The FE package DIANA was used, with user-supplied subroutines to simulate specific constitutive behaviors (TNO DIANA 2011). The construction procedure was simulated in the phased analysis through 23 distinct construction phases, with changes of supports, structural members or parts of the cross section. This small number is a result of grouping together events taking place in close dates, namely those related to the construction of piles and piers. On the other hand, the sequence of construction of the slab and continuity diaphragms was finely modeled based on the recorded construction schedule, because this sequence is determinant for the long-term strain variation. 801
A smeared fixed-crack approach with strain decomposition was used, allowing the simultaneous consideration of creep, shrinkage and thermal effects. The concrete properties were defined based on: (i) the results of compression tests of 150 mm cubes carried out during the construction, at different ages; (ii) the measurements taken in the four shrinkage prisms; (iii) the results of the retro-analysis of a precast girder, considering the deformations measured since the concrete pouring until the continuity connection was made (this retro-analysis provided additional information about the creep and shrinkage properties of the girder concrete). Additional details about the analysis procedures and results, as well as the comparison between calibrated and simplified numerical analyses, can be found elsewhere (Sousa et al. 2012). 4 STATISTICAL MODEL FOR LONG-TERM ASSESSMENT OF THE STRUCTURAL BEHAVIOUR An approach for real time assessment was also developed, consisting in obtaining the normal correlation pattern between the structural response (e.g. rotations or movements of expansion joints) and data related to environmental and material parameters (such as TEMP and concrete shrinkage measurements). Indeed, assuming a healthy behaviour of the structure during the first years after construction, a normal correlation pattern between different measurements can be established. Moreover, these correlations can be used to assess the adequacy of the structure behaviour in the future – prediction models – Equation 1. The model prediction, ymodel (t), for a given parameter (e.g. rotations or movements of expansion joints) at a given time t is calculated as a linear combination of measurements, xmeasure,i (t), weighted by wi . The following conditions must be guaranteed regarding the independent variables, x: (i) the structure response must be sensitive to them; (ii) they must be independent of the structure behaviour, i.e. they represent material properties or environmental parameters; (iii) minimal correlation should exist between the predictor parameters, i.e. avoiding multicollinearity that would lead to imprecise determination of the coefficients wi (Montgomery and Runger 2003). Under the scope of this work, the predictor parameters are TEMP and concrete shrinkage measurements, taking into account their strong influence in the long-term response of the structure.
According to Equation 2, the problem is solved by calculating the problem unknowns, i.e. the weights wi which minimize the differences, ε, between the measurement ymeasure (t) and the model prediction, ymodel (t). The optimal solution is the one that minimizes the function R, i.e. by setting the gradient to zero. It should be stressed that due to the fact that ymodel (t) is a linear combination of a set of measurements, R is a quadratic function and therefore this problem has a unique solution. Another important aspect is the observation period (i.e. the set of times ti ), over which the minimization problem is formulated, which must also be established. Commonly, it is used a fraction of the total available period of observation – ‘training window’. This training window can be fixed by making a set of calculations with increasingly larger dimensions. In this context, the best time window is the one that leads to the same set of weights wi that would be obtained with larger training windows. It is also worthy of note that the size of the training window depends on the phenomena under analysis. Under the scope of long-term structural analyses, where the TEMP and delayed deformations have a seasonal variation with a yearly period, a minimum size
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of one year is recommended. Finally, once the set of weights is determined, the model can be used to calculate the expected structure response ymodel (t) regarding the remaining time window. For the case of a healthy behaviour, the magnitude of the error, ε, shows the robustness of the prediction model. Additional details about the prediction model and results can be found elsewhere (Sousa et al. 2013b).
5 RESULTS AND DISCUSSION Figure 4 shows illustrative results regarding the comparison between measured strains and results of the FE analysis, for one of the mid-span cross sections. Numerous results regarding the comparison between long-term FE analyses of phased constructions and laboratorial tests are reported in the bibliography. However, the comparison with experimental results in real structures made with this construction method is scarce. Therefore, the data recorded in this structure provided valuable information for validation of such long-term analyses. The figure shows experimental results for three different depths: at the bottom of the precast girder; at the top of the girder (average of two sensors, STR-6 and STR-7, visible in Figure 2); in the slab. The frequency of acquisition was equal to 1 measurement per hour. As mentioned in section 3, a retro-analysis was made, based on the deformations measured before the erection of the precast planks, for derivation of the concrete creep and shrinkage deformations in this period of time. The corresponding time interval is identified in Figure 4. The figure shows a notable agreement between measurements and
Figure 4. Time variation of the concrete strain in one of the monitored mid-span cross sections (alignment B, between the transition peer and pier P1). Comparison between experimental results (dots) and FE analysis (line).
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Figure 5. Predictions vs. measured: a) strain at the bottom slab of the cross section TPP1 (V1S), b) Anomaly detection in bearing displacement JDTP15 (V14S).
FE analysis, for the three different measurement locations. The strain variation along the various construction phases is closely reproduced. It is also important to note that the seasonal variation of concrete deformations, during the service phase, with a period of one year, is clearly visible both in the experimental and in the numerical results. This effect was included in the calculations by adding this seasonal variation to the shrinkage model, according to the data recorded in the shrinkage prisms. It could be concluded that the adopted analysis strategy is able to reproduce the observed strain variations, and therefore, the model can be used in the study of the structure behavior. Besides that, the fact that all the relationship between the various measured deformations is in accordance with the theoretical values predicted by the FE model validates the results obtained with each of the strain sensors. Figures 5-a presents the predictions against the measurements for the strain at the bottom slab of the cross section TPP1 (V1S). The training window size was taken as one year, which corresponds to 60% of the complete time range. In addition, the time variation of the error ε, i.e. the difference between measurements and model predictions, is also depicted in the graph. As it can be observed, the prediction model is able to reproduce the real measurement. The relative error concept, defined as the ratio between the calculated error at a given time t and the range of variation of measurements over the complete time window, enables to compare the quality of the predictions. Based on this, the smallest relative errors occur in the case of the joint displacement, with a maximum relative error of 4.6% whereas 95% of the calculated errors are lower than 3.2% for this case. Moreover, the relative error average is 0.5%, which reveals that the model prediction follows the actual measurements. On the other hand, the correlation coefficient between predictions and measurements is 0.99, which confirms a good correlation. Figure 5-b shows a real example, i.e. for one of the monitored bearing displacements, where the usefulness of the prediction model is demonstrated. Interestingly, an abrupt variation at a specific time (9 May 2008 6 am) is observed in the time variation of the calculated error, which indicates some anomaly. In order to explain this fact, it was found that this instant of time corresponded to maintenance operations related to a temporary removal of the LVDT. Upon removal and replacement of the transducer, the reference was lost, which justifies the detected deviation in the calculated error. After investigations, it was concluded that this maintenance operation was carried out without communication with the authors. Indeed, this event was detected only due to the application of the prediction model to the results of this transducer. Therefore, the usefulness of the prediction model is demonstrated, not only restricted to the detection of structure deficiencies, but also to detect anomalies in the monitoring results. 804
6 CONCLUSIONS This paper presents the approach proposed for the surveillance of the structural behaviour of the South Approach Viaduct of the Lezíria Bridge, in Portugal, supported by monitoring-based techniques. Some relevant conclusions could be drawn: – The adopted analysis strategy, based on a phased, non-linear, FE analysis, and considering realistic material properties and construction sequence, was able to reproduce the observed strain variation during construction and in the first years in service. – The calibrated FE model is a useful instrument for interpretation of the measured long-term strain variations and assessment of the structure response. – The prediction model presented in this work can be used to establish normal correlation patterns between material and environmental parameters (e.g. temperature and shrinkage strains) and the observed structural response (e.g. strains, rotations or movements of expansion joints). The model can be employed to calculate the expected response of the different transducers, provided that the problem unknowns are determined by considering an initial time window in which healthy behaviour is assumed. – A real example was presented where it is shown the usefulness of the prediction model in which a maintenance operation with implications in the monitoring results was detected. – The best agreement between measurements and model predictions were obtained or the relative displacements at the expansion joints with relative errors lower than 5%. ACKNOWLEDGEMENT The authors acknowledge the support from the Portuguese Foundation for Science and Technology through the Research Project PTDC/ECM/68430/2006 and the PhD grants SFRH/BD/29125/2006 and SFRH/BD/25339/2005 attributed to the first and second authors. Support from the contractor consortium, TACE, and the infrastructure owner, BRISA, is also gratefully acknowledged. REFERENCES Barr, P.J., Kukay, B.M. & Halling, M.W. 2008. Comparison of prestress losses for a prestress concrete bridge made with high-performance concrete. Journal of Bridge Engineering, 13: 468–475. Cruz, P.J. & Wisniewski, D.F. 2004. Ave River Bridge – A Major Precast Prestressed Concrete U-Girder Bridge in Portugal. PCI Journal, 2004: 72–85. Ghali, A., Favre, R. & Elbadry, M. 2002. Concrete Structures: Stresses and Deformations, 3rd. Ed., London and New York: Spon Press. Lounis, Z., Mirza, M.S. & Cohn, M.Z. 1997. Segmental and conventional precast prestressed concrete I-bridge girders. Journal of Bridge Engineering, 2: 73–82. Ma, Z., Huo, X., Tadros, M.K. & Baishya, M. 1998. Restraint Moments in Precast/Prestressed Concrete Continuous Bridges. PCI Journal, 43: 40–57. Miller, R., Castrodale, R., Mirmiran, A. & Hastak, M. 2004. NCHRP Report 519 – Connection of Simple-Span Precast Concrete Girders for Continuity. Washington: Transportation Research Board. Mirmiran, A., Kulkarni, S., Castrodale, R., Miller, R. & Hastak, M. 2001. Nonlinear Continuity Analysis of Precast, Prestressed Concrete Girders with Cast-in-Place Decks and Diaphragms. PCI Journal, 46: 60–80. Montgomery, D.C. and Runger, G.C., 2003. Applied statistics and probability for engineers. 3rd ed. New York: John Wiley. Sousa, C.F., Sousa, H., Serra Neves, A. & Figueiras, J. 2012. Numerical evaluation of the long-term behavior of precast continuous bridge decks. Journal of Bridge Engineering, 17: 89–96. Sousa, C.F., Fonseca, M., Calçada, R. & Serra Neves, A. 2013a. New Methodology for Calculation of Required Prestressing Levels in Continuous Precast Bridge Decks. Journal of Bridge Engineering, 18: 1219–1226. Sousa, H., Sousa, C.F., Serra Neves, A., Bento, J. & Figueiras, J.A. 2013b. Long-term monitoring and assessment of a precast continuous viaduct. Structure and Infrastructure Engineering, 9: 777–793. TNO DIANA 2011. Diana User’s Manual – Release 9.4.4. Delft: TNO DIANA BV. Valdés, M. 1997. Behavior during construction and under permanent loads of continuous precast concrete bridges. PhD Thesis, Universitat Politécnica de Catalunya (in Spanish).
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Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
Implementation of a B-WIM system in a centenary steel truss bridge F. Cavadas, B.J. Afonso Costa & J. Figueiras CONSTRUCT, University of Porto, Faculty of Engineering, Porto, Portugal
ABSTRACT: This paper presents a B-WIM system implemented in a centenary double-deck steel truss bridge, the Luiz I Bridge. The objective is to characterize the traffic of the upper deck, which holds a double tram-train railway line. The implemented B-WIM system, comprising only two strain gauges installed in two cross beams, is able to identify the crossing direction, the speed and the vehicles loads. In addition, this system is used to assist the collection of accurate influence lines of the structural response under in-service metro traffic for later application of damage detection methods.
1 INTRODUCTION Weigh-in-motion (WIM) systems can be used to estimate vehicle parameters such as speed, number of axles, axle loads, the distance between axles, etc., without affecting the traffic. These parameters and the classification of vehicles are useful information for the entities responsible for the infrastructures management, as it enables the identification of vehicle overloads, the optimized planning of maintenance works and the control of transport operating companies (Silva & Karoumi, 2015). On the other hand, the record of the structure response combined with the knowledge of the corresponding inducing loads, is of great utility in monitoring the structural condition, not only on issues related to fatigue but also for other types of damage. The early detection of damage followed by the application of the necessary corrective actions prevents further deterioration of the bridge, leading to lower maintenance costs and to the extension of the structure life. WIM systems were initially developed to be applied in road bridges (Silva & Karoumi, 2015). However, in recent years, mainly due to the separation between the managing and operating companies of the railway infrastructures, there has been an increasing need to deploy these systems in railway bridges. There are two types of weigh-in-motion systems: track-based and bridge-based (B-WIM) systems (Karoumi et al., 2007). The first type requires the installation of sensors in the rails in order to capture their response during the crossing of the vehicles, whereas the second one takes advantage of the bridge behavior itself by monitoring the strain acting in key elements. The main objective of this work is to characterize the traffic of the Luiz I bridge, namely in terms of speed and vehicles loads. An additional objective is to obtain vehicle influence-line data with the purpose of applying damage detection methodologies based on moving-load responses (Cavadas et al., 2013). In this context, this work includes a section in which the Luiz I Bridge is described. The third section is devoted to the description of the algorithm to obtain vehicle influence-line data, within which the speed of the vehicles is determined. In the fourth section the procedure to estimate the vehicles loads is presented. In the final section the main conclusions drawn from this work are summarized. 2 THE LUIZ I BRIDGE 2.1 General remarks The bridge is constituted by a metallic double hinged arch, supporting simultaneously two decks at different levels over the river crossing (Fig. 1). Besides the pedestrian loads, the lower deck carries 807
Figure 1.
General view of the Luiz I Bridge.
a roadway whereas the upper deck holds a double tram-train railway line and allows the circulation of emergency road vehicles. In 2005 the strengthening and rehabilitation of the bridge aiming at the integration of its upper deck in the infrastructure of the Porto Metro Network was concluded. During this process a fiber optic based monitoring system was implemented to permanently assess the service performance and structural safety of the bridge. The B-WIM work presented in this paper relies on the structural response of the upper deck, which carries the metro traffic, taking full advantage of the installed instrumentation. By default, the acquisition program collects the sensors signals at intervals of 5 minutes. These data are suitable for the characterization of the bridge response induced by ambient actions, namely temperature, and allows capturing significant behavior variations in time caused by degradation or gradual damage. However, in the context of this study a specific area of the upper deck was selected, which holds sensors distributed among four optical branches, and whose signals were recorded with a frequency of 50 Hz. The data has allowed to properly characterize the traffic in terms of speed, travelling direction and bogies’loads, as well as the collection of influence lines for later application of damage detection methods. 2.2 Structural characterization of the upper deck The upper deck is 391.25 m long and comprises two continuous truss girders, 5 m high and 4.65 m apart between axes, supported by the arch crown, seven piers and the abutments, either made of steel or masonry, materializing a total of 13 spans. These truss girders are laterally braced by diagonal and transverse beams at the chords’ levels and in each alignment of the verticals. Only in the area where the upper deck and the arch merge, these girders have variable height, being replaced by I-profiles at the two central spans. The 10 m wide floor system is supported on a main grid of stringers and cross beams, which transversely extend by means of cantilevers to comprise the 1.75 m wide lateral sidewalks. Its structure comprises two parallel load transfer systems. The first one consists of a set of secondary stringers and transverse beams carrying a steel cladding plate to which the pedestrian and vehicle emergency loads are applied. The second one is formed by the rails and the wooden sleepers, and receives the loads of metro vehicles. The loads are transferred by both systems to the main stringers supported on the cross beams, and these elements deliver the corresponding forces to the girders’ upper chords at the nodes of each vertical alignment. In this way, it is assured that the upper deck loads are not directly applied to the girders’ chords, therefore preventing the appearance of adverse additional local forces to the original steel. 2.3 Monitoring system The monitoring system of the bridge relies on a fiber optic network with a tree configuration, in which a main optical cable derives into 14 branches containing multiple sensors connected in series. It enables the measurement of the following parameters: (i) strains in selected truss elements of the arch, upper deck girders, steel piers and suspension ties; (ii) relative horizontal displacements of the expansion joints at the abutments and between the bridge upper deck and the masonry piers; and (iii) temperature of the ambient air and experienced by the steel, both in arch and in upper deck elements (Costa & Figueiras, 2012). 808
Figure 2.
Scheme of the side view of the upper deck depicting the instrumented cross sections.
Figure 3.
Scheme of the instrumented cross sections.
Figure 4.
Scheme of a metro vehicle.
The data collected by the monitoring system has been used for long-term continuous observation of the bridge behavior due to changes in the surrounding environment, namely temperature dependent, as well as variations induced by slow deterioration phenomena. To this end, the sensors signals are acquired every 5 minutes, which constitutes a low reading frequency for fully capturing the structural response as result of the metro vehicles crossing. Therefore, in the context of the work herein presented, a region of the upper deck instrumented by sensors allocated to four channels was selected, for which the data acquisition rate was set at 50 Hz. Without jeopardizing the primary objective of the monitoring system, this upgraded setting has allowed a complete characterization of the traffic in terms of speed, travelling direction and bogies’ loads, as well as the collection of influence lines for later application of damage detection methods. Figure 2 depicts the location of the instrumented cross sections in the upper deck spans under study and Figure 3 presents the corresponding layout of the applied strain sensors. 2.4 Loading system The upper deck of the Luiz I Bridge holds the Yellow double railway Line (D) integrated in the Porto Metro Network, in which the travelling vehicles are Eurotram trains produced by Bombardier Transportation, with a capacity for 80 seated places and capable of reaching 80 km/h. The compositions may be constituted by one or two coupled vehicles. However, this work has targeted only the first ones. The major features of this vehicle are depicted in Figure 4, as well as the loads corresponding to its self-weight. 809
Figure 5.
Measurements taken during a day of observation.
3 PROCEDURE TO OBTAIN INFLUENCE LINE DATA 3.1 General remarks In general, the vehicles geometry, namely the number and the distances between axles, are not known, either in road or railway bridges. However, in this case study this information is known a priori. Therefore, with the objective of characterizing the vehicles, one may take advantage of this knowledge. The sensor B of both cross sections S26 and S29, instrumented in two cross beams about 21 m apart, were selected to characterize the metro traffic on the upper deck of the bridge. The results of the measurements collected for different vehicles (including 1- and 2-vehicles compositions), travelling in both directions, showed that the effects of the individual axles of a given bogie cannot be separated. Furthermore, for 2-vehicles compositions the response of the central bogies cannot be separated either. Therefore, the analysis of the response pattern, namely the identification of the number of peaks, enables to determine the type of composition. As aforementioned, this work focuses on 1-vehicle compositions. However, the procedure described herein can be easily adapted for targeting the characterization of 2-vehicles compositions. This section aims to describe the procedure to detect vehicles traversing the bridge, to identify their travelling direction and to estimate the corresponding speed. In addition, with the objective of later application of damage detection methods, the process of obtaining influence-line data is described. 3.2 Data acquisition and smoothing The first step of any procedure to characterize traffic based on the structural response of a bridge is to acquire data at a rate that enables an adequate characterization of the structural response. As the metro crosses the structure at a speed lower than 10 m/s, an acquisition rate of 50 samples per second (which leads to at least five samples per meter in terms of the vehicle influence line) was found appropriate. Nevertheless, in addition to the phenomenon under observation, the signals include errors, which, regardless of their source, are typically described as noise. In order to remove this noise as much as possible without, at the same time, unduly degrading the underlying information, a Savitzky-Golay filter (Savitzky & Golay, 1964), using a fourth degree polynomial curve, with a moving window of 45 points of equal weight, was applied to smooth the acquired signals. 3.3 Vehicle detection Figure 5 depicts the measurements taken with sensor S26-B along a day of observation, in the period in which the metro is in-service (from 6 a.m. to 1 a.m.). During the day, the wavelength shift related to temperature variations (general trend) is clearly higher than that related to the metro crossings (several peaks). Consequently, in order to identify the passage of vehicles the effects due to temperature variations must be removed. In general, this is done by subtracting the best fit of a low order polynomial curve. Then a max-to-min difference over a short period of time is used to identify the sections of the time signal that correspond to passing vehicles (Silva & Karoumi, 2015). 810
Figure 6.
Measurements taken during a 40-seconds period in which a 1-vehicle composition is detected.
Herein an alternative procedure that consists in finding measurements above a given threshold in relation to the general trend, in a window that moves along the data, is adopted. The window size was chosen as small as possible in order to ensure that the general trend within that period is approximately constant. Provided that the window size is at least twice the period in which the signal of the instrumented element changes due to the vehicles crossing, the general trend may be defined through the median value. Figure 6 depicts the measurements taken with sensor S26-B for a 40-seconds period in which a 1-vehicle composition is detected. The window size (30 seconds long), the median value of the measurements comprised in the window, and the threshold (0.010 nm above the median value) are marked in the figure. A vehicle is detected when measurements are above the threshold. Moreover, as mentioned above, the identification of the number of peaks enables to determine the type of composition. Once a vehicle is detected the set of time-points in correspondence to the peaks are recorded for later use. 3.4 Crossing direction In the previous step, the presence of vehicles over the bridge was identified based on the measurements taken with a single sensor (S26-B). However, information concerning the crossing direction and the speed requires additional data. The procedure described for detecting vehicles with sensor S26-B data was, thus, also carried out for data acquired with sensor S29-B. Hence, two groups (one for each cross section) of time-points sets (corresponding to vehicles detected) are obtained independently. It is noteworthy that a vehicle identified in a given cross section may not be identified in the other, namely if two vehicles traversing the bridge in opposite directions cross the sensor under analysis approximately at the same time. Consequently, prior to any further step, the time-points sets of each group must be matched. As the distance between the instrumented cross beams (about 21 m) is lower than the end-axles spacing (about 30 m for a 1-vehicle composition), within a range of vehicle positions the signals of both sensors change due to the vehicles crossing. Hence, irrespectively of the crossing direction, two conditions may be imposed in order to match the time-points sets identified in each cross section: the instants of the first peaks detected in sensors S26-B and S29-B must be lower than the last peaks detected, respectively, in sensors S29-B and S26-B. Once the time-points sets related to vehicles crossings identified with each sensor are matched, the identification of the crossing direction is straightforward. If the vehicle first reaches cross section S26, the vehicle moves in direction 1, otherwise, it travels in direction 2. 3.5 Travelling speed In a general situation the axles distance is not known a priori. Therefore, the time-points corresponding to the passage of the bogies over each cross section under analysis as obtained above are used to estimate both the speed and the axles distance (Liljencrantz et al., 2007). However, in this case study the axles distance is known and thus the estimates of the travelling speed may take advantage of this knowledge. The analysis of the results showed that successive peaks observed in each sensor under analysis (S26-B and S29-B) occur when the center of the bogies is over the instrumented structural element. Therefore, the time instant in which the vehicle center is aligned with the instrumented structural 811
Figure 7. Vehicle influence line of the strain measured with sensor S26-B.
element is approximately the average of the time-points included in the set of the corresponding cross section. Thus, assuming constant speed between the instrumented cross sections, the average speed is given by the ratio between the distance and the elapsed time. The estimates of the vehicles speed obtained during a day of observation are shown in Figure 8. 3.6 Vehicle influence line As mentioned above, each peak observed in the response occurs when the center of a bogie is over the instrumented structural element. Consequently, to each time-point detected, a position of the vehicle may be assigned. Thereafter, the position of the vehicle along the crossing may be estimated by adjusting a polynomial curve to the data, in which the x-data are the time-points corresponding to the peaks and the y-data are the successive positions of the vehicle. Two different situations were assumed. In the first, the vehicle speed was assumed constant and, thus, a first degree polynomial curve was used. In the second, the vehicle speed was assumed variable with constant acceleration and, thus, a second order polynomial curve was used. Figure 7 depicts the influence line of sensor S26-B obtained for a crossing of a 1-vehicle composition. 4 VEHICLE LOADS ESTIMATES 4.1 Description of the approach The estimates of the vehicle loads are based on the strains collected on the cross sections of the cross beams (S26 and S29) during the vehicles crossing on the upper deck. In the section above it was shown that the passage of the vehicles bogies may be distinguished in the pattern of the structural response. Furthermore, each of the four peak strains occurs when each bogie is over the instrumented structural element. However, as for most bridges, the effects of individual axles of a bogie cannot be separated. Consequently, the algorithm proposed herein is only able to identify the bogies loads. In a structure whose behavior is linear and elastic, the response may be obtained by the superimposition of the effects of each load computed separately. Thus, the response of a vehicle with several axles may be given by the sum of the responses caused by each axle. Taking advantage of the influence line concept, the response of a given axle is obtained by multiplying the axle load by the ordinate value of the influence line at the axle position. According to the information given by the operating company, the axles loads of a bogie of the vehicle are equal, irrespectively of the vehicle occupation. In addition, as the axles distances of every bogie are equal (1.40 m), one may build the influence line of a bogie and obtain the structural response of a vehicle by multiplying the bogie load by the ordinate value of this influence line instead of using the individual axles loads and the unit influence line. As described in section 2.4, the geometry of the vehicles crossing the bridge is known. Moreover, the bogies distances are equal (denoted as d). Therefore, as the vehicle moves forward d, the positions of bogies B1 , B2 and B3 , in the former, and bogies B2 , B3 and B4 , in the latter, are, respectively, the same. Consider, for instance, that in the former, bogies B1 , B2 , B3 and B4 are, respectively, at the abscissas x(Bi ), x(Bi+1 ), x(Bi+2 ) and x(Bi+3 ) – equally spaced by d. As the vehicle moves forward d, the bogie B1 moves to the abscissa x(Bi−1 ) and bogies B2 , B3 and B4 , move, respectively, to the abscissas x(Bi ), x(Bi+1 ) and x(Bi+2 ). Consider two additional positions in which bogies B3 812
and B4 move to the abscissa x(Bi ). The strains obtained in each of these positions, denoted as ε1 , ε2 , ε3 , and ε4 , are given, in the matrix form, as follows:
where B1 , B2 , B3 and B4 are the bogies loads and y(Bi+3 ) to y(Bi−3 ) are the ordinate values of the bogie influence line at the abscissas referred above. With the objective of estimating the bogies loads based on the strains collected as a vehicle crosses the positions pointed above, the ordinate values, y, must be obtained previously. The equations expressed above may be re-written as follows:
Consider a test in which the strains under a vehicle of known loads is performed. Substituting the values of the strains, εj , and the bogies loads, Bk , the expression written above leads to a problem of four equations and seven unknowns. Consequently, an additional test, in which the vehicle load must be distinctly distributed over the bogies, is required in order to obtain at least as many equations as unknowns. Still, with the goal of estimating ordinate values as accurate as possible, further tests should be performed in order to obtain a higher number of equations. The ordinate values are then obtained by solving a minimization problem based on the minimum least squares method. 4.2 Field results The equations presented in the section above refer to a general case in which every ordinate values of the bogie influence line may have a non-negligible contribute to the peak deformations, εj. However, in some situations, some of these ordinate values may be assumed null. Consider the measurements taken with sensor S26-B as a 1-vehicle composition crossed the bridge, shown in Figure 7. The abscissas of the vehicle center corresponding to positions in which each of the bogies is over the instrumented cross section are marked with cyan dotted lines. In addition, the abscissas of the vehicle center corresponding to positions in which the first and the fourth axles are d m apart the instrumented element are also marked (red dotted lines). In the former position, the first axle is at the same abscissa of the second axle for a position in which the first axle is over the instrumented element. Conversely, in the latter, the fourth axle is at the same abscissa of the third axle for a position in which the fourth axle is over the instrumented element. It is, thus, shown that the strains for these both positions are very low. Furthermore, when the vehicle is at farther positions the effects on the instrumented element may be assumed null. Therefore, in this situation, only the ordinate value of the bogie influence line at the abscissa of the instrumented element – y(Bi ) – seems to be significant. Still, the adjacent ordinate values – y(Bi+1 ) and y(Bi−1 ) – were also taken into account to estimate the bogies loads. The remaining ones were assumed null. Hence, the non-null ordinate values may be obtained as expressed in equation 2 by solving a system of four equations with three unknowns. Figure 8 depicts the estimates of the total loads of the vehicles that crossed the bridge upper deck during a Sunday. 813
Figure 8.
Estimates of the vehicle loads and travelling speed.
5 CONCLUSIONS This paper has presented a work with the main objectives of characterizing the traffic of the Luiz I Bridge, namely in terms of speed and vehicles loads, and obtaining vehicle influence-line data for later application of damage detection methodologies based on moving-load responses. Without jeopardizing the primary objective of the existing monitoring system, an upgraded setting applied to a key instrumented region of the upper deck has allowed to collect the fundamental strain data. Taking into account that the structural response is due to several simultaneous actions and factors, a procedure for the adequate separation of the effects associated to the crossing of the metro vehicles has been presented. The process of detecting their passage and crossing directions was also described, as well as the technique adopted for computing their travelling speed, an utmost aspect to obtain accurate vehicles influence lines by shifting the data domain from time to space. The general approach to estimate the bogies’ loads for this case and in other similar bridges was detailed, which is based on the principle of the superimposition of the effects assuming linear and elastic structural behavior. Preliminary results have shown a good quality in terms of speed and load estimates for the 1-vehicle compositions, being already planned the extension of the traffic characterization to 2-vehicle compositions, ultimately aiming at the assessment of the number of passengers using the metro line. ACKNOWLEDGMENTS The financial support provided by the Portuguese Scientific Foundation (FCT-MCES) to the first author through the PhD grant SFRH/BD/42315/2007 and to the second author trough the Post-Doc grant SFRH/BPD/98963/2013 is acknowledged. REFERENCES Cavadas, F., Smith, I. F. C. & Figueiras, J. 2013. Damage detection using data-driven methods applied to moving-load responses. Mechanical Systems and Signal Processing, 39: 409–425. Costa, B. J. A. & Figueiras, J. A. 2012. Fiber optic based monitoring system applied to a centenary metallic arch bridge: Design and installation. Engineering Structures, 44: 271–280. Karoumi, R., Liljencranz, A. & Carlsson, F. 2007. Assessement of Actual Traffic Loads Using B-WIM, Site Specific Characteristic Load from Collected Data & Statistical Evaluation of Dynamic Amplification Factors (Background document D4.3.2). Sustainable Bridges. Liljencrantz, A., Karoumi, R. & Olofsson, P. 2007. Implementing bridge weigh-in-motion for railway traffic. Computers & Structures, 85: 80–88. Savitzky, A. & Golay, M. J. E. 1964. Smoothing and differentiation of data by simplified least squares procedures. Analytical Chemistry, 36: 1627–1639. Silva, I. J. G. & Karoumi, R. 2015. Traffic monitoring using a structural health monitoring system. Proceedings of the ICE – Bridge Engineering, 168: 13–23.
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A novel inspection method for orthotropic steel decks using phased array ultrasonic testing T. Makita & H. Sakai Central Nippon Expressway Co., Ltd., Nagoya, Japan
T. Suzuki MM Bridge Co., Ltd., Hiroshima, Japan
N. Yagi Mitsubishi Hitachi Power Systems Inspection Technologies Ltd., Yokohama, Japan
ABSTRACT: Over the last decade, fatigue cracking has been observed at the weld joint between the trapezoidal rib and the deck plate in many orthotropic steel deck bridges in Japan. Fatigue cracks initiated at the rib-to-deck weld root can cause pavement damage and thus impair transportation safety. It is necessary to detect and repair such fatigue cracks at the early stage of development. This paper presents a novel inspection method for the rib-to-deck weld of orthotropic steel decks with the aid of phased array ultrasonic testing. Automated scanner holding twin phased array ultrasonic probes is developed to simultaneously inspect the deck plate and the weld bead for internal part-through cracks. The developed inspection method is implemented at the Meiko-Chuo Bridge, a steel cable-stayed bridge carrying the Isewangan Expressway over the Port of Nagoya, and its performance is verified. 1 INTRODUCTION For bridge decks, fatigue loading is one of the most detrimental actions because of rather high live-to-dead load ratio and hence large stress range is caused in the components of bridge decks. Among various types of bridge decks, orthotropic steel decks are supposed to be subjected to the most severe fatigue loading due to the light weight for which, however, those have been commonly used in long-span bridges where the minimisation of dead load is of highest importance. The orthotropic steel deck is a welded structure consisting of a flat and thin steel plate stiffened by a series of longitudinal ribs with support by orthogonal crossbeams. Since its development in 1950’s, the orthotropic steel deck system has been utilised throughout the world. Over the last few decades, lots of reports have been published concerning fatigue cracking in orthotropic steel decks resulting from the complicated welded details and minimised plate thickness to reduce weight (Wolchuk 1990, Boersma & de Jong 2003, BASt 2005, Miki 2006, Battista et al. 2008). In Japan, more than 2,300 orthotropic steel deck bridges have been constructed so far (JBA 2015) and fatigue cracking in orthotropic steel decks has been observed mainly in heavily trafficked urban expressways. In 1980’s and 1990’s, fatigue cracking at the rib-to-crossbeam connection and the welded rib splice was a major issue, while during the last 15 years, fatigue cracking at the rib-to-deck weld has been increasingly observed where trapezoidal ribs are used (Mori 2012). Four types of fatigue cracking at the rib-to-deck weld are shown in Figure 1 (Xiao et al. 2008). Crack 1 and 2 are initiated at the weld toe located on the rib wall and the deck plate, respectively. Crack 3 and 4 are initiated at the weld root and grow in the weld bead and the deck plate, respectively. Location of crack initiation depends on various parameters relating to the connection. These four types of cracks can cause pavement damage and thus impair transportation safety. Therefore, detection and repair of those fatigue cracks at the early stage of development are 815
Figure 1.
Four types of fatigue cracks at the rib-to-deck weld of orthotropic steel decks (Xiao et al. 2008).
highly needed. However, while Crack 1 and 2 can be detected by magnetic particle inspection or hands-on visual inspection from the underside of the orthotropic steel deck, Crack 3 and 4 cannot be detected by such conventional inspections until those cracks propagate through the weld bead and the deck plate. In order to detect internal part-through Crack 3 and 4, inspection methods using Non-Destructive Testing (NDT) have been investigated so far and several methods have been used in practice. The most widely studied and employed is ultrasonic testing method to detect internal part-through fatigue cracks in the deck plate (Bekker & de Jong 2003, Murano et al. 2008, Murakoshi et al. 2012). The ultrasonic testing is carried out from the underside of the orthotropic steel deck and thus traffic disruption and pavement removal are avoided. As an advanced method of ultrasonic testing, application of phased array ultrasonic testing was investigated where its high crack detection ability was expected, and it was found that phased array ultrasonic testing can detect internal part-through fatigue cracks in the deck plate whose minimum length is 2 mm (Sugiyama et al. 2010). Concerning detection of internal part-through fatigue cracks in the weld bead, few research results have been reported so far in the literature. By applying infrared thermography from the underside of the orthotropic steel deck, detection of internal part-through cracks in the weld bead was attempted in which temperature difference between cracks and the weld bead is supposed to indicate the existence of cracks in the weld bead (Mizokami et al. 2014). This literature review reveals that fatigue cracks in the deck plate (Crack 4 in Figure 1) are addressed more carefully than fatigue cracks in the weld bead (Crack 3 in Figure 1). This is because repair of through cracks in the deck plate is time-consuming and cost-intensive. However, fatigue cracks in the weld bead need to be dealt with as deliberately as fatigue cracks in the deck plate because when through cracks in the weld bead develop to a certain length, the trapezoidal rib doesn’t stiffen the deck plate sufficiently and rather large deflection can be caused in a part of the deck plate, eventually leading to pavement damage and decrease of transportation safety. The objectives of the present paper are to describe a novel inspection method using phased array ultrasonic testing for the rib-to-deck weld of orthotropic steel decks. Automated scanner holding twin phased array ultrasonic probes is developed to simultaneously inspect the deck plate and the weld bead for internal part-through fatigue cracks initiated at the rib-to-deck weld root. The developed scanner is applied to the orthotropic steel deck of the Meiko-Chuo Bridge to evaluate its performance. 2 PHASED ARRAY ULTRASONIC TESTING Phased array ultrasonic testing is a relatively new NDT technique in the field of civil engineering. In this section, basic principles and advantages of phased array ultrasonic testing are explained according to (Olympus NDT 2007 & 2010, Ditchburn et. al 2009). Phased array ultrasonic testing has been used for medical diagnosis for more than forty years. Meanwhile, application of the phased array ultrasonic testing to industries wasn’t so popular until 1990’s because acoustic properties and dimensions of materials were varied extensively. After that, the rapid technology development encouraged the use of phased array ultrasonic testing in the industrial field. 816
Figure 2.
Simple transducer element arrangements (a) linear array, (b) annular array and (c) matrix array.
Figure 3. Three major beam scanning patterns (a) linear scanning, (b) dynamic depth focusing and (c) sectorial scanning.
A phased array ultrasonic probe consists of multiple transducer elements (usually between 16 and 256), each of which can act as a single ultrasonic transducer. The transducer elements are arranged in various patterns where individual and independent driving of an element must be attainable without inducing vibration in nearby elements and the performance of each element must be as close as possible for formation of a homogeneous beam. Examples of simple transducer element arrangements are shown in Figure 2. Due to the ability to control the acoustic characteristic of ultrasonic beams, phased array ultrasonic systems can achieve the electronic scanning techniques that are not made possible by using conventional ultrasonic systems. In the following, three major computer-controlled beam scanning patterns are explained in which the transducer elements are assumed to be arranged in a linear pattern. (1) Linear scanning A group of the transducer elements is pulsed to generate the desired beam shape and then the focal law giving this beam shape is multiplexed along the length of the array (Figure 3a). This is equivalent to a conventional ultrasonic transducer performing a raster scan. (2) Dynamic depth focusing The focal point is moved along the nominal beam axis by varying the focal laws (Figure 3b). (3) Sectorial scanning The beam is swept through an angular range for a specific focal point using the same group of the transducer elements (Figure 3c). By combining the beam scanning patterns, a far more complicated scan and greater coverage are provided than by using conventional ultrasonic systems. In order to exploit the performance of phased array ultrasonic system to the fullest, the following probe parameters need to be determined deliberately. (1) Frequency In general, penetration increases with lower frequency, while resolution and focal sharpness increase with higher frequency. (2) Size of transducer elements As the size of individual transducer elements gets smaller, beam steering capability increases, but large area coverage requires more transducer elements at a higher cost. 817
(3) Number of transducer elements As the number of transducer elements increases, focusing capability, steering capability and physical coverage area increase, but probe and instrumentation costs and system complexity increase as well. (4) Pitch and aperture Pitch is the distance between individual transducer elements and needs to be small for optimisation of steering range. Aperture is the effective size of a pulsing element usually consisting of a group of individual transducer elements that are pulsed simultaneously and needs to be large for optimum sensitivity, minimum unwanted beam spreading and strong focusing. Three main advantages of phased array ultrasonic testing over conventional ultrasonic testing are increased inspection sensitivity, increased inspection coverage and decreased inspection time. Increased inspection sensitivity is attained through optimisation of the beam angle, shape and width for detection of discontinuity. Increased inspection coverage is achieved owing to complex scans allowing a greater volume of material to be investigated from one probe position. Decreased inspection time is possible because of very rapid scan rates. 3 NOVEL INSPECTION METHOD FOR ORTHOTROPIC STEEL DECKS Although not directly leading to collapse, some defects in bridges lower the serviceability. Detection of such defects at the early stages of development is quite important for bridges to function properly and may significantly reduce the high repair cost. Fatigue cracking in orthotropic steel decks, in particular at the rib-to-deck weld is one such defect. In order to detect fatigue cracks initiated at the rib-to-deck weld root before propagating through the deck plate or the weld bead, a novel inspection method using phased array ultrasonic testing was developed. There are two characteristics of the developed method. One is that the deck plate and the weld bead are simultaneously inspected for internal part-through cracks. Another is that the phased array ultrasonic scanner is automated. In the following, the two characteristics are explained separately. 3.1 Simultaneous inspection of the deck plate and the weld bead Although most existing inspection methods for the rib-to-deck weld aim at detecting fatigue cracks in the deck plate (Crack 4 in Figure 1), the developed method inspect both the deck plate and the weld bead simultaneously, and fatigue cracks in either the deck plate or the weld bead (Crack 3 or 4 in Figure 1) are detected if propagating. Because fatigue cracks in both the deck plate and the weld bead can induce pavement damage resulting in decrease of transportation safety, it was determined that both the deck plate and the weld bead are inspected in the developed method. For simultaneous inspection of the deck plate and the weld bead from the underside of the orthotropic steel deck, a scanner holding twin phased array ultrasonic probes was developed (Figure 4a). It was designed that one probe is placed on the deck plate (“the deck probe” hereafter) and another probe is placed on the rib wall (“the rib probe” hereafter) (Figure 4b). The positions of the probes were determined such that ultrasonic beams reflected by the opposite surface of steel members can sweep the area in which fatigue cracks initiated at the rib-to-deck weld root would propagate, where the sectorial scanning is applied. The deck probe examines both the deck plate and the weld bead, while the rib probe examines the weld bead solely. 3.2 Automated scanner When the rib-to-deck weld is inspected from the underside of the orthotropic steel deck by applying ultrasonic testing, it has been performed so far with manually operated probes where it is necessary for probe operators to raise their arms to fix probes to steel members. Considering the orthotropic steel deck is often used in long-span bridges where a great amount of the rib-to-deck weld joint 818
Figure 4. Automated scanner holding twin phased array ultrasonic probes (a) general view drawing and (b) schematic of twin probes fixed to the deck plate and the rib wall.
exists, manual ultrasonic inspection of the rib-to-deck weld is heavy work. In order to reduce labour in ultrasonic inspection of the rib-to-deck weld and increase the inspection efficiency, the scanner holding twin phased array ultrasonic probes was made to be moved by a motor-controlled drive unit. The scanner was designed to fit in the corner formed by the deck plate and the trapezoidal rib wall to which it magnetically adheres. For easy portability, it was made as small as possible: 130 mm wide, 90 mm deep and 115 mm high with weight of 1.0 kg. It is driven by a small electric motor and operated by wired remote control. The probe position is tracked by an electromechanical encoder set up in the scanner and thus ultrasonic data is correlated with the actual probe position, allowing a proportional view to be plotted and data to be matched to specific areas of the part being inspected. 4 APPLICATION The developed inspection method was implemented in the Meiko-Chuo Bridge to evaluate its performance. Fatigue damage in the orthotropic steel deck of the Meiko-Chuo Bridge is briefly described and application of the automated phased array ultrasonic scanner is explained. 4.1 Fatigue damage in the orthotropic steel deck of the Meiko-Chuo Bridge The Meiko-Chuo Bridge is a steel cable-stayed bridge with a main span of 590 m and symmetric side span of 290 m, carrying the Isewangan Expressway dual carriageway with three lanes in each direction over the Port of Nagoya (Figure 5a and b). Its steel box girder is aerodynamically profiled with an orthotropic deck plate (Figure 5c). Designed based on former design codes where the fatigue safety check of weld joints was not required, the Meiko-Chuo Bridge has fatigue-prone structural details in the orthotropic steel deck such as small thickness of the deck plate (=12 mm). Since its opening in 1998, the traffic on the Meiko-Chuo Bridge is increasing year by year and now annual average daily traffic is about 87,000 vehicles in total 38% of which are categorised as heavy vehicle (gross vehicle weight larger than 80 kN). This traffic situation together with fatigue-prone details caused precocious fatigue damage in the orthotropic steel deck in the spring of 2014. Embarrassing is that the fatigue damage in the orthotropic steel deck was not recognised until a part of the pavement was damaged such that the transportation safety was nearly impaired. A 2-m-long through crack was found in the rib-to-deck weld bead just below the damaged pavement (Figure 6a). Development of fatigue cracks in the same structural details was anticipated and ad hoc hands-on visual inspection was conducted for the rib-to-deck weld in the far left-hand lane (for slow traffic) in each direction of the Meiko-Chuo Bridge where the most severe wheel loading was supposed to happen. Fatigue cracks of total length of 11.7 m were detected, 50% of which were developed in the eastbound main span. 819
Figure 5. The Meiko-Chuo Bridge (a) elevation (b) general view and (c) half cross section.
4.2 Inspection of the rib-to-deck weld using the automated phased array ultrasonic scanner In order to evaluate the usability and the crack detection ability, the developed scanner holding twin phased array ultrasonic probes was applied to the orthotropic steel deck of the Meiko-Chuo Bridge (Figure 6b). As for the crack detection ability, internal part-through cracks in the deck plate and the weld bead were expected to be detected by the developed scanner. For verifying if such internal part-through cracks are properly detected, after scanning the rib-to-deck weld, it is necessary to remove specimens from the orthotropic steel deck and compare scan data with actual crack shapes; however, it is not possible to remove specimens from structures in operation. Therefore, in the present investigation, the crack detection ability of the developed scanner was examined by applying it to the rib-to-deck weld with through cracks in the weld bead in the vicinity of which internal part-through cracks were supposed to exist, and those internal part-through cracks were expected to be detected by the developed scanner. Considering the influence of crack shape and paint coating on the crack detection capability, parameters of phased array ultrasonic probes were determined by experimental tests and simulation as follows. 32 transducer elements were linearly arranged with pitch of 0.4 mm where aperture was 12.75 mm long and 8.3 mm wide, and ultrasonic pulse frequency was 10 MHz. A total of 97.5 m long rib-to-deck weld in the far left-hand lane in the eastbound main span was inspected by the developed scanner during three days. Although the scanner is capable of inspecting 90 m long rib-to-deck weld per hour, its inspection speed was 12.7 m per hour in this application. This is because scan data occasionally failed to be acquired and inspection was conducted on the same part of the rib-to-deck weld two times. For better inspection efficiency, the data acquisition function of the scanner needs to be improved. Figure 7a shows 2-D images of ultrasonic data obtained by the deck probe examination for a part of an inspected rib-to-deck weld bead with a 200 mm long through crack. Top view and side view are projections on horizontal plane and vertical plane, respectively, where the horizontal axis represents the probe movement path in the longitudinal direction of the bridge and the vertical axis represents the ultrasonic beam path in the transverse direction of the bridge for top view and the ultrasonic beam path in the vertical direction for side view (Figure 7a and b). In those views, location of the 820
Figure 6. (a) Through-thickness crack in the rib-to-deck weld bead and (b) inspection of the rib-to-deck weld by applying the developed phased array ultrasonic scanner.
Figure 7. (a) 2-D images of ultrasonic data obtained for a part of an inspected rib-to-deck weld bead with a through crack by the deck probe examination and (b) explanation of top view and side view.
through crack in the weld bead is indicated as an area between dotted lines. Ultrasonic indications were observed not only at the through crack location, but also in the vicinity of the right end of the through crack, which was very likely attributed to internal part-through cracks (Figure 7a). From this it may be said that the developed scanner is capable of detecting internal part-through cracks in the rib-to-deck weld bead. The through crack deviated and propagated into the rib wall at the left end where no ultrasonic indications were produced. In the present application, no ultrasonic indications were observed in the deck plate; therefore, it is not possible to discuss the ability of the developed scanner to detect internal part-through cracks in the deck plate. Looking at ultrasonic indications at the through crack location, those were not produced uniformly. This is supposed to be because the through crack was inclined variously and some reflected ultrasonic beams weren’t received and identified as indications. In order to improve the crack detection ability of the developed scanner, it is necessary to understand the relationship between the shapes of fatigue cracks at the rib-to-deck weld and ultrasonic beam reflection, for which further investigation is needed. 5 CONCLUSION The present paper describes a novel inspection method for the rib-to-deck weld of the orthotropic steel deck using phased array ultrasonic testing. 821
Basic principles of phased array ultrasonic testing is explained and advantages over conventional ultrasonic testing are clarified. Considering the fact that fatigue cracks initiated at the rib-to-deck weld root propagating in both the deck plate and the weld bead can induce pavement damage resulting in decrease of transportation safety, a scanner holding twin phased array ultrasonic probes are developed for simultaneous inspection of the deck plate and the weld bead. Because the orthotropic steel deck usually has a significant amount of the rib-to-deck weld to be inspected, the developed scanner is designed to be moved by a motor-controlled drive unit. The developed scanner is applied to the orthotropic steel deck of the Meiko-Chuo Bridge to evaluate its performance. Although it is understood internal part-through cracks in the weld bead are very likely detected by the developed scanner, further investigation needs to be conducted for evaluating the crack detection ability of the developed scanner more precisely. REFERENCES Battista, R. C., Pfeil, M. S. & Carvalho, E. M. 2008. Fatigue life estimates for a slender orthotropic steel deck. Journal of Constructional Steel Research 64(1): 134–143. Bekker, M. C. M. & de Jong, F. B. P. 2003. Ultrasonic Underside Inspection for Fatigue Cracks in the Deck Plate of a Steel Orthotropic Bridge Deck. Heron 48(4): 277–295. Boersma, P. D. & de Jong, F. B. P. 2003. Techniques and solutions for rehabilitation of orthotropic steel bridge decks in the Netherlands. Proceedings of 10th International conference on Structural Faults and Repair on steel structures. Ditchburn, R. J. & Ibrahim, M. E. 2009. Ultrasonic Phased Arrays for the Inspection of Thick-Section Welds (No. DSTO-TN-0911). Victoria: Maritime Platforms Division Defense Science and Technology Organization. Federal Highway Research Institute (BASt) 2005. Expert Meeting Repair of Orthotropic Deck. Proc. Bergisch Gladbach: BASt. (in German) Japan Bridge Association 2015. Bridge Database. http://www.jasbc.or.jp/kyoryodb/index.cgi (in Japanese) Miki, C. 2006. Fatigue damage in orthotropic steel bridge decks and retrofit works. International Journal of Steel Structures 6(4): 255–267. Mizokami, Y., Kobayashi, Y., Izumi, Y. & Sakagami, T. 2014 Crack probing technology on orthotropic steel deck by temperature gap detection using infrared thermography. Expressways and automobiles 57(10): 35–38. (in Japanese) Mori, T. 2012. Fatigue of Orthotropic Steel Deck: Past, Present and Future. Katayama Technical Report 31: 2–10. (in Japanese) Murakoshi, J., Takahashi, M., Koike, M. & Kimura, T. 2012. Study on practical ultrasonic inspection method for fatigue cracks in steel orthotropic deckplates. Journal of JSCE, Division A: Structural Engineering/Earthquake Engineering & Applied Mechanics 68(2): 453–464. (in Japanese) Murano, M., Saitoh T. & Kinomoto, T. 2008. Inspection method for part-through fatigue cracks in steel deck plate using semi-automated ultrasonic testing. Proceedings of Japan Society of Civil Engineers 2008 Annual Meeting 63(6): 141–142. (in Japanese) Olympus NDT 2007. Introduction to phased array ultrasonic technology applications: R/DTech guideline. Waltham: Olympus NDT. Olympus NDT 2010. Phased Array Testing: Basic Theory for Industrial Applications. Waltham: Olympus NDT. Sugiyama, H., Sugioka, K., Tabata, A., Tsukamot, S. & Utsunomiya, K. 2010. Retrofit and advanced investigation on fatigue cracks penetrating orthotropic steel deck plates. In D. M. Frangopol, R. Sause & C. Kuskko (eds), Bridge Maintenance. Safety, Management and Life-Cycle Optimization: Proc. intern. conf., Philadelphia, 11–15 July 2010. London: Taylor & Francis Group. Wolchuk, R. 1990. Lessons from weld cracks in orthotropic decks on three European bridges. Journal of Structural Engineering 116(1): 75–84. Xiao, Z. G., Yamada, K., Ya, S., & Zhao, X. L. 2008. Stress analyses and fatigue evaluation of rib-to-deck joints in steel orthotropic decks. International Journal of Fatigue 30(8): 1387–1397.
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Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
Self-evaluating smart expansion joints of multi-span and long bridges K. Islami & N. Meng mageba SA, Bülach, Switzerland
ABSTRACT: This paper presents the installed structural health monitoring (SHM) systems of three long suspension bridges, which measure high-frequency movements, inclinations, temperature and vibrations and thus enable a proper understanding of the bridges’ behavior to be developed. These case studies, based on both static and dynamic approaches, demonstrate the usefulness and ease of use of such systems, and the enormous gains in efficiency they offer over alternative manual monitoring methods. The availability of such systems has now led to the development of smart expansion joints: expansion joints that feature an integrated advanced monitoring system, already when fabricated.
1 INTRODUCTION Multi-span and long-span bridges are essential links in transportation networks, and must be inspected and maintained accordingly. They are more likely to experience large deck movements than other bridge types, and these movements must be accommodated by deck expansion joints and bridge bearings. The performance and life expectancy of such components are strongly dependent on the movements to which they are subjected, so the movement and vibration data that can be provided by modern structural health monitoring (SHM) systems can play a pivotal role in improving their performance and extending their service lives. This is demonstrated below with reference to the SHM systems of three major suspension bridges. To maximize in particular the ability to monitor the condition and performance of a bridge’s expansion joints – perhaps the parts of the bridge that are subjected to the highest dynamic impacts – an upgrade to the standard monitoring system of a leading supplier has been realized in one of the bridges. A brand new application incorporates sensors at the expansion joints, providing clear information about the condition of the joints and supporting the planning of maintenance activities. The functioning of the new feature is based on the measurement of structure-borne vibrations. Damage can be clearly identified based on general testing and teaching of the system, facilitating very sensitive, highly robust damage identification. As a result, unexpected damage can be immediately recognized and automatically notified, enabling the timing of replacement of components to be optimized. The implications of the development of techniques of monitoring, of statistical modeling of the response of structures, and of gathering and processing data in real time, are important – especially in the context of particularly sensitive structures. It is desirable to verify whether the effects of various environmental variables measured in situ influence the static or dynamic behavior of the structure. Therefore, it is important to eliminate the influence of these factors, so that small changes due to damage can be detected. This is made possible by the use of regression models, which can determine the static variables starting from a predefined input. All of these techniques have been used in the present study, and more importantly, have been applied to important suspension bridges that exhibit high movements. 823
2 MONITORING OF THE TAIZHOU BRIDGE, JIANGSU PROVINCE, CHINA The Taizhou Yangtze River Bridge (Figure 1), constructed at a cost of USD 400 million and opened to traffic in 2012, is the world’s longest-span bridge of its type: The three-tower suspension bridge, with two main spans of 1,080 m each and side spans of 390 m, crosses the Yangtze River where it has a width of 2.1 km. The ambitious construction project represented the first attempt to create a long-span multi-tower suspension bridge. This extraordinary bridge required some extraordinary key components, such as the expansion joints which accommodate deck movements while providing a driving surface for traffic. Modular expansion joints with 18 gaps each (and thus capable of accommodating 1,440 mm of longitudinal movement) were installed at each end of the deck (Figure 1). A SHM system was installed on the bridge (Islami et al., 2014) to provide the type of data on the bridge’s condition that is likely to be of interest to any owner whose structure is exceptional in some way. The basic system measures and records the movements and rotations of the deck at the expansion joints, and thus gives a valuable impression of the performance of the structure at any time, enabling the need for maintenance or adaptation work to be quickly identified and planned. It also reports accumulated sliding movements over time – a key indicator in evaluating and predicting the condition of key mechanical components such as expansion joints and bearings. An example of the recorded data is presented in Figure 2, showing overall displacements and the correlation between displacements of a particular lamella beam (on the surface of an expansion joint) and the overall movements of the bridge. A 45◦ inclination of the correlation graph would indicate that these values are equal. To maximize in particular the system’s ability to monitor the condition and performance of the expansion joints – perhaps the parts of the bridge which deserve the most inspection and maintenance attention – an upgrade to the monitoring system has been applied. A brand new damage detection feature has been installed, incorporating sensors at the joints to provide clear
Figure 1. The new Taizhou Yangtze River Bridge (left) and a modular expansion joint on the Taizhou Bridge, viewed from below (right).
Figure 2.
Overall movements and correlation between lamella beam and bridge movements.
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information about the condition of the joints and support the planning of maintenance activities. The functioning of the new feature is based on the measurement of structure-borne vibrations recorded at a sampling frequency of 25.6 kHz, with even very tiny changes in the joint or its performance being detectable and visually represented by changes in a curve on a graph. The installed “smart” expansion joint (Figure 3) measures high-frequency movements, inclinations, temperature and vibrations enabling a proper understanding of the joint’s and bridge’s behavior. The main purpose of the project is not only to monitor the condition and performance of the expansion joints due to extensive movement or rotation (basic) but also to detect damages
Figure 3. Preliminary vibrational tests and modal analysis on the large expansion joint (left). Layout of sensors and fault simulation (right).
Figure 4. Example of one month of data sent to the server from the smart expansion joint, at four positions (left). The appearance of dense clouds of measurements outside the normal trend would mean failure of a component. Accurate accelerometers fixed at the main supporting beam and ultrasonic displacement sensors check movements (right).
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at an early stage by recording the level of accelerations and natural frequencies (advanced) caused by traffic. As a first step, many artificial failures (Figure 3) were created in order to simulate damages. The different setups of sensors mostly covered all of the joist beams of the expansion joint. During the tests there was a good distinction by the system between damaged and undamaged joint conditions. This enabled the system to be fine-tuned before permanent installation. The permanent monitoring sends data to a remote server, including daily levels of vibrations due to heavy traffic together with the modal frequencies. If established limits are exceeded, this indicates the occurrence of damage, causing an alarm notification to be sent by email and to appear on the system’s web interface. The next step is a site visit in order to verify the damage and avoid further damage or deterioration. This new application will provide better information about the condition of the joints and support the planning of maintenance activities. Damages can be clearly identified based on general testing and teaching of the system. This will allow very sensitive, but also reliable, damage identification. As a result, unexpected damage can be immediately recognized and notified, enabling the timing of replacement of components to be optimized.
3 PLANNING BRIDGE RENOVATION WORKS – THE ALVSBORG BRIDGE, SWEDEN The Alvsborg Bridge (Figure 5) is a suspension bridge across the Göta Älv River in Gothenburg. Built in 1966 with a main span of 417 m, it is one of the few structures connecting the north and south parts of the city across the river, and is therefore one of Gothenburg’s most important traffic arteries. Moreover, with its 107 m pylons, it is one of the city’s most prominent landmarks. During planning of major renovation works to be completed in the coming years, it was decided to use an automated SHM system to provide detailed information on certain key aspects of the bridge’s condition and performance, allowing the works to be optimized. One part of the project involves the replacement of the bridge’s expansion joints (Figure 6), which are in a poor state of repair as might be expected after a respectable service life of over 40 years. Due to the substantial maintenance effort required by the existing steel finger joints to date, questions have been posed as to whether an alternative type, such as a modular joint, might be more durable and cost-effective considering the particular demands of this structure. In order to optimize the selection and design of the replacement joints, and thus reduce future maintenance and replacement effort, an SHM system was proposed. It was designed to precisely quantify the bridge’s actual movements and rotations, which would be likely to differ from the theoretical values estimated at the time of the bridge’s construction. The system, which was installed in 2011, measures absolute longitudinal and transverse movements, horizontal and vertical rotations,
Figure 5. The monitored Alvsborg Bridge.
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and accumulated longitudinal movements (at high measurement frequency to include all micromovements). It also records the structure temperatures needed to form a frame of reference for the movements and rotations, enabling these to be fully understood. In this way, movements due to traffic can be decoupled from temperature effects, for better understanding. Figures 7 and 8 show the correlation between longitudinal deck movements and temperature, and Figure 7 also shows accumulated deck movements (including micro-movements as well as daily thermal cycles) as they add up over time. It can be seen that the total accumulated movements in the 4-week period shown came to approximately 800 m, from which it might be deduced that the deck experiences total longitudinal movements of over 10 km in a year. For yet more precise information, data measured at a frequency of 10 Hz can be used. For instance, Figure 9 shows typical movements experienced by the structure in 24 hours, describing the behavior with nearly 900,000 values. Such a level of information, generated from high-frequency measurements, could never be provided by manual methods alone.
Figure 6. The monitored finger expansion joint.
Figure 7.
Presentation of measured data (graph form) in the web interface.
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More statistical analysis was carried out using histograms (see Figure 9) and regression models. To improve damage detection at a bridge’s expansion joints, environmental effects can be eliminated using a regression model such as that shown in Figure 10, generated from the correlation between temperature or humidity and the movements of the bridge (Islami & Modena, 2013). The resulting movements recorded by the system do not then include movements resulting from temperature or humidity, enabling abnormal influences (due to damage or other unexpected events) to be easily recognized. This is illustrated by Figure 10, which shows on the lower graph, with normalized values, the bridge movements that are due to traffic etc. and not due to temperature. Any abnormal shift on this graph would be immediately recognizable, enabling the need for repair or preventative action to be assessed. These provide the responsible engineer with optimal information to assist in the planning of renovation works for the expansion joints and for the bridge as a whole. The data generated by the system provides a detailed understanding of the structural behavior of the bridge deck, in particular in relation to thermal impacts and secondary bending moments. It can also be used to provide quantification of the new joints’ required movement and rotation capacities, enabling the most suitable type of expansion joint to be selected (finger, modular or other). And it will enable the design of the joints to be optimized, eliminating overly conservative safety margins and facilitating the selection of the most suitable sliding materials. Considering the high accumulated movements which have been recorded to date, it can be concluded that a standard sliding material such as PTFE would be worn away within only a few years, and that an alternative material that offers exceptional durability and performance must be used.
Figure 8.
Correlation analysis with temperature.
Figure 9.
Expansion joint movements in 24 hours (left) and histogram with distribution of displacements.
828
In addition to providing continuous records of selected variables and analyzing them as described above, the system can also be programmed to send an alarm message should any measured variable exceed a predefined value after temperature effects have been eliminated.
4 MONITORING OF THE TAOHUAYU BRIDGE, ZHENGZHOU, CHINA The Taohuayu ZhengzhouYellow River Highway Bridge (Figure 11) is the fourth bridge of the Xixia Wuzhi Highway over the Yellow River in the city of Zhengzhou, China. The structure, which has a full length of 7,691.5 m, opened to traffic in October 2013. The main structure is a two-tower, threespan self-anchored suspension bridge with spans of 160 m, 406 m and 160 m. The stiffening girder was erected by computer controlled one-way, multi-point and synchronous incremental launching. A permanent SHM system was installed with a focus on the horizontal movements of the bridge deck under environmental and traffic influences, with movements measured at a modular expansion joint at one end of the bridge’s deck (Figure 12, left). In particular, the system measures the longitudinal movements of the first, second and last lamella beams of the expansion joint, and
Figure 10.
Displacement, before and after the elimination of environmental effects.
Figure 11. The Taohuayu Bridge over the Yellow River.
829
Figure 12. A modular expansion joint on the Taohuayu Bridge, viewed from above (left) and overall movements of lamella beams and entire bridge movements shown on the web interface (right).
the corresponding changes in the full width of the bridge gap at this location (Figure 12, right). It also measures deck rotations, and air and structural temperatures. Autonomous data processing and alarming is provided, with online access to data and reports via the web interface. The system thus enables the behavior of the newly constructed structure to be continuously monitored, for the purposes of condition assessment, quality control and environmental effects analysis. 5 CONCLUSIONS The integration of sophisticated structural monitoring systems in the expansion joints of newly built and existing suspension bridges can offer great benefits to their asset management programs. Such systems can efficiently provide data required for almost any purpose, at any stage of a structure’s life-cycle. The newly developed “smart expansion joint” application installed on the Taizhou Bridge provides clear information about the condition of the joints and supports the planning of maintenance activities. It enables damage to be clearly identified based on general testing and teaching of the system. As a result, unexpected damage can be immediately recognized and notified, enabling the timing of replacement of components to be optimized. Whether used during installation, inspection, maintenance or replacement works, or to facilitate assessment of unexpected events or planned modifications, automated monitoring systems are thus sure to be increasingly used in bridge construction and maintenance in years to come. REFERENCES Islami K. et al., 2014, Life-Cycle monitoring and maintenance of Bridges: The role of remote SHM Systems, Istanbul Bridge Conference, Istanbul. Islami K. & Modena C., 2013, Life-cycle assessment by dynamic diagnosis and long term monitoring of old bridges in northeastern Italy, ISHMII, Hong Kong.
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Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
Evaluation of fatigue crack formation in cantilever brackets of a multi-span railway steel box girder bridge L.R. Ticona Melo, R.M. Teixeira, A.P. da Conceição Neto & T.N. Bittencourt Structural Concrete Modeling Group, Polytechnic School, São Paulo University, São Paulo, Brazil
ABSTRACT: This paper presents a rational procedure for evaluation of fatigue crack formation that has been repeatedly detected in cantilever brackets which support the region of the guard rails of a steel box railway bridge. The case study herein considered is a horizontally curved steel box girder bridge composed of 9 spans, with length of 35 m. This bridge is located in the VitóriaMinas Railway, one of the main Brazilian “heavy axle load” (HAL) railroads, mainly used for the transportation of iron ore and general cargo. Recently, the structure received interventions after detection of cracks in the vicinity of welds between the cantilever brackets and the primary box girder. An investigation based on field measurements and finite-element analysis has been performed to identify the causes of these cracks. The purpose of the numerical analysis has been also to provide insight on the structural behavior before and after the bridge rehabilitation. The study suggests that cracking is induced by secondary displacements, which occurs mainly in the regions close to the abutments, where the global rotations are unrestrained. The results and the main conclusions validating the applied methodology are presented in this paper.
1 INTRODUCTION In bridge design several assumptions are adopted in an attempt to simplify the process of determining the internal forces and their final designing. These simplifications, in most situations are conservative, but in some cases the designed system can produce, in service, deformations and secondary stresses that are not taken into account at this stage. Additional stresses occur due to compatibility displacements when the members transfer the load spatially. In general, these secondary deformations are small and the stress associated insufficient to use up the static resistance of member. However, it may result a crack initiation process by fatigue due to cyclic loading, since that this phenomenon is characterized by showing damage in stress levels below the yield stress of the material. Along time, the stress cycling process may result in the appearance of premature failures in relation to the expected lifetime of the bridge. As study case, this paper evaluates the railway bridge over the Córrego do Ouro, which has presented problems related to appearance of cracks in a cantilevered steel substructure that supports the walkway of the bridge. These failures can be attributed to the secondary displacements that arise when the bridge is under loading (trains). In these systems, the lower stiffness of elements that support the cantilever are attached to the steel box girder through welds, subjecting sensitive details to fatigue. The fatigue resistance in these regions is substantially lower due to stress concentrations nearby the welded beads. Due to the numerical modeling capacity achieved in recent years, it is possible to simulate the existence of these secondary deformations more accurately through advanced analysis based on the finite element method. In this research it is presented a procedure to identify, through numerical modeling and field monitoring, the mechanism that generates the secondary deflections, the process of fatigue based on the concept of nominal stress and the performance of the repaired details. 831
Figure 1.
(a) View of the structure and positioning of the monitoring station, (b) Sidewalk cantilevered.
2 BRIDGE DESCRIPTION The bridge over the Córrego de Ouro is located in the Vitória-Minas Railroad, close to Barão de Cocais, a city in the State of Minas Gerais – Brazil. It has 9 spans, each one with 35 m and is composed by a continuous superstructure of steel box girder, supported by reinforced concrete columns of H-type sections. The bridge was designed in 1987, according to the documentation available. The structure is traced in curve, with single track and an approximate radius of 1167 m. The box girder sections are transversally stiffened every 2.91 m approximately having, welded elements in the internal walls of webs. At the extremes of the bridge, the steel box girder is supported by reinforced concrete abutments with the aid of metal bearings. The adjacent walkways are supported by continuous cantilever substructure composed by cross beams attached to the steel box girder through welds and bolts (Fig. 1b). The repair procedure was established and executed after the detection of cracks in these elements. The cracks were detected only in elements of the first and the last spans (V1 and V9), after 22 years in service approximately. 3 INSTRUMENTATION AND MONITORING The monitoring was carried out through instrumentation and data acquisition under normal traffic conditions, collecting up measures of deformations, displacements and accelerations in different points of the structure. The instrumentation was carried out in elements of the superstructure, spans V8 and V9 (Fig. 1a), using access platforms. The strategies of instrumentation and monitoring were previously planned and then updated in the field, based on what had been observed in the preliminary inspection. 3.1 Visual Inspection On the preliminary visual inspection of the bridge it was observed that the handrails of the walkways had fracture-cracking problems in the regions of the welds. The upper beams had been replaced and some welds had been reinforced. Some transverse beams that sustain the cantilever and some consoles that receive these beams and connect directly into web of box girder through welds showed problems related to crack propagation and had been replaced too. The sections of the box girder where these consoles currently exist are spaced approximately 2.91 m. However, before repairs there was cantilevered transverse beams bolted on consoles every 5.82 m, i.e. there was a successive alternation between consoles with transverse beam and consoles without transverse beam along all spans. After repairs, the fractured consoles were replaced and a new transverse beams were introduced in the place of the damaged. Some of these beams were also attached by butt weld in some consoles where they did not previously exist. 832
Figure 2. Transverse sections instrumented (dimension in m).
Figure 3.
Section: (a) S1A (b) S1B and (c) S3 in the support AP9 (dimension in mm).
3.2 Instrumentation In total six sections were instrumented, distributed among the V8 and V9 spans (Fig. 2). The sections S1A, S1B and S3 were planned to evaluate the levels of stress in the cantilever brackets on the north side of the bridge (Fig. 3). Due to the effect of stress concentration in the vicinity of the weld beads it was used an appropriate instrumentation technique with the purpose of capture partially the geometric stress gradient (hot-spot stress) in the details of the structure. In Figures 2 and 3, E is used to designate sensor of deformation, DV to identify displacement transducer and Ac to identify accelerometer. In section S1A, the sensors E21, E22 and E23 were positioned to capture the deformations in the console, near the weld beads that connect this element to the side wall of the box girder. It is emphasized that the console is original and after repairs cantilever beam through a butt weld (Fig. 3a) has been introduced. For evaluating the stress levels in this new welding strain gages E24, E25 and E26 were positioned. In the section S1B (P8 column) the strain gages E27, E27A, E28 and E29 were installed to evaluate the stresses in the connection of the substituted console. At section S3, the sensors E31, E32 and E33 were positioned to catch the effect of the original weld on cantilever, where it has not been reported the existence of cracks neither repairs were made. In all the cases, the sensors were installed perpendicular to the direction of the welded bead. The characteristics of trains during the monitoring period are shown in Table 1. The types of some compositions vary from freight trains with platform wagons, bulk carrier trains with hopper wagons, iron ore unloaded train with GDE-type gondola wagons, locomotives in pairs and passenger trains. In all the cases, the observed locomotives were DASH 9W type. 833
Figure 4.
Railway vehicles: (a) Locomotive DASH 9W, (b) Gondola wagon and (c) Hopper wagon.
Table 1. Loading characteristics for different HAL vehicles.
HAL Vehicle
Maximum Gross Rail Load (kN)
Lightweight (kN)
Load Limit (kN)
Load/Axle P (kN)
Track Loading Density (kN/m)
DASH 9W GDE type A GDE type B GDE type C HFE
1,600.00 960.00 1,000.00 1,100.00 1,000.00
– 165.00 165.00 165.00 230.00
– 795.00 835.00 935.00 770.00
200.00 240.00 250.00 275.00 250.00
69.32 98.46 102.56 112.82 58.62
Figure 5.
Numerical model: (a) Overview (b) Before the repairs and (c) After the repairs.
4 FINITE-ELEMENT ANALYSIS The bridge was modeled in three dimensions to provide accurate representations of the global stiffness and the occurrence of secondary deflections and deformations. The degrees of freedom of the supports were modeled by imposing nodal restrictions such that, on the east side rotations and translations (longitudinal and transversal) were allowed. The support of the other extreme (west side) had their rotations released too, but the translations were restricted. For purposes of computing requirements and savings processing time, the elements of the substructure and the permanent track were not modeled. The assumption of linear elastic behavior was adopted since the study is aimed to examine only the response of the bridge under service loading. In total, two numerical models have been made in order to reproduce the behavior before and after repairs (Fig. 5). In the numerical modelling, the steel elasticity modulus was considered 210 GPa. In each model, the analysis simulating the pass of two DASH 9W locomotives to 42.4 km/h was carried out in order to represent the train composition observed during the monitoring. In the experimental case, so that there was validity of the comparisons with the numerical model, we considered only the results 834
Figure 6.
Numerical model that show the distortion of the cantilever brackets.
corresponding to the passages of the first two locomotives when entering into the bridge, situated in front of the train and traveling in the direction Vitoria-Belo Horizonte (Fig. 2). The selection of this train composition, for purposes of comparison with experimental results was due to a reduction of the computational cost analysis. The analysis procedure was based only on quasi-static response of the structure, excluding the dynamic effects generated by the passage of the train. 4.1 Comparisons between numerical and experimental results It was concluded that, in general, the behavior of the structure modeled numerically represents satisfactorily the behavior of the real structure. The computational model also simulates very well the behavior of the cantilever brackets observed experimentally. The numerical answers in the regions of the instrumented points on the transverse beams, in the same way as observed experimentally, suffer sign reversal insofar that the locomotives advance over the neighboring spans. The model is able to capture the distortion suffered in the cantilevers, which can induce a process of crack formation in the main points of concentration of tensile stresses that exist nearby the weld beads. Figure 6 shows a scheme emphasizing one of the distorted transverse beams when the span is loaded by the vehicle used as reference. The magnitudes of these tensions are intensified due to the geometrical intersections and existing weld beads in the connections of the cross beams with longitudinal sections and with the web of the box girder. These regions have potential to promote the fatigue damage process. The secondary deformations are generated by rotation of the longitudinal members. The arrangement of the structural components causes large amplitudes of these rotations nearby the supports in the extremities of the bridge, as shown in the Figure 6. Thereby, it was observed that the cross beams are more stressed in regions adjacent to abutments and pillars, which may explain the location of the damage indicated during the inspection of the bridge. In order to have a better understanding of the behavior of the elements of the walkway supports, the numerical normal stresses were determined at the same points as those instrumented in the real structure; the results are shown in Figure 7. As illustrated, the model represents adequately the behavior of the members along the crossing of the train composition. The discrepancies between the magnitudes of numerical and measured stresses can be attributed to simplifications in the sections of some components (as example the longitudinal beams), the existence of weld beads not represented in the model and the level of mesh discretization in the finite elements. It is emphasized that these simplifications were necessary to make possible the analysis due to the high computational cost that other refinements would generate. In general, the stresses calculated in the numerical model are greater than those detected in the monitoring. For future investigations, the use of submodeling techniques can be the next step to improve estimates of local stresses. The major rotations in the extreme spans are directly reflected in the levels of tensile stresses of the cantilever brackets. For example, when comparing the numerical tensions in the point E27B located 835
Figure 7.
Local stresses in transverse beams: (a) Numerical and (b) Experimental.
Figure 8. Comparison of numerical stress: (a) Between two points in V8 and V9 and two points in V9, section S1B and S3, (b) Before and after repairs.
in the spanV9 (Point 1, Fig. 8a), with the stresses calculated in the same place and element in the span V8 (Point 2, Fig. 8a), it is observed a reduction in the tensile solicitation. This reduction in the stress level would imply a longer service life for members of the cantilevers located in the intermediate spans (V2-V8), and therefore a lower probability of failures in these regions. Indeed this was verified in the preliminary inspection of the structure. Another indicator that justify the incidence of damage essentially on adjacent spans to the abutments is the occurrence of highest tensile stress levels when it is compared, through the numerical results, a point in one of the support beams where cracks were detected before the interventions and another point with the same detail type, in which the failures were not detected visually. This study is illustrated in Figure 8a (S1B-V9, S3-P8). Trying to detect the stress distributions in some points before and after the interventions on the bridge, a brief numerical study was done whose results are summarized in the Figure 8b. According to the results, the model which simulates the condition after interventions shows a decrease in the stress, which probably will prolong the lifetime of the details relative to the previous conditions repairs. However, the introduction of new beams, attached by butt welds to existing consoles, created other points of stress concentration (e.g. the region where the sensor E24 was installed). The stress levels at those points (see Fig. 7) may be sufficient to initiate in the future, a damage process by fatigue. 5 FATIGUE DAMAGE ANALYSIS In order to evaluate the fatigue performance of the structure after interventions, an analysis procedure based on methods of nominal stress and S-N curves (AREMA, 2012) was developed. The 836
Figure 9.
Details adopted to evaluate the fatigue in the weld points.
hypothesis of Palmgren-Miner was assumed to estimate the damage accumulation (Schijve, 2009). For this study, the determination of the stress ranges was based on the monitored deformations. In the calculation of these deformations, the secondary effects as those discussed above are also included. The cycle counting was carried out using the Rainflow Algorithm (ASTM E 1049, 2010), for characterize the load spectrum. The information about the load characteristic repeatability, whose adoption is necessary for the next verification, were obtained during the signal processing of trains observed during the monitoring. 5.1 Classification of the analyzed details The majority of the current codes are based on the classification method that employs the S-N curves together with tables of the detail categories. Despite the advantages of this method, there are several limitations once that the S-N curves are only available for a limited number of details, so that it is not possible to cover a wide diversity of details that can be found in the steel bridges systems. Furthermore, there are no references for cases involving fatigue induced by distortion. For the specific case of the Córrego do Ouro bridge, the sensitive points to fatigue existing in the walkway substructure elements does not exactly match any detail described in the tables recommended by the main international codes such as the AREMA (2012), EUROCODE (2005) and AASHTO (2011). Furthermore, it should be considered that the geometry, the stress distribution, the way of execution, etc. can also change the basic category to be adopted. In this sense it was adopted a classification procedure whose analysis was based on the following considerations: – The superstructure of the bridge has several types of sensitive details to fatigue, but the analysis is directed only to cantilever elements that was instrumented. No consideration was made about the internal components of the box girder since it was not possible to be instrumented and inspected. However, the information available and the preliminary inspection carried out indicated that the bridge is generally in good condition. – The classification adopted for the type of weld in every detail of the guard rails regions, took the project data and the visual inspection as a reference, following the schemes shown in the Figure 9. In the point E21 it was considered fillet type weld, and in the point E27B it was considered butt welds with chamfer on one of the sides. This consideration was based mainly on the project data, since the nature of the weld penetration is difficult to recognize only by visual inspection. – The details in the points E21, E24 and E27B were classified according to the recommendations of AREMA Manual (2012) for welded joints transverse to the primary direction of stress. The E21 point was classified as category “D” and the fatigue limit was set considering a lower penetration scenario for the fillet weld type. Already the E24 and E27B points were classified as category “C”. In all the cases, the stresses generated by the transversal beam distortions are treated as primary. In the classification of the E24 detail, it was considered that the weld surface on the top connections was not treated (reinforcing not removed) according to the observations 837
verified during the visual inspection. This reduces one category in relation to the similar case where reinforcement of the weld is removed. – The detail E31 is not explicitly classified in the consulted codes. It may be close to the “Base metal at short attachments” and the category “E” classifications, by considering that the transverse weld practically does not offer transition radius; that was verified during the inspection. It is noteworthy that the term “attachments” used in this classification is defined as any welded detail in the aim member, which by their mere presence and regardless of their loading, causes a discontinuity in the flow of tensions, thus reducing their fatigue resistance (AASHTO, 2011). Moreover the term “short” is related to the length of the outbuilding in the direction of the tension in the member, represented in this case by the width of the lower flange of the longitudinal beam. It is impossible to say whether the severity levels of the detail in the point E31 is greater or less compared to which is presented by category “E”; actually there are many variables that can interfere significantly in the fatigue resistance. 5.2 Procedure evaluation according to AREMA (2012) The fatigue resistance of components of steel bridges according to AREMA Manual is based on specific S-N curves for each detail class. The AREMA code also provides recommendations about the use of the constant amplitude fatigue limit CAFL, which represents the stress under which the associated damage can be considered negligible in the case of fatigue under constant amplitudes of stress, but suggests the extension of the S-N curve for cases that involves high cycle fatigue with variable amplitude of tension, wherein at least 0.05% (classes A to D) or 0.01% (classes E and E’) of the tensions spectrum exceeds the CAFL. However, to avoid the underestimation of the fatigue life, it is recommended the adoption of a lower limit for truncation of this spectrum. The recommended value for this cut-off limit is 0.5CAFL for details classes A to D and 0.25CAFL for details E and E’. When the maximum tensions of spectrum SMAX are below the CAFL, no crack fatigue is expected for the detail and the infinite life hypothesis can be assumed. The ratio µfat = (CAFL/γMf )/(γFf · SMAX ) provides a preliminary measure of the safety factor for the infinite fatigue life. In our investigation, it is considered γMf = γFf = 1. The number of iterations of the loading group which exhausts the resistance is a function of the average daily traffic ADT. The extrapolation to estimate the annual spectrum, in this study is made linearly assuming that the ADT remains constant during the lifetime of the bridge, this hypothesis neglects the random effects of loading along time. 6 EVALUTION OF RESULTS The Rainflow counting cycles technique was carried out in the traffic samples and the spectra of tension for each detail analyzed are shown in the Figures 10–11. Since the very small cycles produced by vibratory movements and noises do not effectively contribute to the total damage, the counting algorithm was programmed to ignore cycles with amplitudes less than 3.45 MPa (Frangopol & Kwon, 2010). The results shown in the Figure 10 suggest an improvement in the fatigue performance of the details E21 and E27B after interventions. With the reduction of stresses, these points have nominal tensions far below of the CAFL for each class adopted. Therefore satisfies the verification, so that the damage expected in these details can be considered negligible in the new configuration. It is noteworthy that the stress levels in the previous condition were probably greater in these points (conclusion observed in the numerical modeling, comparing the states before and after repairs) and enough to overcome the CAFL and provide a finite life to the details, the failures detected in the console corroborate this situation. In relation to the detail E24 (Fig. 11) introduced after interventions, it is verified that the maximum tension is very close of the CAFL of the category adopted with coefficient µfat around the unit (Tab. 2). It indicates that this detail and others, similars in the adjacent spans to the abutments, will probably present problems of formation of fatigue cracking in the future. It is important to say that, as there is no clear definition about the 838
Figure 10.
Spectrum of ranges of tensions in the points E21 and E27B.
Figure 11.
Spectrum of ranges of tensions in the points E24 and E31. Table 2. Verification of the safety to fatigue. Point
Location
Detail category
SMAX (MPa)
CAFL (MPa)
µfat
E21 E24 E27B E31
section S1A section S1A section S1B section S3
D C C E
15.21 67.87 12.85 152.81
48.30 69.00 69.00 31.00
3.18 1.02 5.37 0.20
detail category, there is a level of uncertainty inherent to evaluate with accuracy the factor µfat that in some probable scenarios may be less than unity, increasing the probability of early crack formation. The detail E31 has the original configuration features and although not presenting cracking signs at the moment of the interventions on the bridge, the verification carried out shows that there is potential to develop the damage process by fatigue in the future. According to the verification based on the nominal tension measured and the AREMA class adapted, the value of the coefficient µfat is much lower than unity, indicating that a finite fatigue life is expected at this point. In order to evaluate the magnitude of lifetime for this detail a prospection was made, creating a reference scenario in which the monitored traffic sample is repeated two times a day (ADT = 2) and remains constant over the years. Thus, the expected lifetime would be approximately 30 years, which exceeds the age that the structure had when the cracks appeared in the cantilever brackets. If the estimate of the fatigue life idealized for detail at the point E31 was extended for a similar detail in the spans V8 before intervention, it would probably result in a lower life expectancy, which would justify the incidence of the failures detected in these regions. This conclusion is backed up by the results of the numerical model of the original configuration (Fig. 8a), which presented higher tensile stress levels when comparing the detail E31 with the corresponding details in the transverse beams of the guard rail located in the V8 span. 839
7 CONCLUSIONS Fatigue cracks in some cantilever brackets have been detected in the bridge over Córrego do Ouro. It is known that similar failures have been detected in other bridges with similar configurations. In 2012, actions were taken to correct the deficiencies. During the bridge repair procedures a monitoring campaign was carried out to investigate the causes of the failures of the original structure and the performance of the structure after intervention. A numerical analysis of the bridge has been developed in order to support the analysis and to determine the most likely locations where the process of crack formation could start. Based on the main results of this study the following conclusions are obtained: – The structural configuration of the bridge, steel box girder section with guard rails supported by welded crossbeams, allows the appearance of secondary displacements in the cantilevers, imposing components of flexure and torsion in these members. The secondary displacements are generated by the global rotation of the box girder. The amplitudes of these rotations are higher in adjacent spans to the abutments. The numerical models reflected adequately this behavior, which was previously obtained experimentally. When comparing the conditions before and after interventions, the numerical results showed stress reliefs at some points. This indicates possible improvement of the fatigue performance relative to the previous condition. – The existence of welded details in the cross beams generates stress concentration that may contribute to crack forming processes. The weld configuration can reduce significantly the fatigue resistance of these points. In order to evaluate the performance of the details, fatigue verifications based on AREMA code have been held. These verifications showed that some details may present a finite life in the new configuration, since the maximum tension exceeds or is near to the fatigue limit (CAFL) in some analyzed points. – According to this study, considering that the weld configuration has significant effect on fatigue resistance, it was not possible to detect whether the degree of severity is higher or lower than the one presented by the resistance class adopted. In the case of the console connections, there is a possibility of failure in the solder material (failure root) due the partial penetration depth of the fillet weld type or the butt weld and the thickness of the plates. This scenario considers the smallest fatigue resistance among the classes recommended by EUROCODE 3 (2005). Other factors associated with the internal defects of the weld beads may also have significant influence in this process. REFERENCES AASHTO. 2011. Manual for Bridge Evaluation. American Association of State Highway and Transportation Officials, Washington, D.C. AREMA. 2012. Manual for railway engineering. American Railway Engineering and Maintenance-of-Way Association, Washington, D. C. ASTM E 1049. 2010. Standard Practices for Cycle Counting in Fatigue Analysis 1. Annual Book of ASTM Standards, 85 (Reapproved 2005): 1–10. Bittencourt, T.N. 2014. Monitoração da ponte sobre o Córrego do Ouro. Relatório Final. Connor, R. J. & Fisher, J. W. 2006. Identifying Effective and Ineffective Retrofits for Distortion Induced Fatigue Cracking in Steel Bridges using Field Instrumentation. Journal of Bridge Engi neering, ASCE, 11(6): 745–752. Eurocode. 2005. Eurocode 3: Design of Steel Structures—Part 1–9: Fatigue. CEN. Kwon, K., & Frangopol, D. M. 2010. Bridge fatigue reliability assessment using probability density functions of equivalent stress range based on field monitoring data. International Journal of Fatigue, 32(8): 1221–1232. Naito, C. J. Asce M, Li X, Hodgson L.C. & Yen B.T. 2013. Fatigue Crack Formation and Repair Strategies for Steel Cantilever Bracket Tie Plates. Journal of Bridge Engineering, ASCE: 516–524. Schijve, J. 2009. Fatigue of structures and materials. Fatigue of Structures and Materials. The Netherlands: Springer Science: 1–622.
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Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
Investigations of post tensioned bridges with critical prestressing steel regarding hydrogen induced cracking (HIC) A.W. Gutsch & M. Walther Braunschweig Civil Engineering Materials Testing Institute (MPA BS), Institute for Building Materials, Concrete Construction and Fire Protection (IBMB), Technical University of Braunschweig, Germany
ABSTRACT: In Germany 120,000 road bridges are in use. 2,000 of them have intensively to be checked regarding their capacity for the actual higher traffic loads. This group of bridges also cover bridges built from 1960 to 1965 (old type of potential critical prestressing steel) and 1965 to 1978 (new type of potential critical prestressing steel) and potentially critical prestressing steel regarding hydrogen-induced cracking (HIC) was used. The paper presents the procedure for investigation of bridges with potentially critical prestressing steel. In the laboratory the prestressing steel was tested regarding cracks. The test methods were magnetic particle inspection (MPI), microstructure investigation, strength test and optionally micro computer tomography (exanimation of the surface).
1 INTRODUCTION In Germany there are about 120,000 road bridges. Due to consistent increase of traffic loads in the last decade, the bearing capacity of about 2,000 of them has to be carefully checked. In this group, several of the bridges were built between 1960 and 1978. During this period the new construction type of prestressing steel was used in many bridges. Already in the construction phase the first damages on the prestressing steel occurred. Only a few hours after finishing the tensioning of the tendons, single wires failed without advance notice, still before the ducts had been injected with grouting mortar. In other cases wire breakage was detected several years after completion and designated use. The analyses of several cases of damage revealed that the wires and strands had collapsed due to hydrogen-induced stress corrosion cracking (HIC). After a closer look to the observations, single manufactures and batches could be named. Today the prestressing steel used in those days is classified into two groups: the old type of potentially critical prestressing steel produced between 1960 and 1965 and the new type of potentially critical prestressing steel produced between 1965 and 1978. Wires for prestressing steel were mostly produced with an oval, or less often also with a round, cross-section, of 30 mm2 to 40 mm2 surface area. The critical prestressing categories are generally classified to the strength class St 145/160 (new classification St 1420/1570) [Lingemann, Leonhardt]. In this paper the procedure and the investigation methods for prestressing steel of bridges concerning their sensitivity for hydrogen-induced stress corrosion cracking (HIC) will be shown. 2 VERIFICATION AND RECALCULATION OF EXISTING BRIDGES The Federal Ministry of Transport, Building and Urban affairs has published two guidelines [Bundesministerium] for the verification and recalculation of old bridges with potentially critical prestressing steel. These guidelines describe, among other things, the procedure how to extract specimens out of bridges and also describes methods for the subsequent analysis of the structure. 841
At first the bridge, its load bearing capacity and the service limit state is to be recalculated under assumptions of the existing geometry, materials, material parameter and admissible loads. Planning and construction documents are essential for this calculation. The verification of the recalculation is structured in 4 steps: 1. Arithmetical verification based on the valid standards (DIN-technical reports 102 – 104, DIN EN 1992 – 1994), 2. Arithmetical verification with additional regulations of the guidelines, 3. Further verification, taking the results of the structural investigation of the building into consideration (concrete compressive strength, real reinforcement content, deformation through loading tests), 4. Calculation with the help of scientific methods. Structural analysis and verification of the load-bearing capacity using probabilistic methods to determine the probability of failure. If the results of the recalculation of the bridge do not show significant prior signs for possible failure, it is recommended to analyse the prestressing steel as to its sensitivity to hydrogen-induced stress corrosion cracking. For this purpose samples of prestressing steel have to be taken from the building. Usually the structural engineer determines areas, where prestressing steel can be taken with a minimum of impairment to the structure of the bridge. 3 LOCALIZATION OF PRESTRESSING STEEL WITH GEO-RADAR AS NON DESTRUCTIVE TEST METHOD (NDT) The first difficulty in analysing the structure of old buildings is often the lack of planning documents or even the fact that real constructions differ from existing planning documents. Due to this fact it is recommended to investigate the points for extraction of the prestressing steel prior to the extraction itself through non-destructive testing (ndt) to localise the reinforcement and the tendons with the prestressing steel. The outer layers of reinforcement can be localised by magnetic inductive methods. This method is not suitable for the localisation of the prestressing tendons steel which are applied in the inner sections of a cross-section with a minimum of 10 up to 20 cm from the surface. Several inspections have shown that the tendons and the prestressing steel can be localised with radar waves (geo-radarmethod) or by ultrasonic-wave-methods under a concrete cover of 30 cm, Figure 1. The different
Figure 1.
Detection of prestressing steel with geo-radar.
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non-destructive test methods are frequently used for various inspections and investigation projects of the MPA Braunschweig/Braunschweig Civil Engineering Materials Testing Institute on bridges and buildings. 4 SAMPLING OF PRESTRESSING STEEL AND GROUT FROM THE TENDONS OF THE BRIDGE DECK After localising the positioning of the tendons and the prestressing steel the structural element can be opened carefully. Any unnecessary structural injury of the reinforcement, tendons and prestressing steel must be avoided. Usually the opening is realized with a pneumatic chisel, or if more caution is necessary, a high-pressure-water-jet is used, Figure 2. When the tendons become visible in the opening of the structure, the duct has to be opened carefully and the grout has to be taken out of the tendon and packed airproof for further investigation, e.g. regarding the chloride-content of the grout due to de-icing-salt. In the next step samples of the prestressing steel (wires, strands) can be extracted. Before cutting the prestressing steel, strain-gauges can be placed on the prestressing steel to determine the strain deformation due to stress loss during the cutting process. In this way the remaining and actual stress in the structure (wires) can be determined and compared to the initial stress range. With the difference between the initial prestressing stress range and the actual remaining stress the relaxation effects of the prestressing steel and the creep effects of the concrete and their effects on the remaining prestressing stress can be estimated, Figure 3.
Figure 2. Visible tendons/sheath (top) and strands after opening of the sheath and removing the grout (bottom).
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Figure 3.
Measurement of the strain while cutting the strand.
This comparison enables the evaluation of relaxation effects of the prestressing steel and the creep effects of the concrete for the previous service life time.
5 LABORATORY TESTS The analysis on the prestressing steel regarding its sensitivity concerning hydrogen-induced stress corrosion cracking is done in the laboratory. At first the extracted specimens are inspected visually for signs of corrosion. When the prestressing steel is well protected by the grout and no local cracks and leakage water has affected the tendon and the prestressing steel, the surface of the prestressing steel is usually only covered by a thin amount of an uncritical rust coat. For more detailed crack-investigations with the magnetic powder procedure the surface of the prestressing steel has to be cleaned with a soft mechanical treatment. For the testing procedure the specimen is first magnetised and then showered with a test solution. The test solution consists of magnetic powder and fluorescent colour particles. Around cracks close to the surface, magnetic stray fields develop and the magnetic powder conglomerates in the crack areas. The cracks become visible under ultraviolet light from the added fluorescent particles to the solution. If the specimens have been damaged during the sampling out of the bridge, the damaged areas can also be made visible by the magnetic powder procedure. In this case a differentiation between a mechanical damage due to sampling and a local cracking has to be determined through a further test analysis. If questions remain concerning the sensitivity of the prestressing steel for hydrogen-induced stress corrosion cracking after the magnetic powder test, the sample can be analysed by the x-ray computer tomography method, as has been done in recent projects of the Braunschweig Civil Engineering Materials Testing Institute. The surface of critical areas can be closely examined and analysed three-dimensionally with an optical resolution of about 30 µm. With the help of such a computer tomography the differentiation between cracks and mechanical damage is easily made, Figure 4. 844
Figure 4.
Picture from x-ray computer tomography of a prestressing steel taken from a bridge.
Figure 5. Thin section with a layer of ferrite, overview (left) and detail (right).
The micro-steel-structure can be examined with a light optical microscope on thin sections of the specimen to detect micro-failures that might occur during the tempering process. Visible cracks on the surface usually result from the production. The cracks are mostly detected in the oxide layer. In the magnetic powder procedure they are not displayed. These cracks form a predetermined breaking point in the tensile test, but do not have a negative effect on the strengths and deformation parameters of the prestressing steel. Due to a missing compensation process a fine layer of ferrite can develop over the whole surface. This layer can be neglected where the strength and deformation parameters are concerned, Figure 5. The material parameters (yield stress, strengths and deformation) are determined in a tensile static load test. The German Federal Ministry of Transport, Building and Urban affairs defines a value for the critical tensile strength of 1,700 N/mm2 . Beyond this strength value a heightened susceptibility of hydrogen-induced stress corrosion cracking is indicated [Bundesminister, Bundesanstalt fuer Wasserbau]. If the analysis and test procedure up to here indicates cracks or a heightened susceptibility to hydrogen-induced stress corrosion cracking, additional corrosion tests according to DIN EN ISO 15630 can give further information. The test set-up is shown in Figure 6. 845
Figure 6. Test set-up for corrosion tests on prestressing steel.
REFERENCES Bundesminister für Verkehr, Bau und Stadtentwicklung (2009), Handlungsanweisung zur Überprüfung und Beurteilung von älteren Brückenbauwerken, die mit vergütetem, spannungsrisskorrosionsgefährdetem Spannstahl erstellt wurden, Germany Bundesminister für Verkehr, Bau und Stadtentwicklung (2010), Stochastische Untersuchung von Spanngliedausfällen bei Brückenbauwerken mit spannungsrisskorrosionsgefährdetem Spannstahl, Forschung Straßenbau und Verkehrstechnik Heft 1049, Germany Bundesminister für Verkehr, Bau und Stadtentwicklung (2011), Richtlinie zur Nachrechnung von Straßenbrücken im Bestand (Nachrechnungsrichtlinie), Germany Bundesanstalt für Wasserbau (2006), Spannungsrisskorrosion von Spannstählen, BAW-Brief Nr. 3, Karlsruhe DIN EN ISO 15630, Steel for the reinforcement and prestressing of concrete – Test methods – Part 1: Reinforcing bars, wire rod and wire, 2011 DIN EN ISO 6892, Metallic materials – Tensile testing – Part 1: Method of test at room temperature, 2009 Leonhardt, F. (1955), Spannbeton für die Praxis – Zusammenstellung von gebräuchlichen Spannstahlsorten in der Bundesrepublik Deutschland, Verlag Wilhelm Ernst & Sohn, Berlin Lingemann, J. (2010), Zum Ankündigungsverhalten von älteren Brückenbauwerken bei Spannstahlausfällen infolge Spannungsrisskorrosion, Technische Universität München
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Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
Fatigue management of the midland links steel box girder decks C.R. Hendy Atkins, Epsom, UK
S. Chakrabarti Independent consultant, UK
ABSTRACT: The Midland Links Motorway Viaducts carry the M5 and M6 motorways around Birmingham. A number of the longer spans comprise multiple steel and concrete composite box girders. Assessment showed areas of very high overstress in the web plate at the locations of internal bracings. These overstresses affected ultimate, serviceability and fatigue limit states and arose because of poor detailing of the cross braces and stiffeners. Plastic redistribution was considered to demonstrate adequate ULS performance and an analysis was undertaken with potential fatigue damage modelled to prove the structure would not collapse with such damage at the ultimate limit state. In-situ strain monitoring was undertaken both under known test vehicle loading and under normal traffic conditions to derive more accurate fatigue stress spectra. These results were used not only to calibrate the finite element results but also to improve the predicted fatigue life and allow preparation of a long-term strategy for managing the viaducts.
1 BACKGROUND TO THE PROJECT Typical steel-concrete composite box girder spans on the Midland Links Motorway Viaducts are shown in Figure 1 with a general arrangement of the cross section and internal diagonal bracing shown in Figure 2. All steel is mild steel; the web and top flange are in accordance with BS 15 (1961) and the bottom tension flange is in accordance with BS 2762 (1956) (Type Notch Ductile IIB). The box decks were initially analysed using grillage models. The end diaphragms were found to be overstressed under the torque produced at the rocker bearings but after the torsion constant for individual load cases was softened in the analysis, to allow for the distortional flexibility of both the box cross section and the internal bracing, this overstress was eliminated. A method for softening the torsion constant in this way is given in Design Guide for Composite Box Girder Bridges (1994).
Figure 1. Typical steel-concrete composite box girder spans.
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Figure 2. Typical box girder cross section.
Figure 3. Bending moment induced in web at bracings due to distortional forces.
The assessment however predicted areas of very high overstress in the web plate at the locations of internal diagonal bracings. This remaining overstress at bracings affected ultimate, serviceability and fatigue limit states and arose because of poor detailing of the bracing and stiffeners. The stiffeners terminate short of the flanges and thus distortional forces acting on the box cause a high local bending moment in the web between stiffener and flange (Fig. 3). The problem is exacerbated by the presence of a fillet weld connection between web and flange giving rise to a Class W fatigue detail in the weld and Class F detail in the web in accordance with BS 5400 Part 10 (1980). Since neither adequate strength nor service life could be demonstrated by using conventional assessment, despite the fact that the viaducts showed no signs of distress, further investigation and assessment was required to ensure safe operation. 2 ACOUSTIC EMISSION MONITORING OF THE VIADUCTS The critical detail for fatigue just above the sealing weld (Figure 3) was located inside the box girders. Since there was no access provision to the box girders it was necessary to find a way of inspecting and monitoring the boxes from the outside. A programme of acoustic emission monitoring (AEM) and ultrasonic inspection was therefore undertaken. There were indications that fatigue activity might be occurring at the bracing locations. However, ultrasonic investigations found no signs of cracks and subsequent visual inspection (by entering the box through newly formed access hatches) found lengths of missing sealing weld dating back to the original construction but no cracks. This cast doubt on the suitability of AEM to locate fatigue defects in a large steel viaduct such as the Midland Links. It clearly detected increased activity at the fatigue-prone details; the emission hot spots coincided with the areas of predicted overstress in the structural analysis. However the AEM was a poor predictor of actual fatigue damage when compared with visual inspection and ultrasonic inspection. 3 VERIFICATION OF ULTIMATE LIMIT STATE STRENGTH 3.1 As-built analysis It was clear that the use of grillage analysis was too crude to investigate this particular problem properly; the bracing had been included in the grillages, but only as shear flexible members. The critical boxes were therefore remodelled using shell finite elements. Complete spans (comprising ten or more box girders and concrete deck slab) were modelled. Beam elements were typically used for the stiffeners and bracing members but the most heavily loaded bracing locations in each box were modelled using shell elements for the stiffeners and bracings and a finer mesh was used in the web in the area of peak stresses at the ends of the stiffeners. An extract of one fish-bellied girder from a typical model is shown in Figure 4, together with a view of the detailed mesh at one of the bracing frames being investigated. 848
Figure 4.
Layout of typical bracing bay in finite element model showing detail at chosen cross brace.
Figure 6. Influence surface for maximum axial force in typical diagonal bracing in a girder on Bescot Viaduct deck.
Figure 5. Typ. vertical stresses in web at stiffener.
3.1.1 Ultimate limit state Stresses at the ultimate limit state were shown by linear elastic finite element analysis to be well in excess of the yield stress of 247 MPa (up to 450 MPa) in the webs as a result of the distortional stresses attracted to the web by the diagonal bracing – see Figure 5. The extent of the material with stresses exceeding yield along the box was very small; only around 100 mm. With plasticity in the bracings or at the web-stiffener connection, these high stresses could potentially be shed safely or redistributed to a larger zone; it was therefore suspected that ULS would not be a governing limit state. In order to verify ULS performance, two separate checks were performed. First, the yielded material was removed and cross-section checks repeated. Second, an analysis was performed removing the cross bracings altogether and linear elastic analysis repeated. This approach was preferred to a full non-linear analysis because of the time involved in conducting such an analysis on such a large model. 3.1.1.1 Removal of yielded material It was calculated that the distortional moment in the part of the web plate where yield was exceeded could be carried by a length of web plate equal to just over 200 mm acting at yield. This was estimated simply by calculating the moment carried by the yielded portion and redistributing the moment over the length required to attain equilibrium with the material stress limited to yield. This calculation ignored the reduction in distortional moment that would arise due to the reduction in stiffness caused by the yielding material. To verify that the cross-section as a whole was adequate for global effects, a reduced effective cross section was used in global stress checks, ignoring the yielded portion of web entirely in the cross section properties. The effects of secondary bending around these idealised holes in the cross section were also considered. This showed that the cross section overall remained adequate. 849
3.1.1.2 Cross bracings removed If excessive yielding were to occur in the web between stiffener and flange, the cross bracing could be rendered ineffective because of the loss of stiffness. To investigate this, the analysis was repeated removing the cross bracings completely to check whether the box girders would still have adequate strength to resist the applied loading. Removal of the cross bracings led to an increase in distortional flexibility which significantly reduced the distortional stresses in the web adjacent to flange and stiffener; the distortional stresses reduced from 450 MPa to just over 100 MPa at ULS. The reduced torsional stiffness however also increased the box girder midspan sagging moments by up to 5% because of the reduced transfer of loads between the boxes. It was therefore necessary to verify that the boxes remained adequate with this moderate increase in moment. The strength of the concrete slab was also checked as the reduced torsional stiffness of the boxes increased the transverse bending moments on the concrete deck. There proved to be adequate reserve of strength in both elements. The conclusion from these two analyses was that the boxes had adequate ultimate limit state strength. 3.1.2 Serviceability limit state The stresses in the web plate at SLS also exceeded yield on Bescot and Oldbury. As serviceability “failures” do not threaten the integrity of the boxes and as there was no sign of actual distress in the corners of the box (for example disruption of the paint), the SLS predicted failures were not considered to be of concern on their own. It was also considered likely that in-situ stresses would actually be much lower than predicted in the analysis because of additional joint flexibility in the bracings that was not modelled and because of composite action with the deck surfacing. The apparent overstress was eliminated when the scale factor from in-situ strain measurement, determined from the load testing described in section 4.1, was applied to the calculated SLS stresses. These findings led to the conclusion that the serviceability performance of the boxes was adequate. It is also noted that it is not normal practice to assess bridges at the SLS in any case. 3.1.3 Fatigue limit state – damaged condition and damage tolerance Fatigue was however of greater concern. Stress ranges determined from the initial grillage analyses were found to be extremely high and the stress ranges determined from the subsequent shell finite element models were only moderately reduced from these values. Figure 6 shows the influence line surface plot for the maximum force in one typical diagonal bracing member; the force is proportional to the distortional stress in the web. This distortional stress range in the vertical direction was far greater than the longitudinal stress range from main beam flexure which, whilst included in the calculation, was relatively insignificant. The plot shows that large influence surface ordinates only occur in areas in close proximity to the cross frame in question. The influence of other vehicles remote from the bracing being considered is small. The estimated fatigue lives for the deck with and without the diagonal bracing was calculated at the locations shown in Figure 2. The case without bracing was investigated by way of a potential “strengthening” scheme as the work in section 3.1.1 showed that the boxes had adequate ULS strength without them. These results were produced using the fatigue method of clause 8.3 of BS 5400 Part 10, which uses a single vehicle to replicate the damage from the real spectrum of traffic. Without bracing, the box had a fatigue life in excess of 120 years. The lowest fatigue life predicted with bracing was just less than 1 year in the web at the toe of the web-flange sealing weld. The fatigue spectrum approach of clause 8.4 of BS 5400 Part 10, where the bridge is traversed by a spectrum of vehicles of different weights and annual numbers, was then used as a refinement; the vehicles were applied in turn and the damage summed using Miner’s rule. The predicted fatigue life based on stresses from the FE model and the Part 10 spectrum was 5 years. It was considered that the spectrum was representative of the Midland Links motorways with little, if any, in-built conservatism. A proposal was however made to collect actual traffic spectrum data during subsequent load testing – see section 4.2. 850
3.1.3.1 Damage tolerance Although it was possible to demonstrate adequacy of the boxes at the ultimate limit state if plastic redistribution was allowed for, the boxes could not be shown to have adequate fatigue life and therefore the plastic redistribution assumed might not be able to occur if a fatigue crack was to first result in fracture. As a result, the consequences of a propagating fatigue crack were considered. A full height crack was introduced in one web of a box. This was done to investigate if the bridge could remain standing, albeit with plastic damage, should one fatigue crack form at a bracing location and propagate in the web until finally tearing through the full height of the web by brittle fracture. The crack was represented as a gap in the mesh over the full height of the web. A vertical crack was placed in the model at the support to investigate the worst shear case. In a separate model, a crack was added at midspan to investigate the worst bending case. 45 HB and HA loading was considered and the live load effects were combined with the dead and superimposed dead load effects and the sections checked to BS5400 Part 3 (2000). The situation was analysed elastically, rather than non-linearly, which is conservative because no allowance was made for plastic redistribution between boxes after the fatigue crack had formed. Web buckling and yielding checks were carried out for the worst bending and shear cases. In checking strength, all load and material factors were removed which is consistent with the accidental combination used in BS EN 1990 (2002) for extreme events, which reduces all material factors to 1.0 except that for concrete which is reduced to 1.2. It was found that the bridge could just sustain the full loading in this condition by virtue of the loss of elastic stiffness of the damaged box and the shedding of load to adjacent boxes. The webs all passed the web buckling checks and their stresses did not exceed the yield stress (247 MPa). The maximum bending case caused partial plastification of the bottom tension flange adjacent to the ineffective cracked web, but the yielded zone was relatively small and equilibrium was still maintained. The conclusion of this work was that propagation of an undetected crack would not lead to sudden collapse of the superstructure and therefore there would be time to intervene in such an event. It was therefore decided that the Midland Links boxes could continue in operation without traffic restrictions provided an ongoing programme of inspection of the key welds at bracings was made, which in turn would require a programme of access provision for internal inspection. This is essentially a “damage tolerant” approach to fatigue design as permitted by BS EN 1993-1-9 (2005) which permits reduced factors of safety for fatigue design provided two main conditions are fulfilled: – the boxes must be inspectable with an inspection regime in place; – there must be redundancy in the event of fatigue crack propagation. The first criterion was partially satisfied in that external inspection of the web was possible, as was ultrasonic testing. Provision of future access to the box voids would enhance inspectability. The second criterion was shown to be satisfied above; a full depth web crack could be sustained without collapse of the box girders. This “damage tolerant” approach is different from the usual design “safe life” approach where the design aims to ensure adequate fatigue life without the need for inspection. In order to assist inspection, fracture mechanics techniques were also considered to determine a critical crack length which would propagate by fast fracture. Knowledge of such a critical crack length would clearly inform the inspection regime and facilitate determination of when a crack should be repaired. However, no records of fracture toughness were available and the range of fracture toughness expected for BS15 steel was very wide. A pessimistic estimate of toughness led to predictions of very short critical crack lengths which would not easily be detected. Consequently, this study was abandoned in the secure knowledge that there was redundancy available in the event of a propagating crack forming. 4 REVISED FATIGUE ASSESSMENT USING IN SITU STRAIN MEASUREMENT Although the damage tolerant approach to fatigue proved that the structure was suitable for monitoring, it did not provide information on the time available to plan a maintenance strategy, nor predict when significant cracking might start to appear in the boxes. In fact, the codified analysis 851
indicated that cracking should have initiated some time ago when inspection had shown no signs of cracking. There were a number of possible reasons identified to explain the difference including: (i) the real spectrum of traffic being less onerous for fatigue than that in the design code; (ii) the in-situ stresses being lower than predicted in the analysis due to incidental composite action with deck furniture and surfacing or due to additional joint flexibility; (iii) better in-situ material properties for toughness than the design values assumed in the assessment; (iv) the possibility that the as-built details on the drawings differed from those actually present. The fatigue assessment was therefore repeated based on in-situ strain measurement for both a single known controlled load and also from measured in-situ strains during operation. The results are discussed below. 4.1 Load test Following the fatigue assessment and recommendations above, a load test was carried out on Bescot and Oldbury Viaducts with strain measurement on the box diagonal bracing and web-flange junctions. Access holes were installed in the boxes to facilitate attachment of the equipment. The testing employed a six-axle 40 tonne articulated lorry with a 40-foot flat barrier trailer loaded with 11 concrete temporary vehicle restraint barriers. This lorry type represents the second most common commercial vehicle configuration, accounting for 27–33% of all HGV types on the M6 (2.3 million vehicles per year) and M5 (1.6 million vehicles per year) respectively. As the response of the structure was likely to involve both composite action with the surfacing and some tension-stiffening in the concrete deck slab and the effect of both of these are load dependent, it was important to ensure that this test vehicle was representative both of common vehicle configurations and also the weight of the heavier ones. Use of a light vehicle could overestimate the benefit of these effects for real heavy vehicle loading and hence underestimate steel strains for heavier vehicles. The testing involved driving the test vehicle in designated paths (load cases) over each span while recording the induced strains in the relevant diagonal bracing bays. The load cases were designed to replicate both normal traffic loadings and the theoretical worst cases. As the stress at the box diagonal bracing locations were very sensitive to vehicle position, it was essential to ensure that the strain measurements were accurately matched to vehicle position. Consequently the vehicle position was accurately recorded during each load test by using a total station which automatically tracked a prism attached to the rear of the trailer. The time on the total station was synchronised with the strain-monitoring computer to enable correlation with the strain results. The vehicle was driven slowly to eliminate any dynamic impact effects. The strain monitoring equipment to be used was given careful thought. Measurement of the highly localised strains in the web with a high strain gradient over its height required gauges of very short length to avoid excessive averaging out of the stresses, thus underestimating the peak values. Uniaxial electrical resistance strain gauges were chosen as being most appropriate for this purpose. Numerous gauges were provided in order to determine the stress gradient. The strain was measured three ways in order to give some degree of cross-checking: (i) By installation of a grid of strain gauges directly on the web in the area of interest between bottom of stiffener and top of bottom flange. A typical layout is shown in Figure 7. This allowed direct comparison with the predictions for the strain in the web in the FE model. (ii) By installation of strain gauges on the diagonal braces. A typical layout is shown in Figure 8. This allowed comparison with the predictions for the strain in the cross bracing in the shell finite element model. The stress in the web is proportional to the axial force in these diagonal bracings. A large number of gauges were placed on the bracings to allow determination of flexural effects and average axial stress. (iii) By use of laser shearing interferometry (LSI) to measure strains directly in the web in the area of interest between bottom of stiffener and top of bottom flange. This was intended to overcome the averaging limitation of conventional strain gauges. LSI uses a green 532nm laser 852
Figure 7. Location of strain gauges on bracing and web.
Figure 8. Typical strain gauge readings.
beam which is expanded and collimated before illuminating the target surface (the web stiffener interface). Saving an image of the unstressed target and then subtracting all subsequent images obtained during loading of the structure produces a strain field, which is visible as fringes in the subtracted video and allows the strain changes under loading to be calculated. The image has a resolution of 1024 × 768 which is equivalent of approximately 786,000 strain gauges over a circular area of 150 mm diameter. Typical strain readings at the base of the web are shown in Figure 8. The in-situ strain measurements for the known vehicle were used to calibrate the finite element model. A series of load cases modelling both transverse and longitudinal test vehicle position were analysed and the results from the FE model compared to the in-situ strain measurements. A scale factor for fatigue stress range derived from the FE model was obtained and the fatigue life revised. Depending on vehicle position and bracing location, and whether the strain in the web or the cross bracing was used to derive the stress in and adjacent to the weld, the ratio of in-situ strain measurement to strain predicted by the FE model was between 40% and 60%. This ratio was then used to scale down the effects of the fatigue vehicle as determined in the finite element model. Two approaches were then taken to calculating the revised fatigue life based on the methods of clause 8.3 (fatigue vehicle method) and clause 8.4 (design spectrum) of BS5400 Part 10. The above calculations extended the fatigue life to 40 years. 5 REALISTIC FATIGUE LIFE AND ASSET MANAGEMENT PROPOSALS 5.1 Realistic fatigue life prediction The revised fatigue assessment based on the use of in-situ strain measurements to modify the results of the previous shell FE model led to a reduction in stress and therefore an increase in predicted fatigue life, measured from the opening of the viaducts, to 40 years. However, it was considered that this codified prediction of fatigue life was still very much on the conservative side for a number of reasons: – The calculation assumed that the current codified spectrum of vehicles has been in operation since the opening of the bridge. It is difficult to quantify this effect due to the lack of suitable historic data. Draft BD77 (1998) contained a means of approximating this effect via a formula for equivalent fatigue age of the bridge, Ye , from the actual age, Ya , and the age at which the traffic reached saturation point, Ys , taken as the bridge’s age in the year 2000:
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For the Midland links boxes, Ya = 38 years, Ys = 30 years so the equivalent fatigue age of the bridge, Ye = 29 years. This is 9 years younger than the bridge’s actual age which would therefore add 9 years to the predicted fatigue life measured from the opening of the viaducts. This implies the fatigue life is 49 years. – The choice of scale factor used in section 3.1.3 was the most pessimistic for the data collected. If the mid-range value was taken, i.e. 0.5, the calculated fatigue life would have increased to 80 years. – The worst 2.3% fractile material properties (two standard deviations from the mean) were also used i.e. there is a 2.3% probability of failure in the reported fatigue design life. If mean properties are considered, the life rises to 91 years without any of the above other considerations included in addition, but there is a 50% chance of a fatigue failure in that time which is an unacceptable probability. However, if access is to be provided for inspection, an increased probability of fatigue cracking could be considered. If properties corresponding to one standard deviation from the mean were used, this would give a 15.9% probability of failure in 60 years. This increased probability of failure when inspection is possible and planned is consistent with the Eurocode’s damage tolerant approach. In view of these further considerations, it was considered that a realistic fatigue life from the day of opening could be taken as 50 years. It was emphasised in the report to the Highways Agency that this still did not mean that cracks would form after this. It only implied that the likelihood of this event occurring will increase beyond this time. These results of the various analyses and the absence of any detected cracks suggested that the Midland Links boxes could be allowed to continue to operate without traffic restrictions provided that: – an ongoing programme of inspection of the key welds at cross frames was maintained; – any weld cracking discovered was repaired; – a programme of providing access for internal inspection was instigated for future maintenance operations unless other monitoring capable of detecting cracks inside the box is installed. 5.2 Intervention required if cracks occur It was also necessary to have a strategy in place in the event of cracks being found by the inspection regime. Cracks in welds could be repaired by grinding out the damaged areas and re-welding. Short lengths of cracks in the web plate could be arrested by drilling through the crack tips. Larger defects could potentially however lead to the need for strengthening (together with crack repairs or arresting measures), for which there were several alternatives including: – Connecting the stiffeners to the bottom flange which eliminates the high web distortional stresses. – Installing external ring frames around the box bracing locations to carry the distortional forces. This would minimise the amount of working inside the boxes. – Removing the diagonal braces to reduce the distortional forces attracted. REFERENCES BS 15, 1961. Mild steel for general construction purposes, British Standards Institution, London. BS 2762, 1956. Notch ductile steel for general structural purposes, British Standards Institution, London. BS 5400 Part 10, 1980. Steel, concrete and composite bridges – Part 10. Code of practice for fatigue. British Standards Institution, London. BS 5400: Part 3, 2000. Design of steel bridges. British Standards Institution, London. BS EN 1990, 2002. Eurocode – Basis of structural design. British Standards Institution, London. BS EN 1993-1-9, 2005. Design of Steel Structures. Part 1.9: Fatigue. British Standards Institution, London. Draft BD77/98. Assessment, inspection and monitoring of existing steel bridges for fatigue life, Highways Agency. UK Design Guide for Composite Box Girder Bridges, 1994. SCI Publication P140, Ascot, UK.
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Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
Improved structural health monitoring strategies for better management of civil infrastructure systems J. Winkler Atkins, Copenhagen, Denmark
C.R. Hendy Atkins, Epsom, UK
P. Waterfall Imetrum, Bristol, UK
ABSTRACT: As our global civil engineering infrastructure, such as rail and highway bridges, ages and the challenge of keeping it serviceable grows, the need for improved condition information on which to make good cost-effective maintenance decisions becomes ever more vital. Gathering this condition information requires structural health monitoring (SHM) and inspection on a grand scale and, for it to be useful, it must be accurate, inexpensive and easy to interpret. Digital image correlation (DIC) is an evolving measurement technique that has been proposed to enhance bridge inspection for the last 25 years or so, but is only recently starting to be used outside of the research community. DIC can be used for monitoring by imaging a bridge periodically and computing strain and displacement from images recorded at different dates or operating conditions. This data can be held to track the deterioration history of a defect and inform its cause and a suitable intervention. Examples of structural monitoring of different bridge components (local analysis) and an entire bridge (global analysis) using DIC are here presented.
1 INTRODUCTION 1.1 Bridge monitoring – challenges The principal difficulty with adding lots of monitoring systems to bridges is that they produce vast quantities of data. Bridge operators typically do not know what to do with this data unless there are very clear trigger levels defined associated with this data and clear interventions defined if they are exceeded. This means that the structural engineer needs to identify what the most likely deterioration mechanisms are for the particular bridge and design bespoke monitoring that can be used to specifically measure performance in such a way that an acceptable level can be set and checked e.g. force in a cable, movement at a bearing or joint, tensile crack lengths/widths at particular discontinuity regions, wire breaks in a cable, fatigue cracking at a steelwork detail, tilt of a column or pylon or deflection of a deck where a bridge is very sensitive to creep variations and other material parameters. This may require the use of many different systems. Instruments such as strain gauges, accelerometers, fiber optic sensors and displacement transducers are becoming increasingly common in structural health monitoring. These types of sensors can however possess drawbacks such as the need for external power and cabling/antenna for data transmission, high data acquisition channel counts and the limitation of only measuring at discrete points or along a line, so it is necessary to have an idea of where to expect damage when placing the instrumentation. These sensors can be used effectively to continuously monitor for abnormalities 855
Figure 1.
Digital image correlation method.
that indicate damage, but the type and severity of the damage can still be difficult to identify from discrete point measurements. 1.2 Digital image correlation (DIC) DIC represents a photogrammetry technique used for accurate measurements of surface deformation. The digitized images (e.g of a bridge deck) are compared to match facets from one image to another by using an image correlation algorithm (Fig. 1a, b). Image analysis involves capturing a reference image of a bridge component surface in its undeformed state. As the load is applied (e.g truck load), additional images are collected. The algorithm (which can either run in real-time or post-process), involves a stage-wise analysis, in which each stage consists of one image resulting in a description of displacements occurring on the surface of the bridge component. The evaluation of a correlation measurement results in coordinates, deformations and strains of the surface (Fig. 1c, d). DIC method allows high precision surface deformation measurement that can reach the accuracy of a few micrometers. DIC measurement can provide information about strain (all directions), vertical and horizontal displacement, crack size, rotation and acceleration. DIC is independent of scale (local & global monitoring) and material (concrete, steel). DIC is especially helpful in monitoring of bridge components with a difficult access. Real time operation, off-line analysis, remote access to equipment and live reporting of results is also possible. As the technology stands, DIC can be operated with as little as one individual and does not have to be in contact with the bridge therefore avoiding any potential conflicts with traffic or difficult geography. Also, DIC takes a fraction of the time to setup compared to traditional instruments. 1.3 Image analysis in bridge engineering Research activity on the application of photogrammetry in bridge-related projects has been minimal and widely dispersed within the last 25 years (Jiang et al. 2008). Early applications of this technique include measurement of the deflection of a continuous three span steel bridge under dead load 856
(Jauregui et al. 2003) and identification of the bridge deformation (Bales 1985). DIC was also employed to measure the geometry of a suspension bridge (Li & Yuan 1988) and deformation of steel connection of a pedestrian bridge (Johnson 2001). It was shown that DIC is a complementary tool of the conventional measuring systems such as LVDTs and strain gages. The last few years have seen an increased application of the DIC in the measurement of the vertical deflections of steel and concrete bridges due to traffic (De Roover et al. 2002, Lee & Shinozuka 2006a, b, SantiniBell et al. 2011, Chiang et al. 2011, Waterfall et al. 2012, Yoneyama & Ueda 2009, Yoneyama et al. 2007) and train transit (Busca et al. 2012). In general, it was concluded the DIC system accuracy is comparable to existing displacement measurement techniques and DIC is an easier way to measure displacement of multiple points at once. DIC was also proposed as a method to asses dynamic characteristics of suspension bridge hanger cables (Kim & Kim 2013). In this study, a non-contact sensing method to estimate the tension of hanger cables by using digital image processing based on a portable digital camcorder was proposed. Moreover, DIC technique has been used to record the strain on a concrete girder during a full scale bridge failure test (Sas et al. 2012) and for the measurement of the displacement field on a cracked concrete girder during a bridge loading test (Küntz et al. 2006). In both cases the photogrammetry method was able to detect a change in loading condition and locate cracks. Finally, DIC was used in fatigue testing of monostrands for bridge stay cables. Here, the vision-based system allowed for the measurement of the interwire movement (fretting fatigue) being the governing mechanism responsible for the fatigue life reduction in modern stay cable assemblies (Winkler et al. 2014). This paper focuses on application of the DIC technique for the enhanced short and long term bridge inspection and structural health monitoring of bridges. Field measurements were made on full-scale bridges and the results revealed that DIC is an effective approach to monitor the integrity of large scale civil infrastructure. 2 MONITORING OF BRIDGE COMPONENTS USING DIC (LOCAL ANALYSIS) 2.1 Riveted girder (displacement) In 2013 Atkins in collaboration with the Technical University of Denmark used DIC technique on the Storstrøm Bridge (Fig. 2a). The bridge was opened in 1937 and is a part of the railway connection between Denmark and Germany. There was a concern over one of the riveted bridge girders that was found to be cracked (Fig. 2b). The DIC system was successfully employed to measure deformations of the rivered girder under train and truck load. Figure 2c shows the DIC measurement and test setup. The vision-based system provided data that were quickly and easily interpreted in almost real-time. Figure 2d shows part of the Storstrøm Bridge girder with the marked square (strain map) that indicates a specific area being measured. Figure 2e shows the subsequent data analysis using virtual gauge placed on the strain map. 2.2 Stay cable anchorage (bending and axial fatigue) In 2014 Atkins successfully used DIC method in first of its kind full scale bridge stay cable fatigue tests. Large amplitude cable vibrations, threatening safety and serviceability of the cable supported bridge (Fig. 3a), have been frequently reported. However, despite the extensive research on the vibration mechanisms a limited work has been done to assess the fatigue characteristics of cables and cable anchorages (Fig. 3b) subjected to cyclic transverse deformations. Here, application of the DIC method enabled the measurement of individual wire strains along the length of a steel monostrand (Fig. 3c) and provided quantitative information on the relative movement between individual wires (Fig. 3d), leading to a more in-depth understanding of the underlying fatigue mechanisms. Novel application of the DIC technique provided previously unavailable information about the internal state of displacement of a bridge stay cable under bending load and helped in a better estimation of service life of bridges employing bundles of steel monostrands (Winkler et al. 2015). 857
Figure 2. The Storstrøm Bridge (a), deck underside (b), DIC measurement (c), specific area of the girder being measured (d) and analysis of DIC data (e).
2.3 Box girder (buckling of steel plates) The Tame Valley Viaduct carries the Aston Expressway between Birmingham city centre and Gravelly Hill Interchange on the M6 motorway. It is a key section of the UK highway network, of regional and national importance. The viaduct is 620 m long and comprises 21 spans of steel box-girder construction (Fig. 4a). An inspection undertaken in 2004 has established that there are significant deficiencies within the structure. The viaduct is constructed from thin steel plates and relies on thousands of stiffeners to prevent these plates from buckling. There was some concern that the traffic loading might induce buckling of the steel plates beyond an acceptable level. As a part of the overall SHM strategy Atkins used DIC system to measure buckling of a number of plates and bracing members on the bridge. The DIC system was used insitu inside the steel box sections, which have an approximate dimension of 3 m × 1.8 m (Fig. 4b). The camera used inside had an integrated infra-red light, providing diffuse, constant illumination. More than 30 virtual gauges and extensometers were providing data. The vision system has been used successfully to measure buckling of panels and bracing plates within the viaduct (Fig. 4c, d).
3 MONITORING OF A MULTI SPAN BRIDGE USING DIC (GLOBAL ANALYSIS) The Avonmouth Bridge is a 20 span twin steel box bridge with steel cross girders and a 164 m central span. It opened in 1974, with 3 lanes of traffic in each direction. It was later reconfigured to carry 4 lanes in each direction. The increase in loading was modelled, and strengthening work undertaken. 858
Figure 3. Schematic sketch of a cable supported bridge (a), cable anchorage (b), DIC measurement (c) and analysis of DIC data (d).
Following resurfacing works, there was a desire to see if the actual global movements fitted with those modelled under normal operating conditions. Here, the DIC system allowed simultaneous monitoring of 15 points along a 120 m section in real time, with the furthest point being 230 m from the camera. 3.1 Synchronized deck and column movement A two-camera configuration was used, allowing synchronized measurements of a bridge pier and the deck midspan (Fig. 5a). The deck and pier movements are coupled in a manner consistent with the central bridge span behaving as an arch (given that the bearings are pinned, not sliding). Figure 5b shows the outcome of the DIC measurement. 3.2 Lateral rotation and bending behavior of the bridge deck The lateral rotation and bending behavior of the bridge deck was also of interest to the client, and this was obtained by generating rotation measurements from points on the western edge, centre, and an eastern edge of the structure (Fig. 6a). The DIC system can track the movement of various points simultaneously, giving their deflection or rotation relative to each other. The DIC data gathered 859
Figure 4. The Tame Valley Viaduct (a), box girder (b), measured buckling of steel plates (c,d).
Figure 5. The Avonmouth Bridge (a) synchronized measurements of a deck and column movement (b).
has determined primary and secondary rotational modes across the deck. The primary mode is of the deck rotating as a whole (Fig. 6b).The secondary mode has the deck flexing, with greater flexure at the Western Edge as would be expected from the layout of the traffic lanes (Fig. 6c). The similarity indicates that the dominant rotational movement is rigid body rotation of the whole deck cross-section. As an additional indication of value, substantial further analysis can be done, such as by comparing the displacements of all the points identified with the single camera setup over time against a structural analysis of the bridge, or by further investigation of the rotation data. For example, by taking the root mean squared (RMS) values of the relative rotations, it was observed that the magnitude of these was on average 6% larger on the Western box than the Eastern box. This fits with the new alignment of the footpath and cycle path (hence lighter loading) on the Eastern side of the bridge. These rotational graphs provided an indication of the data that can be generated for the overall performance of the structure under normal loading conditions. All of this was obtained without any need to gain access to the structure to affix targets, or any need for traffic management. 860
Figure 6. Underside of the Avonmouth Bridge (a) primary rotational mode (b) and secondary rotational mode (c) of the bridge deck.
4 CONCLUSIONS The ability to capture a bridge’s behavior with DIC and calibrate a structural model with the collected data provides bridge designers and managers with an easy-to-collect objective measure of bridge performance. DIC is becoming very versatile and cost effective due to the dramatic improvement over the digital cameras and is therefore a promising solution to low-cost structural monitoring of existing infrastructures. A combination of DIC and other advanced monitoring technologies could finally give us the information we need to make effective whole life management decisions that keep our assets in service in a cost-effective way. REFERENCES Bales, F.B. 1985. Close-range photogrammetry for bridge measurement. Transportation Research Record, Washington, D.C., 950, 39–44.
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Busca, G., Cigada, A., Mazzoleni, P., Zappa and Franzi, M. 2012. Cameras as displacement sensors to get the dynamic motion of a bridge: performance evaluation against traditional approaches. Proc., 6th International Conference on Bridge maintenance, Safety and Management (IABMAS 2012), Stresa, Italy. Chiang, C., Shih, M., Chen, W. and Yu, C. 2011. Displacement measurements of highway bridges using digital image correlation methods. Proc., SPIE – The International Society for Optical Engineering, 8321(1), 83211G-6. De Roover, C., Vantomme, J. , Wastiels, J. and Taerwe, L. 2002. Deformation analysis of a modular connection system by digital image correlation. Experimental techniques, 26(6), 37–40. Jauregui, D.V., White, K.R., Woodward, P.E. and Leitch, K.R. 2003. Non contact photogrammetric measurement of vertical bridge deflection. J. Bridge Eng., 8(4), 212–222. Jiang, R., Jauregui, D.V. and White, K.R. 2008. Close-range photogrammetry applications in bridge measurement: Literature review. Measurement, 41(8), 823–834. Johnson, G. W. 2001. Digital close-range photogrammetry – a portable measurement tool for public works. Proc., 2001 Coordinate Measurement Systems Committee Conf., Coordinate Measurement Systems Committee. Kim, S. and Kim, N. 2013. Dynamic characteristics of suspension bridge hanger cables using digital image processing. NDT&E International, 59, 25–33. Küntz, M., Jolin, M., Bastien, J., Perez, F. and Hild, F. 2006, Digital Image Correlation Analysis of Crack Behavior in a Reinforced Concrete Beam During a Load Test. Canadian Journal of Civil Engineering, 33(11), 1418–1425. Lee, J.J. and Shinozuka, M. 2006a. A vision-based system for remote sensing of bridge displacement. NDT&E Int., 39(5), 425–431. Lee, J.J. and Shinozuka, M. 2006b. Real-time displacement measurement of a flexible bridge using digital image processing techniques. Exp. Mech., 46(1), 105–114. Li, J.C. and Yuan, B.Z. 1988. Using vision technique for bridge deformation detection, Proc., International Conference on Acoustic, Speech and Signal Processing, New York, 912–915. Santini-Bell, E., Brogan, P., Lefebvre, P. Peddle, J., Brenner, B. and Sanayei, M. 2011. Digital Imaging for Bridge Deflection Measurement of a Steel Girder Composite Bridge. Transportation Research Board (TRB 90th Annual Meeting), Washington, D.C , USA. Sas, G., Blanksvärd, T., Enochsson, O., Täljsten, B. and Elfgren, L. 2012. Photographic strain monitoring during full-scale failure testing of Örnsköldsvik bridge. Structural Health Monitoring, 11(4), 489–498. Waterfall, P.M., Macdonald, J.H.G. and McCormick, N.J. 2012. Targetless precision monitoring of road and rail bridges using video cameras. Proc., 6th International Conference on Bridge maintenance, Safety and Management (IABMAS 2012), Stresa, Italy. Winkler, J., Fischer, G., and Georgakis, C.T. 2014. Measurement of local deformations in steel monostrands using digital image correlation. Journal of Bridge Engineering (ASCE), 19(10). Winkler, J., Georgakis, C.T., Fischer, G., Wood, S., and Ghannoum, W. 2015. Structural response of a multistrand stay cable to cyclic bending load. Journal of Structural Engineering International (SEI) (May issue 2015). Yoneyama, S., Kitagawa, A., Iwata, S., Tani, K. and Kikuta, H. 2007. Bridge deflection measurement using digital image correlation. Experimental techniques, 31(1), 34–40. Yoneyama, S. & Ueda, H. 2009. Bridge Deflection Measurement Using Digital Image Correlation with Camera Movement Correction. Materials transactions, 53(2), 285–290.
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Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
Assessment of thermal actions in the steel box girder of the Millau Viaduct L. Defaucheux, H. Desprets, Z. Hajar & C. Servant Eiffage TP, Neuilly-sur-Marne, France
M. Virlogeux MV Consultant, Bonnelles, France
ABSTRACT: This paper deals with the analysis of temperature records obtained for the orthotropic steel box girder of Millau viaduct crossing the Tarn valley in south of France. The analysis carried out by exploiting a series of records made over a period of 7 years from 2005 to 2011, helped to accurately know the thermal behavior of the deck, namely the evolution of extreme values of the average temperature of the box girder, values of thermal vertical and transversal gradients. It was also determined the correlation of these variables with meteorological data, temperature and solar radiation.
1 INTRODUCTION The Millau viaduct (Figure 1) crossing the Tarn valley in southern France is a multi-span cablestayed bridge, with a total length of 2460 meters and 343 meters high at the tallest point of the pylon P2 (Figure 1). Because of its exceptional character, the Millau Viaduct was equipped with a great deal of instrumentation aimed at monitoring the structure’s behaviour over time and at validating the design assumptions (wind, temperature, etc.) used in its design. This paper presents the main results of the study of the thermal actions in the steel deck, carried out by the Eiffage TP design office, under the advice of Michel Virlogeux who oversaw a similar study for the steel box girder of the Normandy bridge [1], the results of which were used in defining the thermal action assumptions adopted for the execution design of the Millau viaduct. The 27.75 meter wide deck is also fitted with heavy restraint barriers and protective wind screens (Figure 2). The superstructure is a 4.20 meter deep orthotropic steel box girder whose deck plate is made from 12 to 14-mm-thick steel plates. The deck is supported by spherical bearings at all piers and is stitched to the piers with prestressing tendons. Each span is supported by eleven pairs of stays cables anchored to each side of the steel inverted-Y pylons to form half planes along the deck centerline.
Figure 1. Viaduct elevation.
Figure 2.
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Superstructure cross-section.
Figure 3.
Location of temperature sensors on cross-section.
Figure 4.
Zones of influence of sensors.
2 INSTRUMENTATION OF THE VIADUCT AND PROCESSING OF MAIN DATA It was decided at the execution design stage to equip one deck section of the viaduct, between piers P2 and P3, with 27 sensors that would take temperature readings approximately every 30 minutes (Figure 3). On the basis of the assumptions used, a temperature map was drawn up with 24 zones of influence, each attributed a temperature Tk measured by sensor number k. These zones of influence are shown in the sketch below (Figure 4): Calculations are performed by discrete analysis of the temperature-sensor zones of influence of sectional area Ak and center of gravity Gk (yk , zk ) for sensor k relative to the center of gravity G (0, 0) of the entire section: 1 – the average temperature: Tav = T(s)e(s)ds SL h T (s)e(s)z(s)ds – the equivalent vertical gradient: GTz = ly L b T (s)e(s)y(s)ds – and the equivalent horizontal gradient: GTy = lz L Atmospheric temperature readings are provided by the Millau Viaduct weather station located on the deck, half way between piers P4 and P5. In addition, weather data were obtained from the national weather Millau Soulobres station 2.5 kilometers north of the viaduct, at an altitude of 712 meters. This includes recordings of the daily global solar radiation (integral of the energy of radiation throughout the day, in watt-hours per square meter) from 2005 to 2012. 3 STUDY OF SHORT-TERM THERMAL EFFECTS 3.1 Summer-time thermal behaviour of box girder 3.1.1 Surface temperatures Figure 5 shows the evolution of the box-girder surface temperatures for a hot day: – The general curve of the surface temperatures basically matches the variation of the atmospheric temperature. – The temperatures of the top faces of the box girder rise much more quickly than the temperatures of the bottom faces. 864
Figure 5. Surface temperature on a hot day (20/08/2011).
Figure 6. Surface temperatures on a cold day (11/04/2011).
Figure 7. Temperatures in central cell on a hot day (02/07/06).
Figure 8. Comparison with Normandy bridge (04/06/98).
– No significant difference is observed between the eastern and western surface temperatures. The temperature difference between a plate on the eastern side and the symmetrically located plate on the western side does not exceed 1.5◦ C. The orientation of the superstructure appears to have little influence on the effects of its exposure to the sun. – There is an offset between the peak in solar radiation and the peak temperature in the superstructure. In general the peak temperature in the orthotropic deck occurs 5 to 6 hours after peak solar radiation (Figure 6). 3.1.2 Temperatures in central cell of the box girder Comparison with the results obtained on the Normandy bridge shows that (Figure 8): – the general shape of the curves is the same for both bridges, – the time offset is shorter for the Normandy bridge 3.1.3 Average temperature and thermal gradients The average temperature of the box girder depends on the atmospheric temperature and solar radiation. The average girder temperature is always higher than the ambient air temperature. Solar radiation acts especially on the difference between the highest air temperature and the highest average box-girder temperature (Figure 9). The vertical thermal gradient is closely linked to solar radiation. The largest vertical thermal gradient occurs 3 to 5 hours after the peak in solar radiation (Figure 10). The horizontal thermal gradient is small and insignificant compared to the amplitude of variation in superstructure and ambient air temperatures. 865
Figure 9. Comparison of the evolution of the average box-girder temperature for different solar radiation levels.
Figure 10. Evolution of the equivalent vertical thermal gradient (14/05/2006).
Figure 11. Evolution of the difference between maximum superstructure temperature and maximum atmospheric temperature over three years.
3.2 Thermal behaviour of box girder in Winter 3.2.1 Surface temperatures In Winter the variations in the surface temperatures of the box girder also match the pattern of variations in the ambient air temperature. In central cell of the box girder, the phenomena observed are similar to those observed in hot weather. 3.2.2 Average temperature and thermal gradients On cold days, solar radiation plays a major role in the temperature variations of the superstructure, as can be seen in Figure 12 which shows the evolution of the average superstructure temperature for cold days with high radiation (Rth = 4,642 Wh/m2 on 01/03/05). In the absence of solar radiation the vertical gradient is slightly negative, which can probably be explained by the effect of wind.
4 STUDY OF LONG-TERM THERMAL EFFECTS 4.1 Average box-girder temperature 4.1.1 Maximum daily values The annual variation of maximum average daily superstructure temperatures is similar to that of the maximum daily atmospheric temperature. 866
Figure 12. Evolution of the difference between minimum average superstructure temperature and minimum atmospheric temperature over 3 years.
Between the end of November and mid-February the maximum atmospheric temperature is generally a few degrees higher than the maximum superstructure temperature. For the rest of the year, the superstructure reaches a maximum average temperature higher than the maximum atmospheric temperature. Between mid-May and mid-July this temperature difference rises to around 8–10◦ C, which gives the following equation:
In the Design basis, the maximum superstructure temperature was estimated with the following equation:
The study of the thermal behaviour of the superstructure over a short period showed that its temperature variations depend mainly on solar radiation.
In his thesis on thermal actions [2], J-M Lucas obtained a similar correlation for Normandy bridge, using readings taken over a three-year period (i.e. 950 readings):
4.1.2 Minimum daily values With the exception of a few days between December and February, the minimum average temperature of the superstructure is always a little higher than the minimum atmospheric temperature. The temperature difference varies between 0.5◦ C and 5◦ C, with seasonal variation in the temperature level. The following equation gives a good approximation of the minimum average temperature of the superstructure:
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4.1.3 Yearly extreme values For the study period (2005–2011), the extreme values of the average superstructure temperature and atmospheric temperature were recorded every year. It can be seen with respect to these extreme values that the equations presented in §4.1.1 and $4.1.2 are on the safe side for the period 2005–2011:
With a reference temperature Tref = 10◦ C, the technical specifications for the Millau Viaduct set the positive deviation at +35◦ C and the negative deviation at −40◦ C for rare thermal effects in the steel. The data gathered at the Millau-Soulobres weather station provides the atmospheric temperature extremes since 1965:
At the end of 2004 it was therefore decided to raise the reference temperature to 13◦ C for definition of the viaduct’s unloaded state. Clause 6.1.3.2(1) of Eurocode 1 Part 5 gives reference values for maximum and minimum shade air temperatures for each region of France. Clause 6.1.3.1(4) specifies the deviation between the shade air temperature and the superstructure temperature for different kinds of superstructure. Application of Eurocode 1 to a steel superstructure in the region of the Millau Viaduct gives:
The two reference temperatures used in the construction design, i.e. 10◦ C, then 13◦ C, can be compared here:
Figure 13. Tables comparing temperature variations.
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Figure 14. gradient.
Day of greatest horizontal thermal
Figure 15. Equivalent vertical linear gradient versus daily solar radiation.
These tables give an indication of the relevance of the thermal assumptions adopted for the Millau Viaduct project and those of regulatory texts: – Maximum positive and negative thermal variations around Tref: ◦ The thermal assumptions of the Design basis presents a margin of about 8◦ C relative to the extreme values observed between 2005 and 2011. – Deviation between atmospheric temperature and average superstructure temperature: ◦ The values recorded between 2005 and 2011 are very close to those in the assumptions report. – The Eurocode 1 regulations, however, are pessimistic for positive temperature deviation. 4.2 Horizontal and vertical thermal gradient The equivalent horizontal thermal gradient varies between the extreme values of −3 and +3◦ C (−5 and +5◦ C for the Normandy bridge) When the evolution of the horizontal gradient is studied for a day of great horizontal thermal amplitude, it is seen that the variation of the gradient is similar to that of the temperature difference between the two vertical edge webs. There is just a slight time offset of about 30 minutes. Examination of the yearly evolution of equivalent vertical thermal gradients and daily solar radiation shows that there is a seasonal pattern and a certain correlation between the two physical quantities. On days with high radiation the gradient can be calculated with the following equation:
For Normandy bridge the equation obtained with readings from between 1996 and 1998 was:
Solar radiation has less effect on the vertical thermal gradient of the Millau Viaduct than it does on the Normandy bridge. This difference certainly comes from the different structural depth of the two bridges; 4,20 meters for the Millau viaduct, and 3,00 meters for the Normandy bridge. The same observation was made, in paragraph 4.1, about the role played by solar radiation in variations of average superstructure temperature. In the absence of precise data on radiation, the design basis considered a severe maximum equivalent vertical thermal gradient of 25◦ C, to be corrected in accordance with radiation readings. For a steel superstructure, paragraph 6.1.4.1 of Eurocode 1-5 gives a value of 18◦ C for the vertical thermal gradient (positive) when the upper surface is hotter than the lower surface. This value matches that of readings for the Millau Viaduct over the period 2005–2011. 869
The inverse gradient (negative), i.e. when the lower surface is hotter than the top, can be as high as 13◦ C according to the Eurocode. For the Millau Viaduct, the gradient is no more than 4◦ C. 4.3 Concomitance Plotting the maximum equivalent vertical thermal gradient against the maximum average temperature shows that the two parameters are relatively well correlated. For absolute temperatures, concomitant values represent about 78% of the yearly maximum value.
For the Normandy bridge concomitance was similar, and the two coefficients were 0.80. For temperature variations relative to the reference temperature, the combination equations become:
These results can be compared to the recommendations of Eurocode 1-5, §6.1.5, for which: – the combination coefficient for the action of the vertical gradient must be ωN = 0.75
– the combination coefficient for temperature variation must be ωM = 0.35
The second equation in the Eurocode does not ensure the same degree of safety. 5 CONCLUSION Statistical analyses of the readings taken over a period of seven years from 2005 to 2011 give precise information on the thermal behaviour of the superstructure, i.e. the evolution of average superstructure temperature extremes, thermal gradients (vertical and horizontal), and the concomitance between the two phenomena. The correlation between these physical quantities and meteorological data, especially the ambient temperature and solar exposure (solar radiation) was also determined. The analyses also confirmed the design assumptions of Michel Virlogeux which were used to design and verify the viaduct’s design. They show that the design of the deck and piers of the Millau Viaduct was based on realistic and slightly conservative assumptions. In conclusion, in the light of the observations and calculations of the study, it was decided, with the agreement of the concessionaire, to keep a set of 8 sensors (instead of the initial 27) in order to reduce the volume of readings while still recording significant temperatures. REFERENCES Lucas J.M., Virlogeux M., Louis C. 2005. Temperature in the Box Girder of the Normandy Bridge – Structural Engineering International, vol. 15, No. 3. Lucas J.M. 2001. Actions thermiques dans un caisson métallique orthotrope. Modélisation et mesures sur le Pont de Normandie. Doctoral thesis, Le Havre University.
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Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
Delayed deformations of concrete structures: The Savines bridge and the Cheviré bridge J.-P. Sellin, J.-F. Barthélémy, G. Bondonet & B. Cauvin Cerema – DTecITM, Sourdun, France
J.-M. Torrenti Université Paris-Est IFSTTAR, France
ABSTRACT: Deformations resulting from delayed behaviour of materials are the source of serviceability troubles for users and may distribute unwanted stresses into structures. This paper illustrates these phenomenon for the Savines and for the Cheviré bridges which suffer from spectacular delayed deformations. Thanks to a structural health monitoring system installed on the Cheviré bridge, theoretical deflections as given by standards are compared with the variation of sag. It shows that the measurements exceed the calculated deflection and that the latter does not enable the prediction of deformation rates. Empirical corrections of the material laws are therefore proposed in order to comply with respect to uncertain, unknown and sensitive parameters. The proposed method enables to evaluate the current condition of the structure and to check its limit states. A discussion is then being held about the relevance of the proposed optimization procedure and the ability to forecast future deformations.
1 INTRODUCTION Since the first findings by Hyatt in 1907 (USA) and by Freyssinet in 1912 on the Veurdre bridge (France), it has become well known that concrete undergoes delayed deformations that may result in structural failures and serviceability problems. Many research results [Pons et al., 2008] have shown that delayed deformation of concrete can be split into two parts: shrinkage (independent of the loading) and creep (depending on the stress state). On the one hand, shrinkage results from the effect of capillary pressure due to self-desiccation (due to hydration) or drying. On the other hand, creep can be explained by two phenomena: basic creep related to viscous deformations at the scale of the hydrates and desiccation creep caused by drying i.e. water diffusion out of the concrete volume. As desiccation shrinkage and creep are generated by water diffusion, their evolution should be governed by the characteristic time of the latter process. Likewise self-desiccation shrinkage should be controlled by the characteristic time of the hydration process. With regards to steel behaviour, the relaxation of tension with time is the result of propagation of micro-cracks or the sliding of grains joints under mechanical external stresses. Despite the origin and the evolution of these phenomenon are still debated, their consequences on structures are quite known. Indeed, deformations of materials are used to appear on large bridges where deflection at mid-span may cause serviceability problems. The Savines Bridge is one of them exhibiting high deflections at mid span as represented in Figure 1. 2 FEEDBACK FROM THE SAVINES BRIDGE STUDY The Savines Bridge is one of the first generation of post-tensioned pre-stressed concrete bridges built in France. It has been constructed between 1958 and 1960 by the free cantilever method. 871
Figure 1.
Figure 2. Comparison of the predicted deflections with experimental measurements.
Present view of the deck.
It is a 77 m long box-girder bridge made of 13 spans. The concrete deck is post-tensioned with internal tendons. At the middle of each span, rotations and horizontal displacements are free of motion thanks to a horizontally sliding hinge which transmits only shearing forces. According to a symmetry argument, each cantilever can be considered in a first analysis as a statically determinate structure independent from the others. The evolution of deflection for a span (modelled as a cantilever) is regarded according to Eurocode standards [NF EN1992-1-1, 2005] and [NF EN1992-2, 2006]. The deflection calculation is then compared with in-situ measurements. These simulations are undertaken with the ST1 software [CEREMA (Structural Calculation and Software Division), 2015] using a beam model and the incremental theory for the calculation of delayed behaviour of materials. This analysis have shown the limitation of standards to estimate delayed deformations of the bridge as shown in Figure 2 where the evolution of the deflection according to standards [NF EN1992-1-1, 2005] [NF EN1992-2, 2006] is compared with in-situ measurements1 . The discrepancies exposed in Figure 2 might come from a non appropriated model used for the simulation, unknown characteristics or unadapted behaviour laws of materials. Moreover, although measurements are precise enough, because of change of baselines, it remains difficult to assess the real long term behaviour. Despite these uncertainties, it is clear that, in the case of the Savines bridge, calculations based on current standards systematically underestimate long term deformations. That is why, in order to predict the evolution of the deflection, it is proposed to reach in-situ measurements by means of an adaptation of compliance functions thanks to weighting coefficients. This approach enables to identify major parameters having a high sensitivity with regards to deflection. Unknown and/or significant parameters are the coefficient of relaxation at 1000 hours and the specific weight of concrete. Weighting coefficients in the compliance function are the basic creep kbc , desiccation creep kdc and a kinetics coefficient kkc allowing to change the kinetics of the compliance function are set as optimisation parameters (see [Sellin et al., 2014] for more details). The process uses the least square method. Results are resumed in Table 1. Despite this manner offers the chance to reach the level of in-situ measurements, the analysis of the evolution of deflection arises several questions. Indeed, even if optimized coefficients are rather high (see Table 1), Figure 3 does not show any stagnation of the deflection. In fact, the continuous evolution of deflection makes arise the decisive role of relaxation of tendons: despite 1 Since
the reference measurement baseline has changed several times without information cross-referencing, the deflections are defined up to an unknown translation from one dataset period to another one. Consequently, for some dates, manual corrections are required in order to estimate likely values of absolute deflections and two trends are therefore imagined.
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Table 1. Savines Bridge: results of optimization for EN1992-2.
Figure 3.
Parameters
Constraints
Trend 1
Trend 2
kbc kdc kkc ρ1000 γc
0.4 ≤ kbc ≤ 1.6 0.4 ≤ kdc ≤ 1.6 0.1 ≤ kkc ≤ 7.5 9.0% ≤ ρ1000 ≤ 12.0% 25 kN/m3 ≤ γc ≤ 27 kN/m3
1.5 1.5 1.4 11.8 26.7
1.6 1.6 1.4 12.0 27.0
Prediction of the optimized computed deflection for the next 50 years.
compliance function for creep and for shrinkage presents an asymptotical behaviour, the deflection does not stop even when the characteristic time of the latter is passed. The explanation comes from formulation of the relaxation function: likewise concrete, evolution of the relaxation function is based and extrapolated from experiments carried on few years. Moreover, the evolution of relaxation as written in Eurocode [NF EN1992-1-1, 2005] is open to interpretation since it is proposed for the calculator to use its value at 500,000 hours for the evaluation of relaxation losses at the end of service life (despite the function is a power law). In other words, it amounts to introduce a constant value from 500,000 hours (about 57 years). This stagnation is not taken into accounts in the study. It is therefore reasonable to wonder about real losses of pre-stressing as our calculations are not a part of a design procedure. 3 THE CHEVIRÉ BRIDGE Built in the 90’s, the Cheviré bridge crosses La Loire river with two access viaducts and a composite central span for a total length 1563 m. The central span (240 m long) is made by a steel free span (160 m long) supported by two cantilevers (40 m long). The access viaducts are composed by two concrete structures joined with hinges. The steel deck is supported at the edge of the North and of the South cantilever by means of hinges respectively named NCSH and SCSH (for North and South Concrete/Steel Hinge). The longitudinal profile is represented in the Figure 4. Some general characteristics of the bridge and information about its construction are available in [Virlogeux et al., 1990]. Despite the Cheviré Bridge is younger than the Savines Bridge, it presents rather high deformations which may cause serviceability troubles for users. Indeed, since the beginning of service life, high deformations are observed at the end of cantilevers. Unlike the Savines bridge, the Cheviré bridge is a non determinate structure. Therefore, delayed behaviours of materials may redistribute unexpected stresses into the structure. For a sake of simplicity and despite other delayed 873
Figure 4.
General longitudinal profile of the structure.
Figure 6. Comparison of the predicted deflections with experimental measurements.
Figure 5. Trend of the deflection since 1991.
deformations phenomenon observed (for instance a global shortening of the deck due to shrinkage), the study is focussed on vertical deflections of the South cantilever. The purpose of the analysis is to perform a evaluation of the capacity of standards to model the delayed response of the structure. Then, a numerical adaptation of standards likewise the previous analysis [Sellin et al., 2014] is performed. 3.1 Monitoring Since its beginning of service life in 1991, the bridge is carefully inspected and the evolution of the deflection is frequently regarded by means of height measurements. Since 2001 the evolution of deformations is analyzed by means of rotational sensor settled in the South cantilever. Indeed, eight rotation sensors and eleven temperature sensors transmit every four hours instantaneous rotations and temperature measurements. Deflection of the cantilever is then obtained with integration of rotation data along the latter. It is consequently necessary to subtract temperature effects from the variation of rotation. Furthermore, likewise the Savines Bridge monitoring, the data treatment requires a manual translation from one dataset period to another one: from the height manual measurements to the automatic treatment by means of sensors in 2001 and for their partial replacement in 2009. The trend since 1991 of the deflection is represented in Figure 5. t0 corresponds to the first date of measurement in order to set a standard comparison date. At that time, deflections are set equal to zero. Even so, the current deflection is about 25 to 30 cm (after only 24 years of service) and the deflection rate does not seem to decrease. An older study (from F. Imberty and J-L. Clement) done between 2002 and 2006 showed discrepancies between measurements and numerical computation using standards. Several experiments on concrete samples removed from parts of the deck and on reconstituted concrete were performed. This enabled to build a PCP numerical model [CEREMA (Structural Calculation and Software 874
Figure 7.
Sketch of the restrained model.
Division), 2015]. In order to estimate more precisely the delayed behaviour of the concrete, several specimens extracted from segments on piles were tested in laboratory [Clement et al., 2006] (confidential study). Experiments carried on almost three years enabled to evaluate and to compare the compliance function of concrete to the theoretical one obtained by the BPEL (old French norm) and by the Eurocode [NF EN1992-1-1, 2005]. These measurements achieved the subsequent conclusions: – in order to correctly represent creep deformations, and according to [NF EN1992-2, 2006], the creep compliance function should be weighted by about 1.2 to 1.7 (depending on the concrete from the segment tested); – concrete has a high variability of its Young modulus along the structure; – the average Young modulus measured is about 26,500 MPa which is very low; – the average compressive strength of concrete at 28 days is 45.7 MPa for the Southern viaduct. In addition, concrete specimens presents a rather high water permeability (about 16%). The concrete used for the Cheviré Bridge comes from two different factories with different type of aggregates. Despite the same formulation was used, the concrete is highly heterogeneous. Mechanical and physical heterogeneities are not taken into account in the numerical model where average values are used. 3.2 The model Only a part of the Southern structure is modelled assuming the effects of delayed deformations are gather. Main assumptions used in the PCP model are summarized bellow: – concrete – average Young modulus: 26,000 MPa – cement type: R – average specific weight of the pre-stressed concrete: 25 KN/m3 – steel (19T15 et 27T15 tendons) – relaxation of tendons after 1000 hours: 2.5% – initial tension: 1400 MPa – site conditions – temperature effects are not taken into account – average relative humidity: 81% PCP automatically calculates the mean notional size for the evaluation of drying effects. For a sake of simplicity and computational time limitation, only a part of the South viaduct is modelled assuming that effects of delayed deformations are limited far from the cantilever. The restrained model is shown in Figure 7. Evolution of deflections and sollicitations analysis performed on this restrained structure are regarded far from the opposite edge where the structure is free of motion and of force. For the following analysis, the distribution of forces and of deflections is regarded only for the first three spans (from S1 to S4, see Figure 7). 875
3.3 Forecasting capability The prediction capacity of standards are compared with in-situ measurements and with adjustments obtained from experiments on specimens and on reconstituted concrete samples [Clement et al., 2006]. Four models are tested: – the original model with the BPEL and the design value of the Young modulus Ecm = 36,000 MPa – EN1992-1-1 with a reduced Young modulus Ecm = 26,000 MPa – EN1992-1-1 modified with a reduced Young modulus Ecm = 26,000 MPa and the creep compliance function weighted by kbc = 1.70 for basic creep (from experiments on concrete samples) – EN1992-1-1 modified with a reduced Young modulus Ecm = 26,000 MPa and the change of kcf coefficient in kcf = 0.53 (originally equal to 0.3) and the creep compliance function weighted by kbc = 1.84 for basic creep (from experiments on reconstitute concrete) The effect of creep and shrinkage writes using weighting coefficients ki in the decomposition of the delayed strains:
where as , ds and bc are respectively denote autogenous shrinkage, desiccation shrinkage and basic creep strains. Eq. B7 in [NF EN1992-1-1, 2005] becomes:
where σc represents the stress applied and Ec the instantaneous Young modulus. βc (j, j0 ) is defined as:
Tests on reconstitute concrete also permitted to adapt the shrinkage evolution by replacing the kks = 0.094 coefficient (originally equal to 0.04 in Eq. 3.10 in EN1992-1-1):
where h is the mean notional size of the member for approximate evaluation of shrinkage h = 2AP c The evolution of deflection for these models is represented in Figure 6. Figure 6 shows that current standards are not able to model correctly deformation of the prestressed cantilever. Moreover, the kinetics of the computed deflection is not representative of measurements in spite of law adaptation thanks to laboratory tests. That is why, in order to reach the level of measurements and predict the trend for the next decade, the above introduced coefficients are optimized by means of the least square method. The high porosity of concrete may intensify effects of shrinkage; more especially of desiccation shrinkage. Hence, weighting coefficients for basic creep and for desiccation shrinkage are manipulated such as the ambient relative humidity and the Young modulus of concrete. Moreover, unlike EN1992-2 [NF EN1992-2, 2006] for High Performance Concrete, EN1992-1-1 [NF EN1992-1-1, 2005] does not distinguish desiccation creep from basic creep. For the same reason as mentioned above, its weighting coefficient kbc is deliberately increased. Kinetic adaptation of compliances function for creep or shrinkage from laboratory tests might also be corrected. Results of optimization are resumed in Table 2 and in Figure 8. 876
Table 2. Cheviré Bridge: results of optimization for EN1992-1-1. Parameters
Constraints
Optimum
Hg Ecm kas kds kks kbc kkc
78% ≤ Hg ≤ 84% 22,000 MPa ≤ Ecm ≤ 30,000 MPa 0.8 ≤ kas ≤ 1.8 2.5 ≤ kds ≤ 3.5 0.8 ≤ kks ≤ 4.0 1.1 ≤ kbc ≤ 2.5 0.2 ≤ kkc ≤ 1.5
78.3% 29,600 MPa 1.6 3.4 3.9 2.0 1.4
Figure 8.
Prediction of the optimized computed deflection.
Figure 9.
Stress rate and deflection along the structure at time t0 and tf .
3.4 Structural effects As a non determinate structure, possible redistribution of forces shoud be regarded. Thanks to the presented process, the stresses are drawn in Figure 9 at the time t0 (corresponding to the 1991’s state) and at time tf (for the current date 2014). Because of the restrained model, the dispersion of stress is regarded in the four right spans. The visco-elastic behaviour of concrete is consistent with the linearity assumption satisfied when stresses do not exceed 0.45 · fck [NF EN1992-1-1, 2005], where fck is the characteristic compressive strength of concrete. Here, the latter is satisfied. However, a drop of stress is observed in the bottom flange (between S3 and S4, see Figure 9). This demonstrate the importance to forecast future deformation in order to prepare retrofitting actions. The analysis of stresses and of sollicitations is still in progress. 877
4 DISCUSSION This study shows and demonstrates at the full scale of these two bridges the inability of current standards to predict accurately delayed deformations of pre-stressed concrete structures. Compliance function of concrete in Eurocodes [NF EN1992-1-1, 2005] and [NF EN1992-2, 2006] can be adapted by means of weighting coefficients. The exposed optimization process enabled to reach current deflections. Nonetheless, the origin of these coefficients is questionable. Adaptation of coefficients from laboratory tests (2002–2005) on concrete samples and on reconstituted concrete enabled to catch up measurements though a limited time interval (during the laps time of measurements). Hence, it remains difficult to guarantee a reliable prediction of the evolution of delayed deformations for a large time laps since the last measurements. This study also demonstrates that it is mandatory to model accurately creep and shrinkage deformations but also the evolution of pre-stressing losses with time. A numerous collection of bridge data measurement presented in [Ba˘zant et al., 2011] summarizes the evolution of deflection of several bridges. It shows that the deflections are linear with time in a logarithm time scale. This observation is used in the formulation of some model particularly in the recent ModelCode MC2010 of the fib [fib, 2012]. The couple of bridge suffering from delayed deformations are a source of data useful for the validation and/or adaptation of standards. This underlines the necessary awareness of monitoring devices, tools, their continuity and their reliability with time. This work using on numerical adaptation of behaviour laws of materials at macro-scale points out the necessity to develop into new laws related to physical phenomenon. 5 ACKNOWLEDGEMENTS The authors wish to thanks Florent Imberty (Razel Bec) who built PCP models and Jean-Luc Clement from IFSTTAR (formerly LCPC) thanks to him, Eurocode laws were successfully adapted from experiments. The authors wish also to thanks PCP and ST1 team developpers [CEREMA (Structural Calculation and Software Division), 2015] who programmed specific laws used. The authors wish also to thanks Jean-Marc Tarrieu for his expert skills and advices for the Cheviré Bridge. Authors appreciates also the Angers Laboratory (Cerema/DTerOuest) for the development and the maintenance of monitoring data devices. REFERENCES Bažant, Z.P. & Hubler, N. & Yu, Q. (2011). Excessive creep deflections: An awakening. In Concrete international, no 33(8), 44-46. Clement, J-L. (2006). Rapport d’analyse des résultats de fluage, IFSTTAR (formerly LCPC), Confidential study. fib, Bulletin 65 (2012). Model code 2010, final draft, fédération internationale du béton (fib), Lausanne, Switzerland. NF EN1992-1-1 (2005). Eurocode 2: Design of concrete structures - Part 1-1: General rules and rules for buildings. NF EN1992-2 (2006). Eurocode 2: Design of concrete structures - Part 2: Concrete bridges – Design and detailing rules. Pons, G. & Torrenti, J-M. (2008). Le retrait et le fluage, La durabilité des bétons: bases scientifiques pour la formulation de bétons durables dans leur environnement, Ollivier J.P., Vichot A., Presses de l’École Nationale des Ponts et Chaussées, pp 167–216. Sellin, J-P. & Barthélémy, J-F. & Torrenti, J-M. & Bondonet, G. (2014). Delayed deformations of segmental prestressed concrete bridges: the case of the Savines Bridge, 1st International Conference on Ageing of Materials and Structures, 266–273, The Netherlands, Delft. CEREMA (2015). structural softwares, website: http://www.setra.fr/html/logicielsOA/logiciels.htm. Virlogeux, M. & Bouchon, E. & Martin, J-C. & Lefevre, J. & Fraleu, P. & Maury,Y. & Guyot, T. & Ryckaert, J. & Pottier, M. & Heusse, A. & Mathivat, J. & Lenoir, B. (1990). The Cheviré Bridge. La technique française du béton précontraint AFGC, fib symposium, 381–415. Germany: Hambourg.
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Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
Laser scanner in identification of pathological manifestations in concrete S. Pavi, P. Gorkos, F. Bordin, M. Veronez & M. Kulakowski University of Vale do Rio dos Sinos, São Leopoldo, Rio Grande do Sul, Brazil
ABSTRACT: Bridges and viaducts are of paramount importance for the development of cities, and the effectiveness of these structures depends on regular maintenance to which they are subject. The ways these inspections are made today are quite rudimentary. Therefore, this study aimed to evaluate the potential use of remote sensing equipment, the Terrestrial Laser Scanning (TLS) to detect pathological manifestations on bridges and concrete viaducts. The methodology consists of scanning areas affected by pathological manifestations and post processing data by a classifier algorithm, which generated images in which it was possible to differentiate the pathological manifestations and concrete area. The results showed that the use of the return pulse intensity information acquired by the TLS identify the areas affected by biodeterioration, humidity and efflorescence on the concrete surface and can therefore be used to complement the current methodology of inspection of bridges and viaducts.
1 INTRODUCTION Among the infrastructure works bridges and viaducts are of paramount importance for the development of cities, both from an economic point of view as of social. These structures called Special Artworks (SA) are essential elements of a road system and their effectiveness depends on regular maintenance to which they are subject. SA’s inspections, the way they are performed today, present as problematic the fact that the presence of the damage is not quantitatively standardized and depends solely on qualitative criteria inspector (González-Jorge 2012). Another limitation is the inspections in structures of hard accessibility that often ends up representing security risk to the inspector. Following what has been proposed by González-Jorge et al. (2012) and Armesto-González et al. (2010); it is believed that the use of remote sensing techniques specifically the use of intensity data obtained from the scanning with the Terrestrial Laser Scanner (TLS) can be of great value to the identification and monitoring of pathological manifestations in concrete bridges and viaducts. In this sense the methodology used in this study aims to evaluate the possibility of using the TLS intensity data to detect the presence of pathological manifestations in bridges and viaducts of reinforced or prestressed concrete as a complementary technique to the methods of SA’s existing inspection. 2 LITERATURE REVIEW The most commonly material used in the construction of bridges and viaducts is the concrete. Thus, in this study it will be mentioned only bridges and viaducts of structural concrete. The structural concrete suffers action of deterioration mechanisms manifested in the structure of a symptom or set of varied symptoms called pathological manifestations. The correct identification of pathological manifestation that affects the structure is of fundamental importance to the choice of materials and the repair correct technique to be employed 879
(Helene 1992). Therapeutic measures for the solution of pathological problems may include minor repairs or even a general recovery of the SA. As the subject of this study is to identify the pathological manifestations in the bridges and viaducts will be addressed biodeterioration and efflorescence, which are manifestations that are visually identify the viaduct in study. 2.1 Efflorescence and biodeterioration Efflorescence is the formation of salt deposits on the concrete surface resulting from their exposure to water infiltrations or weathering (Uemoto 1988). To Cánovas (1988), efflorescence are a pathological manifestation produced by the presence of soluble salts in the mass of concrete that are leached by water to the surfaces of the parameters in which they crystallize producing spots. These constituents may be aggressive salts, cause profound breakdown, and modify the visual appearance of the structure on which are deposited (Pereira 2012). Already biodeterioration can be understood as any undesirable change in the properties of a material caused by the vital activities of organisms (Hueck 2001). The action of these microorganisms primarily affects the concrete contributing to erode the exposed surface to increase the porosity of concrete (intensifying the diffusion coefficient), which facilitates the entry of chloride ions and accelerate the propagation of cracks. All these mechanisms accelerate corrosion of the steel reinforcement causing a reduction in the structural capacity (Sanchez-Silva & Rosowsky 2008). One of the features of biodeterioration is the biofilm formation on the structure. This biofilm is the result of the aggregation of organic substances, microorganisms and products derived from own metabolism of these organisms, which are deposited on the surface of the affected structure in the presence of humidity (Bott 1993). Structural problems related to biodeterioration cost billions of dollars a year in maintenance and repair of structures (Silva & Rosowsky 2008). 2.2 Applications of terrestrial laser scanner in civil engineering The remote sensing technique discussed in this article is the LIDAR – Light Detection and Ranging – specifically the use of the Terrestrial Laser Scanner (TLS), active remote sensing of terrestrial platform. According to Longley et al. (2013), the Remote Sensing measured by electromagnetic radiation, physical, chemical and biological properties of objects without direct contact with them, in other words, it is a non-destructive and non-invasive technique. Lasers scanners enable detection of a large number of data points distributed on the surface observed with a respective intensity with high precision and a fast rate of acquisition. There are several applications of remote sensing techniques with TLS in the field of Civil Engineering. Bosché (2010), in a study applying the TLS to obtain as-built projects concluded that the equipment along with the use of recognition algorithms can be efficient and effective to automatically track the status of 3D construction projects, mainly to control the dimensional conformity of construction. A recent study by Rabah, Elhattab and Fayad (2013) used the TLS in the mapping of cracks in concrete. A computational algorithm mapped the existing cracks and calculated the probability of appearance new cracks in the vicinity producing accurate information about the size, arrangement and the direction of the cracks. TLS intensity data were used to characterize the recognition and pathological manifestations of materials that constitute the construction of historic buildings. Armesto-González et al. (2010) used a methodology that combines the use of scanning the intensity data with classification methods. The intensity information TLS were subjected to rank k-means algorithm that identifies the pathological manifestations incidents mainly lime deposits and humidity in the structure. 880
Figure 1. Viaduct BR 116, São Leopoldo/RS.
González-Jorge et al. (2012) also used the TLS of the intensity data in their studies. The authors describe a method for detecting proliferation of mosses (biofilm) on reinforced concrete pillars of a viaduct by using a classifier algorithm TLS and comparing different algorithms and classification equipment. From the literature review performed, it can be observed that following the studies already carried out previously, the TLS can be successfully used in the inspection of bridges and viaducts and TLS pulse intensity data are useful to classify pathological manifestations incidents in the structures. 3 METHODOLOGY The experiment consisted in scanning the viaduct (located on the BR-116 highway in São Leopoldo, Brazil) using the Terrestrial Laser Scanner Ilris 3D of Optech brand owned by the University of Vale do Rio dos Sinos (Unisinos). Builted in 1992, the viaduct has heavy traffic all day long. According to information of DNIT – National Department of Transport Infrastructure – of São Leopoldo, this viaduct was inspected only once in 2013. The Figure 1 shows an image of this viaduct. In the viaduct of the BR-116 it was performed two scans, which aimed to obtain data of areas whose pathological manifestations had been detected visually in the information gathering stage. Scans were made in a timely manner directly in the focus of visually identified pathological manifestations. Firstly, scanned up the concrete structure in the viaduct junction with the highway, where it was identified a biofilm characteristic of biodeterioration process. The distance of TLS to the target was about 20 meters and the scan time was 40 minutes. The scanning resolution adopted was 2 mm (2 mm distance between the points). The second scanning aimed to observe the efflorescence in the underside of the slab of the viaduct. The resolution of this scanning was 1.1 mm and the TLS was positioned at a distance of 14 meters from the target. The scanning time of this structure was 22 minutes. Figure 2 shows the scanned areas on the viaduct. The processing step of data from scans consisted in transforming the generated file by TLS in a text file through the PARSER software developed by the manufacturer of the TLS. After this stage, the point cloud was manipulated and viewed by POINTOOLS software. In addition, for the classification it was used an algorithm based on k-means algorithm that separates the return pulse intensity values TLS on classes. In each classification it was informed the desired number of classes (estimated according to the number of pathological manifestations expected on that surface). The algorithm classified the points of intensity values in up to four classes and assigned a different color for each class. There will not be displayed the images in five classes or more because they have not represented in satisfactorily manifestations the pathological incidents. 881
Figure 2.
Scans performed on the viaduct.
Figure 3.
Biodeterioration and humidity in concrete cells.
From the data processing generated by the TLS there were obtained images of the scanned areas. These images were analyzed to verify that the algorithm was able to identify/separate pathological manifestations that were visually identified. 4 RESULTS During the information gathering stage, it was observed that the study viaduct had numerous pathological manifestations, being more visible stains caused by humidity, biodeterioration and efflorescence. The areas that were not affected by visible pathological manifestations were called by wholesome concrete and healthy concrete. However, these terms do not represent the actual diagnostic of the concrete structure, since various pathological manifestations can cover in this area and have not been identified in the visual inspection performed. The biodeterioration despite reaching various parts of the structure is more apparent in concrete cells that form the viaduct junctions with the highway. In these areas, due to high incidence of humity, the biofilm layer is fairly thick as shown in the images of Figure 3. Efflorescence, which can be seen in the underside of the slab of the viaduct, is presented in the form of yellowish spots with some cracks and stalactites. The yellow color indicates the presence of efflorescence of iron oxide together with the salts which they were carried to the surface of the concrete. The leached iron oxide is possibly a result of the corrosion process that occurred in the armor of the slab. Figure 4 shows the images of this pathological manifestation. 4.1 First scanning A comparison of the results of the first scanning can be seen in Figure 5, which shows in (a) the image generated by the photographic camera, in (b) the image obtained by the classifier algorithm in two intensity classes, in (c) three intensity classes and in (d) four intensity classes. 882
Figure 4.
Efflorescence on the viaduct.
Figure 5. classes.
Concrete surface photograph (a) and classification by two (b), three (c) and four (d) intensity
In the image (b) of Figure 5, it is possible to detect the materials with different reflectivity of the concrete. Thus, it was possible to identify/separate the area in which the surface is affected by the biofilm of the healthy concrete area. The red color represents the biodeterioration affecting the structure, manifestation identified on visual inspection, and the light blue identifies the wholesome concrete. The classification performed by the algorithm in three intensity classes – wholesome concrete, incidence of humidity and biodeterioration – produced the image (c). As shown in the image (c) on the concrete surface the TLS identified three different reflectivity, which indicates that in this area there are not only one but two pathological manifestations. Starting from the visual inspection and making an analysis with the image obtained by the camera (a), we conclude that the red color of the image is healthy concrete, while green represents biodeterioration and the blue color indicates humidity. In this classification, it was possible to detect the humidity which affects in almost all the scanned area, as is the blue color in the image. This humidity is not capable of viewing on a common inspection and cannot even be detected in the image (a) of Figure 5. To identify through the intensity pathological manifestations that are not possible perception in visual inspections, we adopted yet the classification of TLS data in four intensity classes: healthy concrete, biodeterioration, higher humidity content and lower content of humidity. It is observed from the image (d) further in Figure 5 that the surface area of the concrete is affected by biodeterioration identified in purple color. The parts with increased incidence of humidity where there is a less thick layer of biofilm have been sorted in green, while the region with lower humidity content (compared with the green area) was identified by the color red. The wholesome concrete area was marked by blue. Analyzing and comparing the results it can be seen that the image generated in the classification in four intensity classes is the one that best approximates the actual situation of the affected area in the structure. 883
Figure 6. classes.
Concrete surface photograph (a) and classification by two (b), three (c) and four (d) intensity
In the first scanning, the TLS was effective to detect biodeterioration and humidity in the concrete surface, and the classification of intensity data into predefined classes showed satisfactory results identifying the affected areas of very close so the actual situation of the structure. 4.2 Second scanning The results of the second scanning are shown in images of Figure 6, which compares the image produced by the photographic camera (a) with the classification into two classes (b), three classes (c) and four intensity classes (d). The classification of the data intensity of the second scanning by classifier algorithm in two predefined intensity class – healthy concrete and efflorescences – produced the image (b) of Figure 6 in which it was possible to identify the area where the surface of the concrete is affected by pathological manifestations. The light blue color represents the pathological manifestations that affect the structure identified on visual inspection as efflorescence and humidity. The red color identifies the healthy concrete area. In (c) it can be seen the image obtained by the classifier algorithm based on pre-defined classification of intensity data into three classes: wholesome concrete, efflorescence (greater deposition of salts) represented in blue and efflorescence (lower deposition salts) in green. It was considered efflorescence with greater deposition of salts, the situation in which efflorescence has covered the concrete surface by a layer of crystallized salt deposits. The efflorescence with less deposition of salts is the situation where the salt deposits crystallized layer is less thick compared to the called efflorescence layer (higher deposition salts). As it can be seen in the image on the concrete surface of the bottom slab of the viaduct board, the TLS identified three different reflectivity. The blue color is the thickest layer of a bloom, while the green identified area affected by a thinner layer of bloom (in relation to the thickest layer). The healthy concrete was represented by the red color. Considering that on visual inspection of the in the underside of the slab of the viaduct it was identified that the efflorescence presents three areas of concentration of different salts, we attempted to classify the intensity data generated by TLS in four intensity classes, as shown in image (d) of Figure 6 in order to observe how the algorithm detects this pathological manifestation. The criteria for this classification were: healthy concrete, efflorescence (thicker layer), efflorescence (middle layer) and efflorescence (less thick layer). In (d) also in Figure 6 it was found that the thicker layer of efflorescence was identified by the green color, while the middle layer has been identified by the light blue. The thinner layer of efflorescence, which is the transition zone between the pathologic manifestations and wholesome concrete, has been classified by the red color. The healthy concrete appears in purple. It can be seen in Figure 6 that the classified image in four intensity classes was best approached the real situation of the area affected by pathological manifestations in the structure. In this scanning the TLS was also able to detect by return intensity data efflorescence and humidity in the concrete surface below the viaduct in study. The data classification coming from 884
the scanning by the classifier algorithm enabled the identification of pathological manifestations incidents in SA. 4.3 Discussion of results The results showed that in the two scanning performed in the studied SA the TLS was able to identify pathological manifestations from the different intensities of the return pulse reflected by the materials that were deposited on the surface of the concrete. The classification of the TLS intensity data by the classification algorithm enabled the identification of the areas affected by biodeterioration and efflorescence. Similarly, studies by Armesto-González et al. (2010) and González-Jorge et al. (2012) had already achieved satisfactory results in the identification of pathological manifestations in buildings using the intensity data TLS together with the classification methods. The scanning time, forty minutes for the first case and twenty two minutes for the second one, showed the speed in acquiring information through the TLS system in small areas. 5 CONCLUSION The study demonstrated that the use of pulse return intensity data acquired in the scanning of the TLS when combined with the classifier algorithms, identify satisfactorily the areas affected by the pathological manifestations as biodeterioration, stains caused by humidity and efflorescence on the surface of the concrete SA. Because it allows viewing and scanning of a structure without direct contact with it, and due to the fact that this equipment has capacity range in kilometers, the TLS would be an option in case of inspections SA’s whose pillars, beams or board were inaccessible to contact the inspector. From the analysis of the results obtained in this article, it is concluded that there is a great potential to use the TLS in the identification of pathological manifestations in the concrete. These techniques and methodologies can be used in a complementary way to the current methodology SA’s inspection to detect biodeterioration, humidity and efflorescence on bridges and viaducts. Although applied to bridges and viaducts, the methodology adopted in this study can be extended to other structures or concrete buildings addressing other manifestations pathological incidents in the structures and quantifying them through the development of other computer algorithms. REFERENCES Armesto-González, J., Riveiro-Rodrigues, B., González-Aguilera, D. & Rivas-Brea, T. 2010. Terrestrial laser scanning intensity data applied to damage detection for historical buildings. Journal of Archaeological Science 37(12): 3037–3047. Bosché, F. 2010. Automated recognition of 3D CAD model objects in laser scans and calculation of as-built dimensions for dimensional compliance control in construction. Advanced Engineering Informatics 24(1): 107–118. Bott, T. R. 1993. Aspects of biofilm formation and destruction. In Letanision, R. M. & Eliaz, N. (eds), Corrosion Reviews 11: 1–24. Cánovas, M. F. 1988. Patologia e terapia do concreto armado. São Paulo: Pini. González-Jorge, H., Gonzalez-Aguilera, D., Rodriguez-Gonzalvez, P. & Arias, P. 2012. Monitoring biological crusts in civil engineering structures using intensity data from terrestrial laser scanners. Construction and Building Materials 31: 119–128. Helene, P. R. L. 1992. Manual prático para reparo e reforço de estruturas de concreto. São Paulo: Pini. Hueck, H. J. 2001. The biodeterioration of materials-an appraisal. Internatonal Biodeterioration & Biodegradation 48(1–4): 5–11. Longley, P. A., Goodchild, M. F., Maquire, D. J. & Rhind, D. W. 2013. Sistemas e ciência da informação geográfica. Porto Alegre: Bookman.
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Pereira, L. M. 2012. Avaliação das patologias e da biodeterioração na Biblioteca Central da UFSM Dissertação (Mestrado em Engenharia Civil) – Programa de Pós-Graduação em Engenharia Civil e Ambiental, Universidade Federal de Santa Maria: 1–127. Rabah, M., Elhattab, A. & Fayad, A. 2013. Automatic concrete cracks detection and mapping of terrestrial laser scan data. Journal of Astronomy and Geophysics: 1–6. Silva, M. S. & Rosowsky, D. V. 2008. Biodeterioration of construction materials: state of the art and future challenges. Journal of Materials in Civil Engineering 20(5): 352–365. Uemoto, K. L. 1988. Patologia: Danos causados por eflorescência. In Tecnologia de Edificações, Coletânea de trabalhos da Div. De Edificações do Instituto de Pesquisas Tecnológicas do Estado de São Paulo. São Paulo: PINI.
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Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
Management of the M4 Elevated Section substructures C.R. Hendy & C.T. Brock Atkins, Epsom, UK
A.D.J. Nicholls Connect Plus (M25) Ltd, UK.
S. El-Belbol Highways Agency, UK
ABSTRACT: The M4 Elevated Section in West London is a 1.9 km concrete viaduct structure providing a major arterial route into London. An intervention model bringing together structural assessment and forecast deterioration, corrosion modeling and cracking has been developed to prioritise structural rehabilitation of the substructure crosshead beams. To evaluate residual strength an initial assessment of crossheads was undertaken, which identified a deficiency in tensile capacity at the ends of the crosshead cantilevers compared to current demands. Further assessment has been undertaken, including three dimensional strut and tie, non-linear finite element analysis and plastic analysis, to confirm public safety with continued trafficking of the structure, and to determine the need for strengthening. Additionally extensive monitoring of the crossheads has been implemented including mapping of all cracks, and remote crack monitoring to safeguard the substructures and allow early interventions. The long term maintenance strategy brings together strengthening and cathodic protection concrete preservation methods with removal and repair of concrete delamination. This paper discusses the development of the prioritisation process including deterioration modeling, together with the extended structural analysis to formulate a strengthening programme for these substructures. 1 INTRODUCTION 1.1 Overview of the elevated structure The M4 motorway in West London is supported on an elevated reinforced concrete viaduct 1.9 km in length. The elevated section carries the M4 dual two lane motorway above and parallel to the A4 dual carriageway, forming a major arterial route in to central London. The structure incorporates 103 concrete crossheads supporting the mainline and a further 23 crossheads supporting the Junction 2 slip-roads. The elevated structure comprises simply supported decks, typically with 16.2 m span lengths, formed of pre-stressed concrete beams with an in-situ concrete deck. The substructures are typically single pier reinforced concrete with cantilever crossheads supporting the decks via elastomeric bearings. The crossheads incorporate nibs that extend out from the lower section of the crosshead cantilevers, forming the bearing shelves. The central section of the crosshead extends up to the top of deck level and directly supports the carriageway between spans. Asphaltic plug deck joints are provided between decks above the upstand. 1.2 Strategic management principles The M4 Elevated Section forms part of the M25 Design Build Finance and Operate (DBFO) contract. The DBFO concessionaire is Connect Plus, who have a contract with the Highways Agency to maintain the project network for the 30 year contract duration. 887
Figure 1.
M4 Elevated substructures.
Figure 2.
Existing reinforcement.
The key contractual obligation for Connect Plus is to hand back the M4 Elevated Section at the end of the 30 year contract in a no worse condition than that agreed at a condition assessment completed at contract commencement. The success in achieving these requirements will be measured through expenditure and lane closure tests where it is required that any works in the 15 years following contract completion should not exceed 125% of those measured in the final 15 years of the contract. To achieve these requirements it is essential to understand the current condition of the crossheads, its change with time and external factors such as drainage. It was therefore necessary to develop a strategy for the management of the M4 elevated substructures over a 45 year period. The objective of the strategy was to maximise the operation of the assets and optimise total spend by evaluating the appropriate points at which interventions should be undertaken. Availability of the highway was a key factor because the contract includes financial penalty for periods of unavailability. The best theoretical strategy for the M4 elevated was initially found to be to schedule maintenance works interventions only where they became necessary to maintain structural capacity. However, as discussed in section 2, it soon became clear all crossheads first needed strengthening to meet current demands and this needed to be implemented ahead of operation of the long-term maintenance strategy. Other considerations affected long-term maintenance prioritisation including ease of access to different crossheads and control of other risks to public safety such as that posed by spalling concrete. The time when refurbishment would become necessary was determined. This process brought together corrosion modeling to estimate the degree of reinforcement corrosion and strength loss with the overall structural assessment. From chloride sampling taken from the M4 elevated crossheads the corrosion modeling evaluated the time duration for the level of chlorides to reach 0.3% of cement content at the depth of reinforcement. This threshold level was considered to be the point where active corrosion of the reinforcement was instigated and an assumed rate of corrosion was applied to the affected bars (BRE 2009). To develop the long-term maintenance strategy it was necessary to put in place a system to prioritise intervention. A scoring system was therefore developed considering primary deterioration indicators on which each substructure could be evaluated. This considered in particular: – – – – – –
structural capacity compared to demand predicted deterioration rate (based on chloride levels and half-cell survey results) concrete delamination extents crack extents existence or otherwise of cathodic protection traffic management requirements, access issues and disruption to the travelling public.
Each of these characteristics was weighted depending on the level to which they dictate the intervention prioritisation of the substructures. A threshold level was determined representing a level of deterioration where intervention is necessary and a time to intervention for each crosshead, and an indicative works programme was produced. 888
Figure 3.
Cracks at crosshead cantilever ends.
Figure 4.
Concrete delamination on crossheads.
1.3 Current condition The crosshead substructures have been suffering from extensive corrosion and concrete delamination caused by salt water ingress (from winter maintenance) through deck joints, evidenced by extensive water staining – Figure 4. As part of continued maintenance strategy to the structure, several rehabilitation and prevention methods have been employed including removing large areas of delaminated cover concrete, concrete repair, and the application of cathodic protection and netting. Site testing has confirmed the chloride concentration is significantly in excess of the 0.3% threshold associated with high risk of active corrosion occurring (up to a maximum of ∼8%). The presence of reinforcement corrosion is further supported by half-cell potentials taken from the structure exceeding −350 mV (CET Safehouse 2012), indicating a greater than 90% probability of corrosion occurring. Some cracking has occurred on many of the crosshead ends concentrated at the re-entrant corners – Figure 3. This location acts as a halving joint and is subject to high stress concentration, therefore some degree of cracking is expected. However the crack extents needed to be carefully monitored. In some locations the cracking has propagated between the re-entrant corners to ‘join up’. This behavior indicates a progress loss of concrete tensile strength and subsequent transfer of loading on to the available vertical reinforcement which is of concern as discussed in section 2.1. Due to the shape of the crossheads, it is impossible to inspect the condition of the concrete faces adjacent to the deck ends. Principally testing has focused on the accessible vertical faces and crosshead soffits, both of which are typically in poor condition. It is considered possible that conditions at the inaccessible faces are suffering at least the same level of corrosion as the visible faces.
1.4 Deck joints Water penetration through the deck expansion joints has been reported since the mid 1980s. A programme of expansion joint replacements to all crossheads has recently been completed which involved removing the life expired asphaltic plug joints (APJ) and replacing these with high modulus asphaltic plug joints with improved whole life cost performance. The original design had a discrete APJ above each expansion joint. The replacement joints extend over both expansion joints across the full width of the crosshead. Provision for subsurface drainage in the joints was improved but subsequent surveys have shown significant amounts of water continuing to penetrate the deck joints, probably beneath the surfacing and under the joint, resulting in on-going contamination of the crossheads. An additional problem is that the APJ extends only between kerblines; water can thus penetrate through the verges and central reserve. 889
1.5 Control of concrete spalling Significant areas of concrete delamination are present on the elevated substructures. The potential for delaminated concrete to fall on to the A4 dual carriageway directly beneath the M4 elevated section has been taken seriously and carefully assessed to mitigate this risk. To overcome the risk from falling concrete, a netting system has been installed around the crosshead cantilevers. This netting is designed to catch the full cover concrete without significant deflection or damage to the netting occurring. Additionally a regular regime of delamination surveys has been put in place to identify areas of deteriorated concrete before they become delaminated from the substructures. These delamination surveys are carried out on a three month basis with interim six weekly surveys carried out on crossheads with the worst rate of delamination particularly during the winter months where the risk of delamination is greater due to freeze-thaw action. Further mitigation is provided through regular walk through inspections examining the netting for any captured concrete, which would indicate acceleration in deterioration. Loss of concrete section has also exposed large areas of reinforcement as shown in figure 3, exposing reinforcement to accelerated atmospheric corrosion. 2 SUB-STANDARD STRUCTURE MANAGEMENT STRATEGY 2.1 Initial structural assessment An initial assessment of the crosshead structures was undertaken based on record drawings to determine live load capacity and its sensitivity to reinforcement corrosion. The assessment demonstrated that the majority of each crosshead was able to carry 40 tonne assessment loading, but the light vertical reinforcement in the end 3 m of each cantilever was overstressed. The location of the bearing shelves below the top of the crosshead, where the main flexural reinforcement is provided, necessitates load transfer up to this reinforcement. The vertical reinforcement provided at the ends of the cantilevers was found to be inadequate to cater for current loading demands. Additionally, current codified minimum reinforcement requirements were not met. This made the presence of cracks as shown in Figure 2 a concern. 2.2 Management strategy formulation Following the conclusions of the initial structural assessment a rigorous management regime was implemented in accordance with BD 79/06 (Highways Agency 2006). This regime included a thorough examination of the likelihood and implications of non-intervention, together with instigating a detailed inspection regime focusing on the ends of the crossheads where the start of any potential failure mechanism would first manifest itself. The management regime focused on a risk based approach to protecting public safety, putting in place defined trigger levels in the event any evidence of structural deterioration was identified. The following actions were undertaken to ascertain the optimum management strategy for the M4 elevated substructures: – Extended analysis considering three dimensional strut and tie analysis, non linear concrete modelling and considerations of reinforcement strain hardening. – Assessment of ductility of potential failure mechanisms and hence the amount of warning these would give. – Detailed inspection of the substructures at cantilever ends to identify evidence of initiation of a failure mechanism. – Analysis of the impacts and disruption to the travelling public in the event of emergency closure, including the elevated accident risk associated with traffic diversions. – Design and procurement of contingency measures to be mobilised immediately in the event evidence of accelerated structural deterioration was identified. 890
2.3 Refined structural analysis Refined analysis was undertaken to optimise the predicted load carrying capacity of the crossheads and to inform behaviours mechanism. Extended analysis considering both strut and tie modeling and non-linear concrete modeling was undertaken. The strut and tie analysis considered optimisation between two load paths available at the end of the crosshead. The first load path represented the transfer of load vertically from the bearing locations to the flexural reinforcement at the top of the crosshead section (refer to diagram 5). The second strut and tie load path considered took advantage of concrete compression struts in the lower portion of the crosshead to transfer the bearing loads to the bent up bars closer to the pier. This load path reduces the load required to be carried on the vertical links but requires a tensile tie to be available directly below the bearing shelves. Record drawings indicate there is only one 25 diameter bar located at the top of the nibs which could provide this resistance. However cover meter surveys confirmed that the bars curtailed without end anchorage as indicated on the record drawings. The load capacity of the crosshead cantilever ends was optimised by superimposing the resistances provided by both load paths. However the contribution of the second load path is limited by the contribution of the longitudinal tension bar within the nibs, therefore it was not possible to generate significant benefits over and above the predictions of the initial analysis. In tandem a non-linear concrete model of the crossheads was developed to investigate potential strength benefits and to provide greater certainty as to how a failure mechanism would ultimately develop. The model (Figures 7 and 8) incorporated all reinforcement and considered a rigid connection to the concrete, therefore not taking in to account bond characteristics. The analysis was performed in Abaqus using its ‘concrete damaged plasticity’ model for concrete behavior. The analysis considered tensile cracking, tension softening and compressive crushing of the concrete. This analysis showed that for relatively small tensile concrete characteristic strength (∼1 N/mm2 ) the concrete was able to transfer the bearing loads and remained uncracked at the end of the cantilever. Model failure scenarios occurred at the cantilever root at a much higher load than applicable. This was not considered realistic because the model did not properly take account of the potential brittle fracture of the concrete arising from high stress concentrations at the crack tip, nor the presence of pre-existing shrinkage cracks or voids. To reduce the influence of concrete tensile strength, the analysis was re-run for lower concrete tensile strengths which then gave very similar results to the strut and tie approach, as would be
Figure 5.
Strut and tie analysis load path 1.
Figure 6.
Strut and tie analysis load path 2.
Figure 7.
Non-linear FE crosshead modelling.
Figure 8.
Non-linear FE crosshead analysis.
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expected since it ignores concrete tensile strength. For low concrete strengths, cracks propagate at the crosshead ends under lower loading conditions, resulting in redistribution of force to the vertical bars in the crosshead ends and ultimately ductile failure at a load lower than that required. For higher concrete strengths, failure occurs above the required load but is governed by concrete tensile failure and is hence brittle. This resistance model could not be relied upon and hence adequate strength of the crossheads could still not be demonstrated. 2.4 Consideration of overall failure mechanism and strain hardening The refined analyses in section 2.3 still focused on the crosshead in isolation and not the potential for any redistribution of load to elsewhere in the structure upon localised failure. To understand the additional reserve of strength, the combined superstructure and substructure was considered as a single structural system. A ductile failure, ignoring concrete tensile strength, occurring anywhere in the final 3.0 m of a crosshead cantilever was considered. This part of the crosshead supports the verge, 600 mm hard strip and a proportion of the inside traffic lane (the M4 motorway having no hard shoulder in this location). In the event the cantilever ends displace, and support is lost to the outer deck bearings, the deck itself can span transversely due to the presence of end diaphragms and concrete infill between the beams. In reality, the ductility of the vertical steel in the crossheads means that the two mechanisms can be superposed, thus adding strength in excess of the predictions of section 2.3. Reinforcement test data also concluded the ultimate steel strength is greatly in excess of the yield strength by 45% as a minimum. When strain hardening was additionally considered it was possible to demonstrate that the full traffic loading could be carried at the ultimate limit state, albeit with significant plastic deformation and damage occurring to the crosshead cantilever and adjacent decks leaving the structure unserviceable and very expensive to repair. Whilst this gave confidence overall collapse would not occur, a residual concern that lumps of concrete might spall under the high strains and fall on to the A4 below was considered. This risk was however mitigated by the presence of the protective netting described in section 1.5. In light of this analysis and mitigation, the overall risks to public safety from continued operation in the short to medium term were considered to be low if suitable monitoring was also put in place as discussed in section 3. However, it was concluded strengthening was required to maintain adequate reliability long-term. This is discussed in section 4. 3 STRUCTURAL MONITORING A detailed regime of structural monitoring was put in place to identify early signs of structural deterioration to allow intervention before damage occurs. Monitoring has been implemented on a risk based approach with focus being on the piers showing the greatest cracking. A series of trigger levels was defined based on risk level. DEMEC studs have been installed at all re-entrant crack locations and a 3 monthly programme of re-measurement has been implemented. Where cracking extends the full distance between re-entrant corners the affected crossheads are classed as high risk and real time monitoring devices have been installed over the cracks. These devices take regular measurements of displacements parallel and perpendicular to the cracks and the data can be remotely access through a web based portal. The monitoring data is constantly interrogated to identify any signs of progressive deterioration through steady increasing crack length and width. Typically the monitoring shows a very high correlation with temperature affects, with both daily and seasonal changes observed. Various methods have been used to remove these temperature affects with varying degrees of success. They do not prevent long term trends being identified, which in the case of some crossheads do show progressive crack increase. Trigger levels for crack width (based on localised reinforcement yielding) have been set. If they are exceeded, temporary interim strengthening measures have been designed and fabricated and are available to be installed rapidly. In accordance with the requirements of BD 79, strengthening of all the deficient regions is programmed within three years. 892
4 STRENGTHENING AND PROTECTION To provide adequate reliability over the residual life of the structures, localised strengthening is required to the piers. It was essential to produce a cost-effective solution which would additionally have minimum impact on the travelling public during installation. The strengthening method additionally was developed to minimise any deterioration of the recently replaced expansion joints (thereby not inadvertently leading to greater water ingress of reinforcement corrosion). The strengthening (Figure 9) comprises installing, by drilling and fixing, new reinforcement both transversely and vertically to supplement load path 1, shown in Figure 5, at the ends of the cantilevers. The requirement not to damage the expansion joints or close the M4 during the works meant the vertical bars could not be anchored on the top of the crosshead. Therefore the vertical bars are resin anchored as high as feasible in the concrete section and additional longitudinal reinforcement installed where the vertical bars are fully anchored. This combination of additional reinforcement provides a truss system to transfer the bearing loads at the ends of the cantilevers to the sections of greater reinforcement. To minimise disruption to the traffic on the A4 below, all works were undertaken during nighttime. As the crosshead concrete is contaminated with high levels of chlorides the maintenance strategy needed to take future corrosion of reinforcement in to consideration. To address chloride induced corrosion, stainless steel reinforcement has been specified which has enhanced protection for high chloride environments. Additionally, resin grout has been specified both to take advantage of its greater bond strengths and its chemical isolating properties. Fully grouting the additional reinforcement will minimise the risk of chloride attack and the stainless steel will further mitigate the extent of future deterioration. A key issue in the strengthening installation was the drilling of holes for new reinforcement insertion without damaging the existing reinforcement. This led to the need for both a prescriptive drilling protocol, requiring core holes to be abandoned and relocated in some situations, together with the addition of temporary external longitudinal pre-stress applied by pre-stressing bars anchored to a steel frame at each end of the crosshead. This pre-stressing force supplemented load path 2 shown in Figure 6 (by increasing the resistance of the horizontal tie member) and thus provided mitigation against damaging the sparse existing vertical reinforcement during drilling. The layout of core holes was carefully designed to minimise the potential for hitting reinforcement and 3D modeling proved helpful in this regard as discussed in section 6. 5 BUILDING INFORMATION MODELLING Building Information Modelling (BIM) was used during the interim planning and strengthening design phases and continues to be used for ongoing asset management of the crossheads. The geometry of all the crossheads and the surrounding carriageway below the viaducts was laser scanned and a 3D model created from the point cloud data – Figure 10. This provided valuable measurements of the as-built structure to cross-check against the record drawings. It also allowed standby contingency emergency temporary propping systems (for use in the event that monitoring trigger levels were exceeded) to be visualised and developed such that their founding locations on the A4 carriageway below had the minimum impact on traffic flows and safety.
Figure 9.
Strengthening reinforcement to be installed within the crosshead cantilevers.
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Figure 10.
Laser scan image.
Figure 11.
BIM – variation in chloride levels.
At the detailed design phase for the strengthening, as-built reinforcement was incorporated into the 3D models. These models were then used to route virtual core holes for the new reinforcement and cathodic protection anodes through the crosshead in locations to minimise clashes with existing reinforcement. The BIM model contains numerical and visual attached data relevant to the ongoing asset management of the viaducts to allow maintenance to be prioritized and planned. Each crosshead is split into smaller uniquely referenced surfaces to which condition data is attached including chloride levels, half cell readings, strength utilization, estimated reinforcement section loss, crack widths and presence of spalling and exposed reinforcement. The data can be viewed numerically or by colour scale in Navisworks (see Figure 11) and an embedded deterioration model in the source data allows predictions of this condition data to be viewed with time. 6 CONCLUSION A management strategy in accordance with BD 79 has been instigated, which has considered further analysis including strut and tie and non-linear concrete modeling to understand to extents of this deficiency and the anticipated consequences of not intervening. Plastic analysis incorporating the post-yield behavior of the reinforcement and transverse contribution of the decks was used to justify continued trafficking of the structure together with rigorous monitoring to identify structural deterioration early. However, strengthening is required to maintain long-term performance. A comprehensive prioritisation process has been developed, including deterioration modelling, to formulate a programme both for the short term strengthening and longer term intervention requirements for these substructures. Concrete rehabilitation methods have been evaluated and a programme of cathodic protection combined with concrete repairs has been developed on a risk basis to meet the handback requirements discussed in Section 1.2. A BIM model was developed to assist a standby emergency propping and strengthening design and to provide a visual representation of the maintenance strategy, allowing the user to see how the substructures deteriorate over time and when intervention is required. Additionally the BIM model provides an interactive and intuitive record system. REFERENCES Highways Agency. 2006. BD79/06: Design Manual for Roads and Bridges Volume 3: Section 4: Part 18. The Management of Sub-Standard Structures. Lee, J., & G. L. Fenves. 1998. Plastic-Damage Model for Cyclic Loading of Concrete Structures, Journal of Engineering Mechanics, vol. 124, no. 8, pp. 892–900. Lubliner, J., J. Oliver, S. Oller, & E. Oñat. 1989. A Plastic-Damage Model for Concrete, International Journal of Solids and Structures, vol. 25, pp. 299–329. Nokken, M., Boddy, A., Hooton, R.D., & Thomas, M.D. 2006. Time dependent diffusion in concrete—three laboratory studies, Cement and Concrete Research, Vol. 36, pp. 200–207 Yong Ann, k. & Song, h. 2007. Chloride threshold level for corrosion of steel in concrete, Corrosion Science, vol. 49, pp. 4113–4133.
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Using data mining and numerical simulations for on-line monitoring of long span bridges J. Santos & P. Silveira LNEC, Portugal
C. Crémona Cerema, France
A. Orcesi IFSTTAR, France
L. Calado Instituto Superior Técnico, Portugal
ABSTRACT: Performance control of brides based on structural health monitoring has been carried out as an a posteriori task, where information extracted in situ is compared with reference baselines which must be acquired while the structures are undamaged and unchanged. In addition, the vast majority of structural health monitoring techniques relies either on numerical or data-driven techniques. While the first type is computationally inefficient to be applied in on-line (real-time) strategies, the second type generally provides information with lesser detail, which may allow detecting structural changes but generally does not allow locating or classifying them. The present work presents an innovative on-line methodology with the ability to detect, locate and classify structural changes without resorting to baseline comparison. It consists of combining data-driven techniques and numerical analysis, and its effectiveness is validated, in the present paper, by applying it to time-history numerical simulations of the International Guadiana Bridge, located in the southwest of the Iberian Peninsula.
1 INTRODUCTION Structural health monitoring (SHM) can be defined as the development and implementation of damage identification strategies in structural systems, civil engineering or other (Farrar & Worden, 2007). Ideally, these strategies should allow identifying damage before it reaches a critical state (early-damage), which demands for the development of on-line strategies, capable of real-time identification. This topic has been addressed using two distinct approaches (Doebling, Farrar, Prime, & Shevitz, 1996): forward, which relies on data mining to extract meaningful information from data acquired in situ, and inverse, which is also known as model updating of system identification, and which consists of fitting numerical models to the data acquired on site (Glaser & Tolman, 2008). The first type of approaches generally leads to non-linear and non-unique solutions, and therefore is generally coupled with optimization strategies to deal with computational complexity. Therefore, they are not appropriate for on-line, or real-time damage identification. The approaches classified as forward are generally much simpler than inverse ones and are therefore more appropriate for on-line applications (Giraldo, 2006; K. Worden & Dulieu-Barton, 2004). However, they cannot extract structural information from data, and therefore only allow detecting damage but fail to locate it or classify it (unless unrealistic amounts of sensors are used). The present work describes an original approach for identifying damage in large-span bridges. It consists of combining data mining techniques with numerical simulations so as to obtain a 895
computationally simple strategy which is appropriate for on-line identification but which is also capable not only of detecting damage, but also of locating and classifying it. The strategy is divided into two major tasks, being the first the detection of damage and the second the location and classification of the previously detected damage. Both tasks rely on sequentially applying neural networks and clustering methods to the data acquired in situ so as to obtain increments that are robust enough to be compared with their numerical counterparts in real time. After this brief introduction, section 2 describes the theoretical concepts used herein and section 3 the case study. Section 4 describes the original methodology and, finally, in section 5 several conclusions are drawn. 2 THEORETICAL CONCEPTS 2.1 Data normalization using neural networks On-line strategies must be as statistically robust as possible so that identification can be relied upon and free from outliers and false detections. The first step towards robustness is named in the literature as data normalization and consists of suppressing environmental and traffic effects from the data acquired on site, since changes generated by these effects are frequently of higher magnitude that those generated by damage (Sohn, Worden, & Farrar, 2001; Keith Worden, Farrar, Manson, & Park, 2007). Normalization of monitoring data responses is addressed herein using the feed-forward multi-layer perceptron (MLP) algorithm with a single hidden layer, which is the most used neural network algorithm in fields addressing data-driven analysis (Bishop, 2006). Herein, the MLP is defined as a nonlinear mapping between a set of structural responses (identified as y in Figure 1 and in Eq. (1) and a set of actions’ measurements (identified as t). The choice of this algorithm over simpler ones such as the multivariate linear regression was based on the capability of the MLP in not requiring any assumption about the intrinsic relations between actions and measured effects. Instead, the algorithm is capable of learning them by itself. The MLP estimation models used herein comprises a number of inputs equal to the amount of temperatures measured on site, and one output consisting of a structural response measured on site. Hence a number of neural networks equal to the number of structural measurements was defined, following Eq. (1) and Figure 1,
where u(II ) and u(III ) are the sets of coefficients defining the neural network, T is the input temperatures, y the estimated structural quantities and re the set of residual errors, which are defined as the difference between real data and estimations. Non-linear perceptron units located in the hidden layer are non-linear while the remaining ones are linear. The optimal number of hidden layers is defined by testing multiple architectures and
Figure 1.
MLP graph model.
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choosing the one leading to smaller average estimation error. Over-fitted solutions are avoided using quadratic regularization (Bishop, 2006). The definition of the networks’ coefficients is made through the gradient descent optimization algorithm, repeated 10 times for each variable to avoid ill initialization. 2.2 Information extraction using symbolic clustering methods Clustering methods consist of unsupervised statistical learning algorithms capable of classifying data objects as members of different subsets (or clusters), such that those within each cluster are more similar to one another than those assigned to different clusters. The aim of a clustering method can be mathematically posed as (Hastie, 2011) the attempt to minimize the dissimilarity between data objects assigned to the same cluster (within-cluster dissimilarity) and, consequently, maximize the dissimilarity between objects assigned to different clusters (between-cluster dissimilarity). Considering a given partition containing K clusters, PK = {C1 , . . . , CK }, the overall within-cluster dissimilarity W(PK ) can be defined as (Cury, Crémona, & Diday, 2010)
where c(i) is a many-to-one allocation rule that assigns object i to cluster k, based on a dissimilarity measure, dij , defined between each pair of data objects, i and j. The overall dissimilarity of a data set, OD, is defined in Eq. (3), where N is the total number of objects. The between-cluster dissimilarity is obtained by subtracting the other two, B(PK ) = OD − W(PK ).
The most well-known and used clustering algorithm is the k-means (Hastie, 2011), which addresses the problem of minimizing the overall within-cluster dissimilarity, W(Pk ), of a given data partition, Pk , through an iterative optimization scheme. Its popularity and reported effectiveness have motivated its use herein. The k-means requires that the number of K < N clusters be initially defined (Hastie, 2011) along with a randomly defined set of K clusters’ prototypes, which are objects of the same type as those being clustered. This task is known as initialization and is exemplified in Figure 2a, where the larger dots consist of the randomly defined prototypes while the smaller dots represent the objects being clustered. Following the initialization phase, each iteration starts by allocating the objects to the clusters according to an allocation rule, c(i). This step
Figure 2. K-means clustering algorithm: (a) initialization (three cluster prototypes to describe eleven objects); (b) iteration 1, allocation; (c) iteration 1, representation; (d) iteration 2, allocation; (e) iteration 2, representation, (f) final set of clusters’ prototypes.
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is represented in Figure 2b, where the dashed lines show the allocation of each object into each cluster. The k-means allocates each data object to the less dissimilar cluster prototype,
where k is the index identifying one of the K clusters belonging to Pk , and dik is the dissimilarity between object i and cluster prototype k. The second step of each k-means’ iteration is named representation and consists of finding the set of K prototypes that best represents the clusters defined in the allocation phase. The k-means is based on representing the clusters by their centroids (new prototypes shown in Figure 2c). These two steps, allocation and representation, are afterwards repeated in Figure 2d,e) until an objective function, which depends on cluster compactness and separation, reaches its global minimum value (Hastie, 2011). The k-means takes the squared within-cluster dissimilarity measured across the K clusters as objective function (Hastie, 2011). Clusters’ dissimilarities are generally defined as distance metrics. Among these, the Euclidean (square root of the sum-of-squares) is, by far, the most used. In the realm of SHM, these metrics allow quantifying the dissimilarity between pairs of SHM measurements acquired at distinct timeinstants. However, the application of clustering methods to the measurements themselves leads to poor robustness and lesser ability to identify changes in real-time. To overcome these difficulties, the Gowda-Diday symbolic dissimilarity measure (Billard & Diday, 2006; Gowda & Diday, 1991) was used herein. Instead of quantifying the dissimilarities between pairs of n single measurements, this metric allows quantifying the dissimilarities between pairs of measurement samples. By clustering the samples instead of the measurements themselves, the influence of outliers and the computational complexity are greatly reduced. Considering the samples acquired from p sensors (1);T
(1)
(r)
(r)
(p)
(p)
during the time period i, the p intervals are Ti = (Ti,inf i,sup ) . . . (Ti,inf ; Ti,sup ) . . . (Ti,inf ; Ti,sup ). Bearing this notation in mind, the Gowda-Diday distance (Gowda & Diday, 1991) between two sets of p samples, i and j, is given by
where ϕr is obtained for each of the p interquartile intervals as
Additional element on this dissimilarity are thoroughly described in (J. Santos, Crémona, Orcesi, & Silveira, 2013; J. Santos, Orcesi, Crémona, & Silveira, 2015; J. P. Santos, 2014). In the present work, clustering methods inputted with the symbolic dissimilarity were used as powerful and effective tools to extract very compact and robust information from bridge monitoring data, following the works presented in (J. Santos et al., 2013, 2015; J. P. Santos, 2014). After temperature suppression, using the neural networks presented in the previous subsection, clustering methods are applied to reduce an entire data set, composed of an arbitrary number of samples described by the interquartile intervals, into a set of cluster centroids, each describing the data allocated to the corresponding cluster. Afterwards the average distance between all prototypes, DC,
is computed as a robust and compact descriptor of the variations observed in the measurements of a single sensor or of multiple sensors combined (Eq. 6 combines data from p sensors), being k ¯ j ) the symbolic dissimilarity between clusters the number of cluster within a data partition, d(C¯ i , C ¯ i and j, and Ci the centroid of cluster i. 898
3 CASE STUDY The structural system studied herein is the 666 m long cable-stayed International Guadiana Bridge (Figure 3a), located in the south-west of the Iberian Peninsula. It comprises a central span of 324 m and two lateral and transition spans of 135 m and 36 m, respectively. The deck is a prestressed concrete box girder 18 m wide and 2.5 m high (Figure 3d), suspended by one hundred and twenty eight stay cables (Figure 3a) with lengths varying from 48 m to 167 m. The A-shaped pylons (Figure 3e) are 95.1 m and 96 m high and support the deck at a height of 35 m. The support devices located on the abutments, pylons and piers allow for horizontal movement and T-shaped steel devices connect piers P1 and P4 to the deck, to prevent uplift generated by important live loads travelling on the central span. Each pylon’s foundation is composed of 13 concrete piles rigidly connected by concrete caps (Figure 3e). The piers P1 and P4 as well as the abutments are supported by shallow foundations (Figure 3a). Cables are numbered from Spain to Portugal, from 1 to 128. The structural health condition of the Guadiana Bridge has been controlled, since the beginning of its life cycle (1991), through periodic visual inspection and manual measurement of strain and temperature, using embedded sensors installed during the construction. Up to the present, no damage was identified. However, this conclusion is based on quarterly measurements/inspections and limited to physical phenomena of important magnitude. The ageing of the structure and its increasing social and economic importance have motivated the installation of the monitoring system presented in Figure 3, which was developed so as to allow controlling three types of damage scenarios to which the structure is considered particularly vulnerable: scour on the pylons’ foundations (due to the soft soil and important river stream), damage in the stay-cables (generated by corrosion or vehicle collisions) and increase of friction forces in the support devices (due to their ageing). Data is being synchronously acquired from the Guadiana Bridge at each hour, since the 1st of February 2011, and sent over cellular network at 3G rates to a database server which checks its integrity and consistency upon receive, as described in (J. Santos, Silveira, Santos, & Calado, 2010; J. Santos & Silveira, 2012). Since no damage was observed in the bridge throughout its life-cycle, the testing and validation of the methodologies proposed herein was conducted by simulating damage using the three-dimensional numerical model show in Figure 4d.
Figure 3.
Guadiana International Bridge: overview and sensorial system.
Figure 4. Numerical simulation of damage scenarios: (a) temperature measured on site, (b) tilt-meters noise, (c) displacement noise, (d) numerical model, (e,f) simulated rotations and displacements.
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Figure 5. DI distributions over 100 noise simulations and for stiffness reductions in stay-cable 71 of: (a) 0%, (b) 1%, (c) 2%, (d) 5%, (e) 10%, (f) 20%.
The numerical simulations of damage scenarios were conducted on time-histories and aimed at obtaining the structural quantities being measured in situ on the exact locations of the real sensors. To truthfully reproduce these structural responses, only experimental data was used as input: the temperature action (obtained using the embedded thermometers installed in situ, Figure 4a) and the noise outputted by tilt-meters and displacement sensors were inputted in the simulations, Figure 4b,c). These noise distributions were measured in situ and added to the series outputted by the numerical simulations. The dynamic effects of traffic and wind action were not considered in the simulations since these are being filtered, in real-time, on the Guadiana Bridge (J. Santos & Silveira, 2012). For the present work, damage scenarios consisting of stiffness reduction in six stay cables considered representative of the entire cable set were simulated. To ensure that the methodologies developed are capable of detecting early-damage, 1% to 20% stiffness reduction in a single stay-cable was simulated at a time. All damages were simulated instantaneously on the 1st of February 2013 (day 365 of monitoring). 4 DAMAGE IDENTIFICATION 4.1 Damage detection Damage identification was divided, in the proposed strategy, in two steps. The first aims at detecting damage and relies only on data-driven techniques and statistical hypothesis tests. It consists of obtaining samples at predefined time periods (assumed to be in accordance with the requirements of damage identification swiftness) and in computing the distance between clusters, DC, using data from all the sensors installed on site. The results obtained from this strategy in the present work can be observed in Figure 5, which shows values of a damage indicator (DI) defined herein as the difference between the cluster distance, DC, and a 99.9% confidence interval of the same distance (assuming Normal distribution). According to its definition, positive DI values indicate damage while negative values suggest an unchanged structural system. The values shown in Figure 5 are the result of one hundred noise simulations per damage scenario simulation, and are therefore presented in box-and-whiskers plots. From these it can be observed that damage as small as 5%, 10% and 20% of stiffness reductions is detected in the 24 hours following its occurrence, while smaller changes of 1% and 2% can be detected in the 48 hours following the damage occurrence. Samples of 24 hours were considered for clustering, thus providing a time resolution of 24 hours for the on-line detection procedure. 900
Figure 6. Increments defined as the distance between cluster (DC) of the data acquired from each sensor separately, and for damage scenario of stiffness reductions in stay cables (a) 66 and (b) 71.
Figure 7. MAC indices obtained between sets of DC obtained from time-history numerical simulations comprising real temperature and noise and sets of increments comprising the database of damage scenarios.
4.2 Damage localization and classification The procedure adopted for localizing and classifying damage is similar to the one described in the previous subsection concerning the processing of data acquired in situ. It consists of obtaining the distance between clusters from the monitoring data (considering samples of 24 hours of measurements). However, it differs from the previous in the fact that these distances are not obtained for the ensemble of sensors but for each sensor separately. Hence, instead of outputting a single indicator capable of describing the changes in an entire structure, the output consists of a set of indicators which are regarded as robust and reliable increments of the quantities measured in each sensor separately. For the case of damage scenarios in stay cables 66 and 71, these increments are shown in Figure 6, for stiffness reductions of 1% to 20%. The robust increments obtained from the data using the combination of neural networks and cluster analysis are then compared with a database of numerically generated scenarios. Unlike the truthful simulations described previously comprising noise and real temperature measurements, those conducted to define the database of damage scenarios consist of single-step static simulations and their output consists of a set of increments similar to those shown in Figure 6. Each different damage scenario considered results in a different increment set. Once a damage is detected, the corresponding increments are compared to all those included in the database. This comparison is made by computing the MAC index (J. P. Santos, 2014) between each DC set and each increment set included in the database. For the time-history simulations of stiffness reductions in stay cables 66 and 71, the MAC indices are shown in Figure 7, where it can be observed that the highest MAC indices were obtained for the database entries corresponding to the stay cables 66 and 71, respectively for Figure 7a and b. The database was defined with 128 damage scenarios, each comprising the reduction of one stay cable of the Guadiana Bridge.
5 CONCLUSIONS This paper presents an original SHM strategy for conducting on-line detection, localization and classification of early-damage. It consists in sequentially applying the MLP neural network and the 901
k-means clustering algorithm to the monitoring data, discretized in samples of predefined timelengths, so as to extract robust indicators capable of accurately describing the changes observed in situ, and generated by damage. The proposed strategy was applied to numerical data generated in time-history analyses including noise and temperature measured on the Guadiana cable-stayed bridge. These simulations allowed concluding that damage with small magnitude and local character can be detected using the singlevalued indicator DC to describe an arbitrary amount of sensors. From the simulations, damage as small as 5% of stiffness reduction in a single stay cable was detected in the 24 hours after its occurrence, while 1% was detected in the 48 hours after. Localization and classification of damage was proposed herein through the comparison of sets of DC with increments comprising a damage database. By evaluating the fitness between both it was observed that once a damage occurrence is detected it can also be instantly located and classified. REFERENCES Billard, L., & Diday, E. (2006). Symbolic DataAnalysis. Merrill-Palmer Quarterly (Vol. 52, p. 321). Chichester, UK: John Wiley and Sons. Bishop, C. M. (2006). Pattern Recognition and Machine Learning. Pattern Recognition (p. 748). Berkeley, USA: Springer. Cury, A., Crémona, C., & Diday, E. (2010). Application of symbolic data analysis for structural modification assessment. Engineering Structures, 32(3), 762–775. Doebling, S. W., Farrar, C. R., Prime, M. B., & Shevitz, D. W. (1996). Damage Identification and Health Monitoring of Structural and Mechanical Systems from Changes in Their Vibration Characteristics: A Literature Review. Distribution (p. 134). Los Alamos, USA. Farrar, C. R., & Worden, K. (2007). An introduction to structural health monitoring. Philosophical Transactions. Series A, Mathematical, Physical, and Engineering Sciences, 365(1851), 303–315. Giraldo, D. F. (2006). Damage Detection Accommodating Varying Environmental Conditions. Structural Health Monitoring, 5(2), 155–172. Glaser, S. D., & Tolman, A. (2008). Sense of Sensing: From Data to Informed Decisions for the Built Environment. Journal of Infrastructure Systems, 14(1), 4–14. Gowda, K. C., & Diday, E. (1991). Symbolic clustering using a new dissimilarity measure. IEEE Transactions On Systems Man And Cybernetics, 24(6), 567–578. Hastie, T. (2011). The Elements of Statistical Learning, Data Mining, Inference and Prediction (2nd ed., p. 763). Stanford, USA: Springer. Santos, J., Crémona, C., Orcesi, A. D., & Silveira, P. (2013). Multivariate Statistical Analysis for Early Damage Detection. Engineering Structures, 56, 273–285. Santos, J., Orcesi, A. D., Crémona, C., & Silveira, P. (2015). Baseline-free real-time assessment of structural changes. Structure and Infrastructure Engineering: Maintenance, Management, Life-Cycle Design and Performance, 11(2), 145–161. doi:10.1080/15732479.2013.858169 Santos, J. P. (2014). Smart Structural Health Monitoring Techniques for Novelty Identification in Civil Engineering Structures. Instituto Superior Técnico – University of Lisbon. Santos, J., & Silveira, P. (2012). A SHM framework comprising real time data validation. In IALCCE 2012 – 3rd International Symposium on Life Cycle Civil engineering. Vienna, Austria: IALCCE – International Association for Life-Cycle Civil Engineering. Santos, J., Silveira, P., Santos, L. O., & Calado, L. (2010). Monitoring of Road Structures – Real Time Acquisition and Control of Data. In 16th IRF World Road Meeting Lisbon. Lisbon, Portugal. Sohn, H., Worden, K., & Farrar, C. R. (2001). Novelty Detection under Changing Environmental Conditions. Proceedings of SPIE, 4330(4), 108–118. Worden, K., & Dulieu-Barton, J. M. (2004). An Overview of Intelligent Fault Detection in Systems and Structures. Structural Health Monitoring, 3(1), 85–98. Worden, K., Farrar, C. R., Manson, G., & Park, G. (2007). The fundamental axioms of structural health monitoring. Journal of the Royal Statistical Society – Series A: Mathematical, Physical and Engineering Sciences, 463(2082), 1639–1664.
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Monitoring based assessment of fatigue resistance of 40 year old pc bridges H. Weiher & K. Runtemund matrics engineering GmbH, Munich, Germany
ABSTRACT: Monitoring is used to determine real structural response of post-tensioned concrete bridges under thermal and traffic loading. This is to assess the resistance of the prestressing steel at coupling joints of bridges erected in the 1960s and 1970s with span-by-span method. Heating of the bridge deck may lead to transition of the coupling section to cracked state II und subsequently to significantly higher stress range due to traffic. Hence, strain gauge based crack width sensors and temperature sensors are used for data collection. With the help of these real data the fatigue design of old bridges can be done more realistic (often less conservative) and avoid or limit the extent of strengthening measures. This article presents the monitoring concept and two exemplary projects: Köhlbrandbrücke in Hamburg (Germany) crossing the Elbe river and Neue Innbrücke in Schärding/Neuhaus (Austria/Germany) crossing the Inn river.
1 DESIGN OF FATIGUE RESISTANCE 1.1 General Post-tensioned concrete bridges (multiple-span continuous girder) of the 1960s and 1970s often have little rebar and a high grade of prestressing. The steel is subjected to fatigue failure especially in the region of changing moment. This is also the preferred region for span-by-span erection joints with tendon coupling. Temperature gradient has not been taken into account for design in those days. Even if not a big issue for static resistance but this additional moment may lead to significant cracking of the section, especially at 100% coupling because of local losses. In cracked state the stress ranges due to traffic increase significantly. The design usually is done by limiting the stress range under fatigue loading. Due to the age of the bridge usually damage accumulation design is not successful. 1.2 Prestressing systems Most common curved pt-tendons for longitudinal direction formerly used are – – – –
12 mm-wire bundles with central wedge (similar to old Freyssinet system), 6/7 mm-wire bundles with button head (similar to old BBRV system), strand bundles with individual wedges (similar to nearly all actual systems) and sometimes bar (threaded or smooth, prebent)
1.2.1 Prestressing steel at coupling joints The resistance is limited by tendon coupling anchorage. Usually the maximum stress range is taken from approval tests with 2 million cycles, e.g. σmax = 110 N/mm2 . The allowable stress range under frequent loading (e.g. T = 7 K) is σfrequ = 0.7 · σmax , e.g. σfrequ = 0.7 · 110 N/mm2 = 77 N/mm2 . 903
Figure 1. Exemplary M-σ curve for bottom tendon at a coupling joint with 100% and 70% of prestressing force (30% extra losses locally at coupler).
1.2.2 Prestressing steel The resistance for the prestressing steel is significantly higher outside anchorages. It might be limited by transversal pressure at curvatures (fretting fatigue). The allowable stress range under frequent loading is σfrequ = σmax , e.g. σ = 110 N/mm2 (or 140 N/mm2 for straight tendons). 1.3 M-σ curve Illustrating the calculated stress range under fatigue traffic model against the bending moment (without traffic part) gives the M-σ curve (see Fig. 1). There are two curves, one representing 30% local losses of prestressing force at the couplers (must be used for theoretical design; “70% V”) and another one without local losses to get an idea about the range of real structural behavior. For the exemplary curve “70% V” the allowable stress ranges of 77 N/mm2 /110 N/mm2 at frequent/maximum temperature gradient of 7 K/12 K are exceeded significantly and design proof is not possible. 2 MONITORING CONCEPT With the help of monitoring the real M-σ curve might be found or at least approximated. Variables to the theoretic curve are – Local losses of prestressing force at couplers – Any differences in bending moment due to creep between construction and final state, modelling, permanent loading, stiffness (cracks) etc. By proper measurement the point of transition from uncracked state I to cracked state II can be found. All permanent bending moments are constant over time but the temperature gradient part. Hence, crack width amplitudes (→ σ) must be measured together with T (→ MT → M) to get a measurement curve that corresponds to the theoretical M-σ curve. The bending moment difference at the distinct crack opening can be seen as corrective element. 2.1 Crack instrumentation Crack width amplitude sensors are strain gauge based. They should detect crack opening/closing induced by trucks. It should be taken into account that the crack width itself is not subject of investigation and data should be treated carefully since strain gauge sensors are drifting over time. 904
Figure 2.
Crack with measurements at joints (sensor, position, data).
Figure 3. Temperature (field) measurement.
Nevertheless, for the given task crack width amplitude depends on the basic bending moment (only permanent loads incl. temperature) and can be measured within very short time (one truck passing). For crack width measurement a frequency of minimum 50 Hz is recommended, better are 100 Hz. Figure 2 illustrates a typical sensor type, location of arrangement and recorded signal. 2.2 Temperature measurement Ordinary Pt 100 sensors are used (see Fig. 3). To find out a very realistic vertical gradient T quite a number of sensors are necessary to detect the temperature field very well. This field can be evaluated by finite element model to get T. Just measuring top and bottom slab is not enough in most cases and problems arise because of reduced correlation (mistake) between measured and real T. Especially the deck slab that is supposed to sun radiation should be detected in different thickness levels. Temperature change is slow and measurements can be done with very low frequency, e.g. 1/min. The data need to be directly linked to the crack width data (e.g. by date/time). 2.3 Calibration Optionally one may detect the structural response under defined traffic loads, e.g. typical truck fully loaded with some 40–44 tons. Especially for that situation the very detailed knowledge of T 905
Figure 4.
Köhlbrandbridge (from Weiher et al. 2014, Ullerich 2010 and Wittfoht et al. 1975).
Figure 5.
Coupling of tendons at Köhlbrandbridge (from Weiher et al. 2014 and Polensky & Zöllner 1969).
is of great importance. This allows later analyze of the individual weights without explicit measures (e.g. weigh-in-motion).
3 CASE STUDIES 3.1 Köhlbrandbridge, Hamburg The bridge has been built 1970-1974. It is some 3.800 m in length and consists of two main pc ramps and a cable stayed bridge crossing the Elbe river (Fig. 4). It has been designed for 24.000 vehicles per day. The latest numbers are some 36.000 vehicles per day of which there are some 13.000 heavy trucks on their way to or from Hamburg harbor. The fatigue investigation focusses on the eastern ramp with 15 spans of double box section. The ramp has been built with movable scaffolding span by span coupling all tendons at the construction joints (Figs 4 and 5). Theoretical design by WTM engineers has shown that all joints are subjected to unacceptable stress ranges. Cracks had been detected at the joints at the connections between bottom slab and webs. Hence, monitoring has been done in 2013 and 2014 at three construction joints to get better knowledge about the fatigue situation. Figure 6 illustrates some impressions of the installation of sensors and monitoring system. 906
Figure 6.
Installation of sensors at Köhlbrandbridge (courtesy of C. Ullerich).
Figure 7.
Innbridge.
Figure 8.
Coupling of tendons at Innbridge.
3.2 Innbridge, Neuhaus/Schärding The bridge has been built 1972–1973. It is some 420 m in length with six pc box girder spans (maximum span length 77 m). The bridge crosses the river Inn representing the border between Austria and Germany near Passau, see Figure 7. The bridge has been built span-by-span with a special coupling zone at the construction joints with fully reanchoring of anchored tendons stepwise, Figure 8. Theoretical design by matrics engineering GmbH has shown that most joints are subjected 907
Figure 9.
Figure 10.
Installation of sensors at Innbridge.
Exemplary results of T [K] and w [mm].
to unacceptable stress ranges. Numerous significant cracks had been detected directly at the joints and especially in the short region of “missing” tendons. Hence, monitoring has been done in 2014 at three construction joints to get better knowledge about the fatigue situation, Figure 9.
4 ANALYSIS AND RESULTS 4.1 Determination of ‘real’ structural resistance Figure 10 illustrates an exemplary distribution of T during the monitoring period (autumn). Right hand side the crack width amplitudes induced by traffic is displayed depending on the respective T at that moment. It can clearly be seen that the crack width amplitude and subsequently the stress range increases significantly T (see mean value curve and 10% curve). In this case a calibration measurement had been done with a 28 ton truck. During the measurement period some 10% of the vehicles had been heavier than 28 tons. Transition of uncracked state I to cracked state 2 occurs around – 1 to 0 K. Now, one may take the difference between the transition points T of crack width and theoretical stress range curve (e.g. 2 K) and derives a correction moment M (T = 2 K) to calibrate the M-σ curve (dashed line, Fig. 11). The design can be done now with the calibrated M-σ curve, which represents very well the existing permanent bending moment at the construction joint. For the design the actions like fatigue traffic model and T values are still taken according to codes and guidelines (e.g. Eurocode). 908
Figure 11.
Calibrated M-σ curve (dashed line).
Figure 12.
Effect of strengthening on (calibrated) M-σ curve.
4.2 Determination of ‘real’ actions on structure If quantity and quality of the data set is good, that means representative temperature measurement to cover hourly, daily, seasonal and annual variations and/or large crack width amplitude database covering real traffic one may modify the design actions with the help of probabilistic methods. Special care needs to be taken to compensate for the sample size. 4.3 Effect of monitoring based assessment In any case by applying that concept the knowledge of real behavior of an existing bridge under fatigue loading could be achieved. Especially at tendon couplers the design usually can be improved for the structure since many theoretical approaches and assumptions are quite conservative. So either the design can be done now successfully or the quantity of strengthening (preferably by external post-tensioning) can be reduced significantly, see Figure 12. Instead of increasing the structural resistance by strengthening measures one may reduce the actions on the bridge. Limiting traffic is mostly no option, but T can be influenced by special pavement (thickness, color) or actively by cooling the top deck by piping between sealing and pavement. Calculations have shown that even if only small regions of pavement are equipped with 909
cooling the bending moment from T drops significantly. The same technology could be used for heating in winter to avoid ice. 5 CONCLUSION Monitoring is used to determine real structural response of post-tensioned concrete bridges under thermal and traffic loading. This is to assess the resistance of the prestressing steel at coupling joints of bridges erected in the 1960s and 1970s with span-by-span method. With the help of these real data the fatigue design of old bridges can be done more realistic (often less conservative) and avoid or limit the extent of strengthening measures. Partners within these projects had been Hamburg Port Authority, Staatl. Bauamt Passau, Amt der oberösterreichischen Landesregierung (clients/authorities), KMP Ziviltechniker and WTM Engineers. REFERENCES Penka, E. 2005. Beurteilung der Ermüdungssicherheit von Koppelfugenquerschnitten bestehender Spannbetonbrücken durch Langzeitmessungen. Dissertation, Berichte aus dem Konstruktiven Ingenieurbau, Technische Universität München. Weiher, H., Ullerich, C. & Runtemund, K. 2014. Monitoring an den Koppelfugen der Köhlbrandbrücke (Rampe Ost) in Hamburg. In Proceedings of 1. Kolloquium Brückenkolloquium – Beurteilung, Ertüchtigung und Instandsetzung von Brücken, Esslingen, Germany. Ullerich, C. 2012. Permanentes Echtzeit-Monitoring von Verkehrslasten auf der Köhlbrandbrücke. Bauingenieur, Vol. 87 (2012), Heft 10, S. 433–42. Wittfoht, H., Bilger, W. & Steffen, W. 1975. Die Spannbetonüberbauten der Köhlbrandbrücke. Beton- und Stahlbetonbau 70 (1975), Heft 6, S. 133–142. Polensky & Zöllner 1969. Zulassungsbescheid Spannverfahren Az. II B 2 – 2420 Zul. 95, 1969.
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Maintenance method for cable-stayed and extradosed bridge with composite main girder H. Sakai Central Nippon Expressway Co. Ltd., Nagoya-shi, Japan
ABSTRACT: Cable-stayed bridges and extradosed bridges are composed of the same types of structural members, which are main girders, pylons and stay cables. These two structures of bridges are deemed to be in the same category. In Japan, composite structures have been adopted in cablestayed and extradosed bridges to reduce the construction costs and the dead load for the purpose of increasing the span length. For example, these composite structures are concrete-steel mixed girder and hybrid box girder with prestressed concrete slabs and corrugated steel webs. Therefore, it is necessary to clarify the appropriate maintenance method in consideration of structural properties of these bridges. In this paper, the appropriate maintenance method for cable-stayed and extradosed bridge with concrete-steel composite main girder was investigated. And maintenance method for these bridges was proposed.
1 INTRODUCTION This study was targeted to cable-stayed and extradosed bridges in composite structure that has a concrete pylon and steel-concrete combined main girder (incl. corrugated steel web). In Japan, composite structures have been adopted in cable-stayed and extradosed bridges, e.g. Ibi River Bridge (6-span Prestressed concrete box girder and steel box girder hybrid continuous extradosed bridge, Figure 1, Figure 2) and Toyota Arrows Bridge (4-span Steel box girder and hybrid box girder with prestressed concrete slabs and corrugated steel webs hybrid continuous cable-stayed bridge, Figure 3, Figure 4). However there were no criteria for proper maintenance of these bridges. Therefore appropriate maintenance method was studied. The key results of this study are as below. – Maintenance method of these bridges consists of establishment of a maintenance plan, diagnosis of bridge’s conditions, implementation of repair or strengthening, and recording of relevant data. The maintenance operator of the bridge shall establish a maintenance plan so performance of these bridges can be maintained more than the required performance. – The procedure of diagnosis for these bridges consists of: inspection, estimation of deterioration mechanism, deterioration prediction, performance evaluation of the bridge or its members, and judgment of whether or not any repair or strengthening is required.
Figure 1.
Side view of Ibi River Bridge.
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Figure 2.
Ibi River Bridge.
Figure 3.
Side view of Toyota Arrows Bridge.
Figure 4.
Cross section of Toyota Arrows Bridge
– When girders are in the hybrid structure of concrete and steel, the connection between corrugated steel web or steel truss and concrete floor slab is an important part of the structure. Since this connection needs to maintain sufficient durability and safety during the design service life, it is important to conduct inspection with extra attention. 2 MAINTENANCE PROCEDURE The maintenance operator of a cable-stayed or extradosed bridge must ensure that its performance level keeps more than the required performance over its planned service life. For this purpose, the maintenance operator must draft a maintenance plan first, and then accordingly implement a system of maintenance implementation, to maintain the bridge. This includes inspection, deterioration prediction, diagnosis consisting of evaluation and judgment, implementation of repair/strengthening and recording of data. 912
Figure 5.
Procedure for the maintenance of cable-stayed or extradosed bridge.
The principles of the maintenance of a cable-stayed or extradosed bridge are no different from that of general bridges. The distinctive characteristic of a cable-stayed or extradosed bridge, however, is that it requires maintenance of such critical members as stay cables and pylons, which do not exist in other types of bridges. Maintenance of a cable-stayed or extradosed bridge consists of establishment of a maintenance plan, diagnosis of bridge’s conditions, implementation of repair or strengthening, and recording of relevant data. Diagnosis as part of maintenance involves inspection, estimation of deterioration mechanism, deterioration prediction, evaluation of bridge’s performance and judgment of whether some repair or strengthening needs to be implemented. Figure 5 shows a workflow of the maintenance of a cable-stayed or extradosed bridge. In order to carry out appropriate and timely maintenance of a cable-stayed or extradosed bridge, its status must be examined at the time of inspection. Based on the inspection results, the current or possible deterioration mechanism must be estimated to make predictions for the future deteriorations. Furthermore, it should also be evaluated whether the safety, serviceability or durability will fall below the performance requirements before the end of service life. Taking this evaluation into account, it should be judged whether or not some repair or strengthening needs to be implemented. Causal factor, description of the state and indicator for each deterioration mechanism in concrete structures and steel structures shows in Table 1. In this table, the deterioration indicator means the index to evaluate the progress of deterioration or severity of deterioration. 3 INSPECTION Inspections carried out as part of diagnosis on cable-stayed or extradosed bridges shall be appropriate to the purpose of the diagnosis. Initial inspections are carried out, as part of an initial diagnosis, 913
Table 1. Relationships among causal factor, description and indicator for each deterioration mechanism. Deterioration mechanism Concrete structure
External cable, Stay cable
Steel structure
Causal factor
Description of the state
Indicator
Carbonation
Carbon dioxide
Due to carbonation reaction in hydrated cement caused by carbon dioxide, pH of concrete is lowered and steel members are corroded, leading to cracks and peeling of concrete surfaces and to reduced sizes of steel members
Carbonation depth Amount of corrosion in steel member
Chloride induced deterioration
Chloride ions
Corrosion of steel members within concrete is induced by chloride ions, leading to cracks and peeling of concrete surfaces and to reduced sizes of steel members
Chloride ion concentration Amount of corrosion in steel member Corrosion cracks
Alkali-silica reaction
Reactive aggregate
Aggregate containing alkali-silica-reaction causing minerals reacts with an alkali solution and expands abnormally, leading to cracks in concrete
Amount of expansion (cracks)
Frost attack
Freeze and thawing action
When moisture in concrete freezes and thaws, concrete deterioration escalates with the development of scaling and cracks
Freezing depth Amount of corrosion in steel member
Chemical attack
Acidic materials Sulfate ions
Concrete is decomposed by acidic substances and sulfate ions, or deteriorated by expansion pressure generated when chemical compounds are formed
Penetration depth of Deterioration factor Carbonation depth Amount of corrosion in steel member
Fatigue of RC slab
Repeated loading
Cracks are generated in floor decks of bridges due to repeated loading of vehicle weight through wheels, leading to steel member corrosion and cave-ins
Crack density Deflection
Wear
Abrasion
Abrasion by flowing water and vehicle wheels wears concrete over time
Amount of wear Speed of wear
Corrosion
Chloride ions Acidic substances & oxygen Rainwater (dews) etc.
Anti-corrosion effect is lost and corrosion advances due to penetrating rainwater and oxygen, leading to cable breakage
Effectiveness of anti-corrosion effect Breakage of prestressing steels
Fatigue
Repeated loading Winds & vibrations
Repeated loading applied by vehicles and the repeated stresses caused by winds or traffics lead to breakage of prestressing steels
Tensile force Breakage of prestressing steels
Corrosion
Ultraviolet rays Chloride ion Acidic substances & oxygen Rainwater (dews) etc.
Ultraviolet rays cause chalking where coating substances are decomposed into powdery state and lost its shiny finish. Macro-cell corrosion is developing under coating membrane first by water penetration and oxygen, then coating membrane is perforated (corrosion spots), and lastly those corrosion spots expand further due to micro-cell corrosion. As they develop into larger corrosions, the sizes of steel members will be thinned out, which is escalated further by chloride
Discoloring & peeling Area of corrosion Thinning of steel Reduced steel size
Fatigue
Repeated loading Concentration of stress
Stresses below yielding point are repeated, leading to concentration of stress in the end of welded parts (particularly, around boxing weld joint) and the root of fillet welds
Cracks
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Figure 6.
Procedure for the maintenance of cable-stayed or extradosed bridge.
to grasp initial conditions for the maintenance purpose. Routine and regular inspections are carried out, as part of a regular diagnosis, to grasp any change in the conditions of the bridge. When an extraordinary diagnosis is required, an extraordinary inspection or emergency inspection shall be carried out depending on the purpose of the diagnosis. Inspection shall basically include standard survey with its frequency, target structure and method as prescribed in the maintenance plan. If any degradation is identified through standard survey and further details need to be found out, detailed surveys are carried out. If the inspection concluded that some emergency measures are required, they should be carried out immediately. An overview of inspection types, inspection methods and inspection frequencies stipulated is shown in Figure 6. In carrying out inspections, relevant inspection items, target members, inspection frequency and inspection method must be determined appropriately, depending on the purpose of each inspection. Surveys are carried out as part of inspection to obtain some specific information about the conditions of a bridge or its members in a method that would be able to achieve the purpose of inspection, with the inspection targets determined appropriately. Survey items are selected appropriately taking into consideration such factors as the type and purpose of inspection, status of the bridge to be inspected, required information, and cause of deterioration.
4 FOCAL POINT OF INSPECTION FOR CABLE-STAYED AND EXTRADOSED BRIDGE WITH CONCRETE-STEEL COMPOSITE MAIN GIRDER The maintenance of a cable-stayed or extradosed bridge is conducted mostly in the same way as that of general bridges. The distinctive characteristic of a cable-stayed or extradosed bridge, however, is that it requires maintenance of such critical members as stay cables and pylons, which do not exist in other types of bridges. The inspections and surveys of these, therefore, are particularly important in maintaining the performance required of the bridge in terms of safety, serviceability and durability. When girders are in the hybrid structure of concrete and steel, the connection between corrugated steel web or steel truss and concrete floor slab is an important part of the structure. Since this 915
Figure 7. Joining structure with concrete slab and steel corrugated web (Type ‘d’).
connection needs to maintain sufficient durability and safety during the design service life, it is important to conduct inspection with extra attention. In addition, when girders are in the joining girder of prestressed concrete box girder and steel box girder, the joint girder between the concrete girder and the steel girder is also an important part of the structure. 4.1 Girder (hybrid structure with prestressed concrete slabs and corrugated steel webs) The joint between the steel corrugated web and the lower slab is an element of the structure that can be directly affected by rain and condensation, so it is important that appropriate measures are taken for this joint to protect against corrosion, including measures to drain off and stop water. In particular, with embedded joints, the steel corrugated web is embedded directly in the concrete of the lower panel, so it is vital to take measures against rust, such as sealing the joints so that any stormwater or condensation acting on the interface does not trigger a reduction in durability. To date, the five different joining structures listed below have been used to ensure such performance in Japan. Diagram of the type ‘d’ is shown in Figure 7. Type ‘d’ was adopted at Toyota Arrows Bridge. a. b. c. d. e.
Joining structure by implantation of corrugated steel web Joining structure by stud dowel Joining structure by angle dowel Joining structure by perfobond strip and stud dowel Joining structure by twin perfobond strips
The case of focal points for inspection of the joining structure at Toyota Arrows Bridge is shown in Figure 8. 4.2 Joint girder (between hybrid/concrete structure and steel structure) The joint girder between the hybrid/concrete girder and the steel girder consists of shear connectors, bearing plate and filling concrete. Through such components, stresses applied on the joint are transferred from the concrete girder to the filling concrete, from the filling concrete to the shear connectors or bearing plate, and then from the shear connectors or bearing plate to steel girder. There is a great difference in the rigidity of steel girder and that of the hybrid/concrete girder. In order to avoid a sudden change in the rigidity in the steel-concrete joint, which causes the edge of the joint surface to bend, the area near the joint between low-rigidity steel girder part and highrigidity concrete part is made into a transitional part. This enables a smooth transition of the level of rigidity from the steel part to the concrete part, and prevents local stress. Further, concrete is prestressed for enhanced strength, and the applicability and the quality of welding are assured for fatigue prevention. 916
Figure 8.
Focal points for inspection of the joining structure at Toyota Arrows Bridge.
Figure 9.
Joint types (Joint girder).
The model of stress transfer in the joint part is shown in Figure 9. Characteristics of the joint part are described below. a. Metal plate type: stress is transferred through a bearing plate to the concrete girder. Shear force is transferred via stud dowels placed in front of the bearing plate. b. Front Plate Concrete filling type: part of axial stress is transferred to filling concrete via stud dowels to even-out stresses in the bearing plate. By doing so, a thin plate can be used as a bearing plate. Shear force is transferred through the stud connector placed in front of the bearing plate. 917
Figure 10.
Focal points for inspection of the joint girder by rear plate concrete filling type e.
Figure 11.
Dual plate concrete filling type in Toyota Arrows Bridge.
c. Rear Plate Concrete filling type: axial stress is transferred from a rear-side bearing plate and stud dowel connectors. Shear force is transferred to concrete filling. d. Dual Plate Concrete filling type: axial force is transferred from stud dowels in concrete filling and on front-side and rear-side bearing plates. Shear force is transferred via the stud connector on the front plate. Dual Plate Concrete filling type was adopted at Toyota Arrows Bridge and Ibi River Bridge. A case of dual plate concrete filling type in Toyota Arrows Bridge is shown in Figure 10. The case of focal points for inspection of the joint girder by rear plate concrete filling type is shown in Figure 11. 5 CONCLUSIONS Several cable-stayed and extradosed bridges with concrete-steel composite main girder have been adopted in Japan. But the maintenance method for these bridges has not been clarified. In this study, maintenance methods of these bridges have been studied. As a result, the guidelines of the maintenance method of these bridges were drawn. The guidelines were clarified the following items. – Procedure for the maintenance of cable-stayed or extradosed bridge with concrete-steel composite main girder was showed. – Inspection methods for concrete girders, steel girders and hybrid girders were clarified. – Focal points for inspection of the joint girder between hybrid/concrete girder and steel girder were clarified.
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Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
Construction control of a long-span single pylon cable-stayed bridge C. Liu, L.J. Sun, Y.S. Ni & D. Xu Tongji University, Shanghai, China
ABSTRACT: Long-span single pylon steel-concrete hybrid cable-stayed bridges have wide range of application in the engineering field because of its own advantages. The line and internal force of hybrid cable-stayed bridge after completion are highly coupled with its construction method and installation procedure, thus the method of construction control being of vital importance. Determination of reasonable completion state of bridge and key technologies of the application of backward calculation method are described on the engineering background of Haihe bridge, so does the analysis of non-closure factors and adjustment of cable force. Measured results show that the stress, line and cable force of final state of hybrid cable-stayed bridge are controlled at standard range by adopting backward calculation method, as well as timely adjustment of cable force and line given non-closure factors.
1 INTRODUCTION The construction of cable-stayed bridge is staging. The stiffness of the structure during construction is much smaller than that of the completion state, geometric nonlinearity outstanding. Due to loads, for example self-weight, construction equipment and prestressing, imposed on the structure step by step, each stage of construction may accompany with configuration change, creep and shrinkage of material, change of boundary constraint, prestress tension and system transformation. The stress states and mechanical properties of the structure have close connection with the earlier structure. Moreover, the cable-stayed bridge is high-time statically indeterminate structure, making the line and stress state of the structure after completion highly coupled with the construction method and installation procedure. Construction control therefore is of vital importance during the construction of cable-stayed bridge. Taking Haihe bridge, Tianjin as an example and considering its actual construction process, this paper aims at discussing the application of backward calculation method in the construction control of long-span single pylon cable-stayed bridge. 2 ENGINEERING BACKGROUND Haihe bridge, Tianjin is double cable plane single pylon cable-stayed bridge, the span arrangement of which is (310 + 50 + 50 + 40 + 40) m. The main bridge is half floating system that sets up 0# cable and longitudinal damping limit device at the pylon. The girder is steel and concrete hybrid structure, of which the main span is steel box girder and the side span is PC box girder. The anchorage zone of stay cables lies on the sides of the girder. The cast in-place PC box girder in the side-span is constructed by full falsework method, while the steel main girder is constructed by segmental cantilever lifting method. The cross section arrangement of main bridge is 1.335 m anchorage zone + 2 m maintenance footpath + 16.5 m traffic lane + 2 m maintenance footpath + 1.335 m anchorage zone. The width of the bridge is 23.17 m. The bridge is of bi-directional cross slope 1.5%, which is formed by setting up cross slope on the top flange of main girder. 919
Figure 1. The vertical plan of main bridge.
Figure 3. The cross section of steel box girder of main bridge.
Figure 2. The cross section of box girder in the side span.
The concrete pylon adopts a type of diamond, located at the top of the cap, the total height of which is 164.798 m and the height above the deck is 126.365 m. The main girder adopts hybrid girder structure, namely, the most part of main span (310 m) adopts steel girder and both side spans and the part within the scope of 10 m near to pylon of main span adopt PC girder. The height of the girder is 3 m equally. The steel box girder adopts single box and three cells flat streamlined section, while the PC box girder adopts single box and five cells section coincident with the shape of steel girder. The stay cables adopt space fan-shaped layout. The distance of intersection in the vertical direction of the cable plane in the pylon is 1.5∼2.5 m. The anchorage points of stay cables in the transverse direction lie in the center line of the pylon. The total amount of stay cables is 37 couples (74 stay cables).
3 BACKWARD CALCULATION METHOD Backward calculation method is one of the simulation calculation methods of construction control. The idea of the method is: appropriate completion state of bridge needs to be determinated firstly. Taking the appropriate completion state of bridge as the initial state in the analysis of construction stage, backward calculation of the structure is conducted according to inverse process of construction sequence. The state of each stage can be gained by accumulating the influence on the remaining structure due to unloading installed elements in the construction stage each time, as well as the control parameters (elevation, cable forces and stresses of controlling section) of each construction state. As long as segmental geometric shape during unloading is under control and the internal forces between unloading part and loading part in the theoretical model is zero, backward calculation method is authentic. The concept of backward calculation method is clear and the stay cables can be tensioned to the prototype at once according to certain construction parameters. What’s more, the influence of the shrinkage and creep of concrete during cantilever lifting of Haihe bridge is small. Therefore, 920
Figure 4. The schematic diagram of grillage division of box girder.
Figure 5. The schematic diagram of the finite element of the whole bridge.
backward calculation method is a very effective calculation method in the construction control of Haihe bridge. 4 NON-CLOSURE FACTOR AND CABLE FORCE ADJUSTMENT 4.1 Finite element model The simulation calculation of Haihe bridge adopts space bar finite element model. Plane grillage model is built based on the division of the web in the girder section, as shown in Figure 4. The torsional stiffness of longitudinal and transversal elements divided is amended in order to maintain the torsional stiffness of the whole close section of original cross girder. For stay cable element, space truss element is adopted. The finite element structure diagram of the whole bridge is shown in Figure 5. 4.2 Temporary falsework and closure of fold segment As concrete parts and side span of Haihe bridge adopt full falsework method, there exists the problem of temporary falsework and the closure of fold segment. Internal forces exist in the falsework and consolidation of the tower and girder when removed in the forward calculation. The external forces acted on the structure need to be unloaded when removed. Therefore, initial forces of falsework and consolidation of the tower and girder need to be specified corresponding to actual state when adopting backward calculation. When ideal completion state of bridge is certain, specifying internal forces actually means changing theoretical unstressed length of stay cables. Due to one-time tensioning method of this bridge, unstressed length of stay cables has never changed during backward calculation. Thus, the stress state of the structure has nothing to do with the construction process as long as final constraint state is the same. The displacement constraint of the girder in the backward calculation model can be 921
Figure 6. The flow diagram in the backward calculation including falsework element.
the same as that of actual construction state by specifying internal forces or rendering compelled displacement to the falsework and consolidation of the tower and the girder under appropriate completion state of bridge. Fold state is the most dangerous state with the longest cantilever during the construction of the cable-stayed bridge. The construction of fold segment which is the last process in the connection of main bridge is subject to temperature, cable forces, self-weight of main girder and temporary loads. The construction quality of fold segment has a direct effect on the stress state and service life of the whole structure, thus being very important. To achieve the closure of the whole bridge quickly by adopting safe and appropriate method is of vital importance for Haihe bridge. Actually, there exists shear force and bending moment in the fold segment when simulating the removal of it in the backward calculation. However, initial shear force and bending moment of fold segment cannot be achieved in the practical welding process. In order to eliminate this non-closure factor, it is necessary to consider complicated state of closure from backward calculation and construction technology. Figure 6 shows the flow diagram in the backward calculation including the state of falsework element. Construction control of fold segment adopting above method has gained good effect. 4.3 Amendment of temperature effect It is found from theoretical calculation and practical observation that the influence on the deflection of main girder caused by temperature variation is very small when the temperature of cable-stayed bridge structure is overall up or down. Therefore, the influence on the line of main girder caused by season temperature variation is not considered in the construction control. Daily temperature variation however has large influence on the deflection of main girder, especially when the cantilever length is increasing. The cantilever length and the temperature variation between day and night of Haihe bridge are large, thus leading to obviously influence on the construction state by temperature. Amendment of temperature effect adopted in this bridge is: the temperature and sunlight change of the whole day is observed against special time. Continuous observation is also conducted on 922
Table 1. Observation results of temperature. Time
Temperature
Time
Temperature
5:55 8:25 10:20 12:45
26◦
14:20 16:10 19:25 20:25
36◦ 33◦ 28◦ 27◦
30◦ 33◦ 35◦
Figure 7. The function diagram of the elevation of hanging points 6∼8 with the temperature in the state of lifting segment M8.
Figure 8. The function of the elevation of hanging points in the midline of M8 with the temperature (unit: mm).
Figure 9. The function of the deviation of the top of the tower with the temperature (unit: mm).
the cable force, deviation of the tower and deflection of the girder. The function between the state of main structure of the bridge and sunlight change, as well as temperature change, in one day is finally obtained. The observation results of temperature variation and corresponding girder change are shown in Table 1 and Figure 7∼9. The deformation and the state of cable forces of Haihe bridge are more stable between 10 p.m. and 6 a.m. after observation. Therefore, the orientation of steel box girder and secondary tension of stay cable are conducted at this time in order to guarantee control effect and accuracy. The actual control result shows that the deviation of construction control caused by sunlight and temperature can be precisely eliminated in this way. The accuracy requirement of construction control is satisfied well. 4.4 Adjustment of cable force Based on the analysis of traditional tensioning method, Du Guohua and Han Zhenyong proposed optimized tensioning method of stay cables by adopting tensioning influence matrix of stay cables and functional extremum principle. Resolution and repeatedly tensioning could be conducted of 923
Figure 10. The comparison diagram of measured result and theoretical result during cable adjustment.
the cable force of local cables if necessary, thus making most cables reach target cable force by only one-time tensioning. The essence of the adjustment of cable force is to adjust the unstressed length of cables. Consequently, the only variation in the whole process of adjusting cables is the unstressed length of cables. Based on the structure model of actual construction state and taking the state of zero-force and zero-displacement as initial state, initial forces of part cables are specified to make bridge structure achieve target construction state after cumulating the deformation, displacement, stress and cable variation of structure with that of actual bridge structure. The stay cable specified initial internal force that produces change of unstressed length is the cable necessary to adjust. This cable force is the theoretical deviation of cable force after adjusting. The stiffness of cable-stayed bridge is mostly provided by stay cables. It is found from the validation results of the stiffness of the cable-stayed bridge when lifting girders that the absolute stiffness and relative stiffness of the structure in theoretical model are reasonable. The deviation of unstressed length between adjusted state and target state can be figured out in the model, serving as control parameter of adjusting cables. Anchorage length of stay cables is controlled to change the unstressed length during actual cable adjustment, while cable force is only serving as a reference. Therefore it is unnecessary to take into account the influence of temperature and change of construction loads for cable adjustment in each stage. Figure 10 shows the comparison of measured result and theoretical result during cable adjustment. Taking anchorage length of stay cables as main controlling object, the adjustment of each anchorage shackle is given out according to above method. The validation result adjusted is very satisfying during execution of cable adjustment of Haihe bridge, proving the accuracy of this method.
5 MONITORING AND CONTROLLING RESULT It is vigorously effective to monitor and control bridge construction by adopting reasonable control procedure, leading to appropriate state of internal force and standard geometric line. Significant attention is paid on the stress of main girder, flip-height, cable force and deviation of the tower in this bridge during construction control. The testing points in the girder are arranged at 14 representative cross sections by theoretical calculation. The arrangement in the cross section is longitudinal stress testing point of top and bottom flange near to the web. Figure 11 shows the arrangement of stress testing point and Figure 12∼13 show the envelope diagram of the stress of main girder. The controlling testing points of the displacement of the tower are shown in Figure 14. Figure 15 shows the comparison of theoretical and measured displacement of the top of the tower, while Figure 16 shows that of final cable force. 924
Figure 11. The elevation diagram of stress testing cross section in the girder.
Figure 12. Normal stress envelope diagram of steel girder (unit: kPa).
Figure 14. The testing points of the displacement of the tower.
Figure 13. Normal stress envelope diagram of concrete girder (unit: kPa).
Figure 15. The comparison of theoretical and measured displacement of the top of the tower.
Figure 16. The comparison of theoretical and measured cable force of final cable force of Haihe bridge.
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It can be found from above measured and theoretical comparison diagram that the stress, line, cable force and the displacement of the tower are controlled in reasonable range for Haihe bridge. The measured and theoretical value anastomoses well, proving the consistence between theoretical calculation and process control perfect. 6 CONCLUSIONS After analysis of process control of a long-span single pylon steel-concrete hybrid cable-stayed bridge—Haihe bridge, conclusions can be drawn as follows: (1) For construction control of long-span single pylon hybrid cable-stayed bridges, non-closure factors can generate non-closure between theoretical state and actual construction state. Nonclosure factors can be amended by specifying initial state of internal force in the computation simulation of bridge, thus producing theoretical model corresponding with actual construction state. This theoretical model can reflect actual construction state well and provide precise construction parameter for construction control. (2) Given bridges that are similar to Haihe bridge in structure and construction character, the influence on the structure caused by temperature and creep and shrinkage of concrete cannot precisely calculated in the theoretical model. Therefore it is necessary to monitor the key deformation and stress state of the structure during construction control for the sake of obtaining experience value. The construction data also needs to be amended according to experience value. (3) Tensioning and adjustment of cable force are actually the process of changing the unstressed length of stay cables in the calculation of theoretical model and actual construction. The unstressed length of stay cables can serve as a control parameter during construction control of cable-stayed bridge. The adjustment of stay cables can be accomplished fast and precisely by adjusting the anchorage length at the anchorage end of stay cables on the premise that theoretical calculation possesses certain accuracy. REFERENCES Du G.H., Jiang L. 1989. Reasonable Cable Force and Construction Tensioning Force for Cable-stayed Bridges, Bridge Construction. Du P.J., Huang C.L., Zhang Z. et al. 2003. Construction Control of Highway Cable-stayed Bridges. Journal of Dalian University of Technology, 43(6):783–786. Han Z.Y. 2007. Optimized Cable Tensioning Method for Cable Supported Bridges. Structural Engineers, 23(2):75–79. Kang, J. 2004. Determination of Reasonable Dead-load and Construction Cable Force for PC Cable-stayed Bridges. Changan University Liang P. Xiao R.C., Zhang X.S. et al. 2003. Practical Method of Optimization of CableTensions for Cable-stayed Bridges. Journal of Tongji University(Natural Science Section), 31(11):1270–1274. Liu L.J., He S.H., Song Y.F. et al. 2004. Analysis of Temperature Stress in Control of Long-span Bridge Construction. China Journal of Highway and Transport, 17(1):53–56. Qin, S.Q. 2008. Unstressed State Control Method for Bridges Constructed in Stages. Bridge Construction, (1):8–14. Xin K.G., Feng Z. 2004. Nonlinear Static Reverse Analysis of Long-span Cable-stayed Bridges during Construction. Engineering Mechanics, Vol. 21 NO. 5. Yan D.H., Li X.W., Liu G.D. et al. 2003. Step-by-step Arithmetic for the Reasonable Finished Dead State of the Concrete Cable-stayed Bridges. China Journal of Highway and Transport, 16(1):43–46. Zhang Z.C.,Ye G.R., Chen H.Z. et al. 2004. Analysis and Computation Method of Structure for the Construction Control of Bridges with Large Span. Journal of Zhejiang University(Engineering Science), 38(2):210–214.
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Effect of cable corrosion on the structural response of cable-stayed bridges O.A. Olamigoke, G.A.R. Parke & M. Imam University of Surrey, Guildford, UK
ABSTRACT: The cables on cable-stayed bridges can fail due to vehicle collision, explosion, and excessive corrosion of the strands which may enhance the vulnerability of this structure, leading to progressive collapse. The structural response of cable-stayed bridges will differ when a combination of factors such as corrosion and extreme events leads to the loss of the cable(s) as will opposure to extreme events occurring with no corrosion present. In existing cable-stayed bridges, corrosion may affect the stays which will decrease bridge performance compared to that of the new structure. Varying the extent of corrosion critical locations, the response of a cable-stayed traffic bridge is analyzed, for different loading combinations, using three dimensional non-linear dynamic analysis. The results show that a cable subjected to 30% or more corrosion before blast will have a significant effect on other cables with no corrosion. The presence of corrosion in a lost cable causes the cables close to it and the nearest backstay to exceed the design limiting stress.
1 INTRODUCTION Cable-stayed bridges have become a very popular option for long span bridges spanning over 1,000 m. They are aesthetically pleasing, self-balancing and a stiff option for long span-bridges but are also highly redundant structures (Gimsing, 1997). Cables on cable-stayed bridges play a key role in the structural stability of its self-balancing mechanism and they also transfer the load applied on the deck to the pylons. The cables on cable-stayed bridges are exposed to intentional events such as terrorist attacks, accidental events such as vehicle collision, fatigue and excessive corrosion of the strands which leads to the vulnerability of these structure with the likelihood of progressive collapse (Starossek, 2006). Other factors that affect the cables are variation in stress and the friction acting between strands (Roura, 2011). Steel as a material is highly susceptible to corrosion and the cables on cable-stayed-bridges in the presence of air and water are no exception. The cable strands on many of the old cable-stayed bridges have to be removed from the strand bundle and be replaced with new strand. Also, due to excessive corrosion, it may be necessary to replace a whole cable which is an expensive and complex procedure. Interestingly, even cables on newly constructed cable-stayed bridges are experiencing corrosion; as seen on the Zarate-Largo Bridge in Argentina which experienced total failure of a stay due to a combination of fatigue and corrosion (Mozos & Aparicio, 2011). Corrosion can be simplistically defined as the destructive attack of a metallic material through a chemical reaction with its environment. Corrosion of cables is usually highly concentrated at the anchorages of the cable (Hamilton III et al., 1995). According to Xu & Chen, 2013 corrosion reduces the cross sectional area of the cable as well as the mechanical properties of the cable such as the yield load, ultimate load, and ultimate strain which affects the ultimate tensile strength of the cable. Therefore the response of a cable-stayed bridge with the cables corroded, to the sudden loss of a cable, will have a significant effect on the mechanical properties of the cables affected 927
by corrosion because the cables are usually designed to operate at a limit of 45% of their ultimate tensile strength. The American Post Tensioned Institute (PTI) guidelines for the loss of a cable only cover for the sudden loss of a cable and recommend a Dynamic Amplification Factor (DAF) of 2.0 with an accidental load combination (PTI, 2007) while the Eurocode recommends an amplification factor of 1.5 (EC3, 2006). However, when corrosion is present before the loss of a cable this amplification factor may not be appropriate. The sudden loss of a cable can be due to a bomb blast which at the point of loss not only drops the cable force to zero over a very short time period but also induces high stress variations in the cable due to the positive and negative phase caused by the detonation pressure and wave refraction respectively (Agrawal & Yi, 2009; Deng & Jin, 2009; Mays & Smith, 2012). A number of research studies have been carried out on the sudden effect of the loss of a cable on various cable supported structures either by obtaining the DAF, varying the type of load, varying the number of cables lost or by obtaining the cable rupture time using quasi-static and dynamic analysis (Bruno, 2008; Gerasimidis & Baniotopoulos, 2011; Mozos & Aparicio, 2010; RuizTeran & Aparicio, 2009; Wolff & Starossek, 2009). Mozos & Aparicio, 2011 derived the stress acting on a damaged and undamaged seven wire strand at different strain rates determining the wire rupture time while Vikas et al. looked into the effect of corrosion on the cable-stayed bridge response (Vikas, et al., 2013). However this paper combines the effect of corrosion before a potential explosion event aiming at determining the effect of corrosion on the response of a cable-stayed traffic bridge model when subject to cable loss triggering extreme events. The corrosion analysis method is validated by modeling the cables as solid elements and introducing corrosion by area loss.
2 VALIDATION OF CORROSION EFFECT ANALYSIS METHOD Corrosion reduction in a finite element model has been accounted for by the reduction of Young’s modulus to cater for the material characteristic associated with corrosion and a reduction in area for the cable corrosion (Vikas et al., 2013). Parallel wire cable is generally used nowadays on cable-supported bridges because it gives an easier option for replacing the cables. In this research, a parallel wire cable has been modeled with solid elements to estimate the effect of corrosion, which leads to reduced cross sectional area, on the cable force. According to Mozos & Aparicio, 2011 the strands in a cable positioned on the side get wet with dew at night but do not dry up totally, unlike the top strands while the strands at the bottom of the cable stay wet day and night. Figure 1 shows seven parallel strands with a diameter of 30 mm each with a space for grout all enclosed in a casing creating a 100 mm diameter cable 40 m long. Two models of parallel wire strands are adopted with varying corrosion effects of 9% and 43% of the area of steel in the cable lost along the sides and bottom of the strands in the cable. The value of 9% is chosen to represent corrosion at its initial stage while 43% is chosen because the cable design limit is set to be 43% of the Ultimate Tensile Strength (UTS) of the cable.
Figure 1.
Parallel cables modelled showing varying corrosion extents on cable cross-sections.
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The cables are pinned at each end and bonded together using a non slip contact in Abaqus 6.13 to represent grouting. The load applied to the cable is a total load of 3146 kN to account for the cable prestress force as well as gravity load for self weight. The force in the cable is obtained from the nonlinear analyisis model. The prestress and self weight is applied to the cable in one step and the area reduction in the next step with both steps appraised using a nonlinear dynamic type of analysis to obtain the force in the cable. Table 1 shows that the results obtained have a linear relationship between the force in the cable and the viable cross-sectional area, therefore the percentage of area lost due to corrosion can be modelled as percentage drop in the cable force. 3 NUMERICAL ANALYSIS A three dimensional nonlinear analysis of a cable-stayed bridge has been modelled in Abaqus 6.13 to obtain its response to extreme loading following the corrosion of a cable. 3.1 Description of model geometry and layout A named cable-stayed dual carriage highway bridge with a total span of 676 m having two planes of stays in a semi-fan cable arrangement pattern stemming from a modified A-shaped pylon was modeled in three dimensions using Abaqus 6.13 software. The cable system was made up of 116 cables with two clear main spans of 230 m each and back spans of 108 m each, as shown in Figure 2. The cables sized 100 mm–200 mm are spaced at 12m centres along the bridge with 14 m between each set of fan cables. The cable-system includes two backstays pairs (CP1, CP2 & CP57, CP58) 6 m apart to help balance the load in the main span without increasing the cable size above 200 mm while the extreme backstays (CP1 & CP58) are 2 m away from the ends of the bridge. The composite plate girder deck has a thickness of 260 mm with the plate girders placed along the bridge edges, with cross beams at 4 m intervals and a total deck width of 22 m. The pylon is a hybrid pylon which is in reality a combination of an H-pylon and an A-pylon having a height below the deck of 33 m and above the deck of 85 m. Figure 2 below shows the general layout of the bridge. The deck is allowed to expand as far as 31 mm on one side and is pinned on the other side. The cables have been modeled as truss elements with a reduced elastic modulus to account for the cable sag and reduced bending stiffness with the values shown in Table 2. The cables have an ultimate tensile strength of 1660 N/mm2 . A summary of the material and geometric properties is shown in Table 2 below. 3.2 Loading The loading applied was in accordance with the Eurocode for highway bridges (EC1-2, 2003; EN 1991-1-2, 2002), for the Strength ultimate limit state (STR) and the Accidental limit state (ACC) in accordance to the Eurocode loading combinations for group 1 as the worst load combination. An unfactored UDL of 5 kN/m2 and wheel load (tandem system) of a maximum unfactored value Table 1. Relationship between lost cable areas and cable force reduction. % of corrosion
Area lost (mm2 )
Total area (mm2 )
Force in cable kN
Force drop to kN*
% Area lost
Force reduction %
9% 43%
1100 5300
12400 12400
119792 119792
109591 65995
9% 43%
9% 43%
*due to lost area.
929
Figure 2.
General layout, deck cross section and pylon section of the model.
Table 2. Material and sectional properties of the numerical model. Element Cable
Truss: tension only (100–200 mm Ø) Deck Shell S84R (260 mm concrete) Main girder Beam B32 (I-girder) Cross beams Beam B32 (I-girder) Pylon Beam B32 (Box section) Pylon Cross Beam B32 beam (Box section)
Density Poisson’s Material E (N/mm2 ) (kg/m3 ) ratio, ν Dimensions Steel
195
7860
0.3
100–200 mm diameter cables
30
2500
0.25
260 mm thick
Steel
210
7860
0.3
2 m deep 60 mm thick
Steel
210
7860
0.3
1.448 m deep by 20 mm
Steel
210
7860
0.3
Steel
210
7860
0.3
Varying hollow box 4 m by 5.5 m × 0.5 m to 5.65 m × 9.65 m × 1.65 m Varying hollow box 4.1 m by 4.3 m by 2 m to 3 m by 7.5 m by 0.8 m
Concrete
of 200 kN was applied. The cable prestresses were obtained by applying the STR load combination with restriction of zero (0) deflection at anchorage points.
where Ed = Design Value of Effect of actions, γ G,j = partial factor for permanent actions, Gk,j = Characteristic value of permanent actions, γ Q,j = Partial factor for variable action,i. ψ0,1 = Factor for combination of a variable action 1, Qk,i1 = Characteristics value for a leading variable action 1, Qk,i = Characteristic value for an accompanying variable action 1. A = Accidental Action (EC0, 1990). 3.3 Numerical analysis method Both static and dynamic analyses were carried out on the model with the cables modeled using tension only truss elements with the Ernst modulus of elasticity to account for the nonlinear behavior of the cable. 930
Figure 3.
Zones defining the variation of amplitude of corrosion and blast applied to the cable force.
The cable-stayed bridge was analysed under static and dynamic conditions to obtain its response to permanent and variable loads as well as accidental loads. The first two mode shapes affected the whole of the bridge are used to obtain the Rayleigh damping coefficients and the natural period of 1.9seconds was obtained using a damping ratio of 2% for the cable. Previous research suggested a cable removal time as 0.05s in modeling the amplitude and time of cable loss (Mozos & Aparicio, 2011). However, this research is using the simplified blast load amplitude recommended by the US Department of Army as shown in Figure 3 (United FAcilites Criteria (UFC), 2008). The corrosion is considered for 0%, 10%, 20%, 25%, 30%, 40%, 50% and 55% loss of area. The boundary of 55% corrosion is governed by the limiting stress of 45% of the UTS of the cable. To apply the corrosion before the blast in Abaqus 6.13, the cable force is reduced first due to corrosion over the natural period of the bridge (1.9 seconds) in a step after which the amplitude of the blast load is applied to the remaining cable force in another step.
4 RESULTS AND DISCUSSION 4.1 Dynamic response to cable loss after varying corrosion extent To obtain the dynamic response of the cable-stayed bridge model to blast, only cable CP21 is subjected to corrosion. The structure is analysed in this condition over the natural period of the bridge (1.9 seconds) after which blast load is applied to the same cable. The stress redistributed into the cable CP1 (backstay) is obtained over the first 7.5 seconds of the analysis period. When a cable is lost, the stress in the cable is redistributed to the other cables on the bridge, causing a stress fluctuation. The results show a significant increase in stress for the corrosion variation of 30%, 40%, 50% and 55% before the blast with a stress increase of up to 50 N/mm2 in the cable CP1. The loss of the cable CP21 due to blast after 10% corrosion gives a stress profile over time slightly higher than that with the profile obtained with no consideration of corrosion before the blast, while the loss of the cable due to blast after 50% and 55% corrosion redistributes the highest stress in cable CP1 as shown in Figure 4. Also, in examining the effect of the loss of the corroded cable CP21 to blast, the bending moment in the cross beam attached to the deck anchorage from which the cable is lost is also investigated at 2.45 seconds which is 0.04 seconds before the blast event occurred. A 9.5% increase in moment is observed in the cross beam for the 50% corrosion before blast scenario when compared to the moment in the cross beam with 25% corrosion before blast as shown in Figure 5a. This shows the tension in the cable reduces with an increase in the corrosion in the cables. However 25% corrosion of the cables before the blast occurs does not cause a significant change in the moment in the cross beam after blast has occurred when compared with the moment when no corrosion is considered before the blast. 931
Figure 4.
Stress in cable CP1 (backstay) due to the loss of the cable CP21.
Figure 5. (a) Moment of cross beam at point of cable loss due to blast after corrosion and (b) Displacement of node at point of cable loss on deck over 7.5 seconds after the blast at 3.5 seconds.
Figure 5b shows the displacement of the node from which cable CP21 is lost at the deck anchorage over the first 7.5 seconds (blast occurs at 1.9 seconds). During the first 1.9 seconds which is before the blast occurs, the displacement increases as corrosion increases reducing the serviceability confidence of the bridge user which will be a major issue on cable-stayed bridges with large cable spacing. After the blast, the effect of the extent of corrosion is noticeable in the higher amplitudes of displacements compared to when corrosion is not considered before the cable is lost to due to blast. 4.2 Effect of location of cable lost to blast on corroded cables To examine the effect of a cable in relation to its location with the other cables on the cablestayed bridge model the cables CP1 (backstay) and cable CP21 (longest cable in the main span) are individually subjected to corrosion before the blast. 932
Table 3. Redistributed stresses (N/mm2 ) in cables due to loss of backstay. Corrosion Rates Cable Lost
0%
10%
20%
25%
30%
40%
50%
55%
CP1 other CP2 CP21 CP61 CP10
449 427 696 482 696
454 430 760 485 696
455 433 761 485 696
456 434 761 485 696
457 434 761 485 696
458 435 761 485 697
474 452 766 498 664
475 453 766 498 664
Table 4. Redistributed stresses (N/mm2 ) in cables due to loss of longest cable in the main span. Corrosion Rates Cable Lost
0%
10%
20%
25%
30%
40%
50%
55%
CP1 CP21 Other CP22 CP58 CP10
740 607 714 282 411
771 612 800 284 414
801 612 854 284 414
901 612 905 285 414
101 613 907 285 414
101 613 911 285 414
101 614 914 286 414
101 614 916 261 414
4.2.1 Loss of backstay (Cable CP1) The redistributed stress to the cable CP2 on the same plane (nearest cable), cable CP1 on the other plane of cable (adjacent cable), cable CP10 on the same plane (shortest cable), cable CP21 on the same plane (longest cable along the main span) and cable CP58 on the same plane (backstay) is summarized in Table 3 for the loss of the cable CP1. The results show that the cable CP10 does not show an increase in stress as the corrosion extent is varied from 10% to 40% primarily due to the short length and higher inclination. Interestingly, cable CP58 also shows a similar behavior to the shortest cable CP10 even though it has a longer length and higher inclination but it is farthest away from the location of the lost cable. However, due to the loss of cable CP1, the stress in cable CP21 reaches a value of 747 N/mm2 which exceeds the 45% UTS limit (1660 N/mm2 ) for all corrosion variations before blast while cable CP1 on the other plane and CP2 on the same plane show a gradual increase in stress as the corrosion percentage is increased without exceeding the limiting stress of the cables. 4.2.2 Loss of longest cable on the main span (Cable CP21) For the loss of the cable CP21 which is the longest cable in the middle of the semi-fan set, Table 4 shows the stress redistributed to cable CP1 on the same plane (backstay), cable CP21 on the other plane (adjacent cable), cable CP22 on the same plane (the nearest cable to the cable lost), cable CP58 (the other backstay) and cable CP10 (the shortest cable). The results show that cables CP1 and CP22 both have stresses greater than 45% of the UTS of the cable for all corrosion variation cases before blast, leading to more than one cable being below the UTS as opposed to when cable CP1 was lost which shows only one cable (CP21) with stresses above the UTS limit. Cables with stresses greater than the limiting stress will fail and will have their forces redistributed to other cables which leads to progressive collapse and can likely cause the bridge to fail. The cable CP10 has the same stress for all corrosion variations which is similar to its response when cable CP1 is lost which implies that the loss of the corroded cable has no effect on the cable CP10. Cable CP22 experiences a gradual increase in stress as the corrosion extent was increased. 933
Table 5. DAF for cables subject to corrosion variation. Stress
Moment
Corrosion %
CP58
CP21 (other)
CP22
CP1
CP10
Deck (M)*
Pylon (M)*
0% 10% 20% 25% 30% 40% 50% 55%
1.89 1.93 1.94 1.95 1.95 1.97 2.11 2.12
1.05 1.47 1.08 1.09 1.10 1.11 2.11 2.12
4.55 4.62 4.63 4.63 4.63 4.63 4.72 4.72
4.52 4.56 4.63 4.56 3.85 4.56 5.11 5.12
3.03 3.03 3.03 3.03 3.03 3.04 2.78 2.78
4.99 4.97 5.05 5.11 5.19 5.25 5.31 5.57
11.43 11.48 11.50 11.54 11.58 11.595 11.61 11.70
(M)* – Moment
When comparing the effect of the blast after corrosion on the other cabes in relation to the location of the lost cable, the worst effect is noticed when cable CP21 (the longest cable) is lost. A difference of 300 N/mm2 with 25% corrosion before blast and 500 N/mm2 with 50% corrosion before blast is obtained when comparing the stress due to the loss of cable CP1 with cable CP21. Therefore, the loss of the longest cable will cause a more critical effect when compared to the loss of the backstay. 4.3 Effect of loss of cable after corrosion on DAF Codes and guidelines recommend the use of a dynamic amplification factors to cater for the sudden loss of a cable. Corrosion before the blast of a cable will affect the DAF, thus the DAF is calculated for all the corrosion loss percentages using the equation presented in Equation 1.
where: Edyn = Dynamic Effects to extreme events, Estatic = Static response to removal of a cable, Ei = Static Effect in the initial state. Table 5 presents the DAF obtained for cable CP21 since this cable has been shown to have the most severe effect when lost with only this cable experiencing corrosion on the bridge. The values obtained shows an average DAF of 4.64 for cable CP22, while the cable CP1 has an average DAF of 4.61 for all corrosion extents before the blast which is twice the value recommended by the code. An average DAF of 1.98 is obtained for the cable CP58 which is lower than the DAF obtained for cable CP1 also a backstay (this cable is closer to the cable lost than cable CP58). This shows that there is less dynamic effect on the cable farther away from the location of the cable loss. The effect of the loss of cable CP21 on the second cable CP21 in the other plane has an average DAF value of 1.39 which is due to the torsion/twist in the deck when the cable is lost. The DAF obtained for the cable CP10 is the same for the 10%, 20%, 25%, and 30% corrosion before the blast (3.03) and a gradual increase in its DAF values as the corrosion extent is increased from 30% to 40% is observed from 3.03 to 3.04 then a reduction to 2.78 for 50% and 55% corrosion before the blast. This increase then decrease in value implies a variation in the static effect in the cables initial state (Ei ), therefore a quasi-static analysis as recommended by the codes and guidelines does not give the same response as the dynamic analysis for the cable CP10. When comparing the dynamic maximum moment in the deck when subjected to blast to its static state; an average DAF of 5.18 is obtained which is higher than the 1.5 and 2 – the code recommended value along the longitudinal center of the deck. The presence of corrosion in the cable will cause the deck to experience some bending due to loss of tension in the cable and when the blast occurs this will further increase the resonance of the deck. 934
An average DAF of 11.54 is achieved when the dynamic and static moment in the pylon is obtained and compared. The moment in the pylon gives the highest DAF value which is due to the pylon been excited by the cable lost but also the forces have been redistributed to the other cables and the deck vibrating. 5 CONCLUSIONS This paper investigates the structural response of cable-stayed bridges when a corroded cable is subject to blast load by undertaking a nonlinear dynamic analysis using a three dimensional model of a cable-stayed traffic bridge. The effect of the loss of the longest cable in the main span (CP21) on other cables, the cross beam and anchorage point is accessed as well as the effect of the location of the loss for the corroded cable to blast on other cables and finally the DAF of the cables, deck and pylon to the loss of a corroded cable to blast is determined. The aim of this paper is to prove that the presence of corrosion in cables will have a significant effect on the structural response of a cable-stayed bridge to blast. The results obtained show that even before the blast occurs the stress redistribution is noticed in the cables due to loss of tension in the corroded cables. This makes blast on a corroded cable have a more significant effect on the other cables, cross beam and anchorage points. It is therefore advised that humidity control to be installed on cables to reduce the rate of corrosion of cables on cable-stayed bridges. No significant effect of the loss of a corroded cable subject to blast is noticed in the short cable with high inclination and the cables farther away from the location of loss, therefore the cables closer to the corroded cable lost to blast should have a different design governing criteria as well as cable with higher stiffness. The loss of the longest cable in the main span of the bridge (cable CP21) gives the worst effect as it is the least stiff cable due to its low inclination hereby redistributing high stresses and causing two cable to exceed the limiting stress for cables and further progressive collapse is likely. The DAF obtained for the cables close to the corroded cable lost is about two (2) times higher than the code recommended value for the sudden loss of a tension member. The deck also has high DAF value while the highest DAF obtained was for the pylon. This should serve as a guide when designing or maintaining these cables. Further studies are in progress assessing the effect of blast on the cables with more than one cable subject to corrosion on a cable-stayed bridge. REFERENCES Agrawal, A. K., & Yi, Z. (2009). Blast Load Effects on Highway Bridges. Transportation Research, (April). Bruno, D. (2008). Dynamic impact analysis of long span cable-stayed bridges under moving loads. Engineering Structures, 30(4), 1160–1177. doi:10.1016/j.engstruct.2007.07.001 Deng, R., & Jin, X. (2009). Numerical Simulation of Bridge Damage under Blast Loads. WSEAS Transactions on Computers, 8(9), 1564–1574. EC0. (1990). Eurocode — Basis of structural design, BS EN 1990:2002 +A1:2005. British Standard Institution. doi:ICS 91.010.30; 91.080.01 EC1-2. (2003). Eurocode 1: Action on structures – Part 2: Traffic loads on bridges. British Standard Institution. doi:ICS 91.010.30; 93.040 EC3. (2006). Design of steel structures. Part 1.11: Design of structures with tension components (BS EN 1993.). British Standard Institution. EN 1991-1-2. (2002). Eurocode 1: Action on structures – Part 1-2: General actions – Actions on structures exposed to fire. (European Committee for Standardization, Ed.). CEN. Gerasimidis, S., & Baniotopoulos, C. C. (2011). Disproportionate collapse analysis of cable-stayed steel roofs for cable loss. International Journal of Steel Structures, 11(1), 91–98. doi:10.1007/S13296-011-1008-4 Gimsing, N. (1997). Cable supported bridges, concept and design. (2nd ed.). John Wiley and Sons, Inc.
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Hamilton III, H. R., Breen, J. E., & Frank, K. H. (1995). Investigation of Corrosion Protection Systemsfor Bridge Stay Cables. Mays, G., & Smith, P. (2012). Blast Effects on Buildings (2nd Editio.). London. Mozos, C. M., & Aparicio, A. C. (2010). Parametric study on the dynamic response of cable stayed bridges to the sudden failure of a stay, Part II: Bending moment acting on the pylons and stress on the stays. Engineering Structures, 32(10), 3288–3300. doi:10.1016/j.engstruct.2010.07.002 Mozos, C. M., & Aparicio, A. C. (2011). Numerical and experimental study on the interaction cable structure during the failure of a stay in a cable stayed bridge. Engineering Structures, 33(8), 2330–2341. doi:10.1016/j.engstruct.2011.04.006 Pranata, Y. A., & Madutujuh, N. (2012). Dynamic time history analysis of blast resistant door using blast load modeled as impact load, XXI (January), 1127–1134. PTI. (2007). Recommendations for Stay Cable Design, Testing and Installation (5th ed.). USA: Post-Tensioning Institute. Roura, P. G. (2011). Structural Safety in Cable-Stayed Bridges when a Failure of a Rope Occurs. Retrieved from upcommons.upc.edu/pfc/bitstream/2099.1/6168/2/01.pdf Ruiz-Teran, A. M., & Aparicio, A. C. (2009). Response of under-deck cable-stayed bridges to the accidental breakage of stay cables. Engineering Structures, 31(7), 1425–1434. doi:10.1016/j.engstruct.2009.02.027 Starossek, U. (2006). Progressive Collapse of Bridges – Aspects of Analysis and Design Previous Research. Symposium A Quarterly Journal In Modern Foreign Literatures, 1–22. United FAcilites Criteria (UFC). (2008). Structures to resist the effects of accidental explosions. (U. . Ar. C. of Command, N. F. A. Engineering, & Air Force Civil Engineer Support, Eds.) (UFC 3-340-., Vol. UFC 3–304-). USA. Vikas, A. C., Prashanth, M. H., Gogoi, I., & Channappa, T. M. (2013). Effect of Cable Degradation on Dynamic Behavior of Cable Stayed Bridges, 3(1), 35–45. doi:10.5923/j.jce.20130301.04 Wolff, M., & Starossek, U. (2009). Cable loss and progressive collapse in cable-stayed bridges. Bridge Structures: Assessment, Design and Construction, 5(1), 12. doi:http://dx.doi.org/10.1080/15732480902775615 Xinzheng, L. U., & Jianjing, J. (2002). Dynamic FEA and Simulation for A Series of Blast-Resist-door. Progress in Safety Science and Technology, Sept, 839–842. Xu, J., & Chen, W. (2013). Behavior of wires in parallel wire stayed cable under general corrosion effects. JCSR, 85, 40–47. doi:10.1016/j.jcsr.2013.02.010
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Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
Fatigue analysis of cable anchorages on cable-stayed bridges N.A.M. Khairussaleh University of Surrey, Guildford, Surrey, UK Universiti Malaysia Pahang, Gambang, Pahang, Malaysia
G.A.R. Parke & M. Imam University of Surrey, Guildford, Surrey, UK
ABSTRACT: The cable anchorage blocks that transfer the deck loads to the cables have complex details which significantly influence the long – term fatigue behaviour of cable – stayed bridges in operation. Determination of the stress ranges at one of the most critical parts in the cable anchorage due to traffic loads passing over the bridge is determined by using the finite element (FE) method. This paper investigates three different approaches namely node stress concentration, average elements and the hot-spot method in order to identify the stress ranges that adversely affect the remaining fatigue life of a cable anchorage. Finally, by using the Fatigue Load Model 4 (FLM4) proposed in the Euro code, the proposed fatigue remaining life in cable anchorage is determined and discussed.
1 INTRODUCTION 1.1 Fatigue in cable anchorages Identifying the fatigue stress range occurring in cable anchorages is becoming important in modern long span bridges in order to fulfil their serviceability requirements and to maintain their structure integrity. This is because the bridge structure is susceptible to the dynamic effect either from the increasing traffic volume or fluctuating wind loads as these bridge structures become lighter and slender to fulfil the design demand. The failure of the cable anchorage boxes on the Fred Hartman Bridge in 1997 has driven researchers to study intensively the stresses causing fatigue not only in cables but also peculiarly in cable anchorages primarily due to traffic loads and/or by other dynamic influences such as wind loading. Consistent findings show that most of the fatigue damage at cable anchorages is observed close to welded details (Xu et al., 2012; Chen et al., 2012; Lin et al., 2012 and Wang et al., 2013). This is because the transition of force between the connected plate elements and the associated change in stiffness result in stress concentrations which are critical locations in a bridge structure with respect to fatigue damage. A fatigue crack may initiate at these locations and through the applied cyclic traffic loading it may propagate and reach a stage where it may lead to sudden failure (Maddox, 1991). Eight critical fatigue points were identified on the upper deck and in the stay cable anchorages of the DongshuimenYangtze River Bridge in China, predicted out by Chen et al. (2012). Furthermore, Lin et al. stated that the stress concentrations were observed at the welded locations in the anchor box-to-girder web connection (Lin et al., 2012). According to Lin et al., the resulting stresses from the direct shear and bending moment due to eccentricity of the cable force centre line and the bridge deck web produce a stress concentration in the area of the anchorage zone Lin et al. (2012). In addition, Wang et al. indicated that the maximum stress and critical stress concentration occurred at the rear of the weld between the gusset plate and the steel anchorage tube under the maximum design force (Wang et al., 2012). 937
1.2 Aim of the study The aim of this paper is to analyse the remaining fatigue life of an anchorage block used in a cable – stayed bridge. The anchorage block of a typical cable – stayed bridge is analysed as a threedimensional finite element (FE) sub-model which was derived by using global forces obtained from the global FE analysis of the same multi-span cable bridge. The Fatigue Load Model 4 (FLM 4) has been used in the global analyses to obtain the fluctuating axial stresses at the cables due to the traffic loads passing over the bridge (BS EN 1991-2:2004). These are then applied to the sub-model to investigate the fatigue behaviour of the anchorage. The difficulty of analysing stress concentrations through the FE method is well-known because of the singularities that these cause. For this reason, three different methods are used to determine the stress ranges namely; the node stress concentration, average elements and the hot-spot method. This is because fatigue damage is very sensitive to the level of the stress ranges applied and, therefore, accurate estimation of these stresses is essential for reliable remaining life predictions. 2 FINITE ELEMENT MODEL (FEM) 2.1 Anchorage model The anchorage block model adopted for the analysis was taken from the detail of a typical stay cable anchorage design. The anchorage is located on the deck and it is connected to the longitudinal girder by fillet welds. The gusset plates and the local stiffeners are also provided at the edge of the girder top flange and girder web to help in the transference of load. From the global analysis of the bridge under traffic loads, it was found that the highest stress range will occur in the outer cable located on the back span which agrees with the findings of Pipinato et al. (2012). Therefore, the outer anchorage block located at the back stay of a cable – stayed bridge was chosen for this investigation. This particular anchorage block is tilted by angles of 27.88◦ and 8.24◦ with the longitudinal (x-axis) and transverse directions (z-axis), respectively. Linear elastic finite element analysis was performed in the commercial software Abaqus 6.11 with pressure applied at the bearing plate connected to the anchorage tube as shown in Figure 1 (Simulia, 2011). A history of twenty three pressures (shown as P0 to P22) was applied to the bearing plate obtained from the axial cable reactions derived from the global bridge analysis. This was to model the fluctuating forces originating from lorries (based on FLM 4) crossing the bridge. 2.2 Mesh sensitivity Three FE models of the anchorage were analysed to investigate the effect of the mesh size on the resulting stresses as a convergence study. Each model has a different number of total elements in the
Figure 1. A 3D block anchorage model with pressure applied at the bearing plate.
938
Figure 2.
Different parts of the mesh generated for the FEM analysis.
mesh in an attempt to refine the accuracy of the analysis of the connection between gusset plates and the anchorage tube as shown in Figure 2. Three different mesh sizes were used, namely; 5 mm × 5 mm (112 811 elements), 10 mm × 10 mm (59 106 elements) and 28 mm × 38 mm (45 107 elements). The mesh sizes of 10 mm × 10 mm and 28 mm × 38 mm were used to conform to the guidance given in determining the hot-spot stress when using the hot-spot method (IIW 2008; DNV-RP-C203 and BS 7608:2014). In parallel to using nodal stresses at the location of the stress concentration (node concentration) and a stress averaged over a number of elements at the vicinity of the stress concentration (average elements), the hot-spot stress is an alternative approach for calculating fatigue damage. Due to the difficulty of evaluating local stress in the connection details required for the application of the S-N curves and to avoid the singularity that usually emanates at notches and sharp corner connections, the hot-spot stress is used as a comparison study to obtain reliable stresses used to calculate local stresses and subsequently, fatigue damage. 3 THE HOT-SPOT METHOD 3.1 Introduction Practically, the hot-spot method is used to determine the hot-spot stress when considering potential fatigue damage from welded joints formed between plates and other structural members (BS 7608:2014). The stress concentrations often arise at the location of the weld toes. As result, the weld toe region has to be modelled with a certain radius to obtain reliable results for notch stresses in the finite element (FE) analysis as well as accounted for in the S-N curves when estimating fatigue damage (Radaj, D., Sonsino, C.M. and Fricke, W. 2006; DNV-RP-C203 and BS 7608:2014). According to Doerk, Fricke and Weissenborn, the notch effected region will cover distances to 0.3 to 0.4t (t = plate thickness) away from the weld toe for relatively thick tubular joints. This is based on the approach of the reference point of stress evaluation and extrapolation at a certain distance away from weld toe (Doerk, O., Fricke, W. and Weissenborn, C., 2002). 3.2 Evaluation of hot-spot stress The hot-spot stress can be estimated by using a technique called surface stress extrapolation (SSE). Normally, two common stress extrapolations are used; the stress distribution depends upon the plate thickness; type ‘a’ and the stresses located at an absolute distance from the weld toe; type ‘b’ (BS 7608:2014). Typical extrapolation methods are either linear or quadratic extrapolation based on the meshing used in the FE model. International Institute of Welding (IIW) and BS 7608 recommend a linear extrapolation over two different points; 0.4t and 1.0t or 0.5t and 1.5t away from the notch (stress concentration) where t is the adjacent plate thickness as shown in Figure 3. Alternatively, for the fine mesh, a quadratic extrapolation is used with three locations from the weld toe; 0.4t, 939
Figure 3.
Recommended stress surface distribution technique by IIW (2008).
Figure 4.
Stress concentrations regions at the cable anchorage analysed.
0.9t and 1.4t (IIW 2008; BS 7608:2014; Doerk, O., Fricke, W. and Weissenborn, C., 2003). For stress analysis that use the type ‘b’ method, a quadratic extrapolation is used at the correspond mesh points 4 mm, 8 mm and 12 mm while a linear extrapolation is used for the coarse mesh size 10 mm × 10 mm at distances of 5 mm and 15 mm (IIW 2008; BS 7608:2014; Doerk, O., Fricke, W. and Weissenborn, C., 2003). 4 RESULTS 4.1 Stress concentrations The regions where stress concentrations occur at the anchorage as shown in Figure 4; these are in agreement with the stress regions mentioned by Wang et al. (2013). This is because many discontinuous elements or an abrupt change occurs at the junction between parts/plates being joined resulting in a non-uniform stress flow which increase the stresses at these particular areas. However, higher stresses were found at locations (a), (b), (d) and (f). Due to the smaller area and the number of parts joining together at the same location, the stress ranges occurring at location (a) were chosen for the fatigue damage analysis. 4.2 Derivation of stress ranges Figures 5 to 7 show the stress histories and the resulting stress ranges obtained from the node stress concentrations with different mesh densities for the three mesh sizes; 5 mm × 5 mm, 940
Figure 5.
Stress history and stress ranges obtained by mesh size 5 mm × 5 mm.
Figure 6.
Stress history and stress ranges obtained by mesh size 10 mm × 10 mm.
Figure 7.
Stress history and stress ranges obtained by mesh size 28 mm × 38 mm.
10 mm × 10 mm and 28 mm × 38 mm respectively. The twenty three pressure load cases (P0 to P22) correspond to the cable forces obtained from the global bridge analyses under the FLM4. Figure 8 to 10 show the graphs of the stress ranges obtained from averaging the elements. Two elements surrounding a node point of stress concentration were considered for the three different mesh sizes; 5 mm × 5 mm, 10 mm × 10 mm and 28 mm × 38 mm respectively. Figure 11 to 13 display the graphs of the stress ranges obtained from the hot-spot method for three different mesh sizes; 5 mm × 5 mm, 10 mm × 10 mm and 28 mm × 38 mm respectively. 941
Figure 8.
Stress history and stress ranges obtained by mesh size 5 mm × 5 mm.
Figure 9.
Stress history and stress ranges obtained by mesh size 10 mm × 10 mm.
Figure 10.
Stress history and stress ranges obtained by mesh size 28 mm × 38 mm.
4.2.1 Stress ranges in 5 mm × 5 mm meshing Compared with the two first approaches; node concentration and average elements, the graphs of the hot-spot method show a different stress history pattern generated by the applied cable forces. The stress ranges in the hot-spot method demonstrate an increasing fluctuation pattern starting at Point 3 (P3) where the highest stress range is 24.91 N/mm2 occurring between point P10 to P11. The graph of average elements depicts the highest stress range at the same location as with the hot-spot method but not at the same location as that predicted using the node concentration approach in 942
Figure 11.
Stress ranges generated in mesh size 5 mm × 5 mm.
Figure 12.
Stress ranges generated in mesh size 10 mm × 10 mm.
Figure 13.
Stress ranges generated in mesh size 28 mm × 38 mm.
which it occurs at point P18 to P19 with a stress range of 34.88 N/mm2 . Despite the stress ranges value, the graphs obtained using the node concentration and average elements approaches show a similar pattern compared to the hot-spot method graph. This is because they indicate stresses that occur in the anchorage block due to the fluctuating tensile forces that are compatible with the cable reactions from the global analysis. 943
Figure 14. Damage accumulation for the average elements approach by using 10 mm × 10 mm meshing where based on BS EN 1993-1-9 and BS EN 1991-2-2004.
4.2.2 Stress ranges in 10 mm × 10 mm meshing The graphs of the stress ranges derived from the global analysis using this mesh size show quite a similar pattern for all three of the methods used. However, further investigation into the node concentration approach shows that the stress occurring in the anchorage block is seen to decline over time. 4.2.3 Stress ranges in 28 mm × 38 mm meshing With this mesh size, only the stress ranges using the average element approach is compatible with the pressure applied. Yet, all the graphs depict the same location of the highest stress range which is between point P18 to P19 (see Figs. 7, 10 and 13). In addition, both stress histories generated using the node concentration approach and the hot-spot method approach illustrate a decreasing stress history pattern starting at Point 3 (P3). 4.3 Evaluation of fatigue stress ranges The derivation of the stress ranges was then carried out to evaluate the resulting fatigue damage. The strength of the particular structural detail category used for estimating the fatigue damage is given in BS EN 1993-1-9 (2005). Each of the detail type corresponds with a particular design S-N curve. In this study, the stress ranges generated using the 10 mm × 10 mm mesh and the average elements approach and the hot-spot method approach were used to calculate the fatigue damage at the anchorage block. 4.4 Remaining fatigue life As has been mentioned previously, the fatigue loads driven from the global analysis was generated from the fatigue load model 4 (FLM 4) (BS EN 1991-2:2004). Therefore, with the stresses resulting from the action effects, the damage accumulation produced from FLM 4 traffic load can be estimated. The Reservoir method was used to carry out the stress range cycle counting (Maddox, 1991). According to Miner’s rule, the limit value for the damage accumulation (Dmax ) for fatigue life is indicated as Dmax = 1.0 (BS EN 1993-1-9) (Palmgren-Miners’ Rule). However, a more conservative value such as Dmax = 0.5 is recommended by IIW (2008) and BS 7608 (2014) due to the large uncertainties present in the fatigue phenomenon. According to Nussbaumer, Borges and Davaine (2011), the use of Dmax = 0.5 is accepted for verifications under both proportional and non-proportional multi – axial stress cases. The results shown in Figure 14 and Figure 15 represent the total fatigue damage accumulation corresponding to the detail category 45 for the average elements approach and the hot-spot method 944
Figure 15. Damage accumulation for the hot-spot method approach by using 10 mm × 10 mm meshing where based on BS EN 1993-1-9 and BS EN 1991-2-2004.
approach both related to the 10 mm × 10 mm meshing respectively. The damage accumulation is computed using the following equation given in BS EN 1993-1-9, Annex A:
where Dd = total damage, Di = damage caused by an applied load cycle, nEi = number of cycles associated with the stress range, NRi = number of endurance cycles for the stress range applied associated with the S-N curve, Dmax = limit value of the damage accumulation. 5 DISCUSSION The basic idea of using Fatigue Load Model 4 (FLM 4) was intended to determine the total damage accumulation using Miner’s rule. This was undertaken by using a conventional number of damage cycles defined by Miner’s rule (e.g. 2 × 106 ) with the set of lorries crossing the bridge in order to simulate the real traffic during the intended lifetime of the bridge. In general, two types of approach were used, namely, the average elements and the hot-spot method in order to predict the possible fatigue damage occurring in an anchorage block. The choice was made based on the compatibility of the stress range patterns with the reaction occurring from the cable forces derived from the global bridge analysis. Furthermore, the results from the same mesh density were used in order to identify which approach best suited the fatigue appraisal as defined in the codes of practice. The computed total fatigue damage using Miner’s rule for both approaches used with the 10 mm × 10 mm meshing; the average elements and the hot-spot method indicated that no fatigue failure would occur in the detail studied. Using the average elements approach, the fatigue life appeared to be infinite because the damage accumulation is very small (Dd = 0.029). On the other hand, the total damage accumulation, Dd obtain by using the hot-spot method is found to be equal to 0.33 which is substantially below the specified limit of 1.0 recommended in BS EN 1993-1-9. This indicated that the detail studied has a fatigue life of 96 years which is 21 years longer than the design life specified by the bridge designers who used the AASHTO code of practice. However, it should be noted that to achieve this design life, maintenance is required as recommended in the codes of practice and design standards. In addition, comparison of the fatigue damage obtained between the two methods also demonstrate how sensitive the fatigue is to the level of the stress ranges, showing the importance of accurate stress estimation in the analyses of similar cases. From the result obtained, if the stress ranges are increased by 10% in order to simulate increasing of traffic volume, the new total fatigue damage calculations reveal that Dd = 0.531 which shows 945
that the connection detail is still safe against fatigue failure. On the contrary, if the Dmax limit value of 0.5, as recommended by the IIW (2008) and BS 7608 (2014), the detail is deemed unsafe during its service life. This conservative value should be considered if the structure is subjected to various axial and bending secondary stress cases, not only from traffic volume increases but also from the flexural loading caused by wind forces which might affect its strength and robustness. Therefore, by calculating this damage accumulation value, a further analysis should be undertaken in order to determine a more reliable assessment of fatigue life particularly for the cable anchorage blocks. 6 CONCLUSION Comparison of the stress histories and resulting stress ranges obtained from the finite element analysis of a cable anchorage by using different mesh sizes and different averaging/extrapolation methods showed their sensitivity to the latter. Comparison of the damage accumulation obtained by using 10 mm X 10 mm meshing for each of two approaches, namely the average elements and the hot-spot method clearly show that the selected structural member used for this investigation remains safe under the traffic fatigue model 4 (FLM 4). The total damage obtained from the hotspot method was found to be considerably higher than the averaging method. A 10% increase in all stress ranges, assumed to represent future loading increases was found to affect the fatigue damage considerably, but still being on the safe side. REFERENCES British Standards Institution 2004, BS EN 1991-2 Part 2: Traffic Load on Bridges, British Standards Institution, London. British Standards Institution 2008, NA to BS EN 1991 Part 2: Traffic Loads on Bridges, British Standards Institution, London. British Standards Institution 2014, BS 7608: Guide to Fatigue Design and Assessment of Steel Products, British Standards Institution, London. Chen, X., Yin, D., Lai, Y. & Liu, X. 2012, “Fatigue Considerations in the Design of Cable-Stayed Road-Rail Bridges with Orthotropic Steel Deck”. Proceeding of the 9th Asia Pacific Transportation Development Conference – Sustainable Transportation System: Plan, Design, Build, Manage, and Maintain American Society of Civil Engineers (ASCE), United States of America, pp. 491-499. DNV – RP-C203 2011, Fatigue Design of Offshore Steel Structures. Doerk, O., Fricke, W. & Weissenborn, C. 2003, “Comparison of Different Calculation Methods for Structural Stresses at Welded Joints”, International Journal of Fatigue, vol. 25, no. 5, pp. 359–369. International Institute of Welding 2008, Recommendations for Fatigue Design of Welded Joints and Components, IIW Commission XIII and XV edn, France. Lin, C., Lin, K., Tsai, K., Jhuang, S., Lin, M., Chen, J., Chen, P., Wang, K. & Lin, S. 2012, “Full-Scale Fatigue Tests of A Cable-To-Orthotropic Bridge Deck Connection”, Journal of Constructional Steel Research, vol. 70, no. 0, pp. 264–272. Maddox, S.J. 1991, Fatigue Strength of Welded Structures, Woodhead Publishing. Nussbaumer, A., Borges, L. & Davaine, L. 2011, Fatigue Design of Steel and Composite Structures, 1st edn, European Convention for Constructional Steelwork, Portugal. Pipinato, A., Pellegrino, C., Fregno, G. & Modena, C. 2012, “Influence of Fatigue On Cable Arrangement In Cable-Stayed Bridges”, International Journal of Steel Structures, vol. 12, no. 1, pp. 107–123. Radaj, D., Sonsino, C.M. & Fricke, W. 2006, Fatigue Assessment of Welded Joints by Local Approaches, 2nd. Edition edn, Woodhead Publishing Limited, Cambridge, England. Simulia, D. 2011, ABAQUS 6.11 Analysis User’s Manual, Dassault Systèmes, USA. Wang, Y., Wang, Z., Weiß, X. & Qiang, S. 2013, “Test and Finite Element Analysis of Gusset Plate Anchorage for Cable-Stayed Bridges”, Stahlbau, vol. 82, no. 4, pp. 313–321. Xu, Z., Wang, X., Zhao, R. & Wang, C. 2011, “Full-Scale Model Test For One Box With Three Rooms Cable-Pylon Anchorage Zone of Cable-Stayed Bridge”, Third International Conference on Transportation Engineering (ICTE 2011), pp. 1970–1976.
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Monitored-based methodology to predict the initiation of corrosion in RC structures E.A. Tantele, R.A. Votsis & T. Onoufriou Department of Civil Engineering and Geomatics, Cyprus University of Technology, Limassol, Cyprus
ABSTRACT: The corrosion of steel reinforcement in RC structures is the main deterioration factor of these structures and the employment of SHM techniques can provide indications on the corrosion activity at its early stages. In this study, a methodology is proposed to predict the initiation of corrosion on RC structures using information from SHM data in order to alleviate the impact of uncertainties currently employed in theoretical corrosion models. The monitored data are obtained from the half-cell potential and the concrete resistivity methods. In the methodology the data processing is combined with condition rating and risk assessment principles in order to assess the current structural condition and predict when corrosion initiation is due. The results show that the proposed methodology can provide information on the timeframe of the corrosion state of RC members and most importantly before its effect are visible and the repairing work is mandatory and costly. 1 INTRODUCTION The deterioration of the reinforced concrete (RC) structures is attributed to many factors but the corrosion of the steel reinforcement bars (rebars) is considered as the leading deterioration mechanism. Every year vast amounts of financial resources are spend by the infrastructures managing authorities on repairs and maintenance in order to address the effects of corrosion. The latter is influencing directly the mechanical properties of the embedded steel rebars (Francois et al. 2012, Du et al. 2005) and consequently the RC member is affected: its total strength is reduced because of the reduction of steel strength due to area loss and the loss of bond between rebar and concrete (Imperatore & Rinaldi 2008). In addition the structural serviceability (appearance of cracks) is affected and researchers are trying to associate the corrosion rate with the appearance and propagation of cracks (Cabrera 1996). These effects are studied extensively and the research effort is focused in determining the corrosion’s progress time path in order to schedule timely interventions which will halt extensive damage and thus reduce the cost of repairs and maintenance. The study and prediction of the corrosion progress is necessary as during the corrosion initiation period there are no visible signs of deterioration; the corrosion effects are visible only when the RC concrete is being subjected to serious damage. Many theoretical models exist in the literature that represent the deterioration of RC structures due to corrosion. The most commonly used is Tutti’s (1982) two stage corrosion model. This theoretical model shown in Figure 1 divides the reinforcement corrosion process into two periods: an initiation period and a propagation period. The initiation period consists of the time from the construction of the structure until aggressive agents (e.g. chlorides) reaches the rebars and depassivate the steel. The propagation period covers the time from the steel depassivation until the structure reached a certain unacceptable level of deterioration. The illustrated profile is adopted by many researchers where in some cases additional stages were introduced to describe in more detail the deterioration process. These models, which basically act as service life estimations, are trying to predict the time of corrosion initiation and its propagation in time until some predefined limits (e.g. reliability index) are reached. 947
Figure 1. Tuuti’s corrosion model.
However, the degree of complexity associated with the chemical and physical processes during deterioration, is making it difficult to describe them within a mathematical framework. Additionally the uncertainties and assumptions used in the theoretical models increases the need for more viable models. The introduction of structural health monitoring (SHM) in the inspection and condition assessment of structures provided an opportunity to improve the performance of the theoretical models but also develop models based either exclusively or partially on monitored data. Cusson et al. (2010) investigated the improvement in service life predictions when monitored field data of the actual structures under study are used, instead of values obtained through literature sources in an attempt to highlight the uncertainty associated with such values. This study proposes a methodology for predicting the initiation of corrosion due to chlorides based solely on data obtained from monitoring activities. These data are used to define the condition state of a structural system at the time of testing and estimate its future performance using subsequent future measurements. The results show that the time for corrosion initiation can be estimated using the monitored based developed methodology. 2 CORROSION MONITORING TECHNIQUES Corrosion is an electrochemical process and as such involves a chemical reaction in the presence of an electrolyte, in which a transfer of electrons takes place between the reactants. Electrochemical principles of reinforcement corrosion in concrete have been described in detail in many textbooks such as Neville (1995) and Mattsson (2001). To achieve an effective condition assessment on RC structures and identify corrosion on time, a variety of electrochemical monitoring non-destructive testing (NDT) techniques have been developed (Song 2007). However, it is stated in the literature that a combination of NDT methods such as the half-cell potential and resistivity can provide adequate information on the probability of corrosion activity of the embedded reinforcement (IAEA 2002). For the current study, data from these two aforementioned methods are employed. A brief description of each method depicted in Figure 2 is given in the following sections. 2.1 The half-cell potential method During the corrosion process the corroding (anode) and the passive (cathode) rebars in concrete have a difference in electrical potential as the free electrons are flow from the anode to the cathode. This potential difference can be measured using the half-cell potential (HCP) method in order to get a characteristic picture of the state of corrosion of the steel surface within the concrete. For this principle a reference electrode (a Cu/CuSO4 half-cell is used in this study) is connected via 948
Figure 2. The half-cell potential (left) and the concrete resistivity (right) principles (Proceq 2009). Table 1. Typical criteria for the HCP values in relation with corrosion risk. Occurring corrosion probability
Value
Low 90%
Source
−200 → 0 mV −200 → +100 mV −200 → +100 mV
−200 to −350 mV
−350 → −600 mV −400 → −600 mV −400 → −600 mV
ASTM C876 RILEM TC 154- 2003 Proceq 2009
a high-impedance voltmeter to the steel reinforcement and is moved in a grid over the concrete surface in an attempt to locate any corrosion activity as shown in Figure 2. The results of the HCP test can be interpret in two ways (Carino 2004): 1) Numeric technique, where the potential values are used to indicate corrosion, and 2) Potential difference technique, using the potential gradients. For the current study the numeric technique is used and Table 1 provides indicative criteria for the interpretation of the obtained measurements from the pertinent literature. The three categories and the values proposed by ASTM C876 (2009) is adopted in this study. The interpretation of those values is that for potentials over an area greater than −200 mV, there is a probability less than 10 percent that steel corrosion is occurring whereas for potentials less than −350 mV there is a probability greater than 90 percent that steel corrosion is taking place in the measured area at the testing time. 2.2 The concrete resistivity method The concrete resistivity (CR), i.e. the ability of concrete to oppose to the flow of electric current, affects the flow of the ions between the anodic and cathodic areas and subsequently the rate of corrosion. As opposed to the HCP method the CR method do not identify the corroded rebars but it gives information about the risk of corrosion damage, i.e. indicates where over a concrete area there are ideal conditions for corrosion to initiate. The most common equipment used to measure CR is the Wenner probe which consists of four electrodes as shown in Figure 2. The indicative values for CR which can be used to determine the likelihood of corrosion, obtained from various researchers and empirical tests are summarised in Table 2. Considering the values proposed in the literature and listed in Table 2, five categories of corrosion risk (ranging from negligible to high risk) associated with CR values have been adapted for the current study as shown in the last row of Table 2. 3 CONDITION ASSESSMENT METHODOLOGY A condition assessment methodology predicting the initiation of corrosion using monitored data obtained at different periods is developed and is illustrated in Figure 3. The methodology is based 949
Table 2. Typical criteria for the CR values in relation with corrosion risk. Corrosion risk (Resistivity, p, values in K.cm) Negligible
Low
Moderate
p > 100–200
100–50 p > 20 100–50 p ≥ 12 – 100–20
50–10 20–10 50–10 8–12 – 20–10
p > 100 p > 100
High 10–5 p < 10 p < 5–10 10–5
Very high
Source
p < 10 p 10)
where R1 , R2 and R3 = additional determinants for linear equation systems solving by Cramer’s rule (which are in relation with Pt ); O = principal determinant of Cramer’s rule for linear equation systems solving. 3. In case of relatively enough amount of statistical data (3 < χ ≤ 10)
It is proposed to use the method of least squares for statistical data processing of reliability Pt = ({t, P}χ ) and Cramer’s rule for linear equation systems solving of principal determinant O = f (Pt ) and additional determinants R1 = f (Pt ), R2 = f (Pt ), R3 = f (Pt ) with parametric values ∀i ∈ χ|tχ > tω according to equations (7–14):
According to mentioned above the main issue for analysis is initial data and present bridge (or element) reliability P which depends on stresses in elements S (according to Building Codes). This could be assumed as:
or as
where σi(1) = permanent loads stress in bridge element i subject to faults, damages and defects absence; σi(1) = f (Hiδ ) = residual stress state effect which depends on information entropy H ; σi(>1) = other loads (not permanent) stress in bridge element i. 1117
Figure 2.
Flow-chart of information entropy changes for system.
In principal flow-chart of bridge service during life period subject to analysis of its technical state and reliability using information theory and entropy looks as follows (Figure 2). Residual stress state effect σi(1) due to rehabilitation and repairing activity can be estimated by fuzzy logic methods according to algorithms which should be developed for certain bridge structure subject to its features and conditions. It should be taken into account after correcting of Pt by maintenance or repairing works (see the Figure 1) the next service stage ω requires test works and applying of adjustment factors to math model for precise determination of reliability Pω with assumed entropy H = 0. The proposed approach allows describing mathematically complicated processes of analysis and estimation of the technical state of bridges in case of damages, defects and failures occurrence during service stages. 1118
REFERENCES Igrickij, V.A. 2012. Structural design using reliability theory. Electronic edition guidelines of Bauman Moscow State Technical University. Moscow. Evaluation of Maintenance Costs in Comparing Alternative Designs for Highway Structures. DMRB BD 36/92. UK Highways Agency. UK Shubin, R.A. 2012. Engineering systems reliability and technogenetic risk. Educational guidelines of Tambov State Technical University. Tambov. Syrkov, A.V. 2012. Life cycle of bridge structures optimization problems. Dorognaja dergava 39(1): 36–41. Syrkov, A.V. 2014. Development trends of automated systems for maintenance of engineering structures. Automation and Remote Control 2(1): 34–38. Vasiljev, A.I. 2008. Traffic structures reliability basis. The Moscow Automobile and Road Construction University. Moscow. Volkov, S.D. 1960. Static theory of strength. MASHGIZ. Moscow.
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Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
Alkali-Silica Reaction, ASR – Review on how to deal with ASR in concrete structures J. Custódio, A. Bettencourt Ribeiro & A. Santos Silva LNEC – National Laboratory for Civil Engineering, Materials Departments, Lisbon, Portugal
ABSTRACT: A significant number of problems related to concrete deterioration have been detected in large concrete structures in Portugal and worldwide, the leading cause being ASR. The importance of these structures, the number of for which ASR has already been identified or is very likely to be diagnosed in a near future, as well as the number of structures that are under or planned for construction, which may also come to develop ASR is why it is still nowadays a major concern and a complex problem. Therefore, a study is being conducted at LNEC to diminish the negative impact of ASR by increasing knowledge on how to reliably control ASR in new structures and on how to properly assess its extent and potential for future development in the existing ones, so that the risks to structural integrity and need for mitigation/remediation actions can be properly assessed. This paper presents methodologies, based on state-of-the-knowledge collected on that study, which may be used by construction industry stakeholders in the prevention of ASR in new concrete structures and on ASR diagnosis and prognosis in existing concrete structures. 1 INTRODUCTION Deterioration of concrete structures by Alkali-Silica Reaction (ASR) has increased dramatically in recent years in Portugal. ASR occurs between alkali hydroxides in concrete pore solution and some siliceous minerals present in certain aggregates, and results in the formation of a hydrophilic gel that expands in the presence of water and, in certain conditions, may disrupt concrete. ASR has important economic implications, since it is normally observed in very large structures (e.g. dams, bridges) and the work necessary to remediate the problem involves large areas of reconstruction and complex and expensive repairing techniques and materials. In addition, ASR diminishes the affected structure service life, may involve the interruption of its function and, ultimately, can lead to its decommissioning and demolishing. In Portugal, ASR is present in tenths of structures. Due to the rhythm at which ASR is being identified in existing structures and the large number of structures under or planned for construction in Portugal, which may also develop ASR, it is predicted that concrete structures deterioration will continue to increase considerably. The reasoning for this is that distress signs appear decades after construction; numerous structures were built with aggregates which are now known to be reactive; concrete formulation considered that the only alkali source was the cement, this is now known not to be true; structures can now be built free from deleterious ASR, but several are being constructed with aggregates for which reactivity tests give unreliable results; large structures require vast amounts of aggregates, so in many cases local rocks are used, thus some new structures use potentially alkali reactive aggregates. 2 NEW CONCRETE STRUCTURES Presently, only two standards deal with ASR, EN 206 (CEN, 2013) and EN 12620 (CEN, 2008), and they just state that actions shall be taken to prevent ASR in new structures using procedures of established suitability. Due to the complexity and multiplicity of factors involved in ASR together with the variability of the materials used (e.g. cements, aggregates, admixtures, supplementary 1121
Figure 1. Methodology for evaluating aggregate alkali-silica reactivity.
cementitious materials), no such established procedures exist, and each country has to rely on national specifications. Because of that, in Portugal, LNEC Specification E 461 (LNEC, 2007) was created in 2004. However, due to the scientific developments occurred since then, a new revised version will soon be published to better support stakeholders on how to prevent ARS in new concrete structures. The main differences between the revised and original version reside in the methodology followed to identify an aggregate reactivity and in the assessment of the level of precaution and choice of appropriate precautionary measures. The methodology envisaged to identify a deleteriously reactive aggregate is presented in Figure 1. It can be seen that in order for a rock be used as an aggregate for concrete in Portugal it is always necessary to assess at the laboratory its ASR reactivity. The petrographic examination shall be made according to LNEC Specification E 415 (LNEC, 1993) and will allow to classify the aggregate as belonging to Class I (very unlikely to be alkali-reactive), Class II (potentially alkali-reactive or alkali-reactivity uncertain) or as Class III (very likely to be alkali-reactive). Although, petrography may be used to classify an aggregate as potentially or likely alkali-reactive, expansion tests are required to determine the extent of the reactivity and appropriate levels of prevention. According to the presented methodology, aggregates may be accepted as non-reactive solely on the basis of petrography but that decision has a certain risk associated; therefore, in some situations it is advisable to perform the expansion tests as well. The expansion tests prescribed comprise an accelerated mortar-bar test – AMBT (ASTM, 2014) and concrete-prism tests – CPT (RILEM, 2014b, RILEM, 2014c, RILEM, 2014d). From all existing tests, it has been considered that those tests produce results more consistent with the ones observed in the field. If the aggregate is considered to be non-deleteriously-reactive (i.e. obtained an expansion lower than the maximum expansion limit defined in tests), it can be accepted for use in concrete with no further consideration of mitigation actions (as long as it complies with NP EN 206-1). If the aggregate is found to be deleteriously reactive, either it is not used or preventive measures shall be used to prevent the development of deleterious ASR in the concrete structure. It has been shown that, for some aggregates, the AMBT might incorrectly identify a deleteriously reactive aggregate as being non-deleteriously-reactive; therefore, in spite of the most reliable 1122
approach for determining aggregate reactivity being the CPT, its duration makes it not feasible for most situations and the AMBT is still essential in the alkali-silica reactivity appraisal. For aggregates that contain porous flint (chert) as a potentially reactive constituent, and in the case of the so called slowly reactive aggregates (e.g. granites and basalts), the AMBT shall not be performed, as it does not provide a reliable result; in this case the route marked with a dashed line in Figure 1 shall be followed instead. The maximum expansion that can be attained by the concrete in the expansion test, which is likely to indicate non-expansive materials, has not received yet a broad consensus. However, it seems that results in the AAR-3.1 test (usually after 12 months) of less than 0.05% are likely to indicate non-expansive materials; whilst, results exceeding 0.10% indicate expansive materials. It is still not possible to provide definitive interpretative guidance for results in the intermediate range of 0.05% to 0.10% and, because of that, aggregates yielding results in this range are considered as being potentially alkali-reactive. For slowly reactive aggregates, a lower criterion of 0.03% at 12 months should be used or, preferably, the test shall be prolonged to 24 months. Independently, of the aggregate evaluated, if expansion is still occurring at 12 months of testing, it is recommended that the test continue until expansion ceases or it has become clear if the criteria will or will not be exceeded. If such continued testing is not feasible, then a judgement will have to be made from the inspection of the shape of the expansion curve up to 12 months, as to whether or not the criteria would be likely to be exceeded during further testing period. In terms of the AAR-4.1 test, it is believed that a maximum expansion in the test of 0.03% at 15 weeks indicates a non-reactive aggregate. Thus, in the case of aggregate producing AAR-4.1 expansion greater than 0.03% at 15 weeks, preventive measures shall be taken to minimise the risk of the development of deleterious ASR in the structure. The findings from the CPT shall always take precedence over the results from petrography examination and AMBT. The assessment of the level of precaution and appropriate precautionary measures is summarized in Figure 2. The process starts by a categorisation of the structure according to the risks associated with ASR occurrence in the structure, namely into three risk categories: R1 – low risk (e.g. non load-bearing elements inside buildings, temporary or short service life structures, easily replaceable elements); R2 – normal risk (e.g. most building and civil engineering structures); and R3 – high risk (e.g. long service life or critical structures where the risk of deterioration from ASR damage is judged unacceptable – nuclear facilities, dams, tunnels, important bridges or viaducts, structures retaining hazardous materials). Then, the service environment to which to structure will be exposed is identified from three possible categories: E1 – the concrete is essentially protected from extraneous moisture; E2 – the concrete is exposed to extraneous moisture; and E3 – the concrete is exposed to extraneous moisture and additionally to aggravating factors, such as sodium chloride based de-icing salts, freezing and thawing or wetting and drying in a marine environment. The previous two categorizations shall be made by the owner or authority responsible for the structure, in collaboration with the designer. Next, the structural and environmental categorisation is combined to provide the level of precaution. Four levels of precaution exist: P1 – no special precautions against ASR; P2 - normal level of precaution; P3 – special level of precaution; and P4 – extraordinary level of precaution. To each precaution level a specific set of preventive measures are then applied. There are four possible measures: M1 – restricting pore solution alkalinity (e.g. achieved by limiting concrete alkali content, using low alkali cement, including sufficient proportion of low lime-fly ash, or other mineral addition that have demonstrated to be effective in the concrete); M2 – ensuring the use of a non-reactive aggregate; M3 – reducing moisture ingress to maintain the concrete in a sufficiently dry state to prevent deleterious expansion of the gel; and M4 – modifying the properties of ASR gel so that it becomes non-expansive. 3 EXISTING CONCRETE STRUCTURES Diagnosis and prognosis of ASR in existing structures are not currently covered by any European standard or regulation. Because of that, LNEC is going to publish a LNEC Specification to provide 1123
Figure 2.
Methodology for selecting preventive measures [adapted from (RILEM, 2014a)].
support to the construction industry stakeholders on how to deal with ARS in existing concrete structures. The methodology that can be followed to appraise an ASR affected structure is summarized in Figure 3. The assessment process can be divided broadly into three stages: (1) initial survey; (2) diagnosis; and (3) prognosis. The first stage consists on the visual observation of the structure, where the visual symptoms of deterioration are annotated and compared to those commonly observed on affected structures (e.g. expansion causing deformation, relative movement, and displacement; cracking; surface discoloration; gel exudations; occasional pop-outs). At this stage, any documents relating to the structure (e.g. exposure conditions; age of structure, details and dates of modifications or repairs; plans, drawings and specifications of the structure; previous surveys or investigations on the structure) and the materials used for the construction (e.g. concrete mix design and details of analyses or tests carried out on concrete constituents) are also gathered and reviewed to assist in the appraisal process and decide upon the likelihood of ASR presence. In the case of being decided that ASR is most likely not present, then a routine inspection plan is defined. If the probability of ASR being present is determined to be significant, then the information collected on the previous stage is used in stage 2 to define the preliminary sampling program to be carried out on a selected number of elements from the concrete members showing visual signs of deterioration. These samples are then subjected a series of destructive tests (e.g. observation by optical and scanning electron microscopes, determination of alkali and cement contents) that allow the diagnosis of the cause(s) of the concrete deterioration and the obtainment of a general assessment of the extent of deterioration. In the case of ASR presence is confirmed and damage to concrete is observed in Stage 2, then the appraisal process advances to Stage 3. In this stage, an extensive inspection programme is devised to allow a structural integrity assessment, and an additional and broader sampling plan is defined to allow a detailed testing program in the laboratory, which quantifies current condition of concrete (i.e. degree of expansion/damage attained to date), evaluates potential for future expansion (i.e. trend for future concrete deterioration), and predicts future structural risks. This sampling plan envisages core extraction in several locations at the structure, including visually deteriorated and 1124
Figure 3.
Methodology for the diagnosis and prognosis of ASR in concrete structures.
sound areas. In the case of a massive structural element, it is also important to extract cores from deep inside the affected element in order to evaluate the degree of the reaction throughout the element thickness. Moreover, cores can be taken also from locations without visible signs of deterioration but that satisfy the requirements to the development of ASR, so that actions to mitigate IER can be undertaken in the initial stage of the reaction, thus minimizing further damage to the concrete. The number of samples required depends on the type and complexity of the structure. The results of the detailed investigation are then analysed and decisions made regarding the need to plan and implement in-situ monitoring programs (measuring expansion and deformation) and repair and/or mitigation strategies. 4 CONCLUSIONS AND FURTHER RESEARCH NEEDS This paper presented methodologies, resulting from LNEC’s accumulated expertise and deriving from the most recent findings of the international scientific community, that may be used by the construction industry stakeholders to better control ASR in new concrete structures, and to more adequately manage affected structures; thus, enabling a more sustainable and effective use of the country’s economic and natural resources. Even though, current knowledge is, in most cases, sufficient to prevent ASR in new structures, several issues still require further investigation. For instance, petrographic examination is still unclear about the potential reactivity of some minerals present in granite like rocks; current expansion tests may produce unreliable results with some aggregate types (e.g. granites), cannot assess concrete compositions with cements other than CEM I at the prescribed alkali level, have 1125
limited comparability with real concrete compositions; and the very long-term effect of mineral additions is still unclear. Concerning existing structures, current knowledge it is clearly deficient, not allowing to assess rigorously the actual condition of an affected structure and to accurately predict the mechanical properties deterioration and the period during which it will effectively perform its function; all essential to determine current safety level and timely and cost-effectively plan eventual mitigation/rehabilitation/reconstruction works. Because of this, LNEC, in collaboration with the international scientific community, is conducting several studies to contribute to the clarification of some of the abovementioned aspects and, consequently, to allow a more effective prevention of ASR in new concrete structures and a more reliable management of ASR affected structures. ACKNOWLEDGEMENTS This work was carried out within the scope of a research project (PTDC/ECM/115486/2009) financed by Fundação para a Ciência e a Tecnologia – FCT (Foundation for Science and Technology, Portugal). The authors wish to acknowledge this financial support. REFERENCES ASTM 2014. ASTM C1260-14 Standard Test Method for Potential Alkali Reactivity of Aggregates (Mortar-Bar Method). West Conshohocken, PA: ASTM International. CEN 2008. EN 12620:2002+A1:2008 Aggregates for concrete. Brussels: European Committee for Standardization (CEN). CEN 2013. EN 206:2013 Concrete. Specification, performance, production and conformity. Brussels: European Committee for Standardization (CEN). LNEC 1993. Especificação LNEC E 415:1993 Inertes para argamassas e betões. Determinação da reatividade potencial com os álcalis. Análise petrográfica, Lisboa, Portugal, LNEC – Laboratório Nacional de Engenharia Civil, I.P. LNEC 2007. Especificação LNEC E 461:2007 Betões. Metodologias para prevenir reacções expansivas internas, Lisboa, Laboratório Nacional de Engenharia Civil, I. P. RILEM 2014a. Alkali-silica reactions in Concrete Structures. RILEM Recommended Specification AAR-7.1 International Specification to minimize damage from alkali reactions in concrete. Part 1 alkali Silica reaction (Unpublished Draft), RILEM-TC219-ACS. RILEM 2014b. Alkali-silica reactions in Concrete Structures. RILEM Recommended Test Method AAR-2 Detection of potential alkali-reactivity – Accelerated mortar-bar test method for aggregates (Unpublished Draft), RILEM-TC219-ACS. RILEM 2014c. Alkali-silica reactions in Concrete Structures. RILEM Recommended Test Method AAR-3.1 Detection of potential alkali-reactivity – 38◦ C test method for aggregate combinations using concrete prisms (Unpublished Draft), RILEM-TC219-ACS. RILEM 2014d. Alkali-silica reactions in Concrete Structures. RILEM Recommended Test Method AAR-4.1 Detection of potential alkali-reactivity – 60◦ C test method for aggregate combinations using concrete prisms (Unpublished Draft), RILEM-TC219-ACS.
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Structural analysis
Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
Non-linear ULS analysis of long-span reinforced concrete arches to EN 1992 J. Nebreda & F. Millanes Ideam, Madrid, Spain
ABSTRACT: Fully non-linear analysis of long-span reinforced concrete arches according to EN 1992-1 and EN 1992-2 requires solid understanding of such concepts as constitutive properties of materials, mean strength, geometric imperfection, buckling modes or safety format. Ideam, as independent checker of two world-record reinforced concrete arches currently under construction in Spain, has had the opportunity of delving deep into the bases of design and safety format underlying EN 1992-1-1 and EN 1992-2. The cases under study are the arch over River Tagus (324 m), and the arch over River Almonte (384 m). Both arches, located almost consecutively and rising nearly 70 m over the waterline, are the major landmarks on the Madrid-Portuguese Border HSRL. An increasingly complex methodology was developed in order to gain deeper knowledge of how a slender concrete arch responds when subject to high compression forces and all possible non-linear effects are taken into account.
1 MOTIVATION 1.1 Problem statement Fully non-linear analysis of reinforced concrete elements under compression, such as bridge piers and arches, is a complex area of study that encompasses solid understanding of a number of concepts ranging from constitutive properties of materials or the difference between eccentricity and geometric imperfection to buckling modes or safety format. Even though most engineers are familiar with each of them, few attempts have been made to bring them together in a systematic way when analyzing long-span arches according to EN 1992-1 and EN 1992-2. This article describes Ideam’s approach and experience. It will be seen that powerful software is not strictly necessary to deal with non-linearity: identifying potentially non-linear areas and using moment-curvature and moment-stiffness diagrams under a given axial force yield reliable results and greater knowledge after only a few hand adjustments. The reason behind putting forward a systematic methodology is identifying the importance of each phenomenon and optimizing analysis efforts by focusing on the critical areas rather than jumping head first into a computationally expensive, blind analysis. 1.2 Cases studied Ideam, as independent checker of two world-record reinforced concrete arches currently under construction in Spain, had the opportunity of delving deep into the bases of design and safety format underlying EN 1992-1 and EN 1992-2. The cases under study are the arches over rivers Tagus and Almonte, with spans of 324 and 384 m, respectively. Both arches, located almost consecutively and rising nearly 70 m over the waterline, are the major landmarks on the Madrid-Portuguese Border HSRL. 1129
In the wake of the studies conducted when analyzing both structures, an increasingly complex methodology was developed in order to gain deeper knowledge of how a slender concrete element responds when subject to high compression forces and all possible non-linear effects (both geometric and mechanic) are taken into account. 2 BASES OF NON-LINEAR ULS DESIGN OF CONCRETE ARCHES TO EN-1992 2.1 Background The latest versions of EN 1992-1-1 and EN 1992-2 deal with the safety format for and analysis methodology of slender elements under compression, with arches being explicitly highlighted (EN 1992-1-1, clause 5.8.2) as sensitive structures. Equilibrium, compatibility and safety checks of structures whose structural behavior can be conditioned by second-order effects must be performed taking into account the deformed shape of the element and with due allowance for uncertainty in all factors. Deformation analysis shall encompass the influence of concrete cracking, time-dependent effects, non-linearity of material laws (constitutive equations) and (perhaps in a lesser degree) foundation flexibility. Non-linear analysis must be performed in all directions in which deformations are expected to occur, through the adequate use of buckling modes and imperfections. These checks are meant to be conducted in the built-at-once structure at the engineering stage. Nevertheless, additional studies shall deal with the impact of the construction sequence, once it is defined. Even though the foundation to non-linear analysis of concrete structures was laid long ago (CEIB-FIP Manual of Buckling and Instability, 1978; CEB-FIP Model Code, 1990), its application was mainly focused on isolated members (such as bridge piers) and columns in building frames. A broader approach to other elements, only put forward in general terms in the latter references, was treated extensively in later normative developments such as EN 1992-1-1 (2004), EN 1992-2 (2005), EN 1993-1-1 (2005) and CEB-FIP Model Code Draft (2010). Furthermore, despite the fact that a number of long-span arches have been erected recently (Maslenica, 1997; KrK, 2003; Infante Don Henrique, 2002; Svinesundbrücke, 2004; Contreras, 2010), a firm consensus as to the implementation of modern codes has not been reached yet. 2.2 Material properties and partial factors Fully non-linear analysis must be carried out using mean material properties and non-linear constitutive (σ − ε) equations. The reason why mean values are used when obtaining section forces and displacements is because global, not local, behavior under ULS loads is being assessed and all materials are meant to perform with the same mean vs. factored strength ratio, 1,10 · γs = 1,27. However, cross-section safety checks are to be conducted as usual, with factored strengths in order to account for local strength drops. 2.2.1 Reinforcement steel Mean yield and mean ultimate strengths shall be taken as fym = 1,10 · fyk = 1,27 · fyd and fum = 1,10 · fuk = 1,27 · fud (EN 1992-2 clause 5.7). A bi-linear σ − ε diagram is accepted instead of a purely elasto-plastic law. Young’s modulus is 200.000 MPa, while ultimate strain is 1%. 2.2.2 Concrete Mean strength and mean Young’s modulus are obtained as per EN 1992-2 5.7:
The non-linear σ − ε diagram defined in EN 1992-1-1 clause 3.1.5 shall be based on the latter properties. 1130
Figure 1.
Effect of eccentricity on an isolated beam (based on EN 1993-1-1, Fig. 5.4).
2.3 Buckling modes Buckling modes must be obtained in order to both identify potentially critical sections and determine the shape of imperfections to be added to actual geometry so that those buckling shapes are adequately triggered. Buckling shapes are associated with a specific factored load configuration, so a set of load scenarios shall be generated depending on which areas are going to be analyzed. For each configuration, a set of buckling shapes may be obtained. Together with each shape, analysis software usually provides the force amplifier for elastic instability, λcr . Moreover, the distance between zero-moment points at either side of the critical section (that with the greatest deflection) is the buckling length, lb . 2.4 Geometric imperfection vs. local eccentricity EN 1992-1-1 states in clause 5.2 that second-order ULS analysis of arches shall take into account unfavorable effects associated with deviations in the structure’s geometry or in the load action line. That eccentricity of axial force shall be applied as proportional to the buckling shape under consideration with a maximum value of:
where lb = buckling length. Curiously enough, EN 1992-2 5.2 does not set a minimum value of α, so the EN 1992-1-1 and EN-1993-1-1 definitions are recommended. Besides, the expression of α in EN 1992-1-1 is conceived for buildings or piers, with H instead of lb /2. Our approach is, in our view, consistent. Figure 1 illustrates Equation 2: If an axial load Nd acts at one end of a kinked or bowed beam with a sag e0 , bending moment at midspan increases by Nd · e0 . However, EN 1992-1-1 and EN 1992-2 do not mention that equation 2 only applies to statically determined elements. When a global model is used, eccentricity shift is not attained by simply applying the aforesaid shape, because part of the bending moment increase is ‘trapped’ by constraints. Therefore, when a structure is studied as a whole, target additional eccentricity must be converted into geometric imperfection, namely, the amount by which actual geometry must be modified in order to induce the eccentricity e0 sought at the critical section. Geometric imperfection in general terms is obtained as per EN 1993-1-1 5.3.2 (11):
where Ncr = Euler’s critical load for the buckling mode considered; EI = bending stiffness of the beam; η and η = deflection and curvature of the buckling shape at the critical section. Since equation 3 only applies to constant-section beams, the target eccentricity must be reached in a more general case by trial and error, that is to say, by adding a random imperfection, assessing the eccentricity shift and inferring the required imperfection. It is noteworthy that eccentricity equals lb /600 whenever buckling length exceeds 18 m, while in steel arches (EN 1993-2 D.3.5) it is generally greater. This does not mean that the bases of design are more stringent in the case of steel. The difference is explained by the fact that steel structures are not only subject to sources of uncertainty such as residual stresses, but also usually analyzed as purely elastic. Moreover, eccentricity can be smaller in concrete because, as it cracks, deformations and second-order section forces increase automatically. 1131
However, it remains a doubt whether such value is reasonable in long-span arches. Imperfections caused by lack of geometric control during the construction stage can be considerably greater than lb /600. A typical case is that of thermal effects prior to the closure of an arch. Therefore, designers’ attention is drawn to the fact that sensitivity studies should be conducted in order to determine whether major safety drops are induced by higher eccentricity values. A complementary analysis with lb /400 is recommended. 3 ANALYSIS METHODOLOGY This section is devoted to setting forth a systematic approach to non-linear ULS analysis of reinforced concrete arches. An increasingly complex process allows one to pinpoint critical sections, narrow down material non-linearity to a few areas and, what is more important, gain greater insight into the relative importance of such phenomena as imperfections, geometric non-linearity, material non-linearity, time-dependent effects or erection sequence. 3.1 Fully elastic linear analysis An elastic linear analysis with mean properties is necessary in order to understand the influence of each load type (dead load, live load either on half arch or on the full span, braking force, wind, etc.), assemble the relevant load scenarios and obtain the buckling shapes and their respective lengths and factors. 3.2 Fully elastic, linear analysis with geometric imperfections Focusing on a particular buckling shape and its associated load configuration, model geometry must be modified so as to induce a bending moment increase of Nd · e0 , that is to say, an eccentricity shift of e0 at the critical section, with e0 as defined in Equation 2. The resulting section forces are defined by EN-1992-1-1 clause 5.8.1 as the first-order structural response. In other words, analysis of compressed members must always include, at least, geometric imperfections. 3.3 Elastic analysis with imperfections and geometric non-linearity First-order response is the base of a geometrically non-linear analysis in which equilibrium and compatibility conditions must be met in the deformed structure. Under the elastic theory, eccentricity and, therefore, bending moment at the critical section are multiplied by Euler’s factor as per EN 1993-1-1 5.2.2 (5)B. The resulting section forces allow one to identify those regions where cracking (and therefore, a stiffness drop) is either bound to or likely to occur. Critical areas are usually located at the springings, around the quarters of span and at the crown, depending on the buckling mode. 3.4 Fully non-linear analysis A fully (material and geometric) non-linear analysis is nothing but the iterative repetition of the previous step after stiffness corrections at the relevant sections have been made. In order to do so, moment-curvature or moment-stiffness under constant axial load are required: given a set (Nd ; Mxd ; Myd ), actual curvature and resulting stiffness are obtained and moments of inertia are modified accordingly in order to start a new run. From our experience, this strategy, although apparently not fine enough to tackle the analysis of a long-span arch, provides enough accuracy when compared to more sophisticated numerical methods. It also allows one to have a better understanding of where sensitive areas are located and how their stiffness changes. It is true that modern software can integrate cross-sectional analysis into global response and adjust stiffness as concrete cracks and rebar yields. However, programs with such features may end up requiring greater model complexity to account for additional, although less explored, non-linear effects such as shear or torsional cracking. A computer’s concept of ‘fully non-linear’ can be too broad and cumbersome. 1132
It must be reminded that sectional stiffness modifications shall be made using mean properties. Besides, the approach taken in this strategy is a one-step, or secant, technique, that is to say, the full ULS load scenario is applied, stiffness (or inertia) is modified, the analysis is re-run, and so on until no more stiffness adjustments are required. A multi-step, or tangent, approach could also be taken, where ULS load is subdivided into smaller steps and the adjusted stiffness after convergence at the end of each step is the base for the next. At the design stage, authors are of the opinion that the secant analysis is enough to fit the geometry of the arch, and the cross-section and rebar layout. As will be dealt with later on, a multi-step analysis is more suitable after the erection sequence has been defined. In any case, after convergence has been reached with 100% of ULS load, cross-sectional safety checks are performed at each section with factored strengths. The load can be increased up to the point of collapse, when a section’s strength (with mean properties) is attained. Technically, whenever a cross-section adjustment is required (greater depth or width, higher rebar ratio, etc.), the whole analysis process should be re-run from scratch due to non-linearity. However, it is up to the engineer’s know-how to determine the extent to which slight modifications require further analysis effort. 3.5 Remarks about load application As stated previously, each load scenario can be input in one step or in many. The latter option is interesting in the follow-up of structural behavior from dead load application to ULS state and, finally, collapse. Therefore, an analysis strategy would be as follows: 1. 2. 3. 4. 5. 6. 7. 8.
Self weight following erection sequence, with time-dependent effects. Superimposed dead loads (wearing course, ballast, barriers, etc). Remaining time-dependent effects until long-term situation (optional). ψ2 times the traffic live load (quasi-permanent situation). (ψ1 − ψ2 ) times the traffic live load (frequent situation). (1 − ψ1 ) times the critical traffic live load (characteristic situation). (γG − 1) times the dead loads and (γQ − 1) times the critical traffic live load. Increase all loads until collapse (when section forces reach sectional mean strength).
Note that step 3 can be neglected when performing a short-term analysis, while steps 4 through 8 can be subdivided into smaller steps for stiffness adjustments, if necessary. 3.6 Remarks about erection sequence Large concrete arches are nowadays erected by the cantilever method (Figures 2 and 3). Therefore, the actual response under self weight is likely to differ from that of the built-at-once assumption.
Figure 2.
Erection of the Arch over River Tagus.
1133
Figure 3.
Erection of the Arch over River Almonte.
Figure 4.
Diagram with load configuration and reference sections, Tagus and Almonte arches.
This situation is all the more relevant the greater the span is, because self weight can account for much of the total load. Special care should be placed when shifting from the design to the construction engineering stage. 4 CASE STUDIES 4.1 General overview Over the last years several remarkable arches have been constructed in Spain: over Contreras reservoir (261 m) on the Madrid-Valencia HSRL, over River Tagus (324 m) and over River Almonte (384 m), the latter two on the Madrid-Portuguese Border HSRL. All of them have successively beaten the world record in railway arch bridges and become major landmarks due to their unparalleled setting and size. The Tagus and Almonte arches (Figure 4) account for an amazing 50% span increase in only one leap, pushing the limits of structural engineering to a new level and making one wonder whether current safety formats were conceived to handle such spans and such slender arches (featuring depth:span ratios of 1/90 and 1/80, respectively). Ideam, in the role of technical advisor to ADIF (the branch of the Public Works Ministry in charge of the railway network), put forward the previously stated bases of design and work methodology and conducted the non-linear ULS analysis as part of the independent check tasks. The following sections describe the main features of and results for both arches. 4.2 Load configuration For clarity’s sake, only results for the first in-plane buckling mode shall be provided in this paper, as it is the critical case. The ULS load scenario comprises the following: – γG times self weight (built-at-once) and maximum superimposed dead loads (30% excess ballast weight). 1134
Table 1. Section forces (ULS) for in-plane buckling analysis with e0 = lb /600 and e0 = lb /400, Tagus arch. 1.00 · qd ; lb /600
1.00 · qd ; lb /400
1.20 · qd ; lb /400
Analysis
N (kN)
M (mkN)
N (kN)
M (mkN)
N (kN)
M (mkN)
P11* S12* P12* S13* P13* S14* P14* S15* P15* S16* P16* S17* P17*
−279235 −280140 −259814 −261746 −249201 −250426 −249376 −250233 −260187 −261381 −260315 −292870 −293397
546280 192430 72326 −397242 −61284 −243566 223494 98103 510424 270216 393015 −372176 −660525
−279377 −280277 −259896 −261911 −249050 −250364 −249482 −250300 −260056 −261197 −260180 −292762 −293321
583407 219402 67786 −427085 −79340 −262817 226733 123433 542815 309442 410833 −392229 −694974
−338576 −339615 −314449 −316837 −300402 −302410 −301621 −302778 −314113 −315413 −313952 −352440 −352927
809828 338890 76513 −586308 −107049 −326652 284154 197823 718184 445134 544973 −465582 −803503
*See Figure 4.
– γQ times 2 UIC71 trains: distributed load acts on half the span and axle loads are located over the maximum deflection section. – γQ times the maximum braking force acting on the deck. – Bearings friction force (taken as 5% of the reaction under permanent loading on the longer side of the deck and 1% on the shorter side, in opposite directions). 4.3 Viaduct over river tagus 4.3.1 Main features This viaduct, designed by CFC S.L., is 1488 m long, with an arch of 324 m between piers 11 and 17 that rises almost 70 m above the springline. The main span is divided into 6 segments 54 m long each (Figure 4). The deck is pin-fixed to the crown at pier 14. The depth ranges from 3,50 m at the crown to 4,00 m at the springing, while the width transitions from 6,00 m to 12,00 m, respectively. Wall thickness is 0,60 m in approximately the central 4 spans, increasing up to 1,20 m at the springing. The arch is made from C70 concrete. 4.3.2 Target buckling mode, local eccentricity and geometric imperfection The first in-plane buckling mode of the arch is asymmetrical with an amplification factor of 4,26 and a buckling length of about 111 m measured horizontally from the crown. The critical section in terms of displacement is S16, between piers 15 and 16, but closer to the former. The resulting eccentricity for first-order analysis is 185 mm, while the required geometric imperfection is 370 mm. Euler’s amplification factor for geometric non-linearity is 1/(1 − 1/4,26) = 1,307. 4.3.3 Summary of results Table 1 summarizes the final results for an in-plane analysis with eccentricity lb /600 and lb /400. In the latter case, results for 120% of ULS loading, when section failure takes place (S13, boldface), are listed too. Vertical displacement at section S16 is 855, 981 and 1490 mm, respectively. Bending stiffness loss is about 32% at P11, 18% at S12, 23% at P15 and 39% at P17 for lb /600 and about 35% at P11, 27% at S12, 26% at P15 and 49% at P17 for lb /400. Results reveal that this arch caters for important force increases caused by both imperfections and total load. 1135
Figure 5.
Section forces in ULS with parallel study, Almonte arch.
4.4 Viaduct over river almonte 4.4.1 Main features The viaduct is 996 m long, while the arch spans 384 m between piers 6 and 15 with a rise of almost 70 m. The main span is divided into 9 segments, the outermost being 45 m long and the rest, 42 m (Figure 4). Two legs, designed to provide global wind stability, emerge from the springings up to piers 8 and 13, where they converge in a single arch. The deck is connected to the crown all along the central segment through lateral walls. The arch is 4,80 m deep at the crown and 6,90 m at the springing, while the width ranges from 6,00 m to 19,00 m, respectively. Remarkable chamfers (2,20 m × 0,75 m), aimed at providing optimized wind stability, create a rather unconventional cross-section. Wall thickness is, on average, 1,40 m between piers 8 and 13, and 0,90 m at the legs. C80 concrete was used in the arch. 4.4.2 Target buckling mode, local eccentricity and geometric imperfection The first in-plane buckling mode of the arch is asymmetrical with an amplification factor of 3,66 and a buckling length of about 120 m measured horizontally from the crown. The critical section in terms of displacement is P12. The resulting eccentricity for first-order analysis is 200 mm, while the required geometric imperfection is 340 mm. Euler’s amplification factor for geometric non-linearity is 1/(1 − 1/3,66) = 1,376. 4.4.3 Summary of results Figure 5 shows the section forces along the arch for the ULS state with lb /600 eccentricity and incremental loading as described in 3.5. The same analysis was performed by Ideam (independent checker), FCC (contractor) and Arenas y Asociados (designer) using different software and stiffness adjustment techniques (manual vs. automatic). The results show that section forces differ in less than 4% at relevant sections. Factored section strength is about 6% greater than ULS forces. Vertical displacement at section P12 is in the range of 1150 mm. 5 CONCLUSIONS After examining both structures, it can be seen that the Tagus arch, more slender than the Almonte arch, allows for remarkable load and eccentricity increases over the values stated in EN-1992-1-1 and EN-1992-2. This is due to its span (about 20% shorter), its greater rise:span ratio (1/4,6 vs. 1/5,5) and the use of a more efficient cross-section in structural terms. Almonte arch, conditioned in a higher degree by wind effects, requires a cross-section with thicker walls and a greater reinforcement ratio in critical areas in order to make up for the loss of bearing section induced by its chamfers. Regarding safety format, it seems that using an imperfection value of lb /400 instead of lb /600 does not lead to overly severe results in rather conventional designs even in the case of slender arches. Another relevant conclusion to be drawn is that the amazing match in results obtained in the case of Almonte arch confirms that a fully non-linear analysis can be conducted cost-efficiently and with manageable tools, and yield the necessary information about the structural response. 1136
Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
Stressing sequence of steel cable-stayed bridges built by cantilevering A. Recupero & M. Calvo Università di Messina, Italy
M.F. Granata & M. Arici Università di Palermo, Italy
ABSTRACT: The construction of cable-stayed bridges by cantilevering implies several changes of geometry, stress and strain patterns during the assemblage of segments. The main target to be satisfied in the construction process is the achievement of the required final geometry and of a convenient state of stress for self-weight and sustained loads. The sequence of stay stressing and the values of prestressing forces at each stage of segment assembling have the main role for reaching the desired result of design, due to the large redundancy of cable-stayed structures. Among the different procedures proposed in the literature for initial cable force determination, the Partial Elastic Scheme (PES) Method has been already applied to cable-stayed bridges of small-medium spans. In this study the same methodology is extended to cable-stayed bridges with larger spans and steel decks, by considering the effects of geometric nonlinearity attributable to the cable sag. In this paper PES method is modified to take into account the axial stiffness of stays and the related nonlinear behaviour of the bridge. Nonlinear analyses are performed with forward staged-construction finite element procedures and different strategies for establishing the initial value of stay stressing at each stage are proposed and compared. A case-study is investigated applying different approaches of stressing sequence. Results show the applicability of PES method to large cable-stayed bridges, when nonlinear staged construction analyses are performed and initial prestressing axial forces are found by linear partial elastic schemes. 1 INTRODUCTION Cable-stayed bridges can have prestressed concrete girders or boxes, steel-concrete composite cross sections and fully steel sections. It depends on the span length covered, which can be from 100 meters to 1000 meters. While prestressed concrete decks can be convenient for the smallest spans, composite sections are widely used for medium and large spans (Fernandez, 2003; Manterola, 2006). The deck typology has direct consequences on the staged construction and on the stress and strain patterns in the final configuration. Moreover, during construction the main problem is the determination of initial cable forces together with the more suitable stay stressing sequence. Steel-concrete composite decks are built in three steps: first, the steel elements are assembled by cantilevering to the previous segments; second the concrete slab is cast over them and third the stay is attached and tensioned. This operation implies that dead load is applied in two subsequent times at each phase of construction and that evaluation of initial cable forces and the stressing sequences of stays are different for concrete and composite decks. The determination of initial cable forces is always related to two different aims: to obtain the desired final geometric profile of deck and pylon and to obtain a convenient bending diagram for the dead load configuration, associated to the required deck shape. This procedure is performed in order to minimize stresses in the steel elements and the concrete slab, avoiding over dimensioned structural elements and concrete cracking. The first is a hard target to be achieved; some cable-stayed bridges present a non-satisfactory shape of the deck (Figure 1) just in the early phases of service life, due to the stressing sequence adopted during construction. 1137
Figure 1.
Non-satisfactory shape of a cable-stayed bridge deck.
The latter aim is deeply dependent on the construction sequence, so a wrong methodology in terms of deck geometry can have also bad consequences on the distribution of internal forces at the construction end. Another important aspect to be considered, in steel-concrete cable-stayed bridges, is the increased importance of cable sag nonlinear effect with respect to concrete bridges. It depends on the span range covered by these bridges and on the stay lengths, for which the geometric non linearity has to be considered (Adeli & Zhang, 1995). A steel bridge emphasizes these two aspects: the optimization of bending moment diagram in construction stages and in the final dead load configuration and the effects of geometric nonlinearity (Freire et al., 2006). In this study, a methodology based on simple partial elastic schemes of construction stages is applied for the evaluation of initial cable forces. This methodology has been already applied by the authors to concrete cable-stayed bridges (Arici et al., 2011) and to arch bridges built by suspended cantilevers (Granata et al., 2012) or by lattice cantilevers (Granata et al., 2013). Here it has been applied to steel decks and to large span bridges with the addition of geometric nonlinear behavior (sag effect). Linear and nonlinear methodologies for the determination of initial cable forces and for the staged construction analysis are compared. The analysis has the target of obtaining a predetermined geometric profile of deck and pylon at the end of construction with acceptable precision and correspondently a convenient distribution of bending moments. The results of the analyses performed on a case-study bridge are reported in terms of displacements and internal forces and commented together with the consequences of nonlinear staged construction analyses. 2 PARTIAL ELASTIC SCHEME METHOD APPLIED TO LARGE SPAN BRIDGES 2.1 Staged construction In the frame of cantilever construction of cable-stayed bridges, the most frequent methodology for establishing the initial cable forces is the so-called backward analysis. It is a deconstruction analysis of the bridge starting from the desired final stage and conducted by dismantling the bridge stage by stage (Chen & Duan, 1999; Gimsing, 1997). In this methodology, initial cable forces are the forces found in the stay at each deconstruction stage. A backward analysis has to be followed always by a forward analysis, which considers the actual construction sequence for obtaining the actual state of stress and deformation. Due to the high grade of redundancy of cable-stayed bridges, the final result found by applying the actual sequence of pre-stress is often different from the expected one. Consequently, it is difficult, especially for cable-stayed bridges that are not completely selfanchored or that are not symmetric (Lee et al., 2008), to obtain the required geometric configuration with a single stay stressing but it is necessary to adjust the value of axial forces for several stays. This is a technological complication and the tendency of designers is that of reducing adjustment operations to the minimum. 1138
Figure 2.
Partial elastic scheme of symmetric cantilevering and displacements recovered by stay stressing.
The most known procedures to obtain the initial cable force are the following: 1) Deflections of stay anchorage points in the deck and the horizontal displacements of the pylon points are nullified by the so-called zero-displacement method (Wang et al., 1993); 2) A pre-determined distribution of bending moments in the deck and the pylon is searched by applying the unit load method (Janjic et al., 2003); 3) An opportune energetic function is minimized (Xiao et al., 2001). In this paper, the first approach has been followed in order to obtain the initial cable forces and then forward analyses are performed to follow the actual construction stages. 2.2 Partial elastic scheme method Initial cable forces are determined by partial elastic schemes of the structure, one for each construction phase (one for each segment assembled by cantilevering and consequently for each stay attached and tensioned). Steel elements of the cantilever segment are assembled and the new stay is attached without modifying its length. The value of initial cable force Fi is found after slab casting on the partial elastic scheme, by evaluating the cable force needed to recover the sum of displacement δI due to steel elements and δI due to concrete cast. At the same time the previous stay, already stressed in the previous stage, is adjusted with force Fi−1 in order to compensate the displacements δI−1 and δI−1 of the related anchorage point, attributable to the actual construction stage (Figure 2). In this way, a compensation of displacements due to construction loads is achieved for each stage at the end of the cantilever, and the final configuration approximates that of a continuous beam on rigid restraints. The disadvantage is that each stay is stressed two times because it implies the stressing equipment is shifted from one stay to another at each adjustment. The evaluation of initial cable forces is done on the elastic scheme of the related stage, by considering contemporarily the displacements of the last segment and those of the previous one. Afterwards a final regulation of all stays is convenient in the last stage, in order to take into account the displacements due to construction errors, superimposed dead loads (pavement, guardrails, footsteps, etc …) and creep effects in construction stages. Moreover, the effects of shrinkage and creep, till construction end, can be partially compensated by a final adjustment of all stays. In fact the axial shortening due to concrete shrinkage, especially in the tower, implies an increased vertical displacement of the deck with respect to the result of theoretical analyses. In order to compensate the pylon axial shortening, it is possible to build concrete elements of an increased length (Schlaich, 2001). 2.3 Nonlinear behaviour For large span bridges, the stay length emphasizes the effects of geometric nonlinearity due to the cable sag. In order to consider this aspect in this study two different procedures are implemented: – linear partial elastic schemes for determining initial cable forces and non-linear staged construction with forward analysis performed by a finite element software; 1139
Figure 3. View of the case-study bridge model.
– nonlinear partial elastic schemes in which the modified elastic modulus of cables according to Dischinger and Ernst hypotheses is considered for determining initial cable forces and a nonlinear staged construction analysis performed by a finite element software. According to the classical theory (Gimsing, 1997), the value of Dischinger reduced elastic modulus can be defined through the following relation:
in which Ep is the steel elastic modulus, γ the weight per unit volume of stay, l the horizontal projection of stay length, σ the average axial stress of the stay (tangent elastic modulus). Naturally, this approach implies an iterative procedure implemented for each partial elastic scheme in order to determine the reduced elastic modulus Ered of stays and the effective value of stay pre-stress. Afterwards, the values found from each partial scheme are introduced in the stages performed in the forward analysis. 3 CASE-STUDY A case study of a bridge with two pylons and three spans is investigated. The main span length is 600 m while side spans are 182 m long. Figure 3 shows the bridge model. Pylons are 100 m high above the deck and the cable system is composed of 58 stays in the main span, attached to the deck with mutual distance of 10 m, with a mixed harp-fan scheme. The cable system is attached in the center of the deck box (one stay plane), the pylons being single concrete columns; side spans present 6 stays anchored to the ground. The bridge model has been implemented in Midas software, performing the partial elastic schemes of each construction stage, with the hypothesis of cantilever construction, for the determination of initial cable forces, as illustrated above. Afterwards an analysis was performed by advancing with construction stages and evaluating the deformations and stresses in the bridge elements. Current stays have diameter 13 cm, while earth-anchored backstays have equivalent steel diameter 18 cm, the steel strength being fptk = 1860 MPa and the elastic modulus Ep = 195000 MPa. Deck is a two-cell steel box with a concrete slab, thickness 30 cm. Four cases are investigated and discussed: – The linear PES method is applied and the linear forward staged construction analysis is performed on the FE model; – The linear PES method is applied and the geometric nonlinear forward staged construction analysis is performed on the FE model; – The linear PES method is applied with the Dischinger elastic modulus and the linear forward staged construction analysis is performed on the FE model; – The linear PES method is applied with the Dischinger elastic modulus and the geometric nonlinear forward staged construction analysis is performed on the FE model. 1140
Figure 4. Results of the analyses performed on the half-bridge FE model. a) linear PES and linear forward analysis; b) linear PES and nonlinear forward analysis; c) Dischinger PES and linear forward analysis; d) Dischinger PES and nonlinear forward analysis.
In all analyses, the PES method is applied with a double stressing procedure (first stressing of the last stay attached and stress adjustment of the previous one). Results are summarized in figure 4 in terms of final deformed configuration after staged analysis and bending moment diagram on the deck, showing the most significant numerical results (upward positive displacement and positive bending moments with bottom fiber in tension) and taking into account the superimposed dead loads due to finishing works (road pavement , guardrails, etc …). It is worth noting that the final result strictly depends on the type of analysis, performed in construction stages, in fact, the deformed shape changes significantly between linear and nonlinear analyses. Although cases (a) and (b) have the same values of initial cable forces found by the linear PES method, the final result is strongly different with higher values of bending moments in the deck for the case of linear analysis, corresponding to a more deflected final configuration in the main span. In the same way, cases (c) and (d) have the same values of initial cable forces found by the PES method applied with Dischinger modulus, but linear and nonlinear analyses lead to different results, according to what found in cases (a) and (b). The worst case, in terms of deformed configuration, is that of PES with Dischinger and nonlinear forward analysis (d), the bending moment diagram being more convenient when nonlinear analysis is performed. By evaluating the initial cable forces found through PES method with and without Dischinger modulus a slight difference has been observed, the maximum value found with Dischinger modulus being slightly higher (about 4%) than that found by the linear PES method. This leads to the consideration that PES method applied with Dischinger modulus is not particularly convenient. The best result seems to be achieved with linear PES method and nonlinear staged construction 1141
analysis, with a higher stiffness of the main span, even though in this case the pylon appears more stressed. The strong variation in the results found by linear and nonlinear staged construction analysis of the FE model depends on the procedure applied through the software used. This fact needs deep investigations on the kind of nonlinear analysis performed by FE procedures, because the actual result can be strongly different from the predicted one, depending on the analysis chosen by the designer, with consequences on the final deformed configuration and on the bending moment distribution. Naturally, the effects of moving loads after construction modify the configuration and the stress state in service life. 4 CONCLUSIONS A study on the stressing sequence of steel cable-stayed bridges has been presented. The Partial Elastic Scheme (PES) method was applied to find initial cable forces in large span bridges, by performing forward linear and nonlinear staged construction analyses on Finite Element models of bridges. Results, found on a case-study of a cable-stayed bridge with main span 600 m long, show a high dependency of the final configuration on the analysis performed. The best results seem to be achieved through linear PES approach and nonlinear staged construction analysis. PES method can be efficiently applied with acceptable final values of displacements in the deformed configuration after construction, but linear and nonlinear analyses show different results especially in terms of bending moment distribution. This leads to the consideration that nonlinear analyses performed with FE procedures have to be evaluated with particular attention and always compared to linear ones, the actual result on the bridge being unknown; predictions of the results achieved with stressing sequences of this type have to be managed by designers paying great attention, needing of further investigations. REFERENCES Adeli H. & Zhang J. (1995). “Fully nonlinear analysis of composite girder cable-stayed bridges”. Computers and Structures, 54(2), 267–277. Arici M., Granata M.F., Recupero A. (2011). “The influence of time-dependent phenomena in segmental construction of concrete cable-stayed bridges”. Bridge Structures, 7(4). 125–137. Chen Wai-Fah & Duan L. (1999). Bridge Engineering Handbook, CRC Press, Washington. Fernandez Troyano L. (2003). Bridge engineering: a global perspective. Thomas Telford. Freire A.M.S., Negrão J.H.O., Lopes A.V. (2006). “Geometrical nonlinearities on the static analysis of highly flexible steel cable-stayed bridges”. Computers and Structures, 84(31–32), 2128–2140. Gimsing N.J. (1997). Cable supported bridges. Concepts and design. John Wiley. Granata M.F., Margiotta P., Recupero A., Arici M. (2012). “Partial elastic scheme method in cantilever construction of concrete arch bridges”. Journal of Bridge Engineering ASCE, 18(7), 663–672. Granata M.F., Margiotta P., Recupero A., Arici M. (2013). “Concrete Arch Bridges built by Lattice Cantilevers”, Structural Engineering and Mechanics, 45(5), 703–722, ISSN: 1225–4568. Janjic D., Pircher M., Pircher H. (2003). “Optimization of Cable Tensioning in Cable-Stayed Bridges”. Journal of Bridge Engineering ASCE, 8(3), 131–137. Lee T.Y., Kim Y.H., Kang S.W. (2008). “Optimization of tensioning strategy for asymmetric cable-stayed bridge and its effect on construction process”. Struct. Multidic. Optim. 35. 623–629. Manterola Armisén J. (2006). Puentes: apuntes para su diseño, cálculo y contrucción. Colegio de Ingenieros de Caminos, Canales y Puertos. Madrid. Schlaich M. B. (2001). “Erection of cable-stayed bridges having composite decks with precast concrete slabs”. Journal of Bridge Engineering ASCE, 6(5), 333–339. Wang P.H., Tseng T.C., Yang C.G. (1993). “Initial shape of cable-stayed bridges”. Computers and Structures, 46(6), 1095–1106. Xiao R., Jia L., Song X., Xiang H. (2001). “Influence matrix method of cable tension optimization for long span cable-stayed bridges”. IABSE Conference on cable-supported bridges. Seoul.
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Dynamic analysis for fatigue assessment of reinforced concrete slabs in railway viaducts J. Malveiro, C. Sousa & R. Calçada Faculty of Engineering, University of Porto, Porto, Portugal
D. Ribeiro Polytechnic of Porto, School of Engineering, Porto, Portugal
ABSTRACT: Detailed calculations of applied stresses and internal forces have to be made whenever a realistic fatigue assessment is necessary. The structural response of reinforced concrete slabs in railway viaducts is governed by local effects, and therefore detailed analysis procedures are required for a proper quantification of the stress records due to train passages. In this context, this paper presents a numerical procedure for detailed finite-element analysis of local effects caused by train passages. The interaction between bridge and train is considered by using an iterative procedure that solves the bridge subsystem by the mode-superposition method and the train subsystem by Newmark’s direct integration method. A static correction procedure is employed to take into account the static contribution of higher-frequency vibration modes (not explicitly considered in the mode-superposition analysis) for a correct quantification of stresses in the reinforced concrete slab. Illustrative results are shown for a real case study.
1 INTRODUCTION Railway bridges are structures particularly sensitive to dynamic effects induced by trains. The dynamic amplification of internal forces and stresses can reach significant values owing to the resonance phenomena originated by the periodic loading associated with the passage of regularly spaced train axle groups. The increase of the train’s speeds, the higher axle loads and the existence of track irregularities and wheels defects can amplify these dynamic effects and give rise to excessive deck vibrations, leading to fatigue damage. Reinforced concrete (RC) deck slabs, in railway bridges and viaducts, can be particularly sensitive to fatigue, depending on the structure and track geometry, and also on traffic and temperature loading, among other factors. A realistic calculation of applied stresses has to be made for a proper fatigue assessment (Sousa et al. 2013). In order to properly evaluate the stresses resulting from train passages, detailed dynamic analyses are required. Besides that, it has been shown that improved numerical results are obtained if calibrated finite-element (FE) models are used (Ribeiro et al. 2012, Malveiro et al. 2014). The authors showed the advantages of using calibrated models, by comparing numerical and experimental responses. As regards the train loading, the numerical methodology that allows obtaining more realistic results is the one that considers the interaction between the bridge and the train, even though it is more complex than the simple approach based on moving loads. Generally, the methodology used in the present work is an iterative procedure according to which, in each time increment, two subsystems are solved independently: the bridge subsystem is solved by the mode-superposition method and the train subsystem is solved using the Newmark’s direct integration method (Ribeiro 2012). The basic mode-superposition method involves a set of uncoupled equations, each one corresponding to a vibration mode of the structure. However, the dynamic behaviour of deck’s slabs are 1143
strongly influenced by the contribution of local vibration modes, which are normally associated to smaller wavelengths and therefore to higher frequencies. Thus, it is common to consider a large number of vibration modes, in the dynamic analysis, to ensure that, at least, the static response of the structure can be ensured. If the structure response is dominated by local vibration modes, there is no guarantee that the correct structure response is obtained, even if a great number of vibration modes is considered in the analysis of the bridge subsystem (when the basic mode-superposition method is used). Therefore, an improved formulation of the mode-superposition method is applied in the present work, in the analysis of the bridge subsystem. It combines the mode-superposition method with a static correction procedure, using the influence surface to obtain the static contribution of vibration modes of higher frequencies, which are not explicitly considered in the mode-superposition method. This methodology guarantees that the static component of the structural response is properly captured, and is correctly added to the dynamic component of the response associated to all the vibration modes considered in the analysis. The relevance of considering a static correction procedure in the calculation of local effects in railway bridge decks has already been pointed out by Pimentel et al. (2008). In the present work, this methodology is applied in the analysis of a real case study: the Alverca viaduct, located in the Northern line of the Portuguese railways. Particular attention is paid to the comparison between calculated time-records of bending moments in the RC slab, considering different analysis methodologies, thus showing the potentialities of the methodologies presented in this work. 2 DYNAMIC ANALYSIS The train-bridge dynamic interaction is a complex problem that involves the use of advanced numerical methods and often demanding from the computational point of view. In this study, the analysis of the train-bridge interaction is performed through an uncoupled methodology that considers the bridge and train subsystems modelled as two independent structures (Calçada 1995). 2.1 Train-bridge interaction methodology In the uncoupled methodologies, the analysis of the train-bridge interaction involves the calculation of the bridge subsystem subjected to the action of the forces transmitted by vehicles, alternately with the calculation of the train subsystem subjected to the displacements of the bridge. The contact points between the two subsystems are the interfaces between the train wheels and the rails. The dynamic train-bridge interaction is considered only in the vertical direction and the loss of contact between the wheel and the rail is not permitted. At each time increment, the methodology uses an iterative process to make compatible the two subsystems, in terms of the dynamic interaction forces and the displacements at the contact points, ensured by the application of a convergence criterion. Table 1 shows, in a simplified manner, the implementation of an iterative uncoupled methodology. A detailed explanation about the operations involved in an iterative methodology are described in Ribeiro (2012). In this table, Fbk corresponds to the wheel load applied to the bridge in each iteration k, Fv,sta is the static wheel load and ubc is the displacement of the bridge under each contact point c. The resolution of the dynamic equation of the bridge subsystem is performed by the modesuperposition method. This method involves the calculation of a set of equations of dynamic equilibrium, each one corresponding to a vibration mode of the structure. The equation relating to the vibration mode j, assuming vibration modes normalized in relation to the mass matrix, is given by Equation 1:
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Table 1. Iterative method to solve the dynamic problem considering the train-bridge interaction.
where Y¨ , Y˙ and Y are the modal accelerations, velocities and displacements, respectively, ζj and ωj are the modal damping and the angular frequency and Fbj are the modal forces, obtained by the following equation:
where φ represents the modal ordinates vector of the contact points for the vibration mode j. Each dynamic equation is solved using the Newmark’s integration method. The displacements in each contact point, ubc , are calculated based on the sum of the different modal displacements with track irregularities, r, as given by Equation 3.
The dynamic analyses considering the train-bridge interaction were performed using a computer application called TBI (‘Train-Bridge Interaction’) developed in MATLAB (2011). The program consists of five routines involving the data introduction (Routine 1), the data import from ANSYS software (Routine 2), the modal analysis of the bridge (Routine 3), the dynamic resolution of the train-bridge interaction through the iterative method (Routine 4) and the post-processing of the results (Routine 5). A detailed explanation of each routine can be found in Ribeiro (2012). In the present work, a subroutine to perform the static correction of the bridge response was developed and included in TBI according to the description made in the following section. 2.2 Static correction procedure The static correction procedure is reported in the literature (Clough and Penzien 1993, Chopra 1995) as a technique to consider the static contribution of vibration modes with higher frequency which are often not considered in the dynamic analysis using the basic mode-superposition method. Clough and Penzien (1993) and Chopra (1995) deal with load regimens whose magnitude vary in time, having a fixed space distribution. The case of loads for which both the magnitude and the space distribution vary in time can be dealt with as described in the following paragraphs. 1145
In the basic mode-superposition method, the time-record Gi (t) for any parameter (displacement, stress, bending moment, etc.) in the node i and in the instant of time t is given by:
where N is the total number of vibration modes considered in the analysis, Gij is the parameter value in the node i and in the j-th vibration mode and Yj (t) the time-record of the modal coordinate for the j-th vibration mode. As mentioned before, this method does not guarantee that the obtained results include the total static response to the applied forces, when the structural behaviour is governed by local effects. The static correction procedure consists of considering, in the improved mode-superposition method, a limited number of vibration modes, N (those which are expected to have a significant contribution in terms of inertial forces, i.e., in terms of the dynamic amplification of internal forces). The static component of the contribution of each of those vibration modes can be calculated, for each instant of time, by dividing the modal forces Fj (t) by the modal stiffness Kj , as shown by Clough & Penzien (1993):
where φj if the vector containing the coordinates of the j-th vibration mode for the nodes of the rail path, fk (t) is the vector of wheel loads applied to the nodes of the rail path in the instant of time t and K is the structure’s stiffness matrix. On the other hand, the total static response of the structure (i.e., the total static component of the parameter G), for each node i, Gi,s (t), can be calculated as a function of the influence surface Aik (value of the parameter G in node i when the structure is subjected to a unit force in node k):
where m is the total number of nodes in the load path. Thus, an improved calculation of the time-record for Gi (t) for any parameter can be determining by adding the total static component to the dynamic component of the response associated to the N modes considering in the improved mode-superposition analysis:
3 ALVERCA RAILWAY VIADUCT 3.1 Brief structure description The Alverca viaduct is a flyover structure in the Northern line of the Portuguese railways. It carries a single railway track and consists of 47 simply supported spans, with lengths of 16.5 m, 17.5 m and 21 m. Each span is composed by a prefabricated and prestressed U shaped beam and an upper castin-place slab, forming a single-cell box-girder deck. The deck is supported by elastomeric reinforced bearings, which are fixed in one extremity and longitudinally guided in the other. Additional details can be found in the paper by Malveiro et al. (2014). 1146
Figure 1. Three-dimensional FE model: overview and track details.
3.2 Numerical modelling In a previously published research (Malveiro et al. 2014), a three-dimensional FE model, for the first three spans near the North abutment, was developed with resource to the software ANSYS (2007), and calibrated based on the experimental results of ambient vibration tests. The calibration was carried out through an iterative procedure, using a genetic algorithm, and considering the following modal parameters: frequencies and mode shapes, for both global and local vibration modes. The calibration procedure allowed obtaining optimal values for geometrical and material properties, which serve as a basis for the FE analysis envisaged in the present paper. The description and results of such calibration can be found in Malveiro et al. (2014). A different FE model was developed for the analyses envisaged in the present work. Bearing in mind that the main purpose is the detailed calculation of time-records of internal forces in the upper deck slab, a finer discretization was adopted in one of the structure’s span (the second span near the North abutment, with a length of 21 m). A refined mesh was adopted for the upper RC slab, the ballast retaining walls, the ballast layer and the sleepers. An extension of the track, with 10 m for each side of the span under analysis, was considered in the FE model, in order to avoid an abrupt entrance of the train in the span. The present study focuses, in particular, in the central region of the upper deck slab, and for that reason, the fact that the neighbouring spans are not considered in the model, is justified. Figure 1 illustrates the adopted mesh. The precast girder, the upper slab and the ballast retaining walls were modelled by shell FEs. The sleepers, the rail pads and the ballast layer were modelled by solid FEs. The compatibility of displacements and rotations between the nodes of the precast beam and the nodes of the upper slab as well as the compatibility of displacements between the nodes of the upper slab of the deck and the lower nodes of the ballast layer were accomplished by rigid finite elements. Each support was regarded as a single point and modelled by a spring element. Beam FEs were adopted to simulate the rails, and were positioned at their centre of gravity. Non-structural elements such as safeguards and kerbs were considered as additional masses. A very good agreement, in terms of natural frequencies and mode shapes, was achieved between the initially calibrated numerical model and the new, more refined, model for one of the structure spans (represented in Figure 1). Different analysis methodologies (identified in Table 2) were implemented, in order to verify the relevance of the static correction procedure, both in analyses where the railway traffic is represented by moving forces and in analyses considering the train-bridge interaction (however, in the present paper, track irregularities are not considered). The Newmark’s direct integration method was considered as the reference methodology for calculation of the correct static structure response in analyses where the inertial forces (i.e., the dynamic enhancement of internal forces) are not considered. As regards the dynamic analyses including dynamic effects, two hypotheses were considered for definition of modal damping ratios in the bridge subsystem: (i) Raileigh damping, for analyses with Newmark’s direct integration, and for the remaining analyses which are to be compared with the former; (ii) a constant modal damping ratio of 1% (consistent with the Eurocode (CEN 2003) recommendations) when different analyses based on the mode-superposition method are to be compared to each other. 1147
Table 2. Different FE analysis methodologies whose results are discussed in the paper.
Analysis
Analysis of the bridge subsystem*
A B C D E F G H I J
M.S. M.S. M.S. M.S. D.I. D.I. M.S. M.S. M.S. M.S.
Train-Bridge interaction
Static correction
Damping coefficients
– –
–
1% 1% 1% 1% ** – ** ** – –
– – – – – –
–
– – –
–
Inertial forces
–
– –
*M.S. stands for mode-superposition and D.I. stands for Newmark’s direct integration method. **Rayleigh damping matrix, setting the damping equal to 1% for vibration modes 1G and 4L, identified in the paper by Malveiro et al. (2014).
Figure 2.
Comparison of bending moments from analyses F, I and J (line F is coincident with line J).
In all the analyses, the railway traffic consists of an Alfa Pendular train travelling at a speed of 100 km/h. All the analyses of the bridge subsystem based on the mode-superposition method are carried out considering all the structure’s vibration modes with a frequency lower than 100 Hz (77 vibration modes). 4 RESULTS AND DISCUSSION In this section, the results of different analysis methodologies are shown and discussed. The results correspond to the bending moment in the upper deck slab, in the transverse direction, in a position which corresponds to the mid-span, both in the longitudinal and in the transverse directions. Figure 2 shows the comparison between analyses F, I and J, in order to verify if the static response is well characterized in analysis based on the basic mode-superposition method. All of these analyses were performed without considering dynamic effects and the static correction procedure was considered in analysis J. It can be seen that the basic mode-superposition method is not able to correctly reproduce the static response. A perfect agreement with the reference results of the Newmark’s direct integration method (analysis F) was achieved when the static correction procedure was considered (analysis J). Figure 3 shows the results of similar analyses with the difference that, now, dynamic effects are considered (analyses E, G and H). By comparing the results of analyses E and G, one can see that the mode-superposition method is not able to properly characterize the RC slab response, obtained through the direct integration method. If the static correction is considered (analysis H), the results 1148
Figure 3.
Comparison of bending moments from analyses E, G and H.
Figure 4.
Comparison of bending moments from analyses A and B.
of the improved mode-superposition method tend to approximate to those obtained through the direct integration, as shown in the detailed view in Figure 3. Figure 4 compares the results of analyses A and B (not considering train-bridge interaction) and Figure 5 compares analyses C and D (taking into account such interaction). The static correction procedure was implemented in analyses B and D. By observing the results in Figures 4 and 5, and taking into account the conclusions drawn from the analysis of Figures 2 and 3, one can realize that the maximum bending forces associated to the passage of each bogie would be overestimated if the static correction procedure was not considered. This fact is particularly important in the case of the more refined analyses considering the train-bridge interaction (see Figure 5). 5 CONCLUSIONS This paper aimed to describe a methodology that combines the mode-superposition method with a static correction procedure, using an influence surface to obtain the static contribution of higher frequency vibration modes (not explicitly consider analyses according to the basic mode-superposition method). This procedure was implemented both in analyses where railway traffic is represented by moving forces and in analyses considering the train-bridge interaction. Different analysis methodologies were applied in the calculation of structural response of a real railway viaduct, in Portugal. A calibrated finite element model had already been obtained for this structure, considering the results of ambient vibration tests. The comparison between different analysis methodologies validated the implemented procedure for static correction, and revealed that this procedure should be considered in the calculation of local effects in railway deck slabs, 1149
Figure 5.
Comparison of bending moments from analyses C and D.
when the mode-superposition method is used in the analysis of the bridge subsystem. Given that the fatigue damage is very sensitive to the applied stress ranges, this procedure is relevant in the context of fatigue analysis of deck slabs in railway bridges and viaducts. The present work is to be further developed by pursuing a procedure for definition of the minimum number of vibration modes which have to be considered in analysis according to the improved mode-superposition method with static correction. ACKNOWLEDGMENTS The present work has been funded by the Portuguese Foundation for Science and Technology (FCT), in the context of the research grant SFRH/BD/79816/2011 provided to the first author. The authors also wish to acknowledge all the information provided by REFER as well as the support provided by CSF, Centre of Competence in Railways of the Faculty of Engineering of the University of Porto. REFERENCES Ansys 2007. Structural Analysis Guide – Release 11.0, Canonsburg, PA, ANSYS. Calçada, R. 1995. Efeitos dinâmicos em pontes resultantes do tráfego ferroviário a alta velocidade. Faculdade de Engenharia da Universidade do Porto. CEN 2003. EN1991-2. Actions on Structures – Part 2: General Actions – Traffic loads on bridges, Europeean Committee for Standardization, Brussels. Chopra, A. K. 1995. Dynamics of Structures: Theory and Applications to Earthquake Engineering, Berkeley, Prentice Hall International. Clough, R. & Penzien, R. 1993. Dynamics of Structures, Second Edition, NewYork, McGraw-Hill International Editions. Malveiro, J., Ribeiro, D., Calçada, R. & Delgado, R. 2014. Updating and validation of the dynamic model of a railway viaduct with precast deck. Structure and Infrastructure Engineering, 10, 1484–1509. Matlab 2011. Getting started guide, Natick, MA, The MathWorks. Pimentel, M., Figueiras, J. & Bruhwiler, E. 2008. Dynamic analysis for fatigue safety examination of existing short span concrete railway bridges. EURODYN 2008 – 7th European Conference on Structural Dynamics. Southampton, UK. Ribeiro, D. 2012. Efeitos dinâmicos induzidos por tráfego em pontes ferroviárias: modelação numérica, calibração e validação experimental. Faculdade de Engenharia da Universidade do Porto. Ribeiro, D., Calçada, R., Delgado, R., Brehm, M. & Zabel, V. 2012. Finite element model updating of a bowstring-arch railway bridge based on experimental modal parameters. Engineering Structures, 40, 413–435. Sousa, C. F., Rocha, J., Calçada, R. & Serra Neves, A. 2013. Fatigue analysis of box-girder webs subjected to in-plane shear and transverse bending induced by railway traffic. Engineering Structures, 54, 248–261.
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Optimized bridge deck design using a genetic algorithm B. Lima & A. Ferreira Streng Engenharia de Estruturas, Porto, Portugal
ABSTRACT: This paper describes a methodology for an optimized preliminary bridge deck design when the span-by-span construction method is used. The formulation assumes that the deck has an infinite number of spans of equal length. Therefore, only two structural systems have to be considered. The first simulates all constructive stages, where the actions of one stage are applied in a semi-infinite continuous beam. In the second model, all the actions applied after the deck execution are considered in an infinite continuous beam. These assumptions lead to a simplification of the problem, being nonetheless the results quite close to those that would be obtained in continuous decks with five or more spans. This is also the case that commonly occurs when the use of a Moveable Scaffolding System (MSS) is economically justified. A Genetic Algorithm based Augmented Lagrangian method (GAAL) is used to search for the optimal solution. In this algorithm, each individual represents a solution, which is defined by a set of variables that characterize the various degrees of freedom of the deck geometry, prestress layout and quantity. The objective function is the materials’ total cost. In order to minimize it, the unitary costs for concrete, prestressing and reinforcing steels have to be assigned. During the algorithm, the best individual evolves towards security and economy. Solutions that do not check for security are penalized in their cost using an Augmented Lagrangian function. Genetic operators used in the algorithm, such as selection, crossover and mutation, are explained in the paper. Finally, an example of a 70.00 m span, with a box section deck designed for a highway bridge is presented. In this example the considered variables are described, and some solutions obtained during the evolutionary process are shown.
1 INTRODUCTION Span-by-span construction is usually adopted in spans ranging from 20 m to 45 m. However, larger spans have been executed, reaching approximately 70 m. One example was the viaduct designed by Jiri Strasky (SHP), built with a BERD’s MSS, over the Hostovsky Creek Valley in Slovakia, with a 69 m typical span. The falsework can be directly supported on the ground, or alternatively, in elements that have already been built, like the piers or the deck itself. When the support lies on the deck, a MSS is usually employed, thus removing the need for other means of positioning on the next span to be built. The construction joint is generally located between 0.20 and 0.25 of the span. If the MSS is supported on the cantilever, the deck becomes subjected to additional forces in relation to the span-by-span method with ground scaffolding. These additional forces don’t disappear at the end of the construction, with a secondary moment remaining installed on the deck (not to be mistaken with the secondary moment of prestress), whose magnitude gradually decreases due to creep. The influence of constructive stages in the distribution of forces is considered given a semi-infinite beam with an infinite number of spans. This consideration greatly simplifies the determination of forces at construction stages, allowing to do it with only one structural system. Likewise, the determination of forces caused by actions on the finished structure is simplified with the consideration of an infinite beam with an infinite number of spans. So, the adopted model is independent on the number of spans of the real structure. In order for proximity to exist between the 1151
Figure 1. Viaduct over Hostovsky Creek Valley, Slovakia (BERD, 2011).
results obtained by the adopted model and the results that would be obtained with a more rigorous design, the real number of spans should be greater than 4. Commonly, this condition is met when the adoption of a MSS is economically justified. In recent years, the structural optimization has been focusing itself essentially on weight, which in a way, is applicable to steel structures. In the case of composite materials like prestressed concrete, the criteria of minimizing weight no longer makes sense (Adeli & Sarma (2006), and the cost becomes the variable to be minimized. In the search for an optimum solution, a Genetic Algorithm based Augmented Lagrangian method (GAAL) is used, in which the objective-function cost depends on the unitary costs attributed to concrete, prestressing and reinforcing steels. This way, an optimized structure that is to be built in a particular country will not be optimized in a different one. Genetic Algorithms (GAs) are evolutionary algorithms that originated in the 60s and are inspired on the origin of species by means of natural selection. Contrary to the traditional optimization techniques, the convergence to a local minima is avoided, handling a set of solutions that are subjected to interrelation and evolution through genetic operators like Selection, Crossover and Mutation. For a better understanding of these algorithms, the reader should refer to Michalewicz (1996) and Haupt (2004).
2 OBJECTIVE FUNCTION DEFINITION This is a f : n → function, in which n is the number of variables, and 75 is its maximum value. Of these 75 variables, 27 pertain to geometry, 24 to the prestress quantity, and 24 to its layout. The output will be, in this case, the total cost of materials. 2.1 Deck geometry As mentioned beforehand, a continuous beam with an infinite number of spans is considered, with all spans having the same length, and a variable box section symmetric to the mid-span. As we will see in the description of the constructive stages ahead, the possibility of the top slab having two casting stages was considered. It is assumed that its preliminary design is concluded, and both its thickness (it depends essentially on the transversal behavior) and initial and final widths are already defined, thus not constituting variables susceptible for optimization. For the longitudinal analysis a constant top slab thickness was considered, although it is generally variable. A good approximation to the geometric characteristics of the sections can be achieved considering an equivalent thickness that equals the slab area (Fig. 2). The deck geometry is defined by the following functions (fig. 3): Deck height H (x); distance between webs exterior faces B(x); webs inclination D(x); webs thickness tw(x) and bottom slab thickness hfb(x). In this case, x represents the abscissa along the span, with its origin in the support. 1152
Figure 2.
Equivalent top slab thickness.
Figure 3.
Deck geometry.
Figure 4.
Deck height interpolation using values at each 1/8L.
These functions may take a multitude of shapes, and so it’s necessary a parameterization procedure on these through a finite number of variables. Thus, only the values at the eights of a span are considered. Outside these sections a cubic spline interpolation is used for H (x), tw(x) and hfb(x). As boundary conditions, the second derivative is set to zero on supports, and the first derivative set to zero in 0.5L. In the case of B(x) and D(x), a linear interpolation is considered. The position of the construction joint Lj , as well as the position of the MSS rear support LP , are also variables with possibility of optimization. This way, the number of variables defining the deck geometry along the whole span, the position of the construction joint and the MSS rear support are: 5 ∗ 5 + 2 = 27. 2.2 Prestress geometry and quantity Prestress is parameterized by means of a cubic spline and 4 parabolas, passing through control points in which the abscissae and the ordinates correspond to variables. The aforementioned spline and parabolas represent the resultant of all prestress tendons in a stage. The spline (tendons family a) is located between supports, where the first derivative is zero, and the parabolas (tendons families b, c, d and e) are located as represented in Figure 5. As can be seen in the figure above, the number of variables that define the position of the control points, and thus the geometry of the layout, is 24. 1153
Figure 5.
Prestress layout parameterization.
The number of strands is variable, but remains constant between two consecutive control points. The possibility of part of the prestress being applied after the top slab second stage casting is considered, therefore acting upon the final structural system. The number of variables defining the total quantity, and percentage of prestress applied in the final structural system is 12 ∗ 2 = 24. According to the authors’ experience, is not difficult to convert the adopted prestress parameterization into a practical disposition of prestress tendons.
2.3 Construction stages, forces and stresses As represented in Figure 3, the possibility of the transversal sections are executed in two stages is considered. Therefore, after the span-by-span construction is completed, a casting of the top slab side cantilevers is performed, with that weight mobilized in the narrowest sections (of W1 width) and in a continuous beam. The sections of W1 width will henceforth be referred to as sections 1. The sections after the casting of the side cantilevers will be designated sections_2. Likewise, the prestress applied to sections 1 will be named prestress 1, and the one applied to the already complete section will be named prestress_2. The steps considered for obtaining the forces on a generic span j before time-dependent effects, taking into account all construction stages, are the following (see figure 6): 1) Applying prestress 1 and self weight corresponding to sections 1, on stage i; 2) Applying the action P on the cantilever of span j, due to the MSS rear reaction on stage_i_b). This reaction has two components: one of them is due to the MSS self weight, and the other is due to the dead weight of the wet concrete between the joint of span j and the joint of span j + 1; 3) Applying the prestress and self weight on stage i + 1, and removal of action P on stage i + 1_a). It should be noted that now the structural system comprehends the hardened concrete between the joint of span j and the joint of span j + 1. The removal of the first component of P, is applied at the same instant in which the prestress and self weight of stage i + 1 are applied. Subsequently, the MSS launching takes place, with the removal of its self weight. The removal of these two components occurs on the same structural system; 4) Repeat the steps 1) to 3), incrementing the counters i and j up to infinite. 5) Applying the self weight of the side cantilevers second stage casting, considering an infinite and continuous beam, mobilizing sections 1; 6) Applying prestress 2 in an infinite and continuous beam, mobilizing sections 2. As can be seen in Figure 6, both the structural system and the actions applied are equal in all stages, since it was considered the model of a semi-infinite continuous beam (obviously, the variable inertia has to be considered in this model). Thus, the forces of a particular stage are obtained by adding the forces from a number of spans, due to the self weight, prestress and MSS rear reaction. In the case of construction with ground scaffolding, the MSS rear reaction forces don’t exist. 1154
Figure 6.
Self weight, prestress (prestress 1) and MSS rear reaction stages.
Simplistically, during constructive stages, neither the prestress time-dependent losses, nor stresses redistribution due to creep and shrinkage are considered. These are only considered after the structure is executed. Forces caused by actions applied upon the final structure are determined in a continuous and infinite beam. The redistribution of stresses due to creep and shrinkage is obtained in a simplified way through the well-known formula of Trost & Wolff (1969). 3 SEARCH FOR THE OPTIMAL SOLUTION Given the materials, the span, the desired deck width, the superimposed dead loads and live loads, an infinitude of combinations of the variables described in 2.1 and 2.2 is possible. Each one of these combinations represents a point in n to which corresponds a cost. This point, called vector x from here onward, has as the inferior and superior bounds of each variable xi , xil and xiu respectively. A point x may lie in the feasible region, which means that the solution checks for security. The security conditions are imposed by constraints gj (x) ≥ 0 of the optimization problem. The j index refers to the number of the considered constraint, and J is the total number of constraints. These constraints should be normalized in order to balance the importance of one towards another. The search for the optimal solution uses the algorithm proposed by Deb & Srivastava (2011). However in this case, elitism is implemented, i.e. from one generation to the next, a number of the best individuals are maintained. The Genetic Algorithm used is hybridized, being the Broyden– Fletcher–Goldfarb–Shanno (BFGS) algorithm implemented, thus allowing a local search in selected generations. Several references to this algorithm are found in literature, as for example Sun & Yuan (2006). In the present case, the constrained optimization problem is:
1155
Figure 7.
Genetic algorithm.
The search for the minimum of f (x) that verifies all constraints is performed using the Augmented Lagrangian P(x, σ t ):
In the previous expression A = A if A < 0, and A = 0 in the opposite case. R and σ j are parameters that are automatically updated in the iterations of the adopted genetic algorithm (Figure 7). There was no stopping rule defined for the algorithm, and this remained dependent on the user’s criteria. For that, at the end of each generation, a generations versus P(x,σ t ) graphic is updated, and the user can then verify if the required convergence was achieved. When convergence is achieved, R and σj allow the calculation of the Lagrange multipliers by λj = −2Rσj , which are useful for sensitivity analyses. 1156
Figure 8.
Normalized constraint function g(x).
3.1 Normalization of constraints Constraints of stresses and shear forces (web crushing capacity) were considered. In the case of prestressed concrete, stresses constraints are imposed both during the constructive stages and in service conditions. Constraints take into account the acting value, here generically designated by Ed , and the maximum (or minimum) value designated by Rd . In the case of the stresses constraints, considering positive sign for tension, and negative sign for compression, there are two types of safety conditions: Ed ≤ Rd and Ed ≥ Rd . Methodologies for normalization of constraints can be found in Datta & Deb (2012) or in Adeli & Sarma (2006). However, in this study the following expression was used:
In which = Rd − Ed if the safety condition is Ed ≤ Rd , and = Ed − Rd if the safety condition is Ed− ≥ Rd . With this expression, both conditions Ed ≤ Rd and Ed ≥ Rd are covered, taking into account the sign of each value, and the possibility of Rd being null, as it happens for example, in the verification of the limit state of decompression. This way, g(x) ∈ ]−1; 1[, ∀ ∈ , with a gradual transition from the unsafety zone to the safety zone. This transition is gradual even when Rd− ≈ 0, due to the parameter r, which, in this case was taken equal to fctk,0.05 . In figure 8 the normalized constraint function g(x) is represented. 103 control sections along the generic span were considered, two of them corresponding to the supports, one to the mid-span, and 100 to the intermediate sections of each piece of L/100 length. Stresses constraints refer to the limits for tensile and compressive stresses on construction stages and in-service combinations (quasi-permanent, frequent and characteristic according to EN19921-1). The constraints of shear forces (web crushing capacity) were also taken into account. So, the total number of constraints considered is J = 103 ∗ (2 ∗ 4 + 1) = 927. 3.2 Binary tournament selection In this operator, two individuals are randomly selected from the whole population, and the one with the less value of P(x, σ t ) (equation 2) is chosen. 3.3 Crossover There are various crossover methods present in GA’s literature. The Simulated Binary Crossover (SBX) (Deb & Agrawal (1995) and Deb & Kumar (1995)) was used, which simulates the crossover operator in binary GA’s. For each variable, a crossover probability of 50% is considered. If crossover exists, and p1 being the value of parent1, and p2 the value of parent2, the values of child1 and 1157
Figure 9. Example of 5000 SBX operations in 2 with: parent 1 = (6.7; 3.0); parent 2 = (13.3; 8.0); 5.3 ≤ x1 ≤ 16.0; 1.0 ≤ x2 ≤ 10.0; xv = 2.
child2 (c1 and c2) are determined as follows:
Were rn is a random number ∈ [0; 1], and xv is the crossover index.
3.4 Mutation In the same line as the previous case, there are variants for this operator in GA’s literature. In this case, the polynomial mutation proposed by Deb & Agrawal (1999) was used. The probability of mutation in each variable depends on the generation counter t, and the generation tmax,mut (user defined value), beyond which the probability of mutation is 100%:
The value p after the mutation of a variable with an initial value p, is obtained with the expressions (8) and (9):
1158
Figure 10. Example of 5000 polynomial mutations in 2 performed at generations 1, 10 and 50 for p = (3.0; 3.5); 1.0 ≤ x1 ≤ 8.0; 2.0 ≤ x2 ≤ 10.0; mv = 10; tmax,mut = 50.
Were rn is a random number, and ηm is the mutation index. This index depends on the generation counter t, and on a user-defined mv parameter: ηm = mv + t. 3.5 Local search At the end of every generation where (t mod v) = 0 and ρ ≤ 0.0001 (see Figure 7), a local search arround OptInd individuals is performed using the BFGS algorithm. The parameter v is a user defined value, defining the generational interval between which the local search can be executed. The choice of individuals to be optimized depends on their place in the ranking when sorted by P(x, σ t ) in ascending order:
For example, if N = 100 and OptInd = 4, the individuals to be optimized will be the ones that occupy the positions 1, 8, 22 and 51. This way, will be optimized more “better” individuals than “worse” ones, always optimizing the best, and the one at the intermediate position. 4 EXAMPLE In this example, the aim is to determine the most cost-effective solution for a highway bridge deck with a 70.00 m span. The unitary costs are the following: concrete, 100 $/m3 ; reinforcing steel, 1.02 $/kg; prestressing steel, 2.82 $/kg. According to the authors’ experience, the expectable reinforcing steel density for this type of highway bridge decks is about 150 kg/m3 . After the spanby-span construction is completed with a width of 7.60 m, a casting of the top slab side cantilevers is performed, having the deck a total width of 22.00 m. These side cantilevers are supported by struts from the bottom slab. The distance between webs exterior faces was fixed in 5.95 m. The top slab preliminary design results in an equivalent thickness of 0.3287 m (Figure 2). A concrete class with fc = 40 MPa at 28 days, and an equivalent fc = 35 MPa at prestress stressing is considered. Prestress strands of 0.6 with 1.40 cm2 , have fpuk = 1860 MPa, and are stressed at 0.75fpuk . The superimposed dead loads amount to 74.20 kN/m. Vehicular live loading consists of an equivalent concentrated load of 932 kN, and an uniformly distributed load of 30.9 kN/m. The temperature gradient is +16◦ C and −8◦ C. Stresses limits are in accordance with AASHTO standard. These 1159
Table 1. Limits to minimum and maximum stresses. Minimum
Maximum
Combination
MPa
MPa
Constructive stages Service I (permanent loads) Service III Service I (all loads)
−21.0 −18.0 – −22.8
3.23 – 2.16 –
Table 2. Variables bounds. Variable
Minimum
Maximum
H hfb(0); hfb(L/8); hfb(L/4) tw(0); tw(L/8); tw(L/4) X2 X4 X5 X6 X7 X8 Yi; i = 1 to 9 Xb; Xc Xd Number of strands between control points % of strands applied in Prestress 1
2.80 m 0.20 m 0.50 m 5.00 m 16.00 m 25.20 m 36.40 m 47.60 m 58.80 m 0.05H 0.10L 0.10L 100 0
5.00 m 1.50 m 1.50 m 12.00 m 25.20 m 36.40 m 47.60 m 58.80 m 65.00 m 0.95H 0.40L 0.20L 700 100
limits during constructive stages, and in service combinations, are the ones indicated in Table 1. Maximum tensile stresses in constructive stages allowed by AASHTO are reduced in 500 kPa, to take into account the restraint that the piers have in the deck. For the same reason, the maximum tensile stress in Service III is reduced in 1000 kPa. Vrd,max values were calculated considering a grouted metal duct of 125 mm by level in each web. For the sake of constructive simplicity, it was imposed that the deck height is constant along the span. The webs inclination are set to D = 10 and also constant along the whole span. It was imposed that tw(L/4) = tw(3/8L) = tw(L/2), and hfb(L/4) = hfb(3/8L) = hfb(L/2). The distance of both the MSS rear support and the construction joint to the pier, were not considered as variables of the problem, considering 9.5 m and 14.0 m respectively. Therefore, out of the 27 available variables described in 2.5, just 7 of them define, in this case, all of the deck’s geometry. Regarding the prestress layout, some conditions were also imposed that reduce the number of variables to be considered: X 3 = 14.0 m; Yb =Yc =Yd = 0.75 m; Xe = 14.0 m and Ye = Y 3. This way, out of the 24 available variables, just 18 of them define all prestress layout. The number of strands is the same on the left and right of each support. So, the number of variables defining its quantity is reduced from 24 to 20. In this example, the total number of variables will then be 7 + 18 + 20 = 45. In Table 2, the lower and upper bounds of each variable are presented, and in table 3, the parameters used in the GAAL. As can be seen in Figure 11, a bigger decrease in the objective function f (or cost) is observed in the first generations. Likewise, it’s also in the first generations that the biggest oscillations in this value occur due to the self-adaptive σ and R parameters. In the last generation, 825 m3 (0.536 m3 /m2 ) of concrete, and 26492 kg (17.20 kg/m2 ) of prestressing steel for each span were achieved. 1160
Table 3. Genetic Algorithm parameters. Population (N )
Elitism
xv
tmax,mut
mv
v
OptInd
720
0.40
2
200
10
25
3
Figure 11. Objective function f versus generations (left) and evolutionary process of prestress layout (center) and geometry (right).
Figure 12.
Shear forces (left) and Service III stresses (right) at generation 1690.
5 CONCLUSIONS The use of GAs in the conception of prestressed concrete decks leads to optimized solutions, as we could verify through the previous example. One of the reasons that justifies the attained level of optimization is the stresses and Vrd,max diagrams, in which many sections are explored almost to the limit. For the definition of the objective function, an infinite number of spans was considered, which vastly simplified the problem, allowing nonetheless, good proximity to the phenomena involved. Some advantages of applying the GAs are the independence of the initial solution and the applicability to multimodal functions. Hybridization with conventional optimization methods (BFGS in this case) allows both a diversified and localized scanning of the search space.
ACKNOWLEDGEMENTS The authors would like to thank BERD Bridge Engineering Research & Design, for the proposed works that lead to the need for the development of an optimization software based on the contents of this paper. We also thank Professor Isabel Espírito Santo, from the Departament of Produção e Sistemas of the University of Minho, and Engineer Soumil Srivastava, of McKinsey & Company. Lastly, our thanks to all our colleagues from STRENG, Engenharia de Estruturas, Lda. 1161
REFERENCES Adeli, H. & Sarma, K. 2006. Cost Optimization of Structures: Fuzzy Logic, Genetic Algorithms, and Parallel Computing. England: Wiley. Costa, L., Espírito Santo, I.A.C.P., Denysiuk, R., Fernandes, E.M.G.P. 2010. Hybridization of a Genetic Algorithm with a Pattern Search Augmented Lagrangian Method. Lisbon: Rodrigues, H., et al. (eds.) Proc. of 2nd International Conference on Engineering Optimization. Datta, R. & Deb, K. 2012. An Adaptive Normalization based Constrained Handling Methodology with Hybrid Bi-Objective and Penalty FunctionApproach. Technical Report, KanGAL Report Number 2012005. Kanpur: Indian Institute of Technology. Deb, K. & Agrawal, R.B. 1995. “Simulated Binary Crossover for Continuous Search Space”. Complex Systems, 9, pp. 115–148. Deb, K. & Kumar, A. 1995. “Real-coded Genetic Algorithms with Simulated Binary Crossover: Studies on Multimodal and Multiobjective Problems”. Complex Systems, 9, pp. 431–454. Deb, K. & Agrawal, S. 1999. “A niched-penalty approach for constraint handling in genetic algorithms”, Proceedings of the International Conference on Artificial Neural Networks and Genetic Algorithms (ICANNGA-99), Springer-Verlag, pp. 235–243. Deb, K. & Srivastava S. 2011. A Genetic Algorithm Based Augmented Lagrangian Method for Accurate, Fast and Reliable Constrained Optimization. Technical Report, KanGAL Report Number 2011012. Kanpur: Indian Institute of Technology. Haupt, R. & Haupt, S.E. 2004. Practical Genetic Algorithms. New Jersey: Wiley. Michaelwicz, Z. 1996. Genetic Algorithms + Data Structures = Evolution Programs. Berlin: Springer. Sun, W. & Yuan, Y.X. 2006. Optimization Theory and Methods: Nonlinear Programming. New York: Springer. Trost, H. & Wolff, H.J. 1969. Zur wirklichkeitsnahen Ermittlung der Beanspruchungen in abschnittsweise hergestellten Spannbetontragwerken. Hannover: Techn. Univ., Lehrstuhl und Inst. für Massivbau.
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Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
Stiffened flanges used in steel box girder bridges P.S. Ferreira Polytechnic Institute of Setubal, Escola Superior de Tecnologia do Barreiro and ICIST, Setubal, Portugal
F. Virtuoso University of Lisbon, Instituto Superior Técnico, DECivil-ICIST, Lisbon, Portugal
ABSTRACT: This paper presents a design method for stiffened flanges used in steel box girder bridges based on a set of design curves that take into account all relevant parameters regarding the ultimate strength of stiffened flanges, including the real boundary conditions for the in-plane displacements at the edges. These design curves were developed and calibrated based on the results obtained with nonlinear analysis using the semi-analytical and finite element methods and they were validated by comparison with the results of experimental tests in accordance with the target failure probability of the European standard EN-1990. A comparison between the ultimate strength obtained using the proposal design curves with that obtained using the nonlinear analysis and the current bridge design rules is presented.
1 INTRODUCTION National standards with design rules for stiffened flanges used in steel box girder bridges were almost non-existent until the early nineteen seventies. These design rules were based on the elastic critical stress (σcr ) obtained by the linear buckling theory of plates with some correction factors to take into account the favorable effect of the post-critical strength reserve (ECCS Committee 8 1976). Research work to establish new design rules for plated structures was strongly increased in the early nineteen seventies as a result of well-known accidents during construction involving box girder bridges (Galambos 1998). Although the behavior and strength of stiffened plates have been extensively studied over the last forty years, there are still some inconsistencies and gaps between the results of some studies and the results obtained through the design rules established in the European and American standards (EN-1993-1-5 2006, AASHTO-LRFD-BDS 2007), which are usually adopted by the bridge designers to solve the problem of predicting the ultimate strength of stiffened plates. In wide flanges and webs of considerable height it is common to use longitudinal stiffeners to reduce the problems associated with the effects of buckling. The study of the behavior and strength of steel plates with longitudinal stiffeners is very complex because they depend on a large number of variables, such as the shape and amplitude of the initial geometric imperfections, the residual stresses, the boundary conditions for the in-plane and outof-plane displacements, the load conditions and the geometric and material data. Recent numerical studies (Ferreira 2012, Braun 2010) showed that the behavior and strength of steel plates are very sensitive to in-plane displacements boundary conditions and the European and American standards (EN-1993-1-5 2006, AASHTO-LRFD-BDS 2007) do not provide a definition and clear guidance for the proper use of the simplified rules available to designers. These rules may give considerable errors, particularly when applied to the safety verification of stiffened plates where the transverse in-plane displacements at longitudinal edges cannot be considered as uniform (in-plane displacements perpendicular to the longitudinal edges constrained to remain straight). 1163
In European standard (EN-1993-1-5 2006) the resistance assessment of webs and internal flanges with longitudinal stiffeners used in steel box girder bridges is performed through the same criterion and independent of the in-plane displacement boundary conditions of the plate. Just apparently both elements may have the same in-plane displacement boundary conditions. Indeed, the webs are usually connected to flanges that possess sufficient rigidity to consider uniform the in-plane displacements at longitudinal edges, while the flanges are connected to webs that usually do not provide this type of constraint and it is more conservative to consider free the in-plane displacements at longitudinal edges, as shown in the work of Ferreira & Virtuoso (2011) through a comparison between numerical and experimental results. The design proposal presented in this paper for stiffened flanges aims to fill a gap in the current European bridge design rules, which should not be applied to stiffened plates with the fully free transverse in-plane displacements at longitudinal edges. 2 DESIGN PROPOSAL The design proposal is a model based on design curves (arithmetic expressions) describing the influence of all relevant parameters on the ultimate strength of stiffened flanges used in steel box girder bridges. It was developed and calibrated based on the results obtained with the semi-analytical and finite element methods and it was validated by comparison with the results of experimental test in accordance with the target failure probability of EN-1990 (2002). These design curves consider that the plate is subject to compressive longitudinal direct stress with a stress ratio higher than 0.5 (lower compressive longitudinal stress to the higher compressive longitudinal stress ratio, ψ > 0.5), the transverse stiffeners provide rigid support lines, the longitudinal stiffeners are fully effective and equally spaced and the plates have the longitudinal in-plane displacements constrained to remain straight at loaded edges (transverse edges) and the other in-plane displacements fully free. This design model does not consider the effects of local buckling of the stiffener elements and tripping of the stiffener. These effects can be avoided by adopting minimum values for the geometric properties of the longitudinal stiffeners as used in European and American standards (EN-1993-1-5 2006, AASHTO-LRFD-BDS 2007). The design proposal is a design approach based on the reduced section concept, where the mean compressive stress at peak load (or ultimate strength σu ) of a stiffened plate is estimated through the yield strength of the effective cross-sectional area according to
where fy = yield stress; ρglob = global reduction factor; Aeff ,loc = effective local cross-sectional area which takes into account the local buckling effects; and A = gross cross-sectional area of the stiffened plate. From Equation 1 it can be noted that the effective cross-sectional area (Aeff = ρglob · Aeff ,loc ) which takes in-to account the local and global buckling effects is obtained by reducing the cross-sectional area of the stiffened plate in two different steps, as illustrated in Figure 1. In the first step the effect of local buckling in the panels between longitudinal stiffeners is taken into account and an effective width of these panels is adopted to obtain an effective local cross-sectional area (Aeff ,loc ) according to
where nst = number of longitudinal stiffeners; ρloc = the local reduction factor; bp = width of the panels between longitudinal stiffeners; t = plate thickness; and Asl = gross cross-sectional area of a single longitudinal stiffener (without the contribution of the plate). The local reduction factor 1164
Figure 1. Reduction of the cross-sectional area of a stiffened plate in the design proposal: (a) the gross cross-sectional area, (b) the effective local cross-sectional area and (c) the effective cross-sectional area.
(ρloc ) takes into account the favorable effects from post-critical strength reserve and the unfavorable effects of the initial imperfections and it is defined for each panel between longitudinal stiffeners by
where λp,norm = relative slenderness of the panels. For a relative slenderness of the panels lower than 0.673 the local reduction factor should be considered equal to 1. Equation 3 is the same criterion used in European standard (EN-1993-1-5 2006). In the second step, which takes into account the effect of global buckling in the stiffened plate, a global reduction factor (ρglob ) is considered to reduce the effective local cross-sectional area (Aeff ,loc ). The global reduction factor is given by
where χc = reduction factor for the column buckling behavior; and ρp,∞ = reduction factor for the plate buckling behavior estimated for a long stiffened plate (stiffened plate where the elastic critical stress σcr no longer depends of the plate length a). Column buckling behavior is considered in stiffened plates with insignificant post-critical resistance. This behavior is modeled by removing the supports along the longitudinal edges of the plate and considering the cross-section of the column equivalent to the plate composed of a single stiffener with an effective width of the adjacent panels to the stiffener. The reduction factor for the column buckling behavior is obtained by
where λc,norm = relative slenderness of the equivalent column; and φc = parameter that depends on the initial imperfections, the relative slenderness of the equivalent column (λc,norm ), the relative flexural stiffness of the stiffened plate γ (second moment of area of the stiffened plate Ist to the second moment of area for bending of the plate corrected by the effect of the Poisson coefficient (bt 3 /[12(1 − ν2 )]) ratio) and the number of longitudinal stiffeners (nst ). The first term in Equation 5 (0.66λc,norm ) is the critical criterion in stiffened flanges with low values of equivalent column relative slenderness (λc,norm ) and it is equal to the reduction factor established in the American standard (AASHTO-LRFD-BDS 2007) for the column buckling behavior of stiffened plates. The other term in Equation 5 is the determinant criterion in stiffened plates with intermediate and high values of equivalent column relative slenderness (λc,norm ) and it is equal to the reduction factor established in the European standard (EN-1993-1-5 2006) for 1165
Figure 2.
Reduction factor for the column buckling behavior obtained using the design proposal.
the column buckling behavior of stiffened plates with an improvement that takes into account the relative flexural stiffness of the stiffened plate (γ) and the number of longitudinal stiffeners (nst ). Figure 2 presents the reduction factor for the column buckling behavior (χc ) obtained using the design proposal. It is also shown the yield, the elastic buckling and the European standard criteria. An equivalent geometric imperfection of 1/400 of the plate length (a) was used. Plate buckling behavior is considered in stiffened plates with post-critical resistance. The design proposal uses this type of behavior to define the minimum ultimate strength of the stiffened plate, which is estimated considering a long plate. The reduction factor for the plate buckling behavior of a long stiffened plate is obtained by
where λp,norm = relative slenderness of the stiffened plate; and ξ = parameter that depends on the compressive longitudinal stress ratio (ψ), the panel slenderness λp (width of the panels between stiffeners bp to the plate thickness t ratio), the relative cross-sectional area δ (cross-sectional area of the stiffeners without any contribution of the plate nst · Asl to the cross-sectional area of the plate (bt) ratio), the relative flexural stiffness of the stiffened plate (γ) and the number of longitudinal stiffeners (nst ). The reduction factor for the plate buckling behavior estimated for a long stiffened plate (ρp,∞ ) is governed by the parameter ξ and Equation 6 becomes the well-known criterion proposed by Winter when this parameter is equal to 0.22. Based on the geometric parameters of the stiffened flanges identified in the survey of stiffened flanges used in real steel box girder brides presented in Ferreira (2012) the minimum and maximum values of the parameter ξ are 0.19 and 0.64 respectively. Figure 3 presents the reduction factor for the plate buckling behavior of a long stiffened plate (ρp,∞ ) obtained using the design proposal with different values of the parameter ξ. It is also shown the yield, elastic buckling and Winter criteria. Figure 3 shows that the application of the design proposal results in a lower ultimate strength (σu ) of the stiffened flange with fully free in-plane displacements at longitudinal edges than that obtained by the current European bridge design rules, which is consistent with the results obtained using nonlinear finite element simulations and with the results of experimental tests presented in the works of Ferreira & Virtuoso (2010, 2011). The complete definition of the procedures and variables used in the design proposal and the statistical evaluation using experimental test and numerical results within the framework of a 1166
Figure 3. Reduction factor for the plate buckling behavior obtained using the design proposal with different values of the parameter ξ.
probabilistic reliability theory in accordance with EN-1990 (2002) can be found in the work of Ferreira (2012). It is noted that the numerical value for the partial safety factor (γM ) obtained from the statistical evaluation is 1.10, which is the recommended numerical value for the partial factor for resistance of members to instability assessed by member checks (γM1 ) on bridges (EN-1993-2 2006). 3 RESULTS, DISCUSSION AND CONCLUSIONS The comparison between the ultimate strength (σu ) obtained using the design proposal with that obtained using nonlinear analyses and current bridge design rules established in the European and American standards (EN-1993-1-5 2006, AASHTO-LRFD-BDS 2007) was performed considering 20 stiffened flanges. The 20 stiffened flanges were obtained based on two types of cross-sections, SF1 and SF2, and considering for each cross-section the plate aspect ratio φ (=a/b) ranging from 0.5 to 5.0 at intervals of 0.5. Both cross-sections have five longitudinal equally spaced and single sided stiffeners, the stiffened flanges with the cross-section SF1 have a relative flexural stiffness γ (=12Ist (1 − ν2 )/(bt 3 )) of 65 and the stiffened flanges with the cross-section SF2 have a relative flexural stiffness γ of 130. All stiffened flanges have a relative cross-sectional area δ (=Asl /(bt)) of 0.5, plate slenderness λplt (=b/t) of 150 and panel slenderness λp (=bp /t) of 25. The nonlinear analyses were performed using the semi-analytical model presented in the work of Ferreira (2012) and considering a yield stress (fy ) of 355 Nmm−2 , Young’s modulus (E) of 2.1 × 105 Nmm−2 and Poisson’s coefficient (ν) of 0.3. The stiffened flanges were considered simply supported under longitudinal uniform compression (σ) with the following two cases for the in-plane displacement boundary conditions: in-plane displacements perpendicular to the edges constrained to remain straight in all edges (case CC) and in-plane displacements perpendicular to the edges constrained to remain straight at loaded edges and free at unloaded edges (case CF). Figure 5 presents the comparison between the ultimate strength (σu ) obtained by the design proposal, the semi-analytical model and current bridge design rules. Figure 5a shows the ultimate strength for the stiffened flanges with the cross-section SF1 and Figure 5b shows the ultimate strength for the stiffened flanges with the cross-section SF2. The results are presented in terms of normalized strength (mean compressive stress at peak load σu to the yield stress fy ratio) for different values of the plate aspect ratio φ (=a/b). From the analysis of Figure 5 it can be noted that the ultimate strength (σu ) obtained by the design proposal presents a much better agreement with the nonlinear analysis results obtained by the semi-analytical model considering the case CF for the in-plane displacement boundary 1167
Figure 4. Comparison between the ultimate strength obtained by the design proposal (DCP), the semi-analytical model (SAM) and current bridge design rules for the stiffened flanges with the cross-section: (a) SF1 and (b) SF2.
conditions than the ultimate strength obtained by the current bridge design rules. This observation is particularly clear in the case of long plates, where the design rules established by the European standard leads to nonconservative results and the design rules established by the American standard leads to extremely conservative results. The design proposal predicts the ultimate strength (σu ) in a quick and simple way and provides viable and conservative results, which thus proves its usefulness as a design tool. ACKNOWLEDGEMENTS This work was carried out in the framework of the research activities of ICIST, Instituto de Engenharia de Estruturas, Território e Construção and was funded by FCT, Fundação para a Ciência e Tecnologia. REFERENCES AASHTO-LRFD-BDS. 2007. AASHTO LRFD Bridge Design Specification. Washington: American Association of State Highway and Transportation Officials. Braun, B. 2010. Stability of steel plates under combined loading. PhD thesis. Stuttgart: University of Stuttgart. ECCS Committee 8. 1976. Manual on Stability of Steel Structures. ECCS publication 22. Brussels: ECCS. EN 1990. 2002. Eurocode – Basis of structural design. Brussels: European Committee for Standardization. EN-1993-1-5. 2006. Eurocode 3 – Design of Steel Structures – Part 1-5: Plated Structural Elements. Brussels: European Committee for Standardization. EN-1993-2. 2006. Eurocode 3 – Design of Steel Structures – Part 2: Steel Bridges. Brussels: European Committee for Standardization. Ferreira, P. 2012. Stiffened compression flanges of steel box girder bridges: postbuckling behaviour and ultimate strength. Ph.D thesis. Lisbon: Instituto Superior Técnico, Technical University of Lisbon. Ferreira, P. & Virtuoso, F. 2011. Efeito do tipo de restrição nos bordos longitudinais no comportamento e resistência de placas metálicas em pontes. In L. Simões da Silva, P. Cruz, N. Lopes, J. Fernandes, and A. Baptista (ed.), Proceedings of the 8th Conference on Steelwork Construction, Guimarães, 24–25 November 2011. Coimbra: CMM and University of Minho. (in Portuguese) Ferreira, P. & Virtuoso, F. 2010. Collapse load of box flanges with longitudinal stiffeners under uniform compression: comparative study of EC3 and AASHTO LRFD bridge design rules with nonlinear finite element analysis. In N.Yardimci, B.Aydöner,Y. Gür’es, and C.Yorgun (ed.), Proceedings of the International Symposium on Steel Structures: Culture & Sustainability, Istanbul, 21–23 September 2010. Istanbul: TUCSA and ECCS. Galambos, T. 1998. Guide to Stability Design Criteria for Metal Structures. NewYork: John Wiley & Sons, Inc.
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Finite Element Modeling of the Fatih Sultan Mehmet Suspension Bridge S.A. Kilic Department of Civil Engineering, Bogazici University, Bebek, Istanbul, Turkey
H.J. Raatschen Fachhochschule Aachen, Fachbereich Maschinenbau und Mechatronik, Aachen, Germany
B. Körfgen Forschungszentrum Jülich GmbH, Institute for Advanced Simulation, Jülich, Germany
A. Astaneh-Asl Department of Civil and Environmental Engineering, University of California, Berkeley, USA
N.M. Apaydin Turkish Directorate of Highways, Kagithane, Istanbul, Turkey
ABSTRACT: This study presents the 3D finite element model of the Fatih Sultan Mehmet Suspension Bridge located In Istanbul, Turkey. All the towers and the deck are modeled with four node thin shell finite elements with the inclusion of internal diaphragms. The main suspension cable, the back-stay cable, and the hanger cables are modeled with two node beam finite elements. An initial nonlinear static analysis utilizing the geometric stiffness is performed in order to obtain the correct pre-stressing forces in the cables. An eigenvalue analysis of the bridge is performed once a converged solution is obtained by the non-linear static analysis. The results of the eigenvalue analysis are compared with the available ambient vibration test measurements and the results of the finite element model of the bridge with only beam elements. The results show that the 3D numerical model utilizing thin shell finite elements can accurately represent the modal periods of the suspension bridge.
1 INTRODUCTION The Fatih Sultan Mehmet Bridge (Figure 1) is the second suspension bridge built across the Bosporus straits in Istanbul. The bridge plays a crucial role in the local commuter traffic as well as in the international trade route. The main span is 1090 m and the deck is 64 m above the sea level. The 39,4 m wide deck carries 8 lanes of traffic. The height of each tower is 111 m above the foundation level. Ambient vibration tests under daytime traffic load were conducted by other researchers (Brownjohn et al. 1992). The modal periods obtained in the ambient vibration tests are compared with the results of the current study. The 3D finite element model will serve as a basis for the future seismic investigations of the Fatih Sultan Mehmet Bridge. Previous numerical studies of the bridge utilized two-node beam finite elements for modeling the cables of the bridge as well as the orthotropic deck (Apaydin 2010). However, the beam element approach is not capable of accurately representing the presence of diaphragms, stiffener plates, and other structural elements that exist in the deck structure. Such beam finite element models of the bridge require an iterative approach to fine tune the model due to the difficulty of simulating the orthotropic nature of the deck system through the use of gross cross-sectional resultant properties. 1169
Figure 1. The Fatih Sultan Mehmet Bridge over the Bosporus straits in Istanbul.
Figure 2.
Finite element model of the Fatih Sultan Mehmet Bridge using only beam elements (Apaydin 2010).
This study presents the eigenvalue analysis of the Fatih Sultan Mehmet Bridge using shell finite elements for the deck and the four towers. Results are compared with the available ambient vibration test measurements. 2 FINITE ELEMENT MODEL OF THE BRIDGE Figure 2 shows the beam finite element model of the Fatih Sultan Mehmet Bridge (Apaydin 2010). The orthotropic deck is represented by longitudinal and transverse beam elements. Resultant stiffness properties of the gross cross-section are provided as inputs for the beam elements. 1170
Figure 3.
Finite element model of the Fatih Sultan Mehmet Bridge utilizing shell finite elements.
Figure 4.
Close-up view of the finite element model showing the towers and the deck.
Figure 5. FE model of the deck segment showing the shell finite elements of the deck plates and the beam finite elements of the hanger cables, and the main suspension cable.
Figures 3–5 illustrate the shell finite element model for the current study. Figure 3 shows the overall view of the model. Figure 4 provides the close-up view of the deck shell elements and the two towers. Figure 5 illustrates the shell elements of a typical deck segment. The hanger cables, the main suspension cable, and the back-stay cables are modeled with beam elements. No parameter fitting or model tuning is required for the shell finite element model. 1171
Figure 6.
First lateral (symmetric) mode with a period of 13.51 seconds.
Figure 7.
Second vertical (symmetric) mode with a period of 6.33 seconds.
3 ANALYSIS PROCEDURE The commercial code LS-DYNA is used as the analysis tool (Hallquist 2006). The calculations are carried out in two steps. In the first step the gravity load and cable pre-straining are applied simultaneously in order to achieve the deformed geometry of the bridge as well as the final tensile forces in all of the cables. A non-linear analysis utilizing the full Newton solver is carried out until convergence is reached. The restart option of the LS-DYNA code is utilized in order to save the converged static equilibrium configuration. The second step involves the eigenvalue analysis of the bridge once the static equilibrium configuration is reached. The solution of the eigenvalue problem is carried out by means of the block shift and invert Lanczos eigensolver of LS-DYNA (Hallquist 2006). 4 RESULTS OF THE EIGENVALUE ANALYSIS Figure 6 shows the top, the side, and the oblique views of the first lateral mode with a period of 13.51 seconds. The ambient vibration test yielded a period of 12.99 seconds for this mode (Brownjohn et al. 1992). The beam finite element result for the same mode is 13.38 seconds (Apaydin 2010). Figure 7 illustrates the second vertical mode with a period of 6.33 seconds. The periods of the ambient vibration test and the beam finite element model for the same mode are 6.45 and 6.26 seconds, respectively. Figure 8 presents the result for the first torsional mode with a period of 3.42 seconds. The ambient vibration test yielded a period of 3.38 seconds (Brownjohn et al. 1992). The beam finite element model result for the same mode is 2.58 seconds. Table 1 provides the comparison of modal frequencies between the ambient vibration test results (Brownjohn et al. 1992) and the calculations of the finite element models for the symmetric (sym) 1172
Figure 8.
First torsional (symmetric) mode with a period of 3.42 seconds.
Table 1. Comparison of the frequencies for the torsional modes. Mode shape no. and type
Frequency from the ambient vibration test (Hz)
Frequency from the shell FE model (Hz)
Frequency from the beam FE model (Hz)
T1 sym T2 asym T3 sym T4 asym T5 sym
0.296 0.352 0.529 0.692 0.867
0.292 0.351 0.536 0.694 0.867
0.387 0.417 0.633 0.799 1.026
and the anti-symmetric (asym) torsional modes T1 through T5. The shell finite element model provides closer results to the ambient vibration test measurements for all the modes presented in Table 1. 5 CONCLUSIONS The finite element model of the Fatih Sultan Mehmet suspension bridge is presented in this study. The deck and the towers are modeled with shell finite elements. Internal structural components such as diaphragms and stiffener plates are modeled with shell elements. The main suspension cable, the back-stay cable, and the hanger cables are modeled with beam finite elements. Comparisons with the available ambient vibration tests as well as the beam finite element model of the bridge show that shell finite elements can accurately predict the modal periods and the eigenmodes. Fine-tuning of the model input parameters is not necessary as the internal structural components are explicitly included in the model. Future work on the bridge model will involve seismic studies of the bridge subjected to ground motions. The shell finite element model is suitable for analyzing the local stress concentrations due to nonlinear geometric and material behavior that might occur in seismic studies. REFERENCES Apaydin, N.M. 2010. Earthquake Performance Assessment and Retrofit Investigations of Two Suspension Bridges in Istanbul. Soil Dynamics and Earthquake Engineering, 30:702–710. Brownjohn, J.M.W., Dumanoglu, A.A., Severn, R.T. 1992. Ambient Vibration Survey of the Fatih Sultan Mehmet (Second Bosporus) Suspension Bridge. Earthquake Engineering & Structural Dynamics, 21: 907–924. Hallquist, J.O. 2006. LS-DYNA Theory Manual. Livermore, California, U.S.A.: Livermore Software Technology Corporation (LSTC).
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Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
Numerical simulation of wind pressure of a continuous fin-back bridge M. Fu, X. Li, Z. Nie & Z. Tang Shanghai Urban Construction Design and Research Institution, Shanghai, China
ABSTRACT: The continuous fin-back bridge is a novel bridge type in China. The change of the cross section along the longitudinal direction of the bridge is very large, so the static force coefficients cannot be obtained by section model test in wind tunnel. The wind field of the full size model of Dazhi River Bridge was simulated by FLUENT, and the influence of barriers and piers on the wind field was also considered. Based on the wind field analysis under total 5 attack angles from −5◦ to 5◦ , the distribution of the pressure coefficient under 0◦ attack angle and the static coefficients under every attack angles were presented. Besides, the static coefficients of the bridge with and without fin under 0◦ attack angle were compared. The results show that the fin increases the windward area of the bridge, which directly causes the large increase of the drag coefficient, the small change of the lift coefficient, and the decrease of the pitching moment coefficient.
1 INTRODUCTION The static force coefficients of the bridge are important parameters for static wind loads and stability analysis. They have direct influence on the precision of wind analysis (Cheng, 2005). Section model of proper scaled ratio was commonly used to test the static force coefficients in wind tunnel. However, disadvantages exist in experimental research such as long period, high cost and complex equipment, etc. Meanwhile, if the cross section varies along the bridge, it is not possible to obtain the static force coefficients using the section model. With the development of computer technology and computational fluid mechanics, numerical wind tunnel (Xiang & Cheng, 2003, Qu & Liu, 2007, Tan & Cheng, 2009, He et al., 2011) has become a new approach to study the static force coefficients of bridges. Compared to wind tunnel test, numerical wind tunnel is not limited by the building size, so full size model could be used. Meanwhile, the mechanism could be directly displayed by visualization of the flow field (Li et al., 2010). FLUENT is used to simulate the response of Dazhi River Bridge under static wind load. Full size model of the bridge was built, and the influence of guardrail and pier was considered. Based on the wind field analysis under total of 5 attack angles from −5◦ to 5◦ , the distribution of the pressure coefficient under 0◦ attack angle and the static coefficients under every attack angle were presented. Additionally, the static coefficients of the bridge with and without fin under 0◦ attack angle were compared. 2 ENGINEERING SITUATION Dazhi River Bridge is a prestressed concrete continuous fin-back bridge with three spans of 92 m, 158 m and 92 m. The bridge width is 35 meters with eight lanes in two ways. The maximum height of the fin is 20 m, while the minimum height is about 2.5 m. The approach bridges are steel-concrete composite beams. The standard span of the approach bridge is 30 m. Figure 1 shows the general layout of the bridge. 1175
Figure 1.
General layout of Dazhi River Bridge (unit: m).
3 PRESSURE COEFFICIENT AND STATIC FORCE COEFFICIENTS 3.1 Pressure coefficient Pressure coefficient is the ratio of the actual pressure on building surface caused by wind load and the wind pressure. It describes the static force distribution under stable wind pressure. It is mainly related to the building’s shape and size (MOC, 2006). According to the regulations for irregular shapes in Chinese load code, the pressure coefficient of wind load is described as follows (Kong et al., 2008):
where ps is the actual pressure on the building surface (N/m2 ), p0 is the wind pressure (N/m2 ), ρ is the air density (1.225 kg/m3 ), u0 is the wind speed (m/s). 3.2 Static force coefficients The cross section is subjected to drag force, lift force and pitching moment in the field of fluid with the speed U . The static force coefficients are defined as (Cheng, 2005):
where D, B and L are the height (m), width (m) and length (m) of the bridge, respectively, FD , FL and MT are the drag force (N), lift force (N) and pitching moment (N·m) of the cross section, respectively. U is the wind speed of the reference point. 1176
Figure 2.
Full size geometry model.
Figure 3. The finite element model.
4 NUMERICAL MODEL 4.1 Geometry and grid model Static wind action of Dazhi River Bridge was investigated under attack angles of −5◦ , −3◦ , 0◦ , 3◦ and 5◦ . The fluid region is a cube of 1750 m long, 1000 m wide and 600 m high. The model of the bridge is located at one third of the fluid region. The full size geometry model is shown in Figure 2, which includes the guardrail and pier. Unstructured meshes are adopted at the center of the bridge, others are structured meshes. The finite element meshes are shown in Figure 3 with 928949 nodes and 2430328 elements in total. 4.2 Boundary conditions The wind speed is served as the boundary condition at the inflow surface. The wind speed profile on the boundary is described by the following equation:
where z0 and u0 are the reference height (10 m) and the corresponding wind speed (33.8 m/s), z and u are the actual height and the corresponding wind speed, α is the terrain roughness factor, which is 0.16 for landform type of Class B at the bridge site. The turbulence intensity is determined by the turbulence parameters at the inflow surface. However, the distribution of turbulence intensity is not ruled in Chinese load code (MOC, 2006), so the correlation formula is taken from AIJ recommendations for loads on building issued by Architectural Institute of Japan (AIJ, 1995). Due to the nearly total development of the effluent, pressure is adopted as the boundary condition at the effluent surface. The second order discrete format (Li et al., 2008) is adopted to solve the problem, while the convergence criterion is defined by the root-mean square residual, which is 10−4 . 1177
Figure 4.
Distribution of velocity vectors.
Figure 5.
Distribution of pressure coefficients.
5 ANALYSIS OF RESULTS 5.1 Results The distributions of wind speed at the middle pier and the middle-span are shown in Figure 4. When the wind passes through the bridge, in the windward region, obvious vortex appears at the highest point of the fin. Meanwhile, flow separation occurs at the top of the fin, which causes the formation of vortex in the leeward region too. As for the middle-span, the vortex in the windward region is smaller than that of the cross section at the pier. Vortex also appears in the leeward region. However, it is extremely smaller than that at the top of the fin. Pressure coefficients in the windward and leeward region of the fin under attack angle of 0◦ are shown in Figure 5. The pressure in the windward region is positive, while it is negative in the leeward region. The greatest positive pressure of the fin is near the middle pier, and the corresponding pressure coefficient reaches 0.95. The smallest positive pressure is near the side pier. The greatest negative pressure is near the middle span, and the corresponding pressure coefficient is around −0.9. The smallest positive pressure is near the side pier, and the corresponding pressure coefficient is around −0.4. 1178
Figure 6. The variation of static force coefficients with attack angles. Table 1. Comparison of the static force coefficients. Items
CD
CL
CM
Fin-back bridge Normal bridge Ratio
3.805 1.726 2.20
−0.462 −0.506 0.91
0.020 0.132 0.15
The variation of static force coefficients with attack angles is shown in Figure 6. When attack angle increases from −5◦ to 5◦ , the drag coefficient and the lift coefficient decreases gradually from 4.23 to 3.63 and from −0.74 to −0.26, respectively, while the pitching moment coefficient varies gradually from −0.06 to 0.11. Under attack angle of −5◦ , the effective windward area is the largest, so that the drag coefficient reaches the maximum. With the increase of the attack angle, the windward area decreases, so that the drag coefficient decreases. The drag coefficient under attack angle of −5◦ is 1.17 times as that under attack angle of 5◦ . The lift coefficients under five attack angles are all negative. The lift coefficient under attack angle of −5◦ is 2.85 times as that of attack angle of 5◦ . The pitching moment coefficient is negative under attack angles of −5◦ and −3◦ , while it is positive under other attack angles. 5.2 Comparison between the bridge with the fin and without the fin The static force coefficients of the bridge and normal bridge (without the fin) under attack angle of 0◦ are listed in Table 1. The fin has significant influence on the drag coefficient and the pitching 1179
moment coefficient, while the effect of the fin on the lift coefficient is very small. The drag coefficient of this bridge is 2.2 times as that of normal bridge. The main reason is that the windward area is increased by the fin. The pitching moment coefficient of this bridge is 0.15 times as that of normal bridge. It is explained by the fact that the height of the cross section of this bridge is close to the width while the cross section of normal bridge is flat, so the pitching moment of this bridge is smaller. However, the fin has very little effect on the lift coefficient because the fin is perpendicular to the wind direction under attack angle of 0◦ . 6 CONCLUSIONS The wind field of the full size model of Dazhi River Bridge was simulated by FLUENT. From the results, the conclusions are drawn as follows: (1) The velocity vector of the cross section varies along the bridge. The greatest positive pressure is in the windward region of the fin, and the corresponding pressure coefficient reaches 0.95. The greatest negative pressure is in the leeward region of the fin, and the corresponding pressure coefficient reaches −0.9. (2) When attack angle increases from −5◦ to 5◦ , the static force coefficients vary monotonically. The drag coefficient and lift coefficient under attack angle of −5◦ is 1.17 times and 2.85 times as that under attack angle of 5◦ , while the pitching moment coefficient varies from −0.06 to 0.11. (3) The fin increases the windward area and makes the height of the cross section to be close to the width, so that the drag coefficient of this bridge is larger than that of normal bridge, while the pitching moment coefficient of this bridge is smaller. The drag coefficient and the pitching moment coefficient of this bridge are 2.2 times and 0.15 times as that of normal bridge, respectively. REFERENCES AIJ (Architectural Institute of Japan) 1995 AIJ recommendations for loads on building. Tokyo: Architectural Institute of Japan. Cheng, Z. 2005 Wind engineering of bridge, Beijing: China Communications Press. He, X., Fang, W., Li, X. & Li, J. 2011 Numerical simulation of static force coefficient on the deck beam of cable-stayed bridge. Transportation Science & Technology 2011(2): 1–4. Kong, J., Jing, M., Yang, N. & Zhong, H. 2008 Theoretical derivation of figure coefficient of wind load in wind tunnel test. West-China Exploration Engineering 20(2): 186–187. Li, X., Zhou, X. & Gu, M. 2008 Study on snow loads on the roof of Beijing south station. Building Structure 38(5): 109–112. Li, X., Zhou, X., Gu, M. & Uematsu, Y. 2010 Numerical simulation on snow drifting around a cube model. Journal of Tongji University(Natural Science) 38(8): 1135–1140. MOC (Ministry of Construction of the People’s Republic of China) 2006 GB5009-2001(2006) Load code for the design of building structures. Beijing: China Architecture and Building Press. Qu, W. & Liu, L. (2007) CFD-based numerical research in the identifying of tri-component force coefficient of bridge. Journal of Wuhan University of Technology 29(7): 85–88. Tan, H. & Cheng, Z. (2009) Application of CFD in calculating static coefficients of bridge section. Engineering Mechanics 26(11): 68–72. Xiang, H. & Cheng, A. (2003) Recent advances in research on aerodynamics of extra long-span bridges. China civil engineering journal 36(4): 1–8.
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Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
Modeling bridge construction phasing by the balanced cantilever method – “Comparison between predicted and real camber values” L.G. Castro Martifer Structural Design Division, Martifer SGPS, Oliveira de Frades, Portugal
R. Bastos Adão da Fonseca Consultores, Porto, Portugal
R.C. Barros FEUP, Dept of Civil Engineering, Structural Division, Porto, Portugal
ABSTRACT: The study focuses on the computer modelling of the bridge “Nova Ponte sobre o Rio Lima” executed by successive advances using the balanced-cantilever method, taking into account the visco-elastic behaviour of the concrete. Such study is divided in two different parts. First, to obtain the camber values for each stage of the bridge construction so that at the end of 25 years the elevation topographic references of the pavement would be the ones predicted by the design engineer. Second, to compare these elevation displacements with the real ones obtained by topographical survey and interpret the deviations occurred.
1 INTRODUCTION The structural design of a pre-stressed concrete bridge executed by the balanced cantilever method is a well-known procedure by the world of civil engineering although it sustains a great deal of complexity, and demands a great care on its modelling and construction analysis. One main aspect of this type of bridges regards the “geometry control”, which provides an accurate construction data so that at the end of the 25 years the topographical elevation of the bridge is the one predicted by the design engineer. So this study focuses on the computer modelling of the bridge “Nova Ponte sobre o Rio Lima” executed by successive advances using the balanced-cantilever method, taking into account the viscoelastic behavior of the concrete. Such study is divided in two different parts: the first part, to obtain the accumulated camber for each stage of the bridge construction so that at the end of 25 years the elevation topographic references of the pavement would be the ones predicted by the design engineer; the second part, to compare these elevation displacements with the real ones obtained by topographical survey and interpret the deviations occurred.
2 STATE OF THE ART Bridges executed by successive advances or successive consoles, with spans between 100 m to 250 m, are used as a structural solution mainly in cases of necessity to cross rugged terrain at high altitude, where the formwork to the ground is no longer an economically viable construction method. The bridge is built by segments, small section elements of the bridge deck, usually in the order of 3 to 6 m, disregarding any formwork with stand supported on the ground. The segments are being pre-stressed as they are built. The connection between consoles coming from consecutive 1181
piers is done via a “lock segment” with a length in the order of 2 m to 3 m in case of streams spans (Reis, 2001). Balanced cantilever constructed bridges begin their life as statically determinate cantilevers, and after connection between parts change their static definition creating a redundant structure. Considering a pair of cantilevers the action of the self-weight creates a downward deflection and rotation at the end of the console. The rear central link between the two consoles at mid span, can transform the two independent structures in hiperstatical structure despite its deformation continues to increase due to creep. Creep phenomena will redistribute efforts, reducing stress on the pillars and increasing them at mid span. (Benaim, 2008) In current continuous structures, extreme spans should be between 0.25 and 0.75 of the intermediate span, in order to allow an appropriate distribution of bending moments. It is usual to assume that the ratio (section height/length of the span) on bridges with variable height can vary between the values h/l = 1/20 in support h/l = 1/50 at mid span. However, these values may vary for several reasons, one being that the resistant section comes from the relationship between the height, the area of the bottom slab and the concrete strength (Manterola, 2000). On bridge phasing analysis, time dependent concrete properties are fundamental aspects to be taken into account so viscoelastic behavior must be well predicted. There are three time-dependent properties and phenomena of concrete that must be considered in order to obtain a structural behavior as perfect as possible which are of very special importance in terms of serviceability limit states and during construction: (1) Hardening, due to elastic modulus variation over time; To estimate the elastic modulus variation over time Eurocode 2 – EC2 (CEN, 2005) uses the value βc,t , at any time “t” with the following equation (with t − tb , in days):
In this equation there are two parameters {s, r} whose values, always positive, can be chosen from the following sets s = {0.38, 0.25, 0.20} to cements {SH, NH/HR, HR + HS} respectively (meaning: SH = slow hardening, NH/RH = normal or rapid hardening, RH + HS = rapid hardening with high strength) and r = 0.5 is taken to be constant. (2) Shrinkage, due to variation of concrete element volume;
The formulation is statistical in nature and divides shrinkage in two elements, the autogenous shrinkage and drying shrinkage. The first is an internal phenomenon but without weight loss with a reduction of volume and the second is the result of water loss. Each one is a product of factors and each formulation of the factors will describe the evolution of this portion in time. (3) Creep, due to deformation in time when subjected to constant stress states. The ϕcc,t,ti creep coefficient, t, (the date t for the load you date) will be the product of all the factors below:
The first creep factor with L subscribed to loading, βccL , reflects the fact that the later application of the load to the date of concreting, the lower the creep phenomena. The second creep factor, with F subscribed to resistance, βccF, reflects the fact that the higher the concrete strength, the lower the creep phenomena. The third factor βccS reflects the fact that the larger the installed tension, the greater the creep phenomena. The fourth factor ϕcc reflects influence of the ambience humidity 1182
Figure 1.
Longitudinal span distribution.
and section properties, in particular the perimeter in contact with the atmosphere, on the creep phenomena. The fifth and final factor βccT reflects the evolution of creep in time.
3 THE BRIDGE NOVA PONTE SOBRE O RIO LIMA The total length of the bridge is 420 meters, divided as shown in Figure 1. The bridge deck is a single-celled box-girder with variable height between 8.0 m on the piers and 3.50 m in the middle span, intermediate supports and extreme spans (Castro, 2014). The construction phasing had the following steps: 1. The constructions begun with the execution of the foundations then start of the piers and their elevations. Simultaneously begun the abutments construction; 2. Upon completion of the piers begun the construction of segments 0 of piers P1 and P4. Then was carried out the assembly of the steel formwork equipment to proceed with the concreting of the following 5 symmetrical segments; 3. After the concreting of the segment 5 the spans were executed near to the abutments with the aid of ground formwork and serving segment 5 as top formwork. Then the lower continuity prestressing of extreme spans was pulled, and the formwork to the ground was removed as the deck goes up; 4. After the execution of the extreme spans, the steel equipment was disassembled and assembled in the central piers: pier P1 equipment mounted on pier P3; pier P4 equipment mounted on pier P2; 5. After the concreting of the 13 segments, the central segments of intermediate spans were executed, the first being the P3-P4 and then the P1-P2. The closing of the deck was obtained with the concrete casting of the central segment with P2 pier formwork; 6. The final stage of the bridge implementation is the execution of the curbs, sidewalks, railings, security guards, paving and expansion joints.
4 GEOMETRY CONTROL The “geometry control” consists on: (1) calculation of the accumulated displacements occurred on the structure until each construction stage considered; (2) definition, for each of those stages, of the camber to give at each concrete element executed, so that at the end of 25 years of usage the elevation topographic references of the pavement would be the ones predicted by the design engineer. The design of structures where the construction stage is a major constraint requires the usage of calculation software that has this modelling functionality. That is, it should allow the addition of internal forces between different calculation models, since there is the addition or removal of finite element models of different stages. Each stage corresponds to a separate calculation model by adding or removing finite elements, forcing the calculation software to assign and add forces and deformations to the same element in different models (Castro, 2014). 1183
So a numerical finite element model was created, using RM Bridge (2011) software, attempting to reflect the geometry, materials and constructive phasing more accurately as possible; the accumulated camber for each stage of the construction were then obtained. The numerical model was created as follows: • Deck – 103 bar elements, the B101 B203; • Piers – 10 bar each finite element (P1 – B5101 B5110 to; P2 – B5201 B5210 to P3 – B5301 B5310 to P3 – B5401 B5410 to) Each bar element corresponds to a segment with variable section along the development of the bridge. The connection between P2 and P3 piers and the deck is simulated by a monolithic linear spring elements of high stiffness in 6 degrees of freedom. The actions considered were: the self-weight of the structure, the self-weight of the metallic equipment formwork and its positioning during construction stages, the pre-stressing action; and the time dependent action-effects on the properties of concrete (creep, shrinkage, maturation) and on pre-stressing steel (relaxation). The analysis of the time-dependent effects of concrete C40/50 – since it was not possible to obtain the results of shrinkage tests, creep and aging of concrete – was made using laboratory results obtained from a concrete of the same strength class but used in another bridge. The only data provided by the site manager concerns the compression tests on cube samples of the used work concrete, which allowed to estimate the modulus of elasticity. As an example of the data used on the studies performed, in the following Figure 2 it can be seen the creep coefficient curves obtained in the tests and the Eurocode 2 – EC2 (CEN, 2005) curves for slow hardening (SH) cements and for rapid hardening with high strength (HR+HS) for three ages of 3,7 and 28 days. As already mentioned above, geometry control reflects in the application of camber to the structure in the concreting stage of each element or each segments. The value of the camber to be applied to the structure at each point is obtained by the sum of their displacements when subjected to planned actions during construction and other permanent loads in the exploration stage. The form traveler is put up or down in order to pursue the camber. The “geometry control” is immutable and cannot be adjusted iteratively during construction. The idea that as the bridge is being built the camber values are adjusted is wrong, since the final result respects the entire construction staging, which means that a slight variation of time intervals
Figure 2.
Creep chart study.
1184
provided, changes the staging form beginning to end, forcing camber values which in the middle of the construction is not possible to perform (Castro, 2014). This enhances the continuous need for a high quality construction and supervision. 5 RESULTS Some of the results obtained for “Nova Ponte sobre Rio Lima” are shown in Figures 3 to 7 that show the value of the displacements accumulated during the work construction. At each construction stage, only three scenarios were assumed as possible: i. Scenario 1: concrete of this stage – corresponds to the concreting of the element; ii. Scenario 2: with pre-stressing – application of pre-stressing of this stage; iii. Scenario 3: with mounting equipment – corresponds to the assemblage of steel formwork advances. In order to clarify the phasing and the expected behavior of the deck, the three graphs below in Figure 3 relate to the execution stage of the segment 4 pier P4. The graphs should be understood as follows: (i) After casting the segment 4, the extremity has a displacement of about 3.8 mm downward; (ii) Then with the pre-stressing the accumulated displacement (i.e. the sum of its own weight with the action of the pre-stressing), is approximately 1.5 mm downward meaning that the deck rises 2.3 mm in this stage; (iii) After the advance of the steel formwork device, has a cumulative displacement at the end of the segment 4 of about 2.25 mm, which means that the deck lowers 0.75 mm due to this operation. In these graphs and figures, the values on the Y axis are displayed in meters. The references in the X axis, PxPiNN, indicate the segment NN counting from pier x to pier i. Figures 4 to 6 show accumulated displacements for stages (23, 30 and 32) of the construction. So the final camber to apply on the deck are of the opposite sign to the stage 32 after 25 years exploration, and present the following configuration and numerical values (Fig. 7).
Figure 3.
Phasing example of segment 4 pier P4.
Figure 4.
Stage 23 – Pier P2-segment 8 and pier P3-segment 9.
1185
Figure 5.
Stage 30 – Close segment P1-P2.
Figure 6.
Stage 32 – Execution of the final permanent load and at 25 years.
Figure 7.
Calculated camber to apply during construction.
6 COMPARISON BETWEEN REAL AND PREDICTED At the time of application of the camber to some concrete element, the elevation data of the bridge was taken topographically and compared with the position predicted for that specific moment or phase of the construction stage. In Figures 8 and 9 are presented, for each segment, the camber expected (black line) and the actual displacements observed (red line), under scenario 1. At the stage when the segment is executed, the camber provided in the calculation was imposed. From that moment onwards the segment was deformed in the course of the construction phasing, and the displacement values still present were recorded. The second instant of surveying occurs after the application of the pre-stress (scenario 2). With the application of the top pre-stress, the deck is subject to a positive moment, forcing it to have an upward vertical displacement. In Figure 10 are compared the values of the vertical displacement previously predicted and the actual displacements observed. That is, under scenario 2 in the prestressing application after casting of each segment, the deck has an upward movement provided by the calculation, as well as a really observed value. One can interpret Figure 10 as follows: for example, a segment 8 on the pier P2 in P1-P2 with the application of pre-stressing was predicted an upward movement on the order of 4 mm, but it was found that it was about 6 mm. 1186
Figure 8.
Camber for P2 and P3 – Segment 11.
Figure 9.
Camber for P2 and P3 – Segment 13 (scenario 1).
Figure 10.
Camber P2 and P3 – Segment 13.
It should be noted that the deviations found between the predicted values and the actual observed values may have several causes, namely: (1) Survey error and topographic marking associated with the error equipment; (2) Concreting error, with irregular face of the upper surface; (3) Different environment temperature during construction; (4) Different planning of the construction in the sequence of concreting due to unforeseen stops; (5) Unexpected changes in the construction phasing; (6) Overload not predicted on the deck; (7) Error in the implementation of geometry control; (8) Improper adjustment of the curves of concrete and steel time-dependent phenomena; (9) Precision error of equipment and mark-up elements used; among other sources. Figure 11 represents the deflections increase after specific periods of time (t), when the time lapses for 5, 10 and 15 days between the concreting of a certain segment and the concreting of the following segment. The offsets are only due to time-dependent effects (Castro, 2014). 7 CONCLUSIONS The geometry control of a bridge executed by the balanced-cantilever method is a technique of paramount importance. For each construction stage, the calculation of the camber to be imposed 1187
Figure 11.
Displacements due to visco-elastic behaviour of the concrete – P2, Segment 13.
to the concrete element being executed is a fundamental data, so that the bridge can be raised with all the necessary comfort and safety for its future users. From the results obtained it can be concluded that: The “geometry control” proved appropriate because there is a real structural behavior similar to that specified; The evolution of the displacements in time follows the expected trend with similar numerical values; The fact that it was not possible to obtain laboratory results of shrinkage, creep and aging of concrete may have influenced the analysis which surely lead to further approximate results; Errors in the survey or marking of the deck, work delays, ambient temperature and different sun exposure on consecutive days are aspects that influence decisively the deformation of the deck and therefore mislead the data. REFERENCES Benaim, R., 2008. The Design of Prestressed Concrete Bridges: concepts and principles. New York. Taylor & Francis. Castro, L.A.G., 2014. Modelação de Faseamento Construtivo da Nova Ponte sobre o Rio Lima e Comparação do Controlo de Geometria Teórico com o Real. Dissertação de Mestrado Integrado, FEUP, Porto. CEN, 2005. Eurocode 2 – Design of Concrete Structures. Part 2: Concrete bridges – Design and detailing rules. Brussels. CEN, 2010. Eurocode 2 – Design of concrete structures. Part 1-1: General rules and rules for buildings. Brussels. Manterola, J.. 2000. Puentes. Escuela Técnica Superior de Ingenieros de Caminos, Canales y Puertos de Madrid, Madrid. Reis, A., 2001. Folhas da Disciplina de Pontes. Instituto Superior Técnico, Lisboa. RM Bridge, 2011. Professional Engineering Software for Bridges of all Types, Bentley Systems, Wien, Austria.
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Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
Fatigue analysis induced by vibrations in stay-cables subjected to along wind turbulence component I. Failla, A. Recupero & G. Ricciardi DICIEAMA, Università di Messina, Messina, Italy
F. Saitta ENEA, Centro Ricerche Casaccia, Roma, Italy
ABSTRACT: Cable-stayed bridges are very well optimized structural systems with light stiffening girder, supported by stays with an efficient transfer of forces to the towers. In recent years, these favorable structural characteristics, as well as aesthetic qualities, contributed to the increase of the length of spans up to 1000 m. The increase of span length brought to the need of investigating new problems: one of these is the out-of-plane cables vibration, which can generate fatigue problems near the anchorages, due to the rise of additional stresses. In this study, the fatigue behavior of stay cables under wind action is investigated, taking into account only the along-wind component of wind velocity. Particularly the fatigue stress analysis at the end zones of bridge stays is considered. The effects induced by the geometric non-linearity of the structure are taken into account; by using the Hamilton’s principle, the motion equations are obtained through an original continuous approach. The problem is carried out in the time domain and the stay is idealized as a bending stiffened cable analyzed in the large displacements field.
1 INTRODUCTION 1.1 Structural system of cable stayed bridge Because of their economic advantage on the middle-large spans and of the aesthetic qualities, cable-stayed bridges have been more and more widely used in the last few decades. New erection technologies (Arici et al., 2011; Granata et al. 2012, Recupero & Granata, 2013), improvement of mechanical properties of the materials used for their construction, allowed this kind of structures to reach span lengths up to 1000 m (Itô, 1996; Virlogeux, 1999). During the last 30 years, starting from the main span of the Saint Nazaire Bridge (France), 404 m long, the record of 1104 m was reached in the Russky Island Bridge (Russian Federation). Nowadays the stay-cable system can be also used in a provisional way to build arch bridges without the use of centerings, supports or temporary towers, for assembling concrete arches, with the classical suspended cantilever method (Granata et al., 2013a) or with the alternative methodology of lattice cantilevers (Granata et al., 2013b). Generally, stay-cables have very low structural damping ratio and are prone to dynamical actions (Caracoglia & Zuo, 2009) and large amplitude oscillations that are in the order of 1–2 m, as it has observed when cables are subjected to wind action in combination with rain action. The exciting sources are various, and they lead cables to large amplitude vibrations in the across wind direction (Gimsing, 1983; Matsumoto et al., 1995; Flamand, 1995; Matsumoto et al., 1998); therefore, the cable oscillates especially in the vertical plane containing its static profile: we will indicate this kind of motion as in-plane, according to Irvine (1981). In many cases, it has been observed fatigue damage nearby cables anchorage. We point out the possibility that, for long stay-cables, such as those used in the most recent realizations, also the out-of plane motion due to along wind action will become important about the evaluation of fatigue damage (Cluni et al., 2007). The fatigue 1189
analysis has been conducted by modelling the cable as a taut rope with not-negligible bending stiffness; the problem has been solved by random analysis approach in the frequency domain and computing the damage through probabilistic criteria. In this paper, a non-linear model of the bending stiffened cable is adopted (Recupero & Ricciardi, 2004), obtained by a continuous approach through the introduction of some simplification, with the aim to derive more precise results, taking into account the cable sag in the initial static configuration; in fact, in the non-linear case, in-plane and out-of plane motions are coupled. Neglecting the effect of turbulence components in the orthogonal direction to the wind one, the instantaneous wind velocity in the along-wind direction can be expressed by two contributions: mean value and fluctuation. The first contribution is modeled as a function of height above ground, while wind fluctuation can be modeled as a multi-variate bidimensional stationary zero-mean Gaussian random field (Simiu & Scanlan, 1996). Then its definition requires tools that are proper of stochastic analysis, such as cross-power spectral density function (CPSD), often expressed in terms of power spectral density function (PSD) and coherence function. The analysis of non-linear systems is usually based on the Monte Carlo simulation of wind loads but this requires an enormous computational effort. However, literature has recently proposed a series of transformation techniques of the wind load very similar to those generally used for structures. They allow the designer to represent the original process as a linear combination of mono-variate uncorrelated stochastic processes (Van Trees, 1968; Ghanem & Spanos, 1991); recent application of such techniques to one-dimensional continuous structures can be found in Carassale & Solari (2002), known as proper orthogonal decomposition (POD). It expresses the multi-variate bidimensional stationary random process by a series of fully coherent components, which can be seen as the modes of the wind process. In the same way of classic modal analysis of structures, a limited number of modes are usually sufficient to approximate the loading process. In the case of non-linear model of the structure, unfortunately, the solution cannot be searched in frequency domain, and the analysis has to be carried out in the time domain; in this work the fluctuating part of each generalized component of wind load is modeled as the output of an appropriate linear digital filter. The selection of the filter equations is suggested by the particular shape of the eigenvalues of the CPSD function of wind velocity fluctuation. In this way, simply, the Gaussian Stochastic Linearization (GSL) provide moment differential equations of the response process. If the PSD model is that formulated by Davenport, all the eigenvalues, obtained by POD, assume a nil value for ω = 0 and therefore generalized loads can be modeled as the velocity of the response of second order linear filters subjected to white noise of suitable intensity (Ricciardi & Sofi, 2003). A stochastic analysis can be performed, determining the Liapunov-type non-linear differential equations in terms of statistic moments. If the system is subjected to Gaussian white noises, by making the assumption of weak non-linearity, the averages appearing in the stochastic differential problem are evaluated by considering a Gaussian probability density function; the adopted technique is the well-known afore mentioned Stochastic Gaussian Linearization (GSL).
2 WIND MODEL AND STRUCTURAL DYNAMIC MODEL 2.1 General structural model Stay cables are composed of wires or strands, generally parallel, encased inside cement grout or flexible epoxy grout in HDPE stay pipe. For a global FEM model of the bridge structure, bending stiffness of stay cables can be neglected, but in the study of the local effects, this hypothesis becomes inconsistent and in general too much restrictive. Then, in the generation of the non-linear stay model, it is necessary to take into account the bending stiffness of the stay cable. In many studies and researches reported in the literature, non-linear models of the stay cable (without bending stiffness) are considered (Irvine, 1981; Luongo et al., 1984), or (Recupero & Ricciardi, 2004) with bending stiffness of cable. In this last paper a complete mathematical 1190
formulation is carried on and for details, it is useful to refer to it. In the following, this formulation is integrally used and it will not be transcribed here. 2.2 General wind model and proper orthogonal decomposition (POD) For the sake of simplicity, let us assume that the variation on time t of wind direction is negligible. The instantaneous wind velocity W (x, t) can be expressed as a sum of a macro-meteorological component W (x), defined as the mean value upon an opportune time range, expressed in terms of ˜ (x, t), defined as height and site topological properties, and a micro-meteorological component W the wind turbulence:
˜ (x, t) along the z axis is a zero mean Gaussian stationary random The turbulent component W process described in the frequency domain by the Cross Power Spectral Density (CPSD) function. For civil engineering purpose, the CPSD of turbulent components can be evaluated by one of the many relations reported in literature (Simiu & Scanlan, 1996). Characterization of CPSD function between different points in the space is performed by the definition of the coherence function. If W (h) is the mean value of wind velocity at medium level h, for the coherence function the following expression is assumed:
where Cx is the exponential decay coefficient. Then the CPSD function is given by:
In the previous equations, the imaginary part (q-spectrum) has been neglected, as custom in wind engineering where the turbulence is considered as a homogeneous and isotropic phenomenon. A further simplification has been done by using a unique height h; in this case, it has been settled as the mean height. Independently from the linear or non-linear behavior of the structure, the above defined wind model is nevertheless onerous. POD can be used as a tool in order to reduce computational effort (Van Trees, 1968; Ghanem & Spanos, 1991; Carassale & Solari, 2002). In this way, a multidimensional variate random process can be expressed through a series of fully coherent uncorrelated components: the process modes. In particular the eigenfunctions and the eigenvalues used in the present work are those proposed in (Carassale & Solari, 2002), different to those proposed in (Ghanem & Spanos, 1991) and used by the authors in a previous work (Recupero & Ricciardi, 2002). For details, the reader could see the previous and another paper of the authors (Recupero & Ricciardi, 2004) 2.3 Determination of wind forces acting upon stay cables If a quasi-static approach is adopted (Simiu & Scanlan, 1996), denoting with fz (x, t) the force per unit length acting on a single stay cable in the z-direction, with fy (x, t) the same quantity acting in the y-direction, with w and v the cable displacements in z and in y directions, respectively, and by taking into account the hypothesis of turbulence in longitudinal direction only, then the wind turbulence load in z direction can be written as follows:
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where ρ is the air density, cD is the drag coefficient and b denotes the diameter of cable cross-section. ˜ 2 (x, t) and w(x, ˜ (x, t), ˙ t) · W Neglecting non-linear aerodynamic and aeroelastic terms w˙ 2 (x, t), W Equation (4) can be written:
where the last term is related to the aerodynamic damping, with coefficient: which supplies a contribution to the total damping: µ ˆ z = µz + µaz . It has been shown (Pasca et al., 1998) that, in the y direction aerodynamic damping is:
and the force acting along y can be expressed in the following way:
The expression of the generalized forces, obtained by the application of the Galerkin method is reported in (Ricciardi & Sofi, 2003). 2.4 Modeling aerodynamic loads as filtered white noises It has been previously said that the PSD of the generalized force are proportional to the k-th eigenvalue of the Power Spectral Density of turbulent component of wind velocity. In the case in which PSD model is the one proposed by Davenport, all the eigenvalues assume null value at ω = 0 and therefore the generalized force can be conveniently modeled as the output velocity of the response of a second order linear filter excited by a white noise of suitable intensity. The expression of the linear differential equation for a second order filter is: where the parameters ζFk , ωFk , and the PSD of the white noise Sk can be determined by the equivalence between the PSD of the k-th generalized force SF˜ zk (ω) and the PSD of the output velocity of the filter response, SY˙k (ω), given by:
The parameter of the filter and the white noise power spectrum can be determined by imposing the equivalence of the PSD at peak frequency value and for another opportune frequency. This procedure of equivalence can be helpful to write the differential equations that govern the moments of the structural response by taking full advantage of the frameworks of Itô stochastic differential calculus. 3 RESPONSE STATISTICS AND FATIGUE DAMAGE EVALUATION 3.1 Fatigue damage model The fatigue damage model, as used here, is the same of a previous paper (Recupero & Ricciardi, 2004) in which some details can be found. As illustrated in literature (Sobczyk & Spencer, 1992; Repetto & Solari, 2001; Recupero & Ricciardi, 2002), in particular, total damage can be calculated by taking into account the occurrence of probability of the fluctuating stress processes linked with different values of the mean wind velocity. Therefore, for an appropriate evaluation of the damage, it is required to have information on the statistic of the stress process. 1192
3.2 The gaussian stochastic linearization (GSL) approach In this work, a Gaussian Stochastic Linearization approach for analysis of wind-excited cables was developed by starting from the Galerkin-reduced model. Differential equations of cables and filters can be rewritten by introducing the first derivatives of modal coordinates and filters as further variables:
where j = 1, . . . , nv , h = 1, . . . , nw & k = 1, . . . , nw ; nv , nw are the considered number of in-plane and out-plane modes, respectively; thus, the total number of equations is equal to n = 2nv + 4nw . For calculating the structural response in term of statistical moments, it is known that in the case of polynomial non-linearity, the differential equations of statistical moments constitutes an infinity hierarchy. Therefore, for solving this system, a closure technique is necessary. If a Gaussian probability density function (PDF) for the response is supposed, the technique is called “Gaussian Closure”. In this case, moment equations of first and second order are utilized. This method is particularly advantageous in weakly non-linear systems, for which the hypothesis of Gaussian response can be accepted. Differential equations, for moments of order k = k1 + · · · + kn , can be rewritten in the following form:
In stationary case, the previous equations become the following algebraic ones:
A subscript that is repeated in a given term is understood to imply summation over the range of the repeated subscript; this is the summation convention for index notation. The technique of “Gaussian closure” consists of supposing as null all the cumulants of order higher than second. The presence of non-linearity of cubic-type, in the differential equations governing the motion, leads to the evaluation of statistic moments of third and fourth order. They must be expressed in function of the moments of first and second order by the relationships that link moments and cumulants. In particular:
A numerical procedure for solving non-linear equations has to be applied. 4 AN APPLICATION In this paper, a stay cable inclined with an angle of 11.31◦ on the horizontal plan and with geometric and mechanical characteristics as shown in tab. 1, is studied. The wind turbulence has been defined by the PSD in the form suggested by Davenport, with the following parameters: shear velocity u∗ = 0.08vref , von Karman coefficient κ = 0.4, roughness 1193
Table 1. geometric and mechanical characteristics of stay cable. m (kg/m)
g (N/kg)
L (m)
H (MN)
E (GPa)
A (m2 )
I (m4 )
74.75
9.8
500
4.00
200
6.66839E–3
5.27877E–6
Figure 1. Power Spectral density of first modal force plotted in three dimensional way versus circular frequency and versus reference wind velocity.
length z0 = 0.01, decay coefficient Cx = 10, air density ρ = 1.225 kg/mc. For the reference velocity vref , 30 values, at intervals of 1 m/sec, in the range 0.5–29.5 m/sec have been considered. For characterization of the mean wind velocity, that is treated as a random variable, the same distribution function for vref , as used in (Recupero & Ricciardi, 2002; Repetto & Solari, 2001) is selected, assuming height of the middle point of the stay-cable equal to 70 m. The cable drag coefficient is Cd = 1.2. Taking into account the damping devices and natural structural damping of stay cable, the adopted damping ratio is 0.004. Simulation has been developed by taking into account the first three in-plane modes and the first three out-of-plane, and accordingly the first three modes of wind. In this case 3 filters have been defined for every wind reference mean velocity for a total of 90 filters. The system of non-linear equations, in virtue of the chosen number of modes, has a number of equations and unknown variables equal to 189. Figure 1 shows the variation of PSD of the first generalized force due to wind turbulence versus the reference wind velocity and versus the circular frequency. The simulation has been performed through the Shinozuka technique, by superposing harmonic components with increasing frequency and random phase. In order to compare the two simulations obtained with filtered white noise and with correct PSD, it has been used the same phase angles for the correspondent samples. Calculus has been carried out for every value of reference mean velocity and by considering the following two cases for structural behavior of stay cables: L linear (dashed line) and NL non-linear with Gaussian Stochastic Linearization (GSL) (continuous line). It has been also developed a MonteCarlo (MC) (continuous line with circles) simulation to obtain the “exact” results in comparison to NL and L cases. In Figures 2 the mean value, the standard deviation and the expected frequency of the stress process (MPa) in a point set far 5 cm from the cross section centre, near the anchorage zone, are reported. In this case difference between linear and non linear structural behaviors is about 7.13% for standard deviation at vref = 29.5 m/sec. It is worth noting that vref values up to 40 m/sec have been considered into next figures, to show the increasing of non-linear behavior for high wind velocity. Finally, Figures 3 report errors on stress in the investigated steel wire near cable anchorage, with respect to the “exact” value derived via Monte Carlo simulation. Results show as for high wind velocity, error reaches considerable values in the linear case, especially on standard deviation. However, for fatigue analysis only values up to 30 m/sec have been taken into account; higher values are in fact not realistic. In Table 2 damage on stay cable near the anchorage zone, calculated for a life of 100 years as reported in international codes on bridges, is reported. 1194
Figure 2. Mean value of stress process, Standard deviation of stress process and Expected frequency of stress process in a selected point of cable cross section for different values of reference wind velocity.
Figure 3.
Stress mean value and Stress standard deviation: error with respect to Monte Carlo simulation. Table 2. Damage on stay cable near the anchorage zone.
NL L
Parallel strands
Parallel wires
3 + 3 modes
3 + 3 modes
5.48% 6.11%
1.66% 1.93%
5 CONCLUSIONS Wind-excited vibrations of stay cables produce fluctuating stresses, in particular near the anchorages, which lead to damage accumulation. The objective of this work has been the estimation of the influence of non-linearities in the assessment of fatigue damage. It has been supposed that wind action was in the transverse direction with respect to the main span of bridge. For this purpose, a mathematical model, able to describe cable vibrations, that is more refined and capable of representing the local effects due to bending stresses near anchorages, has been adopted. Simulations performed on an actual stay cable lead to the following considerations: 1. differences found between non-linear and linear analysis in the reported example reach significant values only for high values of mean velocity of wind; 2. highest values of stresses concern the linear model; in fact, the non linear model shows a decrease of vibration amplitudes and consequently a reduction of bending stresses near the anchorages; 3. in the reported example, the fatigue damage is higher in the linear case, which, therefore, seems to be more advantageous for the sake of safety; 4. the analysis, developed in the present study, emphasizes the importance, for the evaluation of fatigue damage, of an adequate consideration of high-frequency modes, at least up to a certain number: in fact, even if large stress variations do not happen, the expected frequency of the stress process becomes higher, and this causes an increase of the number of stress cycles during the structural life. 1195
REFERENCES Arici M., Granata M.F., Recupero A. (2011). The influence of time-dependent phenomena in segmental construction of concrete cable-stayed bridges, Journal Bridge Structures – Assessment, Design & Construction 7(4): 125. Al Noury S.I. & Ali S.A. 1985. Large-amplitude vibrations of parabolic cables, J. Sound Vibr. 101: 451–462. Caracoglia L., Zuo D. 2009. Effectiveness of cable networks of various configurations in suppressing stay-cable vibration, Engineering Structures 31(12): 2851–2864. Carassale L. & Solari G. 2002. Wind modes for structural dynamics: a continuous approach, Probabilistic Engineering Mechanics 17: 157–166. Carassale L. 2005. POD-based filters for the representation of random loads on structures, Probabilistic Engineering Mechanics 20: 263–280. Cluni F., Gusella V., Ubertini F. (2007). A parametric investigation of wind-induced cable fatigue, Engineering Structures 29(11): 3094–3105. Di Paola M., Muscolino G., Sofi A. 2004. Monte Carlo simulation for the response analysis of long-span suspended cables under wind loads, Wind & Structures 7(4). Flamand O. 1995. Rain-wind induced vibration of cables, J. Wind Eng. & Ind. Aerodyn. 57: 353–362. Ghanem, R. & Spanos, P. D. 1991. Stochastic Finite Elements: A Spectral Approach, Springer-Verlag. Gimsing N.J. 1983. Cable Supported Bridges – Concept and Design, New York: John Wiley & Sons. Granata M.F., Margiotta P., Arici M., Recupero A. 2012. Construction stages of cable-stayed bridges with Composite deck, Journal Bridge Structures – Assessment, Design & Construction 8(3–4): 93–106. Granata, M., Margiotta, P., Recupero, A., Arici, M. 2013a. Concrete arch bridges built by lattice cantilevers, Structural Engineering and Mechanics 45(5): 703–722. Granata M.F., Margiotta P., Recupero A., Arici M. 2013b. Partial elastic scheme method in cantilever construction of concrete arch bridges. Journal of Bridge Engineering 18(7). Irvine H.M. 1981. Cable Structures, Cambridge: The MIT Press. Itô M. 1996. Cable-supported Steel Bridges: Design Problems and Solutions, Journal of Constructional Steel Research 39(1): 69–84. Luongo A., Rega G., Vestroni F. 1984. Planar non-linear free vibration of an elastic cable, Int. J. of Non-linear Mechanics 19: 39–52. Matsumoto M., Saitoh T., Kitazawa M., Shirato H., Nishizaki T. 1995. Response characteristics of rain-wind induced vibration of stay-cables of cable-stayed bridges, J. Wind Eng. & Ind. Aerodyn. 57: 323–333. Matsumoto M., Daito Y., Kanamura T., Shigemura Y., Sakuma S., Ishizaki H. 1998. Wind-induced vibration of cables of cable-stayed bridges, J. of Wind Eng. and Ind. Aerodyn. 74–76: 1015–1027. Pasca M., Vestroni F., Gattulli V., 1998. Active longitudinal control of wind induced oscillations of a suspended cable, Meccanica 33: 255–266. Recupero A., Granata M.F. 2013. A Mixed Approach for the Determination of Initial Cable Forces in CableStayed Bridges and the Effects of ParametersVariability, The Baltic Journal of Road and Bridge Engineering, In Press. Recupero A. & Ricciardi G. 2002. Fatigue effects on stays of cable stayed bridges: Study on protection devices (Fenomeni di Fatica negli Stralli da Ponte: Studio dei Dispositivi di Protezione), Atti del 7◦ Convegno Nazionale di Ingegneria del Vento, Milano. RecuperoA. & Ricciardi G. 2004. Non-linear behaviour of bridge stay-cables in the study of fatigue phenomena (Comportamento non lineare degli stralli da ponte nello studio dei fenomeni di fatica), Atti del 8˚ Convegno Nazionale di Ingegneria del Vento, Reggio Calabria. Repetto M.P. & Solari G. 2001. Dynamic alongwind fatigue of slender vertical structures, Engineering Structures 22: 1622–1633. Ricciardi G. & Sofi A. 2003. Moment equation method for the non-linear stochastic dynamic analysis of a wind-excited suspended cable, Fifth International Symposium on Cable Dynamics, Santa Margherita (Italy) 15–18 September 2003. Ricciardi G. 2007. A non-Gaussian stochastic linearization, Probabilistic Engineering Mechanics 22(1): 1–11. Simiu E. & Scanlan R.H. 1996. Wind Effects on Structures. Fundamentals and Applications to Design, New York: John Wiley & Sons. Sobczyk K. & Spencer B. F. Jr. 1992. Random Fatigue: From Data to Theory, Academic Press. Van Trees, H. I. 1968. Detection, estimation and modulation theory, Part I, Vol. I, New York: Wiley & Sons. Virlogeux M. 1999. Recent Evolution of Cable-Stayed Bridges, Engineering Structures 21: 737–75.
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Rational and practical method for camber control in bridges built by successive segments R.N. Oyamada & H. Ishitani University of São Paulo, Brazil
R.A. Oshiro & A.M.L. Cardoso Outec Engenharia, São Paulo, Brazil
ABSTRACT: In segmental bridges an appropriate camber diagram must be defined in construction (reference camber) to ensure the profile of the bridge. In addition, in curved bridges it is required to control the rotations in cross sections of the superstructure. Several factors can to affect the phenomenon as: loads, creep and shrinkage of concrete, relaxation on the tendons, maturity of concrete, the rate of construction and differential temperature. Usually the effect of differential temperature can be neglected by obtaining the reference points levels in the early hours of the day when the difference between the top and the bottom temperatures are at minimum. The elevations of the superstructure in each stage of construction are compared with the corresponding reference camber. This paper presents a rational and practical method to correct the differences between the reference and field elevations in each stage of construction.
1 INTRODUCTION The final profile of segmental constructed bridges is the result of the combination of several factors that must be taken account to ensure the design elevations and superelevation of the bridge. Firstly, the camber control diagram should consider the effects of dead loads, creep, shrinkage and relaxation on the tendons. In addition to the displacements due to dead load the long-term effects of creep and shrinkage of the concrete and its hardening during the construction have great influence in the determination of an accurate camber control diagram. A parametric study was made in a 71 meters span viaduct comparing four camber control diagrams: – Without creep, shrinkage and concrete hardening; – 7 days cycle (accounting long term effects) and – 14 days cycle (accounting long term effects). The results are plotted in Figure 1, in which it can be seen that the maximum camber is 4 times higher when long-term effects are considered and an increase of 10% in the maximum camber when comparing a 7 days cycle with a 14 days cycle. In site effects such as the weight and deformation of the formwork, live loads (construction) and differential temperature should be considered too. Normally the camber control procedures are done in the early hours of the day, time in wich the difference between the top face temperature and the the bottom is minimal and the sag due to this effect can be neglected (Mathivat, 1980). This paper presents a method to correct the differences detected and effect of differential temperature at the time of concreting the segments. 1197
Figure 1.
Parametric study.
Figure 2.
Segmental bridge composed of n segments.
2 CAMBER DETERMINATION Consider a segmental bridge composed of n segments: s1 , . . . , sn , as shown in Figure 2. According to Podolny (1982) & Oyamada (2004) we have: aij with {j = 1, . . . , n} the displacements corresponding to the construction of the segment si , isolated, where i indicates the segment and j the segments end; bij with {j = 1, . . . , n} the accumulated displacements at the end j until the segment si construction,
bnj with {j = 1, . . . , n} the displacements in each node at the end of construction. The camber for each node is obtained through the sum of all the displacements of this node since its concreting, i.e., the accumulated displacements at the last step considered (n) bnj , see equation 2.
Figure 3 shows an example in a hypothetical 10 segment cantilever. Where: – camber: camber diagram for each segment end; – disp. 8: displacements after s8 construction and the reference level of the formwork for the next segment s9 ; the deformation of the formwork and the form traveler due to the fresh concrete must be considered and incorporated to this reference level; – acum. s7 : accumulated displacements until segment s7 ; – acum. s9 : accumulated displacements until segment s9 . In addition, it is important to apply an adequate camber control on the rotations of the cross section in curved segmental bridges to compensate the torsional displacements due to the vertical loads. 1198
Figure 3.
Hypothetical cantilever.
2.1 Concreting reference levels determination Normally absolute coordinates are used in construction. Consider Figure 4 where: Pa & Pb correspond to the end of constructed segment (segment si ); Pc corresponds to the end of the formwork for the segment si+1 ; δr is the relative camber, accounting the form traveler deformation. It allows the correct definition of the concreting level at any time, since it is relative to the projection of the last segment; Curve 1: design camber reference diagram (Pa1 , Pb1 , Pc1 ); Curve 2: Pa2 and Pb2 are the levels of the segment si taken in site in the early hours of the day to minimize the effects of differential temperature. This curve is coincident to Curve 1 when there were no deviations until the segment si construction; Curve 3: Pa3 and Pb3 are the levels of the segment si taken immediately before the concreting of the segment i + 1, so that the effect of differential temperature is considered, and Pd is the level of the end of the segment to be concreted. Pc1 at Curve 1 is the reference level of the segment i + 1 end including the form traveler deformation. Pc2 at Curve 2 is obtained by the projection of the line (Pa2 , Pb2 ) at the vertical line at the end of the formwork. δr is the relative camber obtained by the difference between the levels Pc1 & Pc2 , which includes the design camber and the correction of any deviation detected to avoid cumulative errors. The correction can be applied in more than one step to smooth the discontinuity. Pc3 at Curve 3 is obtained by the projection of the line (Pa3 , Pb3 ) at the vertical line at the end of the formwork. Pd is the level of the end of the formwork right before the concreting of the segment i + 1 defined by the addition of δr to the Pc3 level. 2.2 Camber control proceeding 2.2.1 Control points surveying Topographical surveying must be done in the early hours of the morning just after each stage (concreting, tensioning of the tendons and advance of the form traveler). 1199
Figure 4.
Concreting reference levels determination.
Figure 5.
Control points location.
2.2.2 Form traveler deformation The displacements of the form traveler are estimated by using the technical specifications provided by the manufacturer. These values are used to determine the first two segments camber, and then calibrated to be used on the next segments. 2.2.3 Formwork levels determination The levels obtained after the advance of the form traveler are used to compute the relative camber value, to be used to correct the formwork placement level at any time of the day: – Estimation of the projected level Pc2 with field reference points survey; – Calculation of the absolute level for the formwork placement (Pc1 ) and; – Calculation of the relative camber δr . 2.2.4 Formwork placement and concreting of the segment With the relative camber, at any time of the day, before concreting each segment, the following procedure must be taken: – Topographical survey of the previous segments; – Calculation of the projected level of the segment to be concreted (Pc3 ); 1200
Figure 6. Viaduct over Castelo Branco Highway, São Paulo, Brazil.
Figure 7.
Matapi River Bridge.
– Determination of the formwork placement level Pd by adding to the projected level Pc3 the relative camber δr and; – Final adjust of the formwork levels and concreting of the segment. 3 APPLICATION The present method was applied with good results in many bridges. Recently on the viaduct over Castelo Branco Highway in São Paulo, with two 75 meter span, and a 105 meters curve radius, shown in Figure 6. Actually, the method is been applied at the bridge over Matapi River, at Amapá State, Brazil, see Figure 7, and had to account some time lag between the end of each span according to the construction schedule. A bridge over Tiete River is under design and the presented method is going to be applied, Figures 8 and 9. 4 CONCLUSIONS The presented method allows a good monitoring of the construction both by the constructor, as by the designer along the bridge construction. This constant feedback allows the designer to detect and correct deviations quickly and effectively. The accuracy depends on which factors are taken account on the model, and the quality of the topographic survey and technology control. As an example, if the material behaves different from the ones taken account on the model the result are not going to be good, and the designer or the constructor should adjust to fit it in the real conditions. 1201
Figure 8.
Bridge over Tiete River (under design).
Figure 9.
Bridge over Tiete River (under design) – Elevation.
A good way to avoid the majority mistakes done along the construction is the use of the relative camber presented in this paper since its value does not depend on a certain time of the day which gives more freedom to the constructor to concrete the segment. This method prevents discontinuity at the closing segment and guarantees the bridge design profile. REFERENCES Ishitani, H. & Oyamada, R.N., 1993. The camber diagram in prestressed concrete segmental bridges: The deflection calculation by finite element method, FIP Symposium, Kyoto, Japan. Mathivat, J., 1980. Construccion de puentes de hormigon pretensado por voladizos sucesivos. Editores Técnicos Asociados, S.A. Oyamada, R.N., 2004. Controle de flecha e adaptação por fluência em pontes construídas pelo método dos balanços sucessivos. São Paulo. Dr. Degree Tesis POLI-USP. Podolny, W. & Muller, J.M., 1982. Construction and design of prestresses concrete segmental bridges. John Wiley & Sons.
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Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
Application of nonlinear FEM to evaluate load bearing capacities – Capability and limitations S. Kattenstedt & R. Maurer TU Dortmund, Dortmund, Germany
ABSTRACT: The NLFEM is a promising method for the assessment of existing bridges as well as other concrete structures and can help to reveal load bearing capacities. Investigations at the TU Dortmund detect that the quality of NLFEM calculations for structures with shear failure highly depend on the implemented material models for concrete. Especially isotropic descriptions can lead to wrong results and do not reflect stresses realistically. The application of the NLFEM requires therefore special caution.
1 INTRODUCTION An evaluation of existing prestressed concrete bridges on the basis of a recalculation according to current standards, e.g. the Eurocode 2, reveals in some cases considerable deficits referring to the load capacity. This especially affects the shear capacity of the superstructures and leads to calculated deficits regarding the required shear reinforcement. A recalculation of prestressed concrete bridges in Germany reveals calculated deficits in 90% of 29 recalculated bridges built until 1966, in 55% of 38 recalculated bridges built between 1967–1973 and in 35% of 40 recalculated bridges built between 1974 and 1980 (Müller & Haardt, 2014). Increasing traffic loads contribute to the deficits of shear capacity, however, no shear failure of existing bridges in Germany has been reported so far. Furthermore, experiments on prestressed concrete beams also show higher load bearing capacities as expected according to common design methods. So the question arises whether the calculated deficits are realistic or the current design methods are too conservative. A promising method to identify existing load bearing capacities is the Nonlinear Finite Element Method, which allows a deeper insight into the stresses of a concrete structure and helps to analyze the structural behavior more in detail. Furthermore, the results, especially the stresses and their redistribution, can be compared with the analytical models to identify the main load bearing mechanisms. Although widely used, the application of the NLFEM on concrete structures requires special attention and knowledge, particularly because in the current codes, the application of nonlinear methods is only covered for bending problems. Usually in the scope of civil engineering a linear Finite Element Analysis is used to obtain internal forces and moments on the action side, which are compared afterwards within a sectional analysis with the internal forces and moments on the resistance side, calculated on the basis of the plasticity theory. With a nonlinear FEM Analysis it is possible to take into account the overall behavior of a concrete structure which leads to a global resistance instead of a sectional resistance. One of the most important issues using NLFEM is to ensure, that the nonlinear behavior of the model represents properly the physical behavior of the material. In the following, the potential and limitations of applying the Nonlinear Finite Element Method on concrete structures to reveal load bearing capacities in shear will be discussed. It turned out, that the quality of simulations highly depends on the implemented material models. Differences especially in terms of crack modeling may lead to considerable varying results. 1203
2 MATERIAL BEHAVIOUR OF CONCRETE Concrete is a highly complex composite with special characteristics, e.g. much smaller tensile strength than compressive strength, which leads to strain induced anisotropy due to cracking. For a Nonlinear Finite Element Analysis a mathematical description of the physical material behavior is needed, which requires the knowledge of the physical behavior of plain concrete as well as of reinforced concrete. A short overview of the most important characteristics is given. 2.1 Material behaviour of plain concrete The nonlinear uniaxial behavior of concrete under compression stresses is widely known. But failure in bending, and especially in shear is more likely caused by cracking due to tensile stresses. So adequate modelling of cracks is one of the main issues in the nonlinear analyses of concrete, especially regarding shear failure. During the cracking process, the strains localize in a single region, the cracking process zone. At the onset of a cracking stress transfer across the crack is still possible. Due to the increase of deformation, former microcracks merge into a macrocrack without any stress transfer capacity in the end whereas the remaining regions of concrete stay undamaged, see Figure 1. The width of the process zone is almost independent of the size of the test specimen which can lead to nonobjective results if the FE model is based on stress-strain relations. For a biaxial stress states a widely used failure surface was detected by (Kupfer, 1973), see Figure 1as well. Here, the failure points of proportional stress paths on panels are represented. The huge difference between tensile strength compared to compressive strength is quite evident. Due to tensile stresses in the second direction the compressive strength is reduced. In three-dimensional stress states the most important issue is the dependency of failure on the hydrostatic pressure, similar to soils, but very different from the behavior of metals. This means that the load bearing capacity increases with increasing hydrostatic pressure. 2.2 Material behavior of reinforced concrete Besides the behavior of plain concrete, the reinforcement has also to be taken into account. Due to the reinforcement several phenomena have to be considered: – bond slip: stress transfer/interaction between reinforcement and plain concrete, – tension stiffening (the increase in stiffness of a cracked member due to the development of tensile stresses in the concrete between the cracks), – dowel action,
Figure 1. Uniaxial behavior of plain concrete under tensile stress [from (Hofstetter, 1995)] and biaxial failure points for panels with proportional stress paths (Kupfer, 1973).
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– aggregate interlock, which is especially important for shear. In this model, based on investigations of (Walraven, 1981), it is suggested that the open crack transmits a certain amount of shear stress via the crack surfaces. In contrast to the former approaches which deal separately with single phenomena, a widely used analytical resp. semi-empirical model used for reinforced concrete was developed by (Vecchio & Collins, 1986) called Modified Compression Field Theory (MCFT). The theory was validated with experiments on 30 reinforced concrete panels which were performed for a better understanding of the behavior of reinforced concrete. Several parameters, such as reinforcement ratio and stress states (combinations of compressive and shear stresses as well as pure shear stress and combinations of compressive and tensile stresses) were varied. It was evident, that with additional shear or tensile stresses in one direction the compressive strength in the other direction decreases. Finally this leads to equations, which determine the resulting load bearing capacity in compression depending on the strain state in the other direction. 3 MATERIAL MODELS AND IMPLEMENTATION FOR COMPUTATIONAL ANALYSIS Most descriptions of the material behavior of concrete used in commercial FEM software are based on the phenomenological approach of continuums mechanics, i.e. starting from observed phenomena and experimental results in mathematical models for the mechanical behavior of the material are formulated. In most cases, this phenomenological approach is equated with a macroscopic scale, although nowadays the methods of continuum mechanics are also used in the mesoscopic or microscopic scale. The material law for concrete should be able to describe both the hardening region after leaving the elastic range and the softening after the peak load. In general, there are several possibilities to describe this nonlinear behavior in the field of continuum mechanics. These material laws deal usually with three-dimensional stress states, although two-dimensional descriptions are possible as well. Common laws are based on nonlinear elasticity, plasticity or damage and have usually the character of stress-strain laws, either in total formulation or rate-based. 3.1 Description of cracking As already mentioned, the cracking behavior of concrete is very important, because it is accompanied with strain-induced anisotropy. In terms of mechanics a crack is considered as a discontinuity in displacement and strains. The different models can be classified by their approach describing this discontinuity as shown in Figure 2. The most common models in commercial programs deal with an approach without discontinuities; the discontinuity, i.e. the jump in the displacement due to a crack, is here distributed uniformly
Figure 2. schematic representation of material models regarding the cracking behavior.
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across the width of a band of a finite thickness h. In literature, sometimes all continuum models with this approach were denoted as smeared crack models, but this term is here reserved for a special class of models without discontinuities. An important issue for models without discontinuity is the transformation of the traction-separation law into a law that links the stress transmitted by the localization band to the average inelastic strain in that band. This is necessary, because usually material models based on continuum mechanics deal with stress-strain laws rather than with traction-separation laws. The link between the deformation based on a traction-separation law and the stress-strain law is made by means of the fracture energy. Plausible approaches to describe cracks are continuum models with discontinuities. The continuum description of a material body is enhanced by displacement discontinuities, e.g. macroscopic cracks or shear bands. One example is the widely known theory of linear elastic fracture mechanics. The behavior of the part of the body which remains continuous can be described by a stress strain law, whereas for the evolution of the crack a governing law, e.g. a traction-separation-law is needed. In the discrete crack approach, cracks are modeled as discontinuities between finite elements into which the structure is discretized. Further developments lead to models with discontinuities placed into the interior of finite elements, like the embedded crack approach and extended finite elements. Although in these models, cracks are treated as discontinuities, the models belong to the class of continuum models with discontinuities and not to discrete models. The reason is that only the crack itself is discrete but the surrounding material is still treated as a continuum. Discrete models don’t consist of a continuum body anymore, but are assemblies of elementary entities of finite size. An example would be a particle model, which deals with particles on microstructure level. Discrete models and continuum models with discontinuities are able to describe the anisotropic behavior after cracking because of their inherent assumptions. This is not necessarily true for models without discontinuity. Whereas smeared models in the narrow sense are also capable to model the anisotropic behavior of cracking, one has to pay attention in using models based on isotropic plasticity or damage. 3.2 Plasticity theory The most widely used form of plasticity is nowadays the flow theory of plasticity formulated in the stress space. Here the strain is decomposed into an elastic and a plastic part and it is assumed that the elastic stiffness remains constant. The main constituents are the yield condition, the flow rule and the hardening law. In the theory of small deformations it is assumed, that the strain can be additively decomposed into the elastic strain and the plastic strain. The stress-strain law for the elastic part is still guided by the linear elastic Hooke’s Law. The yield condition distinguishes the elastic from the plastic region and is not equal to the failure surface of a material though the shape in most cases is similar. For concrete material, flow rules in three dimensional stress states have to be pressure dependent according to the observations in experiments. Well known criteria are the Mohr-Coulomb criterion, developed originally for soils, or the Drucker-Prager criterion, which forms a cone in three-dimensional principle stress space, but also more advanced criteria were established, which fit better the physical shape, like (Ottosen, 1977) or (Menétry & Willam, 1995). When the stress state reaches the yield surface, plastic flow is initiated. During plastic flow, the stress state remains on the yield surface, and an additional rule governing the evolution, i.e. direction of plastic flow, must be postulated. As concrete doesn’t behave in an ideal-plastic manner, hardening variables have to be taken into account, which control the change of the yielding surface during the evolution of plastic flow. The expression “hardening” might be misleading, because especially for tension, there is a distinct domain of softening. For composite materials such as concrete, it is difficult to describe the behavior under an arbitrary loading with a single yield surface. The mechanisms of failure under tension and under compression are quite different, and it is convenient to model them separately, of course taking into account their possible interaction This approach is called Multi-Surface Plasticity, e.g. (Feenstra & deBorst, 1996). These differences in failure can therefore also lead to problems using classical 1206
three-dimensional isotropic models in plasticity, because for stress states with an interaction of tension and compression, like in shear, it is difficult to cover the stress induced anisotropy of the material behavior with an isotropic model. This will be explained later on with an example. 3.3 Damage theory Another approach is made by the damage theory, which is more suitable for the softening region of materials. The idea, originally introduced by Kachanov already in 1958 (Kachanov, 1958), was to describe the phenomenogical degradation of the material properties due to the propagation and coalescence of microcracks, microvoids, and similar defects. In contrast to plasticity, damage models don’t deal with tangent stiffness but with secant stiffness. As in plasticity it is possible to gain isotropic or anisotropic descriptions. For the application on concrete same restrictions are valid: using an isotropic model can lead to problems describing all possible stress states properly. Anisotropic models on the other hand are very difficult to obtain because for anisotropic damage a fourth order tensor with correspondent constants is needed, which cannot be easily determined by experiments. Besides pure plastic or pure damage models, there is the possibility to couple both approaches. 3.4 Smeared crack models Smeared crack models in more narrow sense were the first “engineering” approaches to describe the cracking procedure and deal with the strain-induced anisotropy. In the first formulations, Young’s Modulus of Elasticity was just omitted within the material elastic tensor if the tensile stress reaches the tensile strength. This leads to convergence problems. After some advancements, now most of the models decouple the total strain into an elastic strain and a “cracking” strain, similar to plasticity, and provide a functions based on traction-separation laws to govern the softening after reaching the tensile strength. Three main types of smeared crack models can be distinguished: – Fixed crack models with orthogonal cracks – Fixed crack models with non-orthogonal cracks – Rotating crack models It is claimed, that fixed crack models tend to lead a too stiff material answer whereas rotating crack models behave rather too soft. Within fixed crack models, a shear retention factor or function can be used to govern the transmission of shear stresses over a crack. Rotating crack models rotate related to principle stress axes so no shear transfer is possible. Still these models can give a good prediction of the load bearing capacity of concrete structures. 3.5 Elastic-plastic stress field approach (EPSF) Besides these “classical” approaches to model concrete also other, more pragmatically motivated models exist. One of these was developed at EPFL Lausanne by (Fernandéz Ruiz & Muttoni, 2007). It is a 2D model based on a rigid elastic-plastic approach and abandons any tensile stresses in the concrete, except a small amount needed for convergence reasons. The model is only applicable on reinforced concrete, because the compressive strength is reduced due to transversal strains based on the assumptions of the Modified Compression Field Theory established by (Vecchio & Collins, 1986). The equations for governing the stresses are decoupled by neglecting Poisson’s ratio. Only two constants are needed to describe the material: Young’s Modulus of Elasticity and the yielding stress. For steel, also hardening can be taken into account. 4 APPLICATION OF NLFEM TO BEAMS UNDER SHEAR FAILURE To gain a better understanding the effect of using different material models several models were applied on reinforced and prestressed concrete beams. Within the scope of this paper, two different models are analyzed. A series of reinforced concrete beams with T-shape with shear failure was 1207
Figure 3.
Beams of ET series of Leonhardt.
Figure 4.
Load-displacement curve of simulations with ABAQUS CDP.
tested by Leonhardt in 1969 and varying web thickness. On one side of the beam plain stirrups with lower yield strength were used as shear reinforcements and on the other ribbed stirrups. Failure was obtained by yielding of the stirrups and development of shear cracks into the compressive zone. The beam “ET3” has a web width of 10 cm and a flange width of 30 cm. The reinforcement ratio is ρ = 1.59. Cylinder strength of concrete was measured as well as yielding strength of steel. The tensile strength of concrete was not measured directly, but the splitting tensile strength which can be transformed into the uniaxial tensile strength. This value exceeds the strength which is obtained by estimating the uniaxial tensile strength from the compressive strength by approximately 30%. The beams were calculated with 5 different FE-programs and different material models. For the numerical calculations mean values of material properties were used. 4.1 Calculations with ABAQUS – Simulation of ET3 The beams were calculated with the software ABAQUS using the material model “Concrete damaged Plasticity, a three-dimensional isotropic damaged plasticity model. The yield condition is based on the work of Lubliner (Lubliner, Oliver, Oller, & Onate, 1989) with modifications of Lee and Fenves (Lee & Fenves, 1998). Concrete was modelled with linear 3D continuum elements, the reinforcement was embedded in the continuums elements using the “embedded elements” constraint und consists of 1D truss elements. No dowel action can be taken into account neither shear retention, because within the frame of isotropic plastiticity, shear can only be transferred across the crack due to not coinciding directions of principal stress and principle strains. Within the model some parameters have to calibrated, e.g. the dilatation angle ψ and an artificial damping parameter which is needed to stabilize the calculations. Especially the last parameter influences the result noticeable. The material parameters were set due to experimental tests. With the real set of parameters the peak load of the tested beam was underestimated by about 30%. Further investigations proved the tensile strength fct as the main governing parameter for the value of peak load. The load-displacement diagram shows the result of the calculations of ET3 with 1.0fct and 2.0fct . The cracking pattern of ABAQUS shows a rather different pattern compared to the test. With 1.0fct an almost horizontal second crack occurs, equivalent to a delamination of the flange, which was not observed in the experiment. But also with 2.0fct , differences are present, though there is a 1208
Figure 5.
Comparison of stress fields and cracking pattern with different assumptions for fct .
Figure 6.
Load-displacement curve and stress field for the EPSF method.
good agreement regarding the load displacement curve. Instead of developing several single cracks, only one shear crack occurs, similar to a failure without shear reinforcement. Apparently, in this model the load bearing capacity is not only governed by the strut-and-tie action, but in large part as well by an arching action. Additionally, only 2 stirrups close to the loading area were yielding. The reason of this behavior can be found in the isotropic approach. For mixed stress states (one principle direction under tensile stress, the other under compressive stress), due to isotropy, the stiffness in both directions is reduced after leaving the elastic limit. This means within “cracked” elements not only the tensile strength is reduced, but also the compressive strength. Though a reduction of the compressive strength in concrete due to transversal stress is reasonable and proven by Kupfer, the reduction in this model is too high. Because cracked elements are not able to carry a sufficient amount of compressive stresses anymore, the strut-and-tie model cannot develop properly, and the stirrups were not fully activated. 4.2 Calculations using the EPSF – Simulation of ET3 The simulations with NLFEM based on the elastic-plastic-stress field theory of (Fernandez Ruiz & Muttoni, 2007) show a quite different result. The ultimate load reaches with 239 kN almost 94% of the experimental data and a different stress field pattern. 1209
The load displacement curve differs from the experimental curve, which is explainable by the assumptions of the model. At the beginning, due to neglecting the tensile strength, the EPSF model behaves softer. Later, the curve becomes steeper, because the rigid elastic plastic assumption, governed by reaching the compressive strength, leads to a linear-elastic material behavior up to failure. In contrast to the simulation with ABAQUS, the stress field of the principle compressive stresses shows clearly a strut-and-tie model and all stirrups yielded in the model. 5 CONCLUSIONS The presented example of a T-shaped concrete beam with shear reinforcement shows a significant difference for stress fields and cracking pattern for a calculation using two material models based on different approaches for the description of the cracking behavior. Though by calibrating material parameters resp. internal model variables good accordance of the load displacement curve was obtained with ABAQUS, the stress fields and stirrup stresses do not correspond with the results of the much simpler model using EPSF assumptions. Within a research project (Maurer, Zilch, Kattenstedt et al, 2015) more reinforced and prestressed beams were investigated using four different FE programs. It turns out, that the results highly depend on the used material model. In most of the investigated cases an underestimation of the ultimate load was more likely than the opposite, but no generalization of this can be made. The variation coefficient of the numerical calculation compared with the test results regarding the ultimate load was 0.11. For an application of the NLFE-Method for the assessment of existing bridge structures, safety requirements have to be considered as well. For more details, see (Maurer, Zilch, Kattenstedt et al, 2015). At least, for a safe application of NLFE-Method on real structures more research is necessary, regarding both, the influence of material models and safety format issues. Up to now the introduction of the NLFEM into the design codes is not recommended due to the observed results, though the potential of the method was also obvious. If nonlinear calculations have to be considered, the used material model has to be verified. Numerical analyses of the Kupfer tests as well as of the reinforced panel tests of Vecchio & Collins together with simulations some well known tests of beams under shear failure are strongly recommended to ensure that the material model is capable of describing the shear behavior properly. REFERENCES Feenstra, P. H. & de Borst, R., 1996. A composite plasticity model for concrete. International Journal of Solids and Structures, 33:707–730. Fernández Ruiz M. & Muttoni A. 2007. On Development of Suitable Stress Fields for Structural Concrete, ACI, Structural Journal, Vol. 104 no. 4, pp. 495–502. Hillerborg, A., Modeer, M. & Peterson, P. E. 1976. Analysis of crack propagation and crack growth in concrete by means of fracture mechanics and finite elements. Cement and Concrete Research, 6:773–782. Kachanov. L. M. 1958. Time of the rupture process under creep conditions. Izvestija Akademii NaukSSSR, Otdelenie Techniceskich Nauk, 8:26–31. Menétrey, Ph. & Willam, K. J., 1995. A triaxial failure criterion for concrete and its generalization. ACI Structural Journal, 92:311–318. Maurer, R., Zilch, K., Kattenstedt, S. et al, 2015. Nachrechnung von Betonbrücken – Verfahren der Tragsicherheitsbeurteilung von Betonbrücken im Bestand für die Nachweisstufe 4 der Nachrechnungsrichtlinien, research report FE 15.0523/2011/FRB of Bundesanstalt für Straßenwesen, in preparation. Müller, M. & Haardt, P.: Aspekte der Ressortforschung zur Nachrechnung bestehender Straßenbrücken aus Beton. Bauingenieur, Dezember 2014. Ottosen, N. S. 1977. A failure criterion for concrete. Journal of Engineering Mechanics, ASCE, 103:527–535. Vecchio, F.J. & Collins, M.P. 1986. Modified Compression-Field Theory for reinforced concrete beams subjected to shear. ACI Journal. V. 83, 1986, Bd. No. 2.
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Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
Thermal analysis of a fin-back bridge under sudden drop in temperature F. Tian Tongji University, Shanghai, China
Y. Lu & P. Zhu Shanghai Urban Construction Design and Research Institution, Shanghai, China
ABSTRACT: The surface temperature of concrete bridge is extremely high in summer. If sudden drop of air temperature happens, the surface temperature will drop dramatically, which causes serious nonlinear temperature field and self-stress of the bridge. When the stress exceeds the tensile strength of the concrete, concrete cracking happens. As a novel bridge in China, the finback bridge across Dazhi River in Pudong district of Shanghai was investigated. A 2-D model of the bridge was built and the temperature field of the fin under sudden drop in temperature was studied based on the local meteorological data. The results show that the temperature distribution of the fin exhibits obvious nonlinear characteristics and the maximum temperature difference of the fin reaches 11.5◦ C. The results provide a basis for further stress analysis of similar structures.
1 INTRODUCTION Temperature field of concrete bridge is related to the wind, atmospheric temperature, solar radiation and the material parameters. Because of the strong solar radiation in summer, the surface temperature of concrete bridge is extremely high. If sudden rainstorm occurs at the moment, the surface temperature will drop dramatically. Due to the poor thermal conductivity of concrete (Branco & Mendes, 1993), the inner temperature change of the structure is very small. The serious temperature difference between the surface and interior of the bridge will cause great self-stress. When the stress exceeds the tensile strength of the concrete, concrete cracking happens. This paper mainly studies the temperature field of the bridge under drastic changes of temperature in summer. A 2-D numerical model of the bridge was built based on the local meteorological data. As the main bearing component, the fin of the bridge is a total mass of concrete. It is necessary to do research on the actions of different kinds of negative factors on the fin. Two cases, which include the temperature effects under solar radiation and sudden drop of air temperature, were considered. Firstly, the temperature field of the bridge under solar radiation was obtained. The results were served as the initial condition of the temperature analysis of the bridge under sudden drop in temperature. The influence of the wind speed and the change of air temperature were considered, but the influence of the solar radiation was ignored (Liu, 1991). Lastly, the final temperature distribution under sudden drop in temperature was investigated. 2 HEAT CONDUCTION THEORIES 2.1 Governing equation The ambient temperature and other environmental parameters along the bridge are basically the same, hence the longitudinal variation of temperature is ignored and the temperature field 1211
Figure 1.
Elevation drawing of Dazhi River Bridge (unit: m).
analysis is simplified as a 2-D heat conduction problem described by the following differential equation:
where T is a function of the temperature (◦ C), x and y are Cartesian coordinates (m), t is the time (s), c denotes the heat capacity (J/(kg·◦ C)), ρ is the mass density (kg/m3 ) and λ is the heat conduction coefficient (W/(m·◦ C)). 2.2 Initial condition and boundary condition It is necessary to introduce the initial condition and boundary conditions to solve Equation 1 and obtain the temperature distribution of the structure. The initial condition describes the temperature distribution of the concrete at the initial time, which is expressed by Equation 2.
where T0 denotes the temperature field at the initial time t0 . The boundary condition of heat conduction problem describes the temperature, heat flux and convection on the boundary. In this article, the following boundary condition is adopted:
In which n is the normal direction of outer boundary, αs is heat exchange coefficient of radiation, S is solar radiation intensity (W/m2 ), β is the convective heat transfer coefficient (W/(m2 ·◦ C)), T is the surface temperature of concrete (◦ C) and Ta is the atmospheric temperature (◦ C). 3 ANALYSIS OF THE BRIDGE TEMPERATURE 3.1 Basic data Dazhi River Bridge located at Pudong district of Shanghai is the first long span prestressed concrete fin-back bridge in China. The bridge deck is fully continuous with a main span of 158 m long. Finback with parabolic outline is set above the bridge deck to increase the stiffness of the bridge. The elevation drawing of the bridge is shown in Figure 1. The finite element model of the cross section located at the middle-pier (Fig. 2) was established. The heat conduction coefficient, the heat capacity and the mass density are 2.54 W/(m·◦ C), 1212
Figure 2. The cross section located at the middle-pier (unit: cm).
Figure 3. Atmospheric temperature.
988 J/(kg·◦ C) and 2500 kg/m3 , respectively. Thermal analysis under daily solar radiation was firstly conducted and the stable-state of fluctuation was acquired. The temperature distribution was used as the initial condition to calculate the temperature field of the cross section under sudden drop in temperature. 3.2 Temperature field under solar radiation Under periodical solar radiation, heat exchange will reach equilibrium and the system achieves a stable-state of fluctuation after a period of time, which is served as the initial condition for temperature analysis of the bridge under sudden drop in temperature. In order to obtain the temperature field under solar radiation, atmospheric temperature and solar radiation intensity at the bridge site were measured (Fig. 3 and Fig. 4). In Figure 4, only the solar radiation intensities of the fin are plotted, others are not presented here for simplicity. The boundary condition outside the box girder is described by Equation 3. However, the environment inside the box girder is not the same as that of outside, so the state of heat exchange is different. The radiation inside the box is extremely weak, so Equation 3 is reduced to:
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Figure 4.
Solar radiation intensity.
Figure 5. Temperature distribution of the fin.
Based on the above mentioned boundary conditions inside and outside the box, the transient thermal analysis was conducted for twenty one days, after which the cross-section temperature reached the stable-state of fluctuation. The results show that the maximum temperature difference of the fin on the twenty-first day appears between two and three o’clock in the afternoon. Temperature distribution of the fin at this moment is shown in Figure 5. It shows that the temperature of Face 3 is much higher than that of Face 1 because of the stronger solar radiation of Face 3 than Face 1 in the afternoon. 3.3 Temperature field under sudden drop in temperature 3.3.1 Boundary conditions under sudden drop in temperature Changes of wind speed and air temperature were considered under sudden drop in temperature. According to related literatures (Wang et al., 2007, Zheng & Zhou, 2002), the cooling period and cooling amplitute are 1 h and 10◦ C, respectively. The change of wind speed with time (Wang et al., 2007) is plotted in Figure 6. The effect of wind speed on the convection coefficient (Zheng & Zhou, 2002, Wu, 1995) is listed in Table 1. 3.3.2 Temperature distribution Figure 7 shows the time-history curve of temperature in different depth of the fin. It presents that the surface temperature of the fin drops rapidly within half an hour and subsequently tends to be stable. With the increase of depth from the surface, the temperature variation decreases sharply. 1214
Figure 6. Variation of wind speed with time. Table 1. Wind speed and convection coefficient. v (m/s)
0.0
1.0
2.0
3.0
4.0
5.0
6.0
7.0
8.0
9.0
10.0
β (W/(m2 ·◦ C))
9.4
13.2
16.8
20.6
24.3
28.0
31.7
35.4
39.1
42.8
46.5
Figure 7. Time-history curve of the temperature in different depth of the fin.
When the depth is more than 100 mm, the temperature variation is not obvious. In the middle part of the fin (900 mm depth), the temperature rarely varies. Temperature variation after one-hour cooling was shown in Figure 8a. The maximum drop of temperature is on the surface, which is about 11.5◦ C. Drop of temperature mainly occurs in the range of 0 to 100 mm depth from the surface. In the range of 100 mm to 500 mm, negative drop of temperature occurs, which is caused by the temperature field of the cross section under solar radiation, and it has no relation to the drop of air temperature. Nonlinear temperature in specification of different countries (MOT, 2004, BSI, 2006, AASHTO, 2004) is mainly simplified as polygonal lines distribution. The method is adopted by this paper. The corresponding simplified temperature distribution curve is shown in Figure 8b. 1215
Figure 8. Temperature variation after cooling.
4 CONCLUSIONS The nonlinear temperature field of concrete bridge caused by sudden drop of air temperature is one of the main reasons of concrete cracking. The temperature distribution of a three-span prestressed concrete fin-back bridge was investigated in this paper. The following conclusions are drawn from the results. (1) Under sudden drop in temperature, the temperature drops of the fin mainly occurs within half an hour, afterwards, the temperature variation is very small and can be neglected. Meanwhile, the surface of the fin is very sensitive to the drop of ambient temperature. With the increase of depth from the surface, the temperature variation with time becomes smaller. (2) After 1 h cooling, the temperature variation of the cross section is nonlinear. Drop of temperature mainly occurs in the range of 0 to 100 mm depth from the surface. The temperature variation decreases with the increase of depth, and the maximum drop of temperature is about 11.5◦ C. If the initial temperature field is not considered, in the range of 100 mm to 900 mm depth, the temperature basically stays unchanged. (3) Based on the thermal analysis of the fin-back bridge under solar radiation and sudden drop in temperature, the simplified temperature distribution of the fin is obtained. The results provide a basis for further thermal stress analysis for similar structures. REFERENCES Aashto 2004 AASHTO LRFD Bridge Design Specifications SI Units (Third Edition). Washington, D.C.: AASHTO. Branco, F. A. & Mendes, P. A. 1993 Thermal actions for concrete bridge design. Journal of Structural Engineering 119(8): 2313–2331. Bsi 2006 BS5400-2:2006 Steel, concrete and composite bridge.part2: specification for loads. London: BSI. Liu, X. 1991 Temperature stress analysis of concrete structures, Beijing, China Communications Press. MOT (Ministry of Transport of the People’s Republic of China) 2004 JTG D62-2004 Code for design of reinforced concrete and prestressed concrete bridges and culverts. Beijing: China Communications Press. Wang, C., Zhang, X. & Chen, X. 2007 The change of asphalt pavement temperature field when temperature decreasing suddenly. Central South Highway Engineering 32(3): 113–115+126. Wu, G. 1995 Thermal stress analysis of semi-rigid pavement, Beijing: Science Press. Zheng, J. & Zhou, Z. 2002 Design theory and method of asphalt pavement cracking, Beijing: China Communications Press.
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Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
Comparison study of aeroelastic analysis of a pylon of the Mersey Gateway Bridge with its 2D/3D wind tunnel tests S.B. Kim Samsung C&T, Seoul, Korea
J. Rees Flint & Neill, Gloucestershire, UK
J.Y. Chung TESolution, Gyeonggi-do, Korea
S.H. Jang, G.D. Moir & J.H. Seo Merseylink, Warrington, UK
ABSTRACT: Aeroelastic analysis of a pylon of the Mersey Gateway Bridge is conducted and compared with its 2D/3D wind tunnel tests. Flutter derivatives are approximately estimated from the steady state wind load coefficients obtained from the wind tunnel tests of the 2D section models of the pylon. Since the width of the pylon is varying along the elevation, three different section models are selected as the representative sections of the pylon. Aerodynamic damping effect is considered by including the Scanlan’s flutter equation into the system matrix of the state space equation for the pylon. Spatial coherence of fluctuating wind velocity is modeled with the Karman model, and the effect of the aerodynamic admittance function is considered with the Liepmann model. Aerodynamic stability analysis shows a good agreement with the wind tunnel test result of the 2D section model. Since the parameters used in this analysis are obtained from the 2D section model tests, the aerodynamic stability analysis of the 3D pylon model doesn’t match well with the wind tunnel test result of the 3D pylon model. The static and RMS responses of the pylon from the 2D aeroelastic analysis shows a good agreement with the 2D wind tunnel test results. The aeroelastic analysis results of the 3D pylon model shows a good agreement with the 3D wind tunnel test results, except the cases where the negative damping effect become dominant. Comparison results show the efficacy of the aeroelastic analysis for the prediction of the wind responses of the pylon. 1 INTRODUCTION Pylons of cable-stayed bridges are susceptible to strong wind. During the design process of a cable-stayed bridge, wind tunnel test of the pylon is conducted to verify the stability against the strong wind and to identify the excessive vibrations from the fluctuating wind velocity. Aeroelastic analysis can be used to predict the wind responses and the aerodynamic stability of the pylon. Wind-pylon interaction effects can be considered by self-excitation force term from the Scanlan’s flutter equation. In this research, the flutter derivatives are approximately evaluated from the steady state wind load coefficients as follows:
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Figure 1. Typical section model of the pylon of the Mersey Gateway Bridge.
Figure 2. Wind characteristics at the bridge site.
The structure studied in this research is a 125 m height pylon of the Mersey Gateway Bridge. The width across the bridge is constant with 3.5 m (Fig. 1). The width along the bridge is decreasing with the elevation. Two kinds of model are considered, one is a 2D section model with a constant width and the other one is a 3D free-standing pylon model with the decreasing width along the elevation of the pylon. Both of the analysis models are constructed based on the steady state wind load coefficients estimated from the wind tunnel test of the 2D section models. To consider the 3D effects, 2D section models with three different representative widths are selected and the steady state wind load coefficients are estimated from the 2D wind tunnel tests. Wind velocity, turbulence intensity, length scale, and power spectral density of the fluctuating wind speed are modeled through the wind climate analysis (Fig. 2). Two different construction stages of the pylon are considered: 1) erection stage with free standing pylon, 2) service stage with connected deck and stay cables. Figure 3 shows the mode shapes of the pylon at the erection stage and service stage. Figure 4 shows the steady state wind load coefficients measured from the wind tunnel tests of the 2D section models. 2 WIND RESPONSE ANLAYSIS OF PYLON 2.1 Similarity of turbulence intensity based on the small-scale turbulence The wind tunnel test for the 2D section model of the pylon was conducted at a small-scale wind tunnel. Since the small-scale turbulence at the high frequency range dominates the flow pattern around the structure (Melbourne 1975), the turbulent intensity at the wind tunnel was scaled based 1218
Figure 3.
Mode shapes of the pylon.
Figure 4.
Steady state wind load coefficients.
Figure 5.
Power spectral density of the fluctuating wind velocity in model scale.
on the similarity law for small-scale turbulence (Sangchuwong et al. 2012). Figure 5 shows that the power spectral density (PSD) of the fluctuating wind velocity at the wind tunnel is matched well with the Karman’s PSD model at the high frequency range, while the magnitude of the wind tunnel’s PSD at the 1st natural frequency of the pylon is much smaller than the magnitude of the Karman model’s PSD. 1219
Figure 6.
Static and RMS response of the pylon in erection stage (2D model).
2.2 Wind response analysis of 2D section model of pylon in erection stage Figure 6 shows the static and RMS response of the pylon. Especially the responses at the wind speed of 40 m/s are compared with the results of the wind tunnel tests. Since the static response is dominated by the drag force, the maximum response along the bridge (z-axis) occurs when the wind incident angle is 90 degree and the maximum response across the bridge (y-axis) occurs when the incident angle is 0 degree. Comparison of static responses shows a good agreement between the analysis results and wind tunnel test results. Figure 6e, f shows the abrupt increase of the RMS responses at the zero degree of the incident angle for the z-axis and at the 90 degree for the y-axis, which indicates that the aerodynamic damping is drastically decreased around these regions. Aerodynamic stability of the 2D section model of the pylon will be explained later.
2.3 Aeroelastic analysis of 3d model of pylon in service stage Figure 7 shows the static and RMS response of the pylon. The responses at the wind speed of 40 m/s are compared with the results of the wind tunnel tests. Comparison of static responses shows a small difference between the analysis results and wind tunnel test results. The 3D model used in the wind tunnel test was tuned to match the dynamic parameters of the proto-type model, and only the wind load coefficients of three representative 2D section models at the elevations of 105 m, 85 m, and 65 m are used for the analysis. The RMS responses of the analysis show a good agreement with the results of the wind tunnel tests. Figure 8 shows the RMS response of the pylon as a function of wind speed. In the case of the wind incident angle of 40 degree shows a good agreement between the analysis and the wind tunnel test. However at the incident angle of 90 degree, the RMS response 1220
Figure 7.
Static and RMS response of the pylon in service stage (3D model).
Figure 8.
RMS response of the pylon in service stage as a function of wind speed.
Figure 9. Wind stability analysis result of the pylon.
of the analysis increases rapidly after the wind speed of 33 m/s where the total damping of the wind-pylon system goes negative. 3 AERODYNAMIC STABILITY ANALYSIS Aerodynamic stability analysis of the pylon is conducted and compare with the dynamic wind tunnel tests. Figure 9a shows the results of the aerodynamic stability analysis of the 2D section model of the pylon in erection stage. In the case of the vibration along the z-axis (along the bridge), the instability occurs at the incident angles of 5 and 10 degrees at smooth flow and 0 and 5 degrees at turbulent flow. For the vibration of the y-axis (across the bridge), the instability occurs at the 1221
Figure 10.
Flow pattern obtained from the CFD analysis of the 3D free standing pylon.
incident angles of 85 and 90 degrees at both of the smooth and turbulent flows. These results are almost same as the wind tunnel tests of the 2D section model. Figure 9b shows the results of the aerodynamic stability analysis of the 3D free-standing pylon model in service stage. The instability occurs at the incident angles of 85 and 90 degrees, which is almost same with the results of the 2D analysis of erection stage. Only the critical wind speeds are slightly changed. However, the dynamic wind tunnel test of the 3D free-standing pylon model shows that the pylon is stable for all incident angles. Figure 10 shows the flow pattern of the 3D free-standing pylon from the CFD analysis. We can find a movement of vortex at the rear slit of the pylon along the vertical direction, which indicates that the flow of the 3D free standing pylon has a strong 3D characteristic. 4 CONCLUDING REMARKS Aeroelastic analysis of a pylon of the Mersey Gateway Bridge was conducted and compared with its 2D/3D wind tunnel tests. Aerodynamic stability analysis results show a good agreement with the dynamic wind tunnel test results of the 2D section model. However it should be careful for the application to the 3D model. The static responses of the 2D/3D analysis show a good agreement with the 2D/3D wind tunnel test results. The RMS response of the 3D aeroelastic analysis shows a good agreement with the result of the wind tunnel test for the 3D pylon model, except the cases where the negative damping effects become dominant. Comparison result shows the efficacy of the aeroelastic analysis for the prediction of the wind responses of the pylon. REFERENCES Diana, G., Resta, F., Zasso, A., Belloli, M., Rocchi, D. 2004. Forced motion and free motion aeroelastic tests on a new concept dynamometric section model of the Messina suspension bridge. Journal of Wind Engineering and Industrial Aerodynamics 92: 441–462. Kim, S.B. & Chung, JY. 2013. Mersey Gateway Bridge Project: Evaluation of Turbulence Spectrum on the TESolution’s Wind-Tunnel for Section Model Test, Samsung C&T. Melbourne, W.H. 1975. Probability distribution of response of BHP house to wind action and model comparisons. Journal of Industrial Aerodynamics 1, 167–175. Sangchuwong, P., Yamada, H., Katsuchi, H. 2012. Study on turbulence effects on flow patterns around rectangular cylinders. Seventh International Colloquium on Bluff Body Aerodynamics & Applications. Simiu, E. & Scanlan, R.H. 1996. Wind Effects on Structures: Fundamentals and Applications to Design, John Wiley & Sons Inc. Strømmen, E. 2010. Theory of Bridge Aerodynamics. Springer.
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Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
Numerical models used to simulate the “in situ” testing of a bridge on A1 motorway in Romania I.R. R˘ac˘anel Technical University of Civil Engineering of Bucharest, Romania
ABSTRACT: The final stage in the construction process of a bridge consists in tests performed “in situ” to validate the design solution. For common bridge solutions, simple static tests under the action of loaded trucks placed in appropriate positions of the bridge superstructure are to be performed. For important bridges, the response of the bridge under dynamic loads has to be tested supplementary. The Eurocodes introduced as compulsory design standards around Europe a couple years ago do not state a way to perform the tests on a bridge, because each country has specific conditions for testing. This paper presents the results obtained on numerical models used to simulate the “in situ” testing of a viaduct on A1 motorway in Romania, with respect to the values obtained during the tests. The differences obtained for the measured and calculated parameters are shown in explicit graphs together with appropriate comments.
1 INTRODUCTION 1.1 Problem statement The testing of bridges has the purpose to evaluate the structural response under the applied loads during the tests and through this to check the correctness of the design and the bridge safety in service. Usually, two types of tests are to be performed. The static test consists in measuring the vertical displacements and stresses under the action of groups of loaded trucks placed in several positions on the bridge superstructure. The vehicles position on the structure is established so that the produced effects in terms of stresses and vertical displacement are close to those obtained in the calculation, during the design process. For bridges having large spans, built in special solutions or placed on important roads, in addition to static, a dynamic test should be carried out. Thus, to consider the dynamic effects of the vehicles action on the bridge, caused by several factors, mainly the unevenness of the way layers. This has as results an impulse load transmitted to the road surface, which lead to the increase of the stresses and displacements values. The procedure consists in measuring the time-history of some important parameters, usually stresses and accelerations, when a loaded truck crosses the bridge at several speed levels and pass over an artificially created obstacle at the way level. The measured data are compared with those obtained from the calculations made in the design stage. In our days, the evolution of the computer hardware together with the development of specialized software allow to use complex numerical analysis procedures to reproduce the effects of the tests carried out on the site. This is the aim of present paper, to compare the results obtained both, on the site and through finite elements analyses and finally to outline the differences and comment them. 1223
Figure 1.
General layout of the bridge.
1.2 Description of the bridge structure The bridge analyzed in the paper is located on A1 motorway, section Or˘as¸tie-Sibiu between km 73 + 040.32 and 73 + 285.32 and has a total length of 254.30 m covered with five spans in the following alignment: 32.50 + 3 × 60.00 + 32.50 m (Fig. 1). In plan view, the bridge is placed in a curve having a radius of 2670m and in elevation it has a longitudinal slope between 1.3 and 4.5%. As static scheme, the bridge superstructure is a continuous girder on five spans, expansion joints being foreseen only at each end of the bridge. The bridge superstructure is a composite one (Fig. 2), including two main steel girders with box cross section, cross beams equally spaced along the span (at 5.00 m) and at the upper part of the steel girders a concrete slab with a constant depth of 28 cm. The depth of the girders is 2.60 m and the distance in transverse direction between girders axes is 6.50 m. Final cross beams have a total depth of 2.20 m and the current cross beams have a cross section with variable height: 725 mm in the bridge longitudinal axis and 1.03 m next to the main girders. The connection of the main girders and cross beams with the slab is made through shear studs welded on the top flanges. In the sections above the bearings, inside the steel boxes, concrete was also poured at the bottom part and connected through shear studs with the main girders bottom plate, in order to avoid the occurrence of buckling phenomena. Four piers (P1-P4) and two abutments (C1, C2) form the bridge substructure. The height of the piers is in the range 15.00–21.00 m. The elevation of the piers has “Y” letter shape (Fig. 2) and consists in a reinforced concrete box having the dimensions 2.50 × 5.00 m. The substructure elements have deep foundations, on piles having a diameter of 1.20 m and and variable lengths in the range 26.00–33.00 m. A general layout of the bridge is presented in Figure 1 and a cross section through bridge structural elements is shown in Figure 2. The superstructure sustains a carriage way having a total width of 12.00 m, consisting in two circulation lanes, one emergency lane and other spaces for safety barriers. The transverse slope of the carriage way is 2.50%. For the bearing of the superstructure on the substructure elements unidirectional pot bearings and shock absorbers were used. The fix bearing devices are placed on pier P3.
2 SHORT DESCRIPTION OF TESTING PROCEDURES The response of the bridge under test loads was estimated both by measurements performed on the site and using numerical finite element models and finally the values of measured data were compared. Testing procedures are based on the information included in I.R.S. (1986). For static tests, vertical displacements of the bridge deck, under the action of several group of loaded trucks placed in specific positions on the superstructure, were measured using high precision leveling. Trucks with four axles were measured and weighted and placed on the bridge 1224
Figure 2.
Cross section of the bridge.
Figure 3.
Loading position 4 (plan view – left, cross section view – right).
superstructure. The weight of a loaded truck was about 400 kN. For this bridge, five loading positions were considered. In Figure 3 the loading position 4 in the second span is shown. The values of vertical displacements are measured, for each loading position, in five point across and eight sections along the bridge. For example, for loading position 4, the measurement points were in the range 16–30 (Fig. 3). In the case of dynamic testing, a single loaded truck crossed the bridge at several speed levels without breaking or accelerating, passing over an artificially created obstacle (Fig. 4), usually a wood plank fixed on the road surface. The speed levels were 10, 30, 50, 70 and 90 km/h. Real time on the bridge superstructure induced accelerations were measured in two sections along the bridge, in the middle of the spans 2 and 3. The wood plank was placed first in the middle of the third span and after finalizing all speed levels, in the middle of the second span. Based on registered accelerations values in the post-processing stage of the tests, other important dynamic parameters of the bridge response could be obtained: superstructure speeds and displacements in vertical direction, composition in frequencies of the structural response, real-time oscillations of the bridge superstructure. 1225
Figure 4.
Position of the measuring system and of the wood plank during dynamic tests.
Figure 5. View of the discrete model using shell finite elements.
Same parameters were obtained through numerical analyses and finally compared with measurement values and the other obtained in post-processing stage of the tests. 3 FINITE ELEMENT MODELS USED FOR THE NUMERICAL ANALYSES For numerical analyses two types of finite element models have been used and they will be briefly presented in the following. In the first model, all structural elements of the superstructure were modeled using four nodes shell elements, combining plate and membrane behavior. The element formulation is based on the information given in Wilson (2002). The connection between the concrete slab and main steel girders was done by means of some rigid link elements representing the shear connectors, which avoid the occurrence of relative displacements between the structural elements. The global finite element model is presented in Figure 5 and in Figure 6 detailed views of cross beams meshes are presented. Based on the fact that for nonlinear dynamic analyses, the use of such complex models can lead to a long time necessary to run analyses and post-processing the resulted data, a simplified second model, based on the principles for equivaleting structural response and use of substructures given by Przemieniecki (1968) was built. In this finite element model al components were modeled using two nodes straight frame elements based on the formulations given by Bathe and Wilson (1976). In order to achieve the correct values for axial, shear, bending and torsional stiffness of cross sections of frame elements, some correction constants have been used. The steel box cross sections of main girders were modeled by splitting them in three parts, two of them including the webs and top flanges of steel girders connected to the concrete slab and the third representing the bottom plate of the box with modified geometric characteristics. In order to avoid the relative displacements 1226
Figure 6.
Detailed views of cross beams meshes: left – final cross beam, right – current cross beam.
Figure 7.
Simplified finite element model used for the performed analyses.
of the frame elements forming the main girders, which can occur following the distortion of thin walls cross sections at certain level of stresses, special rigid elements were used. This modeling approach was used in order to improve the mesh density at the level of the top flanges of the girders and concrete slab, forming in this way a plan grillage. A view of this second finite element model is shown in Figure 7. In both models the substructure of the bridge was not considered because it was assumed that it has no significant influence on the computed data. 4 PERFORMED ANALYSES AND RESULTS Consider the loading procedure used for static tests, in order to obtain the values of vertical displacements, linear moving load analyses were conducted. In this way, for each loading position, the deformed shape of the bridge superstructure could be established. From numerical models, the vertical displacements values were checked at positions corresponding to the real positions of the measurement points. Because of lack of space, in this paper only the biggest values for the vertical deflections of the bridge superstructure are presented graphically and that is the case for the loading position 4. All displacements values, that is for all loading positions, are summarized in Table 1. In the charts in Figure 8 a comparison between the values measured on site and obtained from finite element models, in measuring points 16–30, is shown. Comparing the values in the table above, the conclusion is that the use of the simplified model lead to good results, close to those obtained by means of the full finite element model based on plane elements and close to the other measured on site. This conclusion sustains the use of the simplified model further, for the dynamic analyses. The dynamic response of the bridge under the action of a loaded truck crossing the bridge at certain speed levels was obtained numerically by performing linear modal and time-history 1227
Table 1. Comparison of vertical displacements, loading position 4. Vertical displacement [mm] Measuring point 16 Measured Model – shell Model – frame
17
18
19
20
21
22
23
24
25
26
27
29
30
14.9 20.5 27.0 33.2 38.3 23.2 31.0 40.7 49.9 57.4 15.2 20.8 26.8 33.2 39.0 17.6 23.1 28.7 33.6 38.5 28.2 36.0 43.9 51.2 58.3 19.2 24.8 30.4 35.0 40.4 20.1 24.2 30.1 35.4 39.0 31.9 37.7 45.9 53.7 58.8 21.8 25.9 31.9 37.2 40.9
Figure 8.
Deflected shape of the superstructure in cross section, measuring points 16–30.
Figure 9.
First eigenmode, T1 = 0.605 s, f1 = 1.65 Hz.
Figure 10.
28
Second eigenmode, T2 = 0.46 s, f2 = 2.18 Hz.
analyses. These were carried out on the simplified model already presented in the paper. The considered speed levels were: 10, 30, 50, 70 and 90 km/h. The impulse effect of the load induced by the truck on the bridge superstructure was modeled by means of a time-history step function considering the theory presented in Frýba (1996), Yang, Yau and Wu (2004), Paz and Leigh (2004). In the first stage, a linear modal analysis was performed in order to observe the main frequencies of the structure. The first three eigenmodes are presented in Figure 9–11. Following the time-history analysis, the time-history functions of vertical acceleration of the bridge superstructure in two sections along the span were obtained. Using these functions as input 1228
Figure 11. Third eigenmode, T3 = 0.36 s, f3 = 2.77 Hz.
Figure 12. Vertical accelerations of the bridge deck: measured – left, computed – right.
Figure 13. Vertical accelerations of the bridge deck: measured – left, computed – right.
for a specialized post-processing software, the power spectra of induced vertical accelerations can be also obtained. Subsequently, the frequencies range of the damped vibrations of the bridge superstructure could be captured. Finally, the obtained time-history functions of vertical accelerations and the corresponding power spectra were compared with those obtained by post-processing of the “in situ” obtained signals. The position of the points in the finite elements models were the values of the vertical accelerations were pointed out correspond to the position of the accelerometers on the bridge superstructure, that is in the middle of spans 2 and 3. In Figures 12 and 13 the charts of measured and computed vertical accelerations are shown. Figure 14 and 15 contain the power spectra of the registered and computed accelerograms. These charts corresponding to the situation with the wood plank placed in the middle of the third span, with the vehicle crossing the bridge at 10 km/h. 5 CONCLUSIONS The papers presents the results obtained following the testing on site of a viaduct on the A1 motorway in Romania. The measured data in terms of vertical displacements and accelerations are compared with the results obtained using finite element models. Two models were used: a full three-dimensional model based on shell elements and a simplified model composed from frame finite elements, with modified stiffness characteristics, based on the substructures theory. 1229
Figure 14.
Power spectra of the vertical accelerations: measured – left, computed – right.
Figure 15.
Power spectra of the vertical accelerations: measured – left, computed – right.
The numerical results obtained using the simplified model considering the loads in the static tests are in good accordance with those obtained with the complex model, the differences for the maximum displacement being only of 2% with respect to the measured values (as presented in Table 1). Following the dynamic analyses carried out on the simplified model, the time-history function of vertical acceleration of the deck is obtained and by means of post processing with a specialized software, the power spectra of accelerations is also computed. The vertical accelerations of the deck are in the range 0.4–1.3 m/s2 , bellow the allowable limits 0.3–0.4 g recommended in the standards. The performed analyses illustrate a good behavior of the bridge under the test loads. Moreover, the finite elements models used to obtain the bridge response under static and dynamic loads are accurate enough to lead to correct results, based on the good accordance with the measured data. REFERENCES Bathe, K.J. & Wilson, E.L. 1976. Numerical methods in finite element analysis, New York: Prentice Hall, Englewood Cliffs. Frýba, L. 1992 Dynamics of railway bridges, London: Thomas Telford Ltd. I.R.S., 1986 Testing of superstructures with test actions, Bucharest: National Council for Science and Technology. Paz, M. & Leigh, W. 2004 Structural Dynamics. Theory and Computation, Norwell, Massachusetts: Kluwer Academic Publisher. Przemieniecki, J.S. 1968 Theory of matrix structural analysis, New York: McGraw-Hill. Wilson, E.L. 2002 Three-Dimensional Static and Dynamic analysis of structures, Berkeley: Computers and Structures Ins. Yang, Y.B., Yau, J.D. & Wu, Y.S. 2004 Vehicle-bridge interaction dynamics: with applications to high speed railways, London: World Scientific Publishing Co. Pte. Ltd.
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Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
Wind effects analysis on cable stayed bridges decks V.D. Urdareanu & I.R. R˘ac˘anel
Faculty of Railways, Roads and Bridges, Technical University of Civil Engineering of Bucharest, Romania
ABSTRACT: Nowadays, cable supported bridges are great solutions for crossing large obstacles, being both economic and aesthetic. But as stable and safe as the may seem, their decks are in fact very slender and flexible, in order to obtain maximum structural efficiency, thus making them susceptible to various meteorological conditions such as wind or rain and other vibration inducing mechanisms. Underestimating these effects can lead to discomfort due to oscillating displacements, a dramatic shortening of the structure’s lifetime and even its untimely collapse. Perhaps the most famous example of the damage these phenomena can cause is the collapse of the Tacoma Narrows bridge in 1940, due to resonating vibrations caused by low to moderate wind speeds. Currently, the European design codes only cover scenarios for structures with spans of up to 200 m, larger ones requiring more complex analyses. In this context, the paper presents an analysis of wind effects on cable supported bridge decks, done using the finite element method. It contains both steady state and dynamic transient studies for different mean wind speeds, highlighting specific phenomena for each of them. The main purpose of the steady state studies is obtaining drag and lift coefficients under different mean wind speeds and comparing them with the results obtained using the formulas in the euro codes. The purpose of the dynamic transient analysis is to find out whether or not the drag or lift coefficients vary in time under different mean wind speeds and if so, the possibility of resonance occurring. The paper contains case studies for three individual cross sections of bridges designed in Romania, specifically the bridge over the Some¸s river in Satu Mare, made up of three spans of 80 m + 150 m + 80 m, the bridge over the Danube Channel in Agigea consisting of three spans of 80 m + 200 m + 80 m and the bridge over Bucharest ring road at Otopeni with spans of 65 + 160 + 85 m. 1 INTRODUCTION One of the most common concepts in designing long span bridges is using cables to locally transfer the loads from the superstructure to highly resistant piers. These types of structures are commonly known as cable supported bridges and are well known for their economic and structural efficiency, along with very aesthetic shapes. One thing they all have in common is their very slender deck which, as practice has shown, should primarily be designed withstand horizontal or eccentric vertical loads, coupled with oscillation phenomena. Taking all this into account in an early design phase is no easy task, since airflow around the bridge superstructure can cause dynamic instability under a high number of wind mechanisms. There has been a long history of severe wind problems with cable supported bridges. In 1818, a 79 m span foot bridge across the Tweed river at Dryburgh abbey, Scotland, failed as a consequence of deck oscillations caused by wind action. From 1818, until the collapse of the Tacoma Narrows bridge in 1940, there have been 11 recorded examples of bridges that either failed or suffered severe damages. In addition, since 1937 there have been at least 10 other bridges that have been observed to oscillate in the wind. From the evidence available, it is clear that some failures and instances of severe motion had extreme torsional amplitudes that could be attributed to flutter and torsional instability. 1231
The collapse in 1836 of one of the four 78 m spans of the Brighton Chain Pier in England was witnessed and reported in by Lt. Col William Reid of the British Army. His report included remarkable sketches illustrating severe torsional motion and ultimate deck failure. There is a striking similarity between these sketches and the photographs of the Tacoma Narrows bridge failure, 104 years later. The investigation of the Tacoma Narrows Bridge failure by Farquharson has clearly shown that the failure was caused by a severe torsional instability. The Golden Gate Bridge, with a main span of 1280 m, is the largest bridge with a history of wind-excited motion. Its motion was systematically observed and recorded during the period 1943–1951. A particularly severe torsional response occurred during a storm in December 1951 when an amplitude of 1.7 m was reached. Subsequently, torsional stiffness was increased by the addition of a bottom lateral system that has prevented further occurrence of torsional instability. Flutter and torsional instability are phenomena that relate principally to structures with shallow cross-sections where the crosswind dimension is small compared to the longitudinal wind dimension. Their occurrence can cause destructive oscillatory motion of the decks of long span bridges. The best method of prevention against these oscillation mechanisms in having thorough wind tunnel tests done on scaled models of the bridge in the design phase. The problem with wind tunnel tests is they are quite expensive and should not be done without careful consideration of the overall shapes and sizes of the bridges elements. A good approach in establishing an efficient cross section for a cable supported bridge is combining the principles of both structural mechanics and fluid mechanics at the same time, using the finite element method. 2 ANALYSIS 2.1 Structural mechanics Structural mechanics is the computation of deformations, deflections, and internal forces or stresses within structures, either for design or for performance evaluation of existing structures. Structural mechanics analysis needs input data such as structural loads, the structure’s geometric representation and support conditions, and the materials’ properties. Output quantities may include support reactions, stresses and displacements. Advanced structural mechanics may include the effects of stability and non-linear behaviors. Mechanics of structures is a field of study within applied mechanics that investigates the behavior of structures under mechanical loads, such as bending of a beam, buckling of a column, torsion of a shaft, deflection of a thin shell, and vibration of a bridge. There are three approaches to the analysis: the energy methods, flexibility method or direct stiffness method which later developed into the finite element method. As one of the methods of structural analysis, the direct stiffness method, also known as the matrix stiffness method, is particularly suited for computer-automated analysis of complex structures including the statically indeterminate type. It is a matrix method that makes use of the members stiffness relations for computing member forces and displacements in structures. The direct stiffness method is the most common implementation of the finite element method (FEM). In applying the method, the system must be modeled as a set of simpler, idealized elements interconnected at the nodes. The material stiffness properties of these elements are then, through matrix mathematics, compiled into a single matrix equation which governs the behavior of the entire idealized structure. The structure’s unknown displacements and forces can then be determined by solving this equation:
where K is the stiffness matrix, δ vector of the member’s displacements, FI vector of the member’s internal forces and FE vector of the member’s external forces. 1232
2.2 Fluid mechanics Fluid mechanics is the branch of physics which involves the study of fluids (liquids, gases, and plasmas) and the within on them. Fluid mechanics can be divided into fluid statics, the study of fluids at rest; and fluid dynamics, the study of the effect of forces on fluid motion. It is a branch of continuum mechanics, a subject which models matter without using the information that it is made out of atoms; that is, it models matter from a macroscopic point of view rather than from microscopic. Fluid mechanics, especially fluid dynamics, is an active field of research with many problems that are partly or wholly unsolved. Fluid mechanics can be mathematically complex and can best be solved by numerical methods, typically using computers. A modern discipline, called computational fluid dynamics (CFD), is devoted to this approach to solving fluid mechanics problems. Computers are used to perform the calculations required to simulate the interaction of liquids and gases with surfaces defined by boundary conditions. The finite element method is used in structural analysis of solids, but is also applicable to fluids. However, the FEM formulation requires special care to ensure a conservative solution. The FEM formulation has been adapted for use with fluid dynamics governing equations. Although FEM must be carefully formulated to be conservative, it is much more stable than other methods available to this date. In this method, a weighted residual equation is formed and must be solved:
where Ri is the residual at an element vertex i, Wi is the weight factor, Q is the conservation equation expressed on an element basis and Ve is the volume of the element 3 CASE STUDY The case study includes 3 different bridges with composite steel-concrete superstructures as follows: – bridge over the Some¸s river in Satu Mare, made up of three spans of 80 m + 150 m + 80 m, with a cross section consisting of a 20 cm concrete deck and 2 steel “I” shaped steel beams, abundantly reinforced with strong cross frames; – bridge over the Danube Channel in Agigea consisting of three spans of 80 m + 200 m + 80 m, with a cross section made up of a 20 cm concrete deck and 2 steel, trapezoidal box girders with orthotropic plates; – bridge over Bucharest ring road at Otopeni with spans of 65 + 160 + 85 m, with a cross section made up of twin superstructures with a 25 cm concrete deck and a steel, trapezoidal box girder each with orthotropic plates. In order to have a better comparison between the 3, their cross-sections have been scaled so that they all accommodate a 14.80 wide carriageway and twin 5.00 m wide pedestrian sidewalks, whilst keeping their overall height, area and principal moments of inertia the same, with only the shape varying. Three dimensional structural mechanics finite element method analyses have been done for the study of the overall behavior of the structure under dead weight and traffic loads and 2 dimensional fluid mechanics finite element method analyses for the study of the airflow over the cross-sections, and it’s influence on the bridge. The structural mechanics analysis is pretty straight forward and will not be focused on further on, since it is not the topic of this paper. Fluid mechanics models have several governing laws with different hypotheses, in order to accommodate for large scale phenomena or smaller, more local mechanisms. For this analysis, the Shear-Stress Transport k-w model has been used as the numerical calculation method. This method 1233
Table 1. Main parameters of the dynamic fluid flow analysis.
Figure 1.
Mean wind speed [m/s]
Air density [kg/m3 ]
Air dynamic viscosity [kg/(m×s)]
Approximate number of elements nr
5/15/25
1.25
1.983 × 10−5
27,000
Cross sections of the 3 studied bridges.
was developed by Menter to effectively blend the robust and accurate formulation of the k-w model in the boundary region with the free-stream independence of the k-e model in the open field. The transport equations for the SST k-w model are:
where Gk represents the generation of turbulence kinetic energy due to mean velocity gradients, Gω represents the generation of ω calculated as described in the standard k-ω model, k and ω are the effective diffusity of k and ω, Yk and Yw represent the dissipation of k and ω due to turbulence, Dω is the cross-diffusion term, Sk and Sω are user defined source terms. The parameters used in the analysis are shown in Table 1. Different mean wind speeds were used to study whether or not the wind speed has any impact on the airflow and the results. The air density and dynamic viscosity are the ones prescribed in SR EN 1991-4. The top speed is the one recommended for bridge design and the others were chosen arbitrarily. The finite elements are quad shaped, each one with 8 nodes, 4 in the corners and 4 in the midsides. Their shape is approximately squared, with sizes ranging from 1 mm in the center, at the bridge boundary, to about 10 m at the external boundaries. The domain size is large enough so that the external boundaries have no effect near the bridge’s cross-section in the centers and sufficient so that all the phenomena can be covered. The number of elements used was the minimum necessary in order to obtain accurate results. It was obtained by running the same analysis with different mesh densities. Steady state and transient analyses have been run on 3 models, each corresponding to a different bridge. 1234
Figure 2. Streamlines, pressure contours, velocity contours, and vorticity contours for the Bridge over the Some¸s river, in different states of the analysis: initial position, rotated by 1◦ and by 3◦ .
4 RESULTS 4.1 Bridge over the Some¸s river in Satu Mare The airflow around the bridge’s deck is similar to the one around aicrafts wings. The air speed on the bottom side is higher than the one on the upper side, generating drag forces and overall downwards lift forces. Nevertheless, there is an area of high pressure underneath the left cantilever, as well as an area of negative pressure above it, both acting on the deck’s torsional stability, generating excentric upwards lift forces. There is also a large area of negative pressure between the two beams, derived from a sort of suction phenomenae on the superstructure’s inferior fiber. This is also the source of the overall larger downwards forces. These forces generate a torsional moment that tends to twist the section transversely. Structural mechanics analyses showed that a maximum rotation of about 1◦ can be reached by these alone. Adding additional excentric traffic loads led to a total rotation of about 3◦ . As the section twists transversely, the high velocity profiles tend to shift towards the upper side, dercreasing the lift forces values and eventually reversing their direction. This is a dangerous aspect to be taken into account when establishing the section’s torsional stiffness and when considering errection errors. Altough the lift forces don’t jeopardize the superstructure from a static point of view, considering this as a repetitive process could lead to dangerous fatigue stresses. As far as vorticity is concerned, there are 2 areas of high turbulence, one under the left beam and one above the left cantilever. The rotation of the cross section is of little consequence to these area and so can be neglected. A comparison of the lift and drag coefficients in different stages of the analysis can be found in the table below. The manual coefficients were calculated according to SR EN 1991-1-4 and can’t take into account any transverse rotation of the studied cross section. The ones calculated in the F.E.M. model were obtained by integrating all drag or lift forces over surface of the superstructure. 4.2 Bridge over the Danube Channel in Agigea The initial behavior of the wind profile over the bridge’s superstructure is at first, similar to the one of the first case study. The wind speed on the bottom side is still higher than the one at the top, 1235
Table 2. Drag and lift coefficients for case study 1 – Bridge over the Some¸s river. Analysis type
Manual
F.E.M. straight
F.E.M. rotated 1◦
F.E.M. rotated 3◦
Drag coefficient Lift coefficient
4.5 3.1
3.03 −2.71
3.23 −1.53
3.26 2.42
Figure 3. Streamlines, pressure contours, velocity contours, and vorticity contours for the Bridge over the Danube channel, in different states of the analysis: initial position, rotated by 1◦ and by 3◦ . Table 3. Drag and lift coefficients for case study 2 – Bridge over the Danube channel in Agigea. Analysis type
Manual
F.E.M. straight
F.E.M. rotated 1◦
F.E.M. rotated 3◦
Drag coefficient Lift coefficient
4.5 3.1
2.136 −4.953
2.402 −1.21
2.885 3.614
resulting in overall higher downwards forces. The cantilever at the left is still lifted upwards by a pressure gradient, thus rotating the entire section, and the area between the 2 beams is characterized by negative pressure, generating the much higher downwards forces. Although the 2 box girders give a higher torsional stiffness then the “I” beams in the first case, the larger main span led to similar rotation angles of about 1◦ from wind forces alone and 3◦ after adding traffic loads. As the cross section the airflow tends to stabilize, constantly lowering the lift forces up to a point when the upwards forces overcome the downwards ones. This happens at approximately 1.3◦ . From this point on, the speed on the upper side is higher then the bottom one. The pressure gradient on the left cantilever also increases as the section twists. This leads to an ever increasing torsion moment. As in the previous case, there are 2 main turbulence areas that are not affected by the rotation of the superstructure. The biggest one is along a line linking the left cantilever’s end with the bottom of the left girder and the smaller one is located above the left pedestrian sidewalk. As in the previous study, a table containing drag and lift coefficients is provided in Table 3. 1236
Figure 4. Streamlines, pressure contours, velocity contours, and vorticity contours for the bridge over Bucharest ring road at Otopeni, in different states of the analysis: initial position, rotated by 1◦ and by 3◦ . Table 4. Drag and lift coefficients for case study 3 – Bridge over Bucharest ring road at Otopeni. Analysis type Drag coefficient on left deck Drag coefficient on right deck Total drag coefficient Lift coefficient on left deck Lift coefficient on right deck Total lift coefficient
Manual 4.5 0
F.E.M. straight 3.14 −0.12
F.E.M. rotated 1◦ 3.33 −0.2
3.69 −0.67
4.5
3.01
2.87
9.61
2.87
−3.31
−2.55
−1.16
6.30
8.25
7.99
3.1
3.12
F.E.M. rotated 3◦
10.8
3.02 9.15
4.3 Bridge over Bucharest Ring road at Otopeni Studying the airflow around the twin superstructures it can be said that most of the forces from wind effects are concentrated on the left one. The other one is protected by the first and the force inducing mechanisms have little influence on it. This is not necessarily a good thing for the entire structure, as the two are linked together by strong crossbeams. The high pressure area under the left cantilever is much more pronounced resulting in a much bigger torsion moment then the other case studies. This is due to a bad length to height of the left cantilever and the beam itself, generating potentially dangerous flutter and torsional instability effects. The overall rotation of the cross section has little influence when comparing to the much more higher instability phenomena already acting on the structure. A small downwards streamline appears to be forming between the two. Although not very powerful, this is the main source of instability for the right deck and should be taken into account in the design phase. The Table 4 contains drag and lift coefficients for the studied cross section. 1237
5 CONCLUSIONS Although all long span bridges must have wind tunnel tests, optimal cross sections can be first obtained by combining structural mechanics with fluid mechanics in finite element models. This approach however, does not eliminate the need of wind tunnel tests, which are the only way to validate the obtained results. When analyzing cable supported bridges under wind loads, a key aspect, aside from their overall size and shape, is to ensure sufficient torsional stiffness. The excentric forces generated by the varying wind pressure on the surface of the decks tend to rotate the section transversely. Adding excentric traffic loads to the structure leads to a increase the torsion of the bridges superstructure. As previoussly shown, even slight variations of 1◦ to 3◦ in the angle of attack can lead to signifficant changes in the airflow around the bridge deck, having little influence on the drag forces, but generating a dramatic impact in the lift forces. The rotation of the studied bridge sections by only 1◦ lead to an increase of the drag coefficient by 6% to 8%. Further rotation to about 3◦ resulted in the increase of the drag forces acting on the deck by approximately 15% to 20%, depending on the overall shape. When studying the influence of the general angle of the deck on the lift coefficient, the results have proven to be much more complex, with the most important parameter being the exact shape of the superstructure. Nevertheless, the first two case studies (single deck bridges) showed a change in direction of the lift forces by 180◦ , with values ranging from 70% to 85% of the initial ones. If rotating the sections by only 1◦ resulted in decreased lift forces by 50%–75%, showing airflow stability, increasing the rotation of the deck by a further 2◦ lead to an inversion of the lift forces, with values comparable to the initial ones. When designing long-span cable supported bridges, it is a common practice to have decks with enough torsional stiffness, so that they show sufficient stability and safety under eccentric traffic loads, ensuring a sufficient degree of comfort for passengers. The torsional stiffness of the superstructure should also be considered, along with its overall shape, when studying the structures behavior under wind loads. Twin decks should generally be avoided, as they don’t offer the same amount of rigidity to transverse rotation as continuous box girders. Another important aspect to take into consideration is the ratio between the latelar cantilever and the overall height of the deck, depending on the angles of both elements, certain ratios can give more aerodynamic stability the others. Neglecting to have stable bridge sections under wind loads can lead to strong vibration effects that in time, can seriously damage the structure, shortening its lifespan or even collapsing under crytical wind speeds due to resonance phaenomena. REFERENCES Acheson, D. J , 1990. Elementary Fluid Dynamics, Clarendon Press. Batchelor, G. K., 1967. An Introduction to Fluid Dynamics, Cambridge University Press. Chanson, H., 2009. Applied Hydrodynamics: An Introduction to Ideal and Real Fluid Flows, (2009). CRC Press, Taylor & Francis Group, Leiden, The Netherlands. Gordon & Breach, 1973. Lectures on Fluid Mechanics, Shinbrot. Masaru Matsumoto, 2006. Review of Bridge Cable Vibrations in Japan, Kyoto University. Milne-Thomson L.M., 1973. Theoretical Aerodynamics. T Dover Publications. ISBN 0-486-61980-X. Stephen B. & Pope, 2000. Turbulent Flows , Cambridge University Press. Wardlaw R. L., 1994. Flutter and Torsional Instability, Ottawa, Canada, Springer-Verlag Wien. You-Lin Xu, 2007. Wind Effects on Cable-Supported Bridges, The Hong Kong Polytechnic University, Hong Kongm P.R. China. Yunus A. Cengel, John M. Cimbala McGraw-Hill, 2010. Fluid Mechanics: Fundamentals and Applications, 2nd Edition. ***, 2007. Wind Induced Vibration of Stay Cables, U.S. Department of Transportation, Federal Highway Administration.
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Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
“Beam sectional analysis” an innovative technique for analysis of bridge superstructure K. Kashefi & A.H. Sheikh School of Civil, Environmental and Mining Engineering, University of Adelaide, Adelaide, South Australia, Australia
ABSTRACT: A novel finite element-based cross-sectional analysis of beam like slender structures is introduced for the efficient analysis of bridge super-structures. 3D beam analysis is decomposed into a 2D cross-sectional problem and a 1D beam problem through the decomposition of the 3D strain field. The 2D analysis gives the constitutive relationship of the beam accurately without any major assumptions as it is able to consider in-plane and out-of-plane warping of the beam section. Subsequently the 3D stress, strain and displacement field will be recovered accurately. Numerical examples of bridge girders with solid and thin walled sections are solved using this method and the results obtained are compared with those produced by a detailed finite element analysis. The results show a very good performance of the proposed method.
1 INTRODUCTION Beam like slender structures have been used in civil engineering and some other engineering areas for so long now. A common application of such structures is in bridges with box girder deck systems which are basically thin-walled beams having closed or combination of closed and open sections. The behavior of these thin walled structures under an arbitrary loading scenario is quite complex which is primarily due to the cross-sectional warping (out-of-plane warping) and the distortion of the sections (in-plane warping). In the recent years, accurate calculation of warping displacements has been the subject of many research studies since the variation of warping displacements over a cross-section does not follow a standard pattern and it changes from one case to other case depending on the cross-sectional profile and the loading pattern. On the other hand, establishment and development of finite element methods has brought about the opportunity to predict the behavior of these structures in a more detailed manner but this approach seems to have some disadvantages. Firstly, it is very time-consuming and requires rather big computational resources (i.e. computational and storage memory) especially for elaborate geometries which is useless in initial design stages. Moreover, all comprehensive FEM software packages are very expensive compared to simpler beam-modelling tools. The other extreme is to model these structures as beams which makes the analysis highly efficient computationally but a model based on a classical beam theory such as Euler-Bernoulli hypothesis cannot predict the behavior of these structures well because it assumes that a beam cross section will remain plane and perpendicular to beam axis after deformation and so cannot capture the local deformations in the form of warping and the effect of shear deformation. The shear deformation effects are incorporated in Timoshenko’s beam theory but this theory also cannot capture the warping deformation. Saint Venant tried to incorporate the effect of torsional deformation which is absent in the above theories but eventually he could only consider the effect of uniform (unconstrained) warping (out-of-plane) without any shear deformation. A better representation of the deformation of thinwalled beams is made in Vlasov’s (1961) theory which can address the problem of constrained warping (out-of-plane) but it cannot capture the in-plane warping (distortion) of the beam section. A number of researchers (e.g. Bauld and Tzeng (1984), Chandra et al. (1990), Lee (2005) and Sheikh and Thomsen (2008)) used Vlasov’s (1961) theory and made some advancements but these 1239
approaches needed a prior knowledge for the pattern of warping displacements which can change from one case to another. Subsequently, the effect of distortion in thin-walled box girders has been studied by few researchers (e.g. Razaqpur and Li (1991 & 1994) and Kermani and Waldron (1993)) but the techniques they developed for modelling the distortion are based on some ad hoc assumptions and cannot be used in general cases. In this context, the concept of beam sectional analysis proposed by Givatto et al. (1983) and later on extended and employed by Borri and Merlini (1986), Ghiringhelli and Mantegazza (1994) and few others (e.g. Blasques (2014)) seems to be the most attractive of all. In this approach, the complete three dimensional (3D) elasticity problem defining the actual behavior of the beam like structures is decomposed to a two-dimensional (2D) beam section analysis and a one-dimensional (1D) beam analysis without using any ad hoc assumption. The method only assumes that the cross-sectional dimensions are small compared to the beam length which is true for slender beam structures. The 2D beam sectional analysis is carried out using a 2D finite element discretization where the effects of in-plane warping as well as out-of-plane warping are considered. The 2D finite element analysis generates the exact constitutive matrix (or stiffness matrix) of the beam cross-section which ensures a proper coupling between the different modes of deformation. This cross-section stiffness matrix is then used in the 1D beam analysis which can be carried out using a standard 1D beam finite element model. The stress resultants obtained in the 1D beam analysis together with the results of the 2D cross-sectional analysis are used in combination to obtain the warping displacements and finally recover the 3D stress and displacement fields of the beam. The computational efficiency of this approach is remarkable in terms of the prediction of the 3D response of these structures. A similar approach has been employed by Hodges et al. (1992) based on the concept of Variation Asymptotic Method (VAM) which is a method based on a rigorous mathematical foundation. The method (VAM) has been introduced by Berdichevsky (1976) who first applied this method to shell structures (1979) which helped to reduce the 3D problem in to a 2D problem consistently utilizing the shell thickness as the small parameter which is a key concept of this method. Cesnik and Hodges (1997) extended this method for beam analysis and developed VABS (Variational Asymptotic Beam Section analysis) which is quite appealing due to its mathematical consistency but the method is significantly complex with respect to its mathematical treatment. On the other hand, the other approach (proposed by Givatto et al. (1983)) is relatively less complex but it has similar capabilities as that of VABS. It is interesting to note that these methods so far have been only applied in the analysis of aerospace related structures such as helicopter and wind turbine blades, etc. and no one has paid any attention to take advantage of such methods for the analysis of box girder bridge decks and similar structures. In this paper, an attempt has been made to develop Givatto’s technique for the analysis of box girder bridge decks and similar structures having closed or the combination of open and closed sections. The 2D cross-sectional problem is solved using eight-node quadratic isoparametric elements whereas three-node isoparametric linear elements are used to solve the 1D beam problem. A computer code is written in Fortran to implement the different steps associated with the 2D sectional analysis, 1D beam analysis as well as the recovery of the beam 3D response. In order to test the performance of the proposed analysis technique, a number of numerical examples including solid and thin walled box girders with different cross-sectional configurations have been solved which three of which are reported here. Detailed 3D finite element analyses of these beams have been also carried out using ABAQUS and the results obtained are used to validate the results produced by the proposed technique. 2 BEAM SECTIONAL ANALYSIS THEORY 2.1 Basic definitions The theory presented here is valid for a prismatic slender structure “beam” composed of anisotropic and inhomogeneous materials and the section properties remain unchanged along the beam axis. 1240
Figure 1.
Beam reference coordinate system.
The loads are assumed only on the end tips of the beam which means that the local effect of applied loads along the beam is neglected. Also the theory is based on the assumption of small strains. Considering an orthogonal Cartesian reference set like Figure 1, the displacement of an arbitrary point in the cross section is s = [sx , sy , sz ]T where z is along the beam axis. The 3D strain and stress field of a point in the cross section are ε = [εxx , εyy , εxy , εxz , εyz , εzz ]T and σ = [σ xx , σ yy , σ xy , σ xz , σ yz , σ zz ]T subsequently. According to the constitutive law, the stresses and strains are related as σ = Qε where Q is the material constitutive matrix. The vector of cross section forces and moments is also θ = [T , M ]T where T and M are resultant forces and moments respectively. The resultant section forces and moments θ can be calculated as,
where A is the section area and I is the identity matric and x and y are the position of the point in the cross section. 2.2 Cross sectional discretization The displacement s = [sx , sy , sz ]T at a point in the cross section can be divided into rigid body displacement v = [vx , vy , vz ]T as well as in-plane and out of plane warping displacements g = [gx , gy , gz ]T through s = v + g. v can be obtained from v = Zr where r = [χT , ϕT ] is rigid body translations and rotations of the reference point of the cross section and g can be written using finite element discretization as g = Nu where N is shape function matrix and u is nodal warping displacement. So, the total displacement can be rewritten as; Using compatibility relations, the strain field can be written as, where the comma means differentiation with respect to ‘z’ which means that in 3D strain– displacement relation ∂/∂z has been separated. B and S are the following matrixes;
Substituting Eq. 2 into Eq. 3, the 3D strain matrix can be obtained as follows;
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Finally after some manipulations the 3D strain field can be decomposed to a rigid body strain and a warping part as,
where ψ = [τ T ,κT ]T is the roto-translational strain which does not deform the section. τ = [τ x , τ y , τ z ]T consists of transverse shear and axial strain. Accordingly, κ = [κx , κy , κz ]T includes curvatures and twist. 2.3 Cross sectional stiffness matrix If we consider a part of beam with a very small length between two cross sections, the variation of total virtual work for that segment of the beam is comprised of the variation of internal work δW i and external work δW e as,
The variation of internal work or the work done by the elastic strain energy per unit length can be written as,
Also, the variation of the external work considering the tractions on the cross section surface (p=[σ xz , σ yz , σ zz ]) can be obtain as,
Substituting these equations in Eq. 7 the total virtual work for unit length of the beam will be;
The above equation is six times indeterminate. This indetermination can be removed by imposing six constraints on warping displacements as below;
There are two types of solutions for Eq. 10; a particular solution which refers to the central part of the beam and an eigenvalue solution which corresponds to the solutions related to the beam extremities where the loads are applied. The central solution of the beam results in calculation of the cross section compliance matrix Fs . Therefore, the stiffness matrix of the cross section can be obtained as Fs = Ks−1 . This cross section stiffness matrix is defined accurately and can be accounted for any possible geometrical and material couplings. 2.4 Stress/strain recovery The cross section stiffness matrix (Ks ) is then used in the 1D global analysis of the beam to obtain stress resultants. Then roto-translational strain (ψ) and the warping displacement (u) (in-plane and out-of-plane) will be calculated for each section of the beam. Substituting these results into Eq. 6, the 3D strain field will be obtained and finally the 3D stress field can be recovered using the constitutive law (σ = Qε). 1242
3 NUMERICAL EXAMPLES Here, some examples of cantilever prismatic beams of length 20 m with solid and thin-walled box sections are presented, in order to validate the proposed beam model with 3D finite element analysis using ABAQUS. The module of elasticity for the material used (steel) is E = 200 GPa and Poisson ratio is ν = 0.3. Contrary to the huge computational effort involved in the analysis of a 3D FE solid model, the proposed method for section analysis requires much less effort due to the lower number of degrees of freedom (DoFs) as will be shown in the following examples. 3.1 Isotropic solid square beam This first example analyzes a beam with a square cross-section which its dimension is a = 1.0 m. The validity of the cross section beam analysis is checked through the convergence study of various stiffness coefficients per unit length of the beam when using the proposed method and analytical results. The analytical results are based on Timoshenko beam theory given in Timoshenko and Goodier (1951). The complete stiffness matrix is reported in Table 1. Stress/displacement studies are performed using two load cases for this example. Load case-1 is a shear force (F = [0.0, −1.0, 0.0] kN) and load case-2 is a torque (M = [0.0, 0.0, −1.0] kN·m) both applied at the free end. In the presented method, the beam cross section is being discretized by a mesh of 14 × 14 eightnode quadratic plane finite elements for a total of 1941 DoFs and 1D model has been meshed by 100 three-node beam finite elements (1206 DoFs) by introducing field-consistent shape functions. The 3D FE model is generated in ABAQUS using a mesh of 14 × 14 × 80 finite elements with 20 nodes (C3D20) giving a total of 201735 DoFs. The aspect ratio then becomes 3.5:1:1 (L:b:b). Table 2 shows the displacement values of different points of the free-end section and Table 3 shows the angle of twist at the free end provided by the presented method, analytical solution and 3D FEM. As it is obvious from Tables 2 and 3, there is a good agreement between the proposed model and the results of the 3D FEM as well as the analytical results in terms of displacement field. Stress field comparison also reveals a very good agreement between the proposed method and 3D FEM for both load cases. The transverse shear stress (σ yz ) of the beam mid-span cross section for load case-1 and free end cross section for load case-2 are shown in Figures 3–4.
Figure 2.
Example 1, isotropic square section beam.
Table 1. Cross section stiffness components for a solid square section. Component
Analytical
Proposed method
Difference
K11 , K22 (GA (N)) K33 (EA (N)) K44 , K55 (EI (N·m2 )) K66 (GJ (N·m2 ))
6.4102 × 1010 200 × 109 16.6667 × 109 10.8154 × 109
6.4158 × 1010 200 × 109 16.6999 × 109 10.8356 × 109
+0.087% +0.0% +0.2% +0.18%
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Table 2. The displacement components (mm) of example 1, load case 1 evaluated at the free-end section. Component
Analytical
Proposed method
3D FEM
sy (Deflection) sz Point B sz Point C
1.6031 × 10−1
1.5983 × 10−1
5.9928 × 10−3
5.9831 × 10−3
1.5975 × 10−1 5.9875 × 10−3 5.9831 × 10−3
5.9928 × 10−3
5.9826 × 10−3
Table 3. The angle of twist for example 1, load case 2 evaluated at the free end. Component
Analytical
Proposed method
3D FEM
Angle of twist
1.8492 × 10−6
1.8439 × 10−6
1.8464 × 10−6
Figure 3. KN/m2 .
(σyz ) at mid-span section of load case-1; the proposed method (left) and the 3D FEM (right); units:
Figure 4. KN/m2 .
(σyz ) at free end section of load case-2; the proposed method (left) and the 3D FEM (right); units:
Figure 5.
One cell box girder cross-sectional dimensions (m).
3.2 One-cell box girder The results for cross-sectional analysis of a single cell box (Fig. 5) is given in this section. Sectional stiffness matrix components of the girder obtained by the presented model are shown in Table 4. As can be seen there are couplings for shear-torsion (K16 ) and bending-extension (K34 ) of the cross section. The validity of the cross section beam analysis is verified with a 3D solid FE model developed in ABAQUS. 3.3 Two-cell box girder This example considers a two-cell box girder. The dimensions of the beam cross section are according to Figure 6. For this girder, stress/displacement analysis has been performed under a twisting moment of equals to M = [0.0, 0.0, −1.0] kN·m at the tip end. 1244
Table 4. Cross section stiffness components for single box girder. K11
K22
K33
K44
K55
K66
K16
K34
6.068e9
2.698e9
2.796e10
5.376e9
1.237e10
2.437e9
4.025e6
−2.674e4
Figure 6.
Isotropic two-cell box girder; units (m).
Table 5. The displacement components (mm) of the cross section for example 2, at the free end. Component
sx (Proposed method)
sy (Proposed method)
sx (3D FEM)
sy (3D FEM)
Point A Point D
1.5211 × 10−3 −1.6879 × 10−3
6.8154 × 10−3 3.7592 × 10−3
1.4799 × 10−3 −1.7071 × 10−3
6.7688 × 10−3 3.7334 × 10−3
Figure 7.
(σ xz ) at mid-span cross section; the proposed method (up), 3D FEM (down); units (kN/m2 ).
In this case, the beam cross section is discretized by a mesh of 201 eight-node quadratic plane finite elements for a total of 3012 DoFs and 1D model uses 1206 DoFs. The 3D solid model applies a similar grid on using quadratic elements with 20 nodes (C3D20) and it gives a mesh of total of 351051 DOFs. The displacement values of two points at the free-end section are reported in Table 5. These results together with distribution of shear stresses in Figures 7–8 again emphasize the capability of the proposed method to recover the 3D stress/displacement field of the beam when the local load effects are not significant in the plane section distortions. 4 CONCLUDING REMARKS A cross-sectional modeling technique for analysis of straight bridge girders has been presented in this paper. This technique decomposes 3D strain field into roto-translational (1D) strain and 3D warping displacement. From 2D analysis, a cross-sectional stiffness matrix is obtained which can account for all of the geometrical couplings in e.g. multi-cell bridge girders and material couplings in anisotropic beams. Besides, once the 2D stiffness matrix is calculated, it can be used to analyze many beams with various boundary conditions and loading. It is then utilized in the 1D beam analysis to calculate the stress resultants and the 3D warping displacements. The 3D strain, stress and displacement field are then recovered. Numerical results obtained from both solid and thin walled steel girders have revealed a very good agreement with those obtained from 3D FE 1245
Figure 8.
(σyz ) at mid-span cross section; the proposed method (up), 3D FEM (down); units (kN/m2 ).
analysis. Considering the efficiency and the accuracy required for bridge super-structure analysis and design, the presented method seems to have the potential to be used and also developed for static and dynamic analysis of various bridges (e.g. straight, curved, composite, etc.). REFERENCES Bauld, N.R. and L.S. Tzeng, A Vlasov theory for fiber-reinforced beams with thin-walled open cross sections. International Journal of Solids and Structures, 1984. 20(3): p. 277–297. Berdichevskii, V.L., Equations of the theory of anisotropic inhomogeneous rods. in Soviet Physics Doklady. 1976. Berdichevskii, V.L., Variational-asymptotic method of constructing a theory of shells. PMM vol. 43, no.4, 1979, pp. 664–687. Journal of Applied Mathematics and Mechanics, 1979. 43(4): p. 711–736. Blasques, J.P., Multi-material topology optimization of laminated composite beams with eigenfrequency constraints. Composite Structures, 2014. 111(0): p. 45–55. Borri, M., G.L. Ghiringhelli, and T. Merlini, Linear analysis of naturally curved and twisted anisotropic beams. Composites Engineering, 1992. 2(5): p. 433–456. Borri, M. and T. Merlini, A large displacement formulation for anisotropic beam analysis. Meccanica, 1986. 21(1): p. 30–37. Cesnik, C.E.S. and D.H. Hodges, VABS: A new concept for composite rotor blade cross-sectional modeling. Journal of the American Helicopter Society, 1997. 42(1): p. 27–38. Chandra, R., A. Stemple, and I. Chopra, Thin-walled composite beams under bending,torsion, and extensional load. AIAA Journal, 1990. 27(7): p. 619–626. Ghiringhelli, G.L. and P. Mantegazza, Linear, straight and untwisted anisotropic beam section properties from solid finite elements. Composites Engineering, 1994. 4(12): p. 1225–1239. Giavotto, V., et al., Anisotropic beam theory and applications. Computers & Structures, 1983. 16(1): p. 403–413. Hodges, D.H., et al., On a simplified strain energy function for geometrically nonlinear behaviour of anisotropic beams. Composites Engineering, 1992. 2(5): p. 513–526. Kermani, B. and P. Waldron, Analysis of continuous box girder bridges including the effects of distortion. Computers & Structures, 1993. 47(3): p. 427–440. Lee, J., Flexural analysis of thin-walled composite beams using shear-deformable beam theory. Composite Structures, 2005. 70(2): p. 212–222. Razaqpur, A.G. and H. Li, Thin-walled multicell box-girder finite element. Journal of Structural Engineering, 1991. 117(10): p. 2953–2971. Razaqpur, A. and H. Li, Refined analysis of curved thin-walled multicell box girders. Computers & structures, 1994. 53(1): p. 131–142. Sheikh, A.H. and O.T. Thomsen, An efficient beam element for the analysis of laminated composite beams of thin-walled open and closed cross sections. Composites Science and Technology, 2008. 68(10): p. 2273–2281. Timoshenko, S.P. and J.N. Goodier, Theory of Elasticity. McGraw-Hill International Book Company, 1984. Vlasov, V., Thin-walled elastic beams. Office of Technical Services, U.S. Department of Commerce, Washington 25. DC, lT-61-I 1400, 1961.
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Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
Large bridge in pergola for high velocity trains in Spain C. Jurado Director of Civil Department, Construction, Infrastructure and Transport Polytechnic University of Madrid, Spain
ABSTRACT: The present paper exposes the project and construction of one of the most singular viaducts in Spain for high velocity trains which crosses above the existing motorway M-31, near to Palencia, with a very acute angle of 20◦ . Due to the obliquity of the crossing, the solution of the skewed bridge has been made “in pergola” (Spanish terminology) with a length of 153 m.
1 THE SPANISH INFRAESTRUCTURE PLAN 2000-2020 Spain has undertaken with the assistance of European funds a tremendous effort to overcame its deficiencies in transport and infrastructures. From the end of eighties until the mid of the nineties, it has took with France and Germany the first place in the European Union (EU) in the percentage of Gross Domestic Product (GDP) allocated to investment in transport infrastructures. From the threshold of 0.5–0.6% of GDP in the mid-eighties it has gone to values of around 1.7–1.8% in the nineties (Fig. 1). The Spanish Infrastructure Plan 2000–2020 projects to connect all the provinces with the capital of Spain with high velocity trains. The cost of the Plan will be about 114.2 thousand million Euros.
Figure 1. Trends in Public Investment by countries in EU (%GDP).
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Figure 2.
Spanish Infrastructure Plan 2000–2020.
Figure 3.
New railway line Madrid-Valladolid-León.
The finance will be public and private. By the year 2020 all the capital of province will be connected by train with less than four and a half hours from Madrid, in the center of Spain (Fig. 2). 2 INTRODUCTION During the years 2008 to 2011 it has been constructed in Spain the new railway line of high velocity trains (HVT) that will connect Valladolid with León in the northwest of Spain (Fig. 3). 1248
Figure 4.
New line of High Velocity Trains. Section between Palencia and León (122.20 km).
This part of the Spanish Infrastructure Plan has cost 200 million of Euros with a length about 183 kilometers. The speed will be about 220 kilometres per hour, but the railway platform is prepared to reach 350 kilometres per hour. The connection between Palencia and León comprises 122.20 km and in the middle of it there is the area of Cea River-Bercianos del Real Camino with a length of 10.5 kilometers (Fig. 4). The present paper exposes the project and construction of one of the most singular bridges in the actuation of Río Cea – Bercianos del Real Camino, from Palencia to León, which crosses over the existing motorway M-31. The solution of the bridge has been made a skewed viaduct (“in pergola, spanish terminology”) with a length of 153 m and with a width of 14.5 m. 3 THE PROJECT The actuation that has been projected called Cea River-Bercianos del Real Camino, goes along Sahagun’s municipal area and the total length of platform for high speed trains reaches 10.5 km. The work begins, approximately to 3 km to the South of the population of Sahagún, in the P. K. 55+100 of the Informative Study (300+000 of the Construction Project), following with direction to Northwest with a length of 10,501 m, until the P. K. 65+587 of the Informative Study and from 0.6 km to the North of the population of Bercianos del Real Camino. Between the kilometre point of 300+000 and the 310+500 which is the total length of the actuation it has been projected five great viaducts from 36 to 442 meters and also five bridges over the line of high velocity trains, to maintain the permeability of people and animals. Between the kilometre point of 300+000 and the 310+500 which is the total length of the In the table 1 there is a list of the five great viaducts in this actuation called Cea River – Bercianos del Real Camino. 1249
Table 1. The five viaducts included in the actuation of Cea River – Bercianos del Real Camino. Designation
K. Begining
K. End
Total length (m)
Spans distributions (m)
Viaduct over Cea River Viaduct over Valle stream Viaduct over Santiago road
300+072 300+861 306+501
300+188 300+897 306+943
116.00 36.00 442.00
Viaduct over Coso Stream Pergola over A-231 motorway
308+824 308+980
308+976 309+133
152.00 153.00
35.50 − 45.00 − 35.50 36.00 32.00 + 3 × 41.00 + 50.00 + 5 × 41.00 + 32.00 34.00 − 2 × 42.00 − 34.00 76.5 + 76.5
Figure 5.
Situation of the five viaducts.
The first structure nearer to Palencia is the Viaduct over Cea River, and the last one in this actuation nearer to Leon is “Pergola over the A-231 motorway” (Fig. 5). The most important and singular viaduct between the five ones is called “Pergola over the A-231 motorway” with a length of 153 meters and in figure 6 it can be seen a moment of construction. In view of the great obliquity of the crossing of the new railway line of high speed trains over the existing motor A-231, which connect the provinces León and Burgos, with a very acute angle of 20◦ , turns out to be necessary the disposition of retaining walls of great length in the zone of the acute angles North in the abutment E-1 (side Palencia), and South in the abutment E-2 (side León). The deck was solved by means of a hollow slab of constant height of prestressed concrete, with a height in the axis of the deck of 1.52 meters and 1.0 meters in the edges. The pillars of the viaduct have circular form, with diameters of 0.9 and 1.1 meters. Their heights are approximately equal and around of 6 meters. The abutment E-1 (low P. K.) is very singular and it was determined by the presence of the Viaduct over Coso stream immediately before, so it was necessary to arrange a unique abutment for the two structures, that allow in one side to dispose the support of the Viaduct over Coso Stream and on the other side the support of the Pergola Viaduct (Fig. 7). The Abutment E-2 is conventional with walls of containment of lands to avoid the inundation of the motorway A-231 (Fig. 8). The foundation is deep trying to look for the rocky stratum. The abutments are founded by means of circular piles of 1.0 and 1.2 meters of diameter, whereas the intermediate pillars need piles of 1250
Figure 6. Aerial view of the A-231 motorway and the viaduct in Pergola in construction.
Figure 7.
Pergola Viaduct over A-231 motorway.
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Figure 8. Transversal section of the Pergola viaduct.
1.0 and 1.5 meters. In the abutments where are the prestressed nerves of the deck, the foundation is solved by means of piles of 1.0 meter of diameter. The lengh of the piles varies between 20 to 28 meters. The soil under the structure was constituted by the next layers: 1. Anthropic landfills with a maximum of 4.00 meters (called QXP). 2. Tertiary sediments until the depth investigated of 30 meters, composed by: ◦ Clay and slightly sandy clay (called TMA1) ◦ Clayey lime (called TMA2) ◦ Sand and limed sand (called TMA3) So the foundation of the skewed viaduct had to be deep, with piles of diameters 1.20, 1.50 and 1.80 meters which must be fixed at least 6 diameters in the materials with consistency or compacity hard and very hard. In the Figure 9 it must be seen the geological section of the soil in the transversal direction to the Pergola. The project of the viaduct was made by the author of this paper, with a three-dimensional model of finite elements using the program SAP2000N, that it includes all the parts of the structure, such as: foundations made with piles (4,500 m of piles), mat foundations tying the piles, pillars, and the superstructure comprising a prestressed concrete slab, and the two complete triangular abutments. It was generated an integral three-dimensional model that included all the elements of the superstructure and of the infrastructure, such as: – – – – – – – – –
16 piles of 1.00 m of diameter and 20 m of length in abutment E-1 48 piles of 1.20 m of diameter and 22 m of length in abutment E-1 24 piles of 1.00 m of diameter and 20 m of length in central pillar 16 piles of 1.50 m of diameter and 20 m of length in central pillar 16 piles of 1.00 m of diameter and 20 m of length in abutment E-2 48 piles of 1.20 m of diameter and 22 m of length in abutment E-2 Slabs of foundations Abutments Prestressed deck, etc. 1252
Figure 9.
Geological transversal section to the Pergola viaduct.
Figure 10. Three Dimensional Finite Element Model of the Pergola from the NE.
All of this constituted a complete three dimensional model of the whole structure that comprises 8,806 nodes, 6,366 FRAME elements and 2,326 SHELL elements (Fig. 10). This model has permitted to obtain all the responses of the Pergola such as moments, shears, axial forces, etc. in all the elements. 1253
Figure 11.
Bridge on pergola over highway A-231 (year 2010).
4 CONSTRUCTION OF THE SKEWED VIADUCT The process of construction of the structure was: – – – – –
Preparation of platform for piling Construction of piles. Making the walls and the pillars Construction of the abutments Finally the upper slab, etc.
In order to construct the prestressed upper slab it was necessary to cut the traffic in one of the senses (two lanes) of the highway and to deviate the traffic temporarily for the other sense (two lanes), in order to scaffold one of the parts of the viaduct without any risk for the traffic, and to prestress the slab of the deck (Fig. 11). REFERENCES Euro code 1.1. Bases of project and actions in structures. Euro code 2. Project of structures of reinforced concrete. Euro code 3. Project of structures of steel. Jiménez Salas, J.A. Geotecnia y Cimientos. Ed. Rueda. Jiménez Montoya, P., García Meseguer, A., Morán Cabré, F. Hormigón armado. 14th Edition. Ed. Gustavo Gili. MOPU. Spanish rules relative to the actions to consider in the project of railway bridges. IAPF-2007. MOPU. Norm of seismic construction: General part and Building Construction. NCSE-O2.
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Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
Advances of external prestressing tendons in multi-span curved box-girder bridges Y. Shen, T.Y. Song & G.P. Li Department of Civil Engineering, Tongji University, Shanghai, China
ABSTRACT: External prestressing tendon design in a multi-span curved box-girder bridge is presented and compared with traditional internal prestressing tendon design. The application of the principle of prestressing torsion resistance is expanded further in curved box girder on the basis of the flexible spatial layout feature with certain limits. The external prestressing torsion calculation formulas with prior consideration of the horizontal alignment parameters are derived out. The horizontal alignment parameters of external prestressing tendons in a multi-span curved box-girder bridge are optimized for best torsion-resistant ability. The results and comparison indicate that the reasonable space layout design of external tendons instead of internal tendons significantly reduces peak torsion and produces a more uniform torsion distribution, and has almost no impact on the moment and shear. The results also show that the numerical results calculated by the formulas agree well with the finite element analysis.
1 INTRODUCTION Curved box-girder bridges, which can overcome the natural limitations of terrain and ground features and satisfy consistency and fluency of route alignment design, have been widely used in urban viaducts, overpasses and high-grade highways (Nakai & Hong Yoo 1988, Nutt et al. 2008). Multi-span curved girders have several advantages compared with single-span system. Less cost of materials can be performed due to smaller bending moments at control sections. It improves structural rigidity and reduces structural deformation for good, smooth driving comfort. Impact effect on bridges under vehicle loads is also reduced for the absence of expansion joints. Curved girders, under coupled action of bending and torsion, prefer box section which has both good bending-resistant and torsion-resistant ability. In addition, Curved box girders usually adopt prestressing technology to improve structural behave and span of bridges. The main difference between prestressing tendon design of curved girders and that of straight girders is torsion-resistant design by using prestressing tendons (Menn 1984). The basic objective of torsion-resistant design in curved box girders is to adjust torsion distribution, reduce peak torsion and finally make girder sections and bearings uniformly forced. Most of current prestressed curved box girders adopt internal prestressing tendons. However torsion resistance of internal prestressing tendons cannot be adequately brought out with the restriction within concrete section of box girder, and it will increase construction cost if extra tendons are configured for resisting torsion. External prestressing tendons have special feature of flexible spatial layout that can be deviated in both horizontal and vertical directions. In this paper, the application of the principle of prestressing torsion resistance is expanded further in the design of external prestressing of curved box girders. Thus extra tendons are unnecessary to be designed for torsion-resistance in multi-span curved box girders. External prestressing tendons in curved bridges are designed not only to resist bending moment but also to resist the torsion under gravity and some of working loads. The formulas of initial torsion caused by external prestressing tendons are derived out according to space analytic geometry, while the secondary torsion by external prestressing tendons is dominant in most cases. 1255
Figure 1.
Component of prestressing and internal force on the section.
The objective of torsion-resistant design process is to minimize the peak of total torsion. The optimal horizontal alignment parameters of tendons can be obtained by non-linear optimization method. The results show that a rational design of external prestressing tendons in multi-span curved bridges can greatly reduce the peak of total torsion and improve the torsion-resistant ability in girders. 2 PRINCIPLE OF PRESTRESSING TORSION RESISTANCE As shown in Figure 1, the resultant force of prestressing F can be decomposed in vertical, radial and axial directions of curved box section. The three component forces are expressed as Fz s , Fhs and F(R + h)θs , where z, h are the vertical and radial coordinates of F; zs = dz/ds, hs = dh/ds, θs = dθ/ds, where θ is the central angle of curved girder, ds is a small element of prestressing tendon; ds ≈ (R + h)dθ, where R is the radius of curvature. The action of F is applied at the centre of gravity (C.G), where M (F), N (F), V (F) and T (F) are prestressing bending moment , axial force, shear and torsion, and respectively the subscripts M and N represent radial and vertical directions of curved box section. T (F) is established according to the torsion equilibrium condition and higher-order parts omitted, as follows:
where z and h are the derivatives of z and h with respect to θ. From Eq. 1, T (F) depends on its parameters F, z, h, z and h . The parameters of F, z and its derivative z which are the vertical alignment parameters of tendons should be determined according to the requirements of bending and shearing, while h and its derivative h which are the horizontal alignment parameters of tendons are invariable in traditional internal prestressing tendon design where tendons restrained by the concrete section. The prestressing tendons in multi-span curved box girders are usually designed to resist the torsion under gravity and some other loads on the girder. Internal prestressing tendons are the most common form in traditional tendons design for torsion-resistant ability. The design of internal prestressing tendons is generally solved in several ways as follows. (1) The vertical alignments of the tendons in the webs are differently designed between outer side and inner side as shown in Figure 2a (Hui et al. 2011, Liang 2009); (2) The prestressed tensile forces or quantities of tendons, that have the same layout between the webs of outer side and inner side, are differently processed as shown in Figure 2b (Jiang et al. 2011, Li & Song 2003, Li & Ren 2005, Yang & Chen 2011, Yang 2010); (3) Extra curved tendons are oppositely designed in the top flange to that of tendons in the bottom flange as shown in Figure 2c. In the method (1), the desired value of T (F) is provided by the adjustment of the parameter z (refer to Eq. 1). The adjustment can influence the bending strength of the girder, and the adjustable 1256
Figure 2.
Layout of torsion-resistant prestressing tendons in multi-span curved girders.
positions of tendons are limited in the section. The method (2) is the most common method in practice. The different prestressing forces in the inner and outer webs are essentially implemented to adjust the parameter h (refer to Eq. 1). The method (3) adjusts the parameter h (refer to Eq. 1) to resist torsion. The additional internal tendons in the top and bottom flanges bring more crossings and collisions. As a result, it will be more difficult to be designed and constructed in practice. Moreover, for traditional internal prestressing tendon design of multi-span curved box girder, there is an inherent problem that webs may crack during tensioning. Thus extra anti-collapse steel must be arranged in practice. As located outside the girder section and unbounded to the webs, there is enough radial room for the layout of tendons in external prestressing tendon design. The position of F can easily change in the radial direction. The parameters h and h (refer to Eq. 1) have a wide adjustable range. For an externally prestressed multi-span curved box girder, there is no more need for special tendons to resist torsion. Instead, the external prestressing tendons originally designed for bending can achieve the effect of torsion resistance by adjusting the radial position.
3 EXTERNAL PRESTRESSING TORSION CALCULATION 3.1 External prestressing initial torsion From Eq. 1, external prestressing initial torsion T (F) is composed of two parts:
where TP (F) = −Fh z/(R + h) and TS (F) = −Fz h/(R + h) are the torsion caused by the horizontal and vertical curve of tendons. The rectangular coordinate system in horizontal is established as shown in Figure 3. The origin is the centre of curved girder. The y-axis is the centre line of the girder. The horizontal projection equation of external tendons between two deviators or between deviator and end-anchorage is expressed as following:
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Figure 3.
Horizontal projection of external tendons.
Figure 4. Vertical projection of external tendons.
The horizontal projection satisfies the geometrical relations which are expressed as:
The radial coordinate h can be obtained by substituting Eq. 4 into Eq. 3:
The rectangular coordinate system in vertical is established as shown in Figure 4. The origin is the intersection of the C.G-line with the centre line. The x -axis is the vertical projection of C.G-line. The vertical projection equation of external tendons is expressed as:
The vertical coordinate z is calculated by substituting the geometrical relation x = R(π/2-θ) into Eq.6:
Then, the torsion caused by the horizontal and vertical curve of external prestressing tendons is calculated by using the equations from Eq. 5 and Eq. 7 in Eq. 2, which is expressed as:
According to the derivation proposed by SALSE.E.A (Salse 1971), and taking the assumption of ds ≈ Rdθ, Eq. 9.1 is simplified as:
Finally, the external prestressing initial torsion is calculated by substituting Eq. 8 and Eq. 9.2 into Eq. 2, which is expressed as:
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3.2 External prestressing secondary torsion Multi-span curved box girder is a statically indeterminate structure. In addition to prestressing initial internal force, the deformation of the girder under prestressing will generate secondary reaction force in redundant constraints. Secondary reaction force will further generate secondary internal force. Above all, external prestressing torsion calculation should take account of prestressing secondary torsion T (F) in multi-span curved box girders. External prestressing secondary torsion is determined by the position and alignment parameters of external tendons. The same as T (F), T (F) is also the function of F, the slope a, c and intercept b, d of the horizontal and vertical projection of external tendons. Prestressing secondary torsion T (F) can be calculated by the force method. By releasing the redundant bending moment on each pivot, the multi-span curved girder is transformed to be two or more substructures of one degree of indeterminacies. Regarding every one degree of indeterminacy as the basic structure, T (F) can be solved on the multi-basic structures. See the references (Zhang 1991, Zhang 1993) for more details. 4 TORSION-RESISTANT DESIGN METHOD OF APPLYING EXTERNAL TENDONS 4.1 Design concept The optimal design of the layout of external tendons is carried out based on the general design concept in prestressed concrete girder bridges, the principle of prestressing torsion resistance, and the calculation of external prestressing torsion stated above. The goal of the design is to provide an appropriate prestressing torsion resistance for design loads without additional tendons. The optimization objective of torsion-resistant design is to minimize the peak of total torsion under prestressing and other loads (such as gravity). The objective function T is the absolutely largest torsion of critical sections in multi-span curved box girder. The optimization object is radial parameter h of critical sections. T is consist of three parts: (1) torsion caused by loads except prestressing T (p), (2) external prestressing initial torsion T (F), and (3) external prestressing secondary torsion T (F) . When the loads is given, T is the function of F, a, c, b and d, the same as T (F) and T (F) . Among these parameters, F and the vertical projection parameters of external tendons c and d are determined according to the requirements of bending and shearing, so T turns to be the function of a and b only. The optimal horizontal parameters of external tendons [ab]∗ can be solved by non-linear optimization method with certain limits. 4.2 Analysis and comparison An externally prestressed three span curved box girder is shown in Figure 5. The radius of curvature R = 40 m. The double-support bearings are arranged at all piers. The section of the girder is singlecell box section, as shown in Figure 6. There are two and three deviators between side and middle spans respectively. The internal force of the girder under gravity is calculated by using finite element software Midas Civil. The prestress force F and the vertical layout of external tendons are designed according to the bending moment and shear envelopes, as shown in Figure 7. External prestressing initial torsion T (F) and external prestressing secondary torsion T (F) is calculated on basis of the above approach, then the total torsion Ti in each control section i is obtained. The objective function T = max{|Ti |} is optimized by nonlinear method with Fminimax function in software MATLAB. The optimal horizontal parameters of external tendons are calculated for the corresponding horizontal layout of tendons which is shown in Figure 8a. Comparison analysis is processed to estimate the advantages of the optimal external prestressing tendon design over the traditional internal prestressing tendon design in torsion-resistant ability. The two comparison models of internal and external tendon design are established. The F and vertical layouts of tendons are the same in two models, while the horizontal layout of internal tendons is arranged along the girder radial symmetrically, as shown in Figure 8b. 1259
Figure 5.
Layout of curved girder.
Figure 6.
Box section (unit: cm).
Figure 7. Vertical layout of external tendons (unit: cm).
Figure 8.
Horizontl layout of tendons (unit: cm).
The bending moment M and shear V under gravity and prestressing is shown in Figure 9. The date is output at the middle of elements. As the vertical layouts of tendons in two models are the same, the bending moment and shear have almost the same distribution. The difference of absolute peaks is within 2% and 9%. Above all, bending moment and shear-resistant ability is basically the same between two models. The total torsion T under gravity and prestressing is shown in Figure 10. The peak torsion in the internal tendon design is 5.78 × 103 kN · m which occurs in the pivots. The peak torsion in the external tendon design is 2.28 × 103 kN · m which fells 61% compared with the internal tendon design. And the distribution of torsion is more uniform than the internal tendon design. The results indicate that the torsion-resistant design with external tendons has an obviously optimal effect. The total torsion T is decomposed into T (p), T (F) and T (F) , as shown in Figure 11. It can be inferred as follows: (1) The torsion under gravity T (p) has little difference between two models which indicates that different prestressing tendon designs have little impact on T (p). 1260
Figure 9.
Bending moment and shear under gravity and prestressing.
Figure 10. Torsion under gravity and prestressing.
Figure 11. Torsion under gravity or prestressing.
(2) The distribution of prestressing secondary torsion T (F) in two models is basically the same, and both peak torsions occur in the pivots. The directions of T (F) are the same as that of T (p) and therefore the secondary torsion has a negative effect on the structure. T (F) in the middle span is nearly linear while T (F) in the side spans has a significant nonlinear distribution, resulted by the interaction of bending and torsion. As a whole, it infers that the distribution of T (F) mainly depends on its support system. 1261
(3) If the internal tendons in the model placed symmetrically (h = h = 0), there would not be any prestressing initial torsion T (F) in theory, namely T (F) = 0. The prestressing loss of the internal tendon in outer side is larger than that in inner side. It is this small difference that makes the prestressing force on sections is unsymmetrical and causes a little T (F). The direction of T (F) is the same as that of T (p) that has a negative effect on the structure. In the external tendon design, as the tendons move inward at the midspan and move outward at piers, prestressing initial torsion T (F) is in the opposite direction to T (p) and T (F) . T (F) is beneficial for the behavior of the structure. Thus it can be seen that through the radial position design of external tendons, the optimized T (F) is playing an important role in the torsion resistance. 5 CONCLUSIONS (1) External tendons with the flexible spatial layout feature outside the box sections, have a good effect on bending and also torsion resistance. Compared with internal tendons, application of external tendons is more efficient and economical for torsion-resistant design of multi-span curved bridges. (2) The calculation of external prestressing torsion and the torsion-resistant design process are deduced and established. The optimal horizontal parameters of external tendons are obtained by nonlinear method. The horizontal layout of tendons with inward deviation at midspan and outward deviation at piers minimize the peak of total torsion in the multi-span curved box girder. (3) The results and comparison indicate that external tendon design instead of internal tendon design significantly reduces peak torsion and produces a more uniform torsion distribution, and has almost no impact on the bending moment and shear. REFERENCES Hui Z, Qin W H & Zhao J S, et al. 2011. Experimental study on a new type of tendon layout in prestressed curved girders. China Civil Engineering Journal 44(1): 121–127. Jiang B, Zhang K & Hu D L, et al. 2011. Influence of prestressing tendon layout on torsion moment of curved beam bridge. Journal of Hefei University of Technology (Natural Science Edition) 34(11): 1706–1710. Li J H & Song X M. 2003. The adjustment of torsion moment for single-column supported continuous curved prestressed beam bridge. Journal of Heilongjiang Institute of Technology 17(1): 21–24. Li X P & Ren H. 2005. Simulation Analysis of the Prestressing Effect on the Curved Box Girder Bridge. Journal of South China University of Technology (Natural Science Edition) 32(12): 47–50. Liang T T. 2009. Test and Research on Pre-stressed Effect about Curved Girder. Chongqing: Chongqing Jiaotong University. Menn, C. 1984. Prestressing of Curved Bridges. Analysis and Design of Bridges. Netherlands: Springer. Nakai, H. & Hong Yoo, C. 1988. Analysis and design of curved steel bridges. Nutt, Redfield & Valentine, et al. 2008. Development of design specifications and commentary for horizontally curved concrete box-girder bridges. Transportation Research Board. Salse, E. A. 1971. Analysis and design of prestressed concrete circular bow girders for bridge structures. ACI Special Publication 26: 714–740. Yang M G & Chen G. 2011. Influence of Prestressing Tendon Layout on Torsion Moment of Long Span Carved Rigid-Frame Bridge. Bridge Construction 3: 15–18. YangY X. 2010. Discussion on Torsion-resistant Design Method for Small-radius Curved Prestressed Concrete Girder Bridges. Technology of Highway and Transport 5: 45–48. Zhang L X. 1991. A Practical Analytical Method and the Torsional Behavior of Curved Continuous Beam Bridges. Journal of Shijiazhuang Railway Institute 2: 15–23. Zhang L X. 1993. Analysis of Secondary Interior Force in PS Curved Continuous Beam. Bridge Construction 3: 25–29.
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Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
Reinvestigation of post–tensioned bridge over Bitlis river B.D. Öztürk, E. Löker, E. Ökte & E. Talıblı Civil Engineering Department of Middle East Technical University, Ankara, Turkey
ABSTRACT: This study aims to reinvestigate the design of one of the longest post–tensioned bridge structures in Turkey. The present study is inquiry of the portion of the bridge with segmental construction procedure. The part of the bridge has a length of 600 m and built using post–tensioning. Thanks to this bridge, transportation difficulty due to long and wide watercourse of Bitlis River will be overcome. The biggest importance of the mentioned project is the application of post– tensioning technique in such a large design to get a thinner decks and longer spans which is an uncommon approach in the construction history of Turkey. By using post tensioning, material, time and money could be saved highly when compared to classical methods. In this structure, segmental balanced console management technique is implemented because of long pier heights. Although this project is a rare sample of post–tensioning technology, hopefully, it will be an important step for mountainous eastern region of Turkey once it is built. 1 INTRODUCTION In the mountainous geography of Eastern Anatolia, several rivers traverse deep valleys and reach the Arabic Gulf. The Euphrates Rivers is a major stream of this network and assemblies a number of smaller affluents. One of the affluents traverses Eastern Anatolian provinces and it is known with the name of Bitlis River. The construction of a long bridge to be installed in the large valley of river parallel to the flow of the river had been planned years in advance and the design project has already been ready. MEGA Consulting Office Ltd. was responsible of the design work actively contributed by FREYSAS Engineering Company Ltd. which is a well – established Turkish – French bridge engineering societies. The present paper has been obtained as a result of an engineering evaluation work formed by the students of the fourth year structural engineering program and guided by the program responsibilities. The work is consisted to realize a structural evaluation of the design project which had been kindly related to MEGA Company. This is a partly study of reinvestigation and evaluate the design from structural view. The regional importance of the large bridge is taken into account which constitutes the purpose of this report has been considered useful and meaningful. 2 ANALYZES AND DESIGN WORK To begin with, the analyses and design work of this bridge is the perfect combination and production of comprehensive hard-working from many different construction resources and applications. In this study French, American and Turkish codes and sources were mainly referenced. As computer programs, MIDAS CIVIL, SAP2000 and OASYS ADSEC were majorly used in computations. These are utilized for the complicated conditions and limitations of field which were given in Figure 1. 2.1 Used materials and design criteria a. As Used Major Materials: concrete, steel reinforcements and steel post–tensioning tendons were chosen as: I. Concrete Grade: C50 according to international standards 1263
Figure 1. The sideview of area in which the bridge is planned to construct.
II. Steel grade: S420 reinforcement steel according to international standards, #32 for longitudinal steel bars and #16 for lateral steel bars and tie bars. III. *Post–tensioning tendons are chosen as: C15 based on (Freyssinet – Standard: prEN 10138-3). *(Nominal Tensile Strength of Tendons: 1860 MPa Tensile Strength of Tendons is considered: 1400 MPa, using Safety Factor 0.75) b. Design Criteria I. In the approach of post–tensioning technique in Bridge Balanced Segmental Cantilever Method, there are a large variety of different views. At this point, French based, Sétra’s Design Guide for Prestressed Concrete Bridges Build Using the Cantilever Method is a significantly reliable source of information. For both the preliminary and ongoing design stages, the mentioned guide is used for the decision of segmental lengths, cross-sectional areas of bridge deck at different points and the criteria of the arc structure for the lower deck. II. Considering the planned project is to be built in one of the Turkey’s first degree earthquake zone on Bitlis River, it has an utmost importance that the piers and bridge deck should be designed according to the requirements of earthquake design criteria, also. For the practice of this significant issue, Design Guide for Improvement in Design and Construction in Turkish Bridge Engineering has been taken as a reference for the issues such as response spectrum and the delivered loads towards the piers. This Reference Guide of Earthquake has also been useful for the decisions of the load combinations on the bridge. As the most critical issues for the earthquake zones, it is suggested by this guide to use Fd = 1.0 Q + 1.0 G + 1.0 E load combination. III. Another approach for moving load cases has driven by the Midas Program, where the analysis for such cases are done according to the AASHTO Standard Design Criteria. IV. To determine the application points of the vehicle loads upon the deck, specifications of the service road should be known such as lane width, length of the refuge, the shoulders, vehicle gaps and truck sizes. As it is in the rest of the Turkey, codes in the General Directorate of Highways is used in this matter. V. To be precise on the serviceability which in such long spanned bridge, the moments and forces greet by the tendons should be carefully evaluated. Thus, Freyssinet C-Range Standard: prEN 10138-3 code is used for the calculations. For the moments to be covered by the tendons are calculated for upper and lower decks differently. 2.2 Design results of members of the bridge based on codes and criteria calculations For such a long bridge span, it is the best interest to construct the lower deck of the spans in arch formation. The shape of the arch at the bottom of the decks is decided to be a parabola since in the most cases the parabola shape is proved to be adequate in practice. To determine the heights of the each segment from both directions, and function of parabola with respect to horizontal length should be found. The heights for this function are decided with a number of simple arithmetic equations according to the code “Setra”. 1264
Figure 2.
Sideview of a half-span with curvature shape and deck dimensions at edges.
Table 1. Deck heights of a half span. Deck No.
h1 (m)
h2 (m)
Deck No.
h1 (m)
h2 (m)
1 2 3 4 5 6 7 8 9 10
8.2 7.7515 7.326 6.9235 6.544 6.1875 5.854 5.5435 5.256 4.9915
7.7515 7.326 6.9235 6.544 6.1875 5.854 5.5435 5.256 4.9915 4.75
11 12 13 14 15 16 17 18 19 20
4.75 4.5315 4.336 4.1635 4.014 3.8875 3.784 3.7035 3.646 3.6115
4.5315 4.336 4.1635 4.014 3.8875 3.784 3.7035 3.646 3.6115 3.6
Figure 3.
Representation deck dimensions & representation pier dimensions and axial reinforcements.
In order to minimize the dead load, the cross-section elements should be chosen after careful evaluations which would be end up with the least load for decks. Considering such understanding, the dimensions found at the design stage were to keep in minimal limits. In the same manner, cross-section dimensions were also calculated according to the mentioned code. In the column design, the primary concern was to build an approximate pier column which would be able to manage to stand high-scaled moment forces occurred by the earthquake and vehicular loads. The compression force by the load of the bridge was not a concern about determining the dimensions, because for carrying the bulk weight of the bridge deck, the necessary cross-section area was significantly smaller. Here are final determined dimensions of a sample deck and pier in Figure 3. 2.3 Loads There were three major loads that active on the structure: Dead loads, Live loads and Seismic loads. In the preliminary design, dead load was taken as 1.18 times of the whole weight of upper part of bridge. In the final design stage, dead load was considered as 1.05 times. The reason of adding 1265
Figure 4. the road.
LRFD design of a heavy truck and load distribution by wheel ranges & the representation of
of 5% is asphalt coverage and some extra material will be placed on the bridge. Live loads were taken into consideration according to vehicles which will pass on the bridge. In this case, normal passenger cars will pass on interior and trucks will pass on exterior lanes of both sides. Gaps among vehicles and trucks sizes were identified based on AASHTO. Finally, seismic loads was taken as Fd = 1.0Q + 1.0G + 1.0E load combination. Analyses were executed for obtained results. Truck size identification & load distribution and the form of the road is given below based on the code as in Figure 4. 2.4 Analyses of the system during construction stage Due to the complicated conditions of the field, balanced cantilever method will be used for constructing this bridge. According to the design, after piers are finished, the construction of the span segments will be started. Each symmetrical segments against the pier are finished simultaneously for not to have an additional excessive unbalanced moment effect. In this project, MIDAS Civil is used for load analysis. For these reasons, in the construction stage, only the dead load is considered and the effects in each construction step are observed. While the construction is being continued, after each step, the shear force and the moment values are seen as increasing. as estimated, the most critical values are analyzed from the last step of the construction, just before the key segment is constructed. Due to the negative moments on the moment distribution diagram of each cantilever, there observed a critical tensile region on the top of the segments. Therefore, post tensioning tendons are placed on those regions, in order not to have any cracks and failure in concrete. The tendons are tensioned and anchored to the crosssection to have a compression effect. In addition, due to the eccentricity, an additional compression effect is observed on the top region of the segment. The moment forces of the dead load of the cantilever from the last stage of the construction are balanced by these two compressive stresses. The calculation is done as follows:
where, M stands for the absolute moment value that created by dead load, I is the moment of inertia of the segment, A is the area of the segment, c1 is the distance between the centroid and the top of the section, e is the eccentricity of the post–tensioning force and P is the post–tensioning force. On the other hand, because of the moment distribution, a compressive region is observed on the bottom of the segment. The values are seen as not critical, yet should be checked as follows:
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Table 2. Required tendon amounts for upper side of decks in half span in the construction stage. Segment
Moment (t*m)
Post tensioning force (t)
Tendon amount
Total tendon amount
1 2 3 4 5 6 7 8 9 10 11 12 13 14 15 16 17 18 19 20
−75362.25688 −67211.16514 −59654.75025 −52669.91437 −46230.69827 −40306.16616 −34867.94597 −29897.21203 −25376.18858 −21281.54332 −17612.16514 −14366.73191 −11495.90724 −8989.650357 −6838.804281 −5035.75739 −3573.87156 −2449.520897 −1653.964322 −1191.295617
12012 11466 10374 9828 9282 8736 7644 7098 6552 6006 5460 4914 4368 3822 3276 2730 2184 1638 1092 546
2 × 13C15 4 × 13C15 2 × 13C15 2 × 13C15 2 × 13C15 4 × 13C15 2 × 13C15 2 × 13C15 2 × 13C15 2 × 13C15 2 × 13C15 2 × 13C15 2 × 13C15 2 × 13C15 2 × 13C15 2 × 13C15 2 × 13C15 2 × 13C15 2 × 13C15 2 × 13C15
44 × 13C15 42 × 13C15 38 × 13C15 36 × 13C15 34 × 13C15 32 × 13C15 28 × 13C15 26 × 13C15 24 × 13C15 22 × 13C15 20 × 13C15 18 × 13C15 16 × 13C15 14 × 13C15 12 × 13C15 10 × 13C15 8 × 13C15 6 × 13C15 4 × 13C15 2 × 13C15
where, c2 is the distance between the centroid and the bottom. Iterative method is used in excel in which the optimum P values are found. The calculated amounts of tendons are shown in Table 2. It should be noted that the most critical moment value of construction stage which is achieved on the top of segments at edges is higher than the most critical one obtained in the case of completed spans. However, for the bottom of segments the highest value belongs in the case of completion of spans. In each situation, tendon calculations and placements are fulfilled based on higher results. 2.5 Cable detailing a. Service life analyses After the key segments are constructed, the adjacent cantilevers are bounded together and form a continuous structure. The behavior of the whole structure and the load effects are observed as altered from the construction stage which are analyzed by Midas Civil. Unlike the construction stage, positive moment is seen in the moment distribution diagram. These regions are the portions close to the middle part of the spans where the critical tension zone is observed at the bottom region. Therefore, for those parts, posts–tensioning cables are required to be placed to the lower deck. A similar calculation to the upper deck is done and the numbers of required tendons for the lower decks are determined. Please note that initially 2 tendons for each side are already placed in order not to have any continuity problems. b. Transverse loading A third scheme forming a critical loading combination in the transverse direction should be studied. The analysis is done by considering a tank crossing incident, which is taken as 62 tons. As far as the longitudinal flexural moments with estimation of 2 tanks on each span, the total load on the deck would not exceed the value of loads at the case of civil vehicles (truck and smaller vehicles). But, a tank could occupy a contact surface on the deck in rather a far distance from central refuge area. In that case, on the upper part of the box section transverse flexural moments will occur. 1267
This will necessitate an additional study of installation of a transversal cable through of the upper slab of the box. The tank that is used for calculations is Leopard-2 which has a length of 9.97 m and a width of 3.75 m while having a weight of 62 tons when fully loaded. The calculations are done per one meter of the span. Therefore, since the weight of the tank is 62 tons or around 620 kN over a length of 10 m which is the length of the tank, the loading is assumed to be 6.2 tons/m. This loading is acting 1.8 m from the support point, which corresponds to the location of the outer lane. Moment diagrams were derived thanks to SAP2000 program. Please note that under these circumstances the most critical negative bending moment turns out to be 231.3 kN*m per one meter of the span. The most critical positive bending moment occurs when there are no vehicles on the bridge. The maximum bending moment under this case is 47.3 kN*m per one meter of the bridge. For both cases, Freyssinet F-Range cabling is considered as specification. Per 1 meter, 3 continuous A4F15 type tendons are used which passes through above to avoid the negative moment and below to avoid the positive moment cases. The spacing is considered as 33.33 cm since 3 tendons are used in 1 m. In this case the post tensioning force is 252 tons per meter which is safe for both cases. 2.6 Details of lateral actions Since Bitlis is a very seismically active area, the seismic analysis dictates the design of the pier section and reinforcements. The selection of the return period of earthquake is very important. A simple risk analysis is performed to select the most suitable return period with the following formula:
T is the return period, t is the life of the structure and p is the probability of the event occurring. Since it is known that the structure is design to have a lifespan of 100 years, a return period of 475 years results in 19% probability of occurrence. Similarly for a return period of 1000 years, the probability decreases to 9.5%. Accepting the risk, an earthquake with a return period of 475 years is used for design and analysis, keeping in mind that structure should not fail under an earthquake with 1000 years of return period even if major damages occur. A response spectrum is created specifically for Bitlis using “Design Guide for Improvement in Design and Construction in Turkish Bridge Engineering”, a guide that was prepared in Middle East Technical University. The obtained response spectrum will be reduced by a reduction factor of R = 1.5, again as a result of recommendations in the same guide. Please also note that, in all computations, cracked section analysis is used by assuming that the moment of inertia of the piers decreases in half once the cracks form inside the piers. For the final design, MIDAS computer software was used with the same response spectra. MIDAS software automatically generates the stiffness matrix, thus the mode shapes. For the analysis, enough mode shapes were used to ensure that the percent mass participation ratio was over 95% for accurate results. After carrying out response spectrum analysis, for an earthquake with a return period of 475 years, the support reactions were collected from the structure. Note that the x direction is along the bridge and y direction is the transverse direction. In the light of the rather recent earthquake with a magnitude of 7.0 in Van, Turkey in 2011 near Bitlis, our team decided to apply the Van 2011 earthquake to our structure as time history loading to compare the results with the response spectrum analysis. The NS direction of the earthquake which has a higher PGA then EW direction is given to the weak x direction of the bridge along the span where the EW direction with rather smaller PGA is given in the transverse y direction. Observe 1268
Figure 5.
Obtained response spectra – Mode periods & the reprsesentation of second mode of the structure.
Table 3. Comparison of Van Earthquake & Design Earthquake. Van earthquake
Design earthquake
Moment demand (kN*m)
Shear demand (kN)
Piers
in y dir.
in x dir.
in y dir.
Pier 1 Pier 2 Pier 3 Pier 4
2.7*10ˆ5 2.47*10ˆ5 2.54*10ˆ5 2.93*10ˆ5
3.36*10ˆ5 7826 5.03*10ˆ5 15563 4.7*10ˆ5 14126 3.62*10ˆ5 10145
Moment demand (kN*m)
Shear demand (kN)
in x dir. in y dir.
in x dir.
in y dir.
in x dir.
14705 12331 13423 14879
5.46*10ˆ5 8.89*10ˆ5 9.25*10ˆ5 4.73*10ˆ5
13376 21311 18624 13722
16554 12960 15377 17419
4.36*10ˆ5 3.66*10ˆ5 4.07*10ˆ5 4.36*10ˆ5
that all of the cases that are created as a result of the time history analysis of Van Earthquake results in less critical support reactions and moment diagram. Therefore, piers should be designed for the more critical response spectrum analysis, carried out with an earthquake with a return period of 475 years. Here are the comparison table with the exact values as in Table 3. 2.7 Pier design After achieving results (shear force and moment value diagrams) by carrying out analysis with the load combination of Fd = 1.0Q + 1.0G + 1.0E in MIDAS Civil, reinforcement issue is solved according to obtained moment demands on the piers. The longitudinal steel bar distribution is assumed for the placement of the reinforcements in Figure 3. For the distribution, in x direction longitudinal bars are placed with 30 cm gaps and in y direction they are placed with 15 cm gaps as demonstrated in Figure 3. In each side 5 cm concrete cover was released for both directions. The size of the reinforcements are determined by analysis and deducted as 32 mm steel bars. Once the reinforcement size is found, OASYS ADSEC software is used to plot Axial load & Moment capacity graphs. Consequently, the capacity results are obtained as higher values even in critical cases in piers. 2.8 Transverse reinforcement design In addition to calculations above, we located lateral reinforcement of piers to make them stable and resistive against shear forces of earthquake impact which are derived from load combination Fd = 1.0Q + 1.0G + 1.0E. During computing of required lateral bars and tie bars, ASCE Standard for Seismic Design of Piers Wharves, AASTHO LRFD Standards and related Turkish Codes were 1269
Table 4. Indication the most critical – pier 2 to overwhelm design earthquake demands with its capacities. Capacities Moment Capacity (kN*m)
Shear Capacity (kN)
Piers
in y dir.
in x dir.
in y dir.
in x dir.
Pier 2
4.92*10ˆ5
11*10ˆ5
45029.24
20146.2
mainly taken in to the consideration. After placing of lateral bars for each longitudinal bars series, tie bars were added due to uniting perpendicular located bars in order to strength structure against lateral impacts comes from both x and y directions. The final results were found for spacing of transverse reinforcements which are chosen 16 mm bars as: 1. Sconfined zone = 15 cm (confined zone is taken constant for all piers and it is 10 m from distance upper and lower ends) 2. Smid span = 20 cm (it is 20 cm for mid spans of piers) 2.9 Foundation Since our study embraces only upper – structure part of the bridge, foundation computations were not fulfilled. As a brief addition, the soil profile of the valley of the mentioned river has different soil types with mainly including sand particles. Consequently, to place this structure on this soil profile safely, liquefaction, pile existence and similar crucial tests should be applied by foundation designers. 3 CONCLUSION In this paper, it was tried to explain most important points of structural details, analyses and design of the enhanced report of reinvestigation of one of the largest post-tensioned bridge in Eastern Anatolia over Bitlis River. All conditional consequences were solved by the agency of improved engineering techniques. One of the latest structural methods such as pre-stressed post-tensioning technique was successfully implemented in the design. By using post-tensioning, material, time and money could be saved when compared to classical methods. Although this project is one of the rare samples of post tensioning technology, hopefully, it will be an important step for mountainous eastern region of Turkey once it is built. REFERENCES Chopra, A.K. 1995. Dynamics of Structures – Theory and Application Earthquake Engineering. New Jersey: Prentice Hall. Civil Engineering Department of METU. 2007. Design Guide for Improvement in Design and Construction in Turkish Bridge Engineering. Ankara: METU publishing. Ersoy, U., Özcebe, G., Tankut, T. 2008. Reinforced Concrete and ItsApplications. Ankara: METU Development Association Publishing. Hewson, N.R. 2003. Post–Tensioned Concrete Bridges: Design and Construction. London: Thomas Telford Ltd. Karaesmen, E. 2012. Post–Tensioned Concrete. 2013. Ankara: METU Publishing. Tonias, D.E., Zhao, J.J. 2012. Bridge Engineering. New York: McGraw Hill Ltd.
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Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
A graphic exact method for analyzing hyperstatic spatial pergolaes A.G. Lacort
University of the Basque Country, Donostia, Guipúzcoa, Spain
ABSTRACT: This study presents a modification of an earlier graphic procedure that proposed an exact method for analyzing continuous beams in First Order Theory without resolving equation systems. The modification introduced enables other, more spatially complex hyperstatic models to be calculated more easily than via conventional, manual calculation methods. Via graphic operations on some determined surface areas with the data from the model, and using drafting instruments, stresses and deformations may be calculated simultaneously. It is also observed that the results obtained without using such instruments are also sufficiently accurate. The procedure could therefore be used in the initial stages of design of such structures to determine their mechanical behaviour. This study also suggests that other graphic methods could be drawn up that would enable different types of building structure to be analysed exactly in just a few operations.
1 INTRODUCTION The graphic procedure set out here is based on an earlier equilibrium procedure (Lacort, 2015) that enabled certain hyperstatic structures to be analyzed exactly and semi-graphically working with nodal actions in the elastic regime and in First Order Theory. That procedure, inspired by Cross’s method, was iterative and obtained exact results for problems without resorting to matrix algebra. To facilitate its application in the manual analysis of continuous beams and some models of spatial pergolas (Figs 2a,b), the present paper proposes an entirely graphic version of the method that uses the hypotheses of manual deformation without considering the torsional stiffness of the beams. All the graphic operations are carried out considering the information obtained in a graph G from the original semi-graphic procedure, and are traced on regular areas using drafting instruments. The results show that if the procedure is applied without using such instruments the analysis time is greatly reduced and the results are still highly accurate, since the perimeter of the surface area where the lines are drawn reduces the possibility of error. Thus, drawing freehand the calculation speed is increased at the expense of only a minimal error that does not depend on the procedure but rather on the tool used. This facilitates the use of the method in the initial stages of design. By way of example a spatial pergola is drawn freehand and the outcome is compared with the exact results. This investigation suggests that it may be possible to calculate graphically and exactly certain hyperstatic structures widely found in building. Lines of research are also observed to open up into the development of new graphic methods that enable other types of structure to be analyzed directly and accurately. Seeking to justify the graphic operations suggested in this paper, the bases of the original semi-graphic procedure are set out below along with some of the formulations that have been adapted for the graphic analysis of the types of structure considered. 2 SEMI-GRAPHIC METHOD 2.1 Approach & general application The original procedure considers that the deflection of a structure is the sum of partial deflections that can be calculated immediately, each of them produced by a set of loads known as the “primary 1271
Figure 1. Graphs for continuous beams & applications: a) continuous beam 1; b) unit primary state; c) deflection; d) graph G; e) resulting G and calculation of the deflection; f), g) continuous beam 2 and alternative graph.
state”. In a primary state (Fig. 1b) there is an “active” action that causes the node where it is applied to move, and also one or more “restrictive” actions that prevent the remaining nodes from moving. Considering that the external loads are the sum of primary states, the method proposes an exact way of determining those states and thus the deflection by means of numerical operations carried out on a graph G obtained from another graph G related to the model, such as those in Figures 1d,e associated with a continuous beam (Fig. 1a). Graph G represents the links between the unknowns in the problem. Its vertices show the rotations θN of the nodes N , each of which is due to a primary state whose active action is valued at one. The value of θN varies depending on its orientation in space and is obtained with (1), (2) or (3). The coefficients of the denominators depend on the bars joined to N and are as shown in Table 1. Their superindexes show the spatial orientation of their bars. If the structures are pergolas (Figs 2a,b), two of the vertices of G can represent the movements in OX or OZ caused by unit primary states, valued at (4) or (5). The coefficients of the denominators depend on the bars joined to the lintel that moves. Finally, each arc of G represents the link between two nodal movements and is valued at the coefficient b1 , b2 , c1 or c2 . Moreover, it is possible to change the value of graphs G to determine deflections and stresses simultaneously. For instance Figure 1f shows a larger continuous beam decomposed into a hyperstatic part and an isostatic one. With the alternative graph G (Fig. 1g) rotations and bending moments can be obtained. XC and XD are defined with conditions of compatibility. The stresses and rotations due to a specific load state are obtained with mA , mB and the rotations θC and θD from the external loads. The positive sign of θC is shown in (6). By applying certain numerical operations to G a new graph G is obtained. With this it is possible to determine any deflection semi-graphically because the values of its vertices and arcs
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Table 1. Bar coefficients. Parallel bar
b1
b2
b3
c1
c2
c3
To OX To OY To OZ
2EI z /L 2EI x /L 2EI y /L
3b1 /L “ “
2b2 /L “ “
2EI y /L 2EI z /L 2EI x /L
3c1 /L “ “
2c2 /L “ “
Figure 2. Semi-graphic analysis approach for structures that can be calculated graphically: a) pergola 1; b) pergola 2; c) criteria with positive signs for actions, reactions and movements; d) graph G of pergola 2; e) link between sub-graphs; f) link between unknowns in movement; g) graph G .
stablish a link between the movements of all the nodes. By way of example, the deflection of the continuous beam in Figure 1a is calculated. It is obtained from the partial rotation θA1 . Multiplying this rotation by an “ascending” coefficient with a value of ABθB it is possible to determine the partial rotation θB1 that goes with θB2 to form θB . Finally, multiplying θB by a “descending” coefficient with a value of ABθA the rest of the rotation of A is obtained. These operations can be related to the mechanical performance of the model. They are equivalent to those used in the Gauss method to resolve the system of equilibrium equations. It can become extremely difficult to obtain G when the form of G is complex. That is why the graphic method is applied here to structural forms that may be hard to analyze manually but are linked to simple G graphs. The procedure is easy to apply considering G, and may therefore prove useful for the initial stages of the creation of structures. Moreover, it is a visual method, as are other graphic methods available. According to Huanyun & Chunhui (2013) such methods can play an irreplaceable role 1273
in developing the skills of future technicians. Applied to continuous beams, the procedure set out here also formally improves on the method proposed previously (Lacort, 2014) because it requires fewer lines to obtain the same results. Lines are traced on regular surfaces where there is an area referred to as the “deflection area” on which the deflection is obtained. 2.2 Application to pergolas Figure 2d shows the form of the graphs G linked to the most complex types of structure envisaged in this study: corner pergolas (Fig. 2b). The graphs for the other types considered (Figs 1a, 2a) are included in this form. The vertices of G aligned horizontally represent the rotations of the nodes oriented in the same directions in space, as shown in sub-graphs S x , S y and S z . The vertices aligned vertically represent the movements of a single node, and those connected by vertical arcs show the movements of the corner node. The following section seeks to obtain G and the deflection graphically by using the semi-graphic procedure without increasing the number or arcs. To that end G is interpreted as a combination of sub-graphs in the form of branches (Fig. 2d). The first step is to determine the deflection areas for S x , S y and S z . Next it must be determined what influence these sub-graphs have on movements (Fig. 2e). This determines sub-graph Sδ (Fig. 2f), whose deflection area is then also obtained. Finally, a graphic link is established between all the unknowns (Fig. 2g) which is equivalent to graph G . All the operations are obtained by tracing dotted lines sequentially on regular surfaces, reproducing the numerical calculations performed previously. The results are accurate if drafting instruments are used. 3 GRAPHIC METHOD 3.1 Calculating the deflection area of a sub-graph S in the form of a branch The deflection area of a sub-graph S is an area delimited by two unit squares where a set of lines are drawn. The lines are the diagonals α, β of these squares and other vertical straight lines where the movements of the nodes are calculated. These straight lines are obtained from operations performed on squares I , II, III and IV considering the arcs r and the vertices θi of G, arranged around the centers O (Fig. 3a). The procedure is explained below by setting out the calculations for the deflection area for the beam in Figure 1a and its deflection. The procedure is conducted four stages based on the layout in Figure 3b. 3.1.1 Stage 1 (in areas I & II): Obtaining the vertices of G (Fig. 3c) θA is valued at θA because it is the first vertex considered, and θB is obtained from it and from AB as follows: θA is located in area I , and the dotted line is drawn for vertices a, b, c and d. Dimension bd is the product θA θB AB2 and is located in the top left corner of area II. This serves to determine the straight line fg, whose parallel through h intercepts the left side of this area at i, thus determining θB . When there are more than two rotations arc and rotations θi from the model are positioned and the above procedure is applied gradually between two consecutive rotations of G. Figure 4a shows the operations when there are three rotations and when the values of the arcs and vertices of G are not the same. If the vertices of G represent movements δx , δz , (Fig. 2f) their values are several times the unit value, which makes it difficult to carry out these operations. To avoid this, we suggest varying the proportions of the initial data so that the operations can be performed more conveniently without changing the results. This can be done by dividing δx and δz by a coefficient K greater than one, multiplying arc XZ by 0.5 ABK 2 in I and expanding XZ in III and IV (Fig.4c). 3.1.2 Stage 2 (in areas III & IV): Obtaining ascending and descending coefficients (Fig. 3d) Position θA and θB , obtained from squares I & II. From them, draw straight lines sA and sB which, together with AB, determine the segments mn and kq, whose lengths match the value of the coefficients sought. 1274
Figure 3. Obtaining the deflection of a sub-graph S : a) areas of application for a general case; b) areas considered for a continuous beam with two nodes; c) obtaining θB ; d) obtaining the ascending and descending coefficients; e) resulting deflection area; f) calculating a deflection (values depend on EI ).
When there are more than 2 unknown rotations, the values of the arcs & vertices of G must be positioned as shown in Figure 4b. In IV the ascending coefficient between A and B is determined, which works out to the length of segment kq. From this, the ascending coefficient between B and C is obtained: the result is k q . Similarly, in III the descending coefficients mn and m n are gradually obtained. If the values of θi and ri are very small it may be hard to calculate these coefficients. To prevent this problem the first coefficients calculated in II and IV, i.e. kq can be enlarged and the rest can be obtained from there (Fig. 4d). To offset the change of scale the rotations of the ends on the deflection area (Fig. 4e) can be divided by K, and this can be taken into account in the final results. It appears that sub-graph deflections such as S i (Fig. 2d) can be calculated with great accuracy even doing away with the operations in stage 1. Thus the ascending and descending coefficients are obtained with the vertices of S i . 3.1.3 Stage 3: Obtaining the deflection area (Fig. 3e) This is obtained by locating the straight lines A1 , A2 , B1 and B2 on the diagonals. These lines indicate the places where the partial rotations θA1 , θA2 , θB1 and θB2 , respectively, are calculated for any load state. A1 and B2 are determined by shifting the segments kq and mn horizontally from areas III & IV to diagonals α & β. Since the sign of AB is negative, the signs of θB1 and θA2 are opposite to those of θA1 and θB respectively. 3.1.4 Stage 4: Obtaining the deflection (Fig. 3f) Start by taking θA1 = mA θA in A1 . Draw s and its interception of B1 gives θB1 , which is negative. Then, in the top square, calculate the rest of the rotations in the same way. 1275
Figure 4. Operations when the sub-graph has three unknowns: a) in stage 1; b) in stage 2; c) change of scale of the initial data; d) and e) change of scale of the operations in stage 2.
3.2 Calculating the deflection area of Sδ To show the procedure more clearly, the deflection area of the Sδ associated with the pergola in Fig. 2b is determined below, formed by equal bars of 6 m in length with square cross-sections, the graph G for which is shown in Figure 6a. The coefficients c2 & b2 of the bars are valued at s and are designated in the graph by s1 , s2 , s3 & s4 . x (Fig. 2f) is valued in (7) and its terms εi represent y the influence of S z and S y (Fig. 2e). ε2 is valued in (7) and θi represents the rotations obtained in y y y the deflection diagram for S when partial rotations θi1 , valued at θi , are applied. x z Figure 5a shows the area for graphically determining , and the arc XZ, whose value is obtained numerically with (9). The area is formed by six squares on whose sides the values of the arcs si and the vertices δx & δz are arranged. They can all be modified without altering the final result by applying a coefficient K greater than one when δx and δz are too large for graphic y operations to be carried out. Begin by arranging the sum of the rotations θiz , θi and θix modified by K on the straight line rr . Those that make up each sum are like those that appear in (8). Using them in I , II , II and I it is possible to determine ε1 , ε2 , ε3 and ε4 respectively (Fig. 2e). For example, ε2 has the value of segment db (Fig. 5b) obtained in II by drawing the line a-b-c-d. Also in II XY is obtained, which has the value of b-e. In III and III x /K & z /K are obtained using a procedure similar to that used in areas II (Fig. 5c). With these values and using the procedure described in subsection 3.1 the deflection diagram of Sδ (Fig. 2f) is determined using the areas in Figure 5d. The vertices and arc of Sδ can also be modified with K so that the results match those of G .
The deflection may be determined as in Figure 1e by calculating each nodal movement as the sum of two ones. Begin by determining the partial rotations θi1 in the deflection diagrams of S x , S y , S z if there are external moments. These rotations produce forces on OX and OZ which are obtained on the upper sides of areas I , II , II and I and which are added to the external actions 1276
Figure 5. Graphs to analyze a pergola (depending on EI ): a) to obtain x , z and XZ; b) detail 1; c) detail 2; d) to obtain x , z and the movements; e), f) supplementary diagram equivalent to G .
to determine the movements in the deflection area in Figure 5d. Finally, calculate the rest of the deflection by means of another diagram that connects all the sub-graphs (Fig. 2g). The diagram is made up of a triangle ABC whose side BC is of unit length (Fig. 5f). Between AC and AB arrange y z and θiz , θi and θix as they appear in Figure 5a. Add a square to the triangle where the value s of the arcs is located. By way of example Figure 5e shows the calculation of partial rotations depending on z . 4 EXAMPLE OF CALCULATION In Figure 6 a freehand calculation of the vertices and arcs of graph G of the above pergola based on graph G (Fig. 6a) is shown. All the results depend on EI. The vertices of G linked to S z and S x are determined in Figure 6b and those linked to S y in Figure 6c. In both cases it is necessary to increase the size of the operations via the procedures shown in subsections above. In line with these results the arc XZ∼0,004EI and the vertices of Sδ (Fig. 6d) are determined using a coefficient K = 4 to reduce the original values of δ . Finally, Figure 6e obtains z ∼ 4.25/EI and the deflection area of Sδ . The exact values of XZ and z are 0.045EI and 4.19/EI, respectively. 5 CONCLUSIONS Based on this research, the possibility emerges of calculating certain hyperstatic structures commonly found in building graphically and exactly. 1277
Figure 6. A freehand calculation of a pergola: a) graph G; operations performed: b) in S z & S x ; c) in S y ; d) in Sδ ; e) obtaining of x , z & XZ.
Possible lines of research are also pointed to in the developing of further graphic methods that can analyze other types of structure directly and accurately. ACKNOWLEDGEMENTS I would like to thank my friend José Múgica, Manager of TRADUTECNIA and Chris Pellow for the help that they have provided in translating this manuscript into English. REFERENCES Huanyun, W., & Chunchui, P. 2013. Teaching reform of Engineering Graphics on the relationship between hand drawing and computer drawing. International Conference of Information, Business and Education Tecnology, Beijing, March 2013. Lacort, A.G. 2015. Control semigráfcio de análisis elástico para ciertas estructuras de edificación. Inf. Tecnol. 26(3), in press. Lacort, A.G. 2014. Algunas estrategias para analizar gráfica y exactamente ciertas estructuras de edificación. Congreso Internacional de ACHE. Madrid, julio 2014. Madrid.
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Multi-Span Large Bridges – Pacheco & Magalhães (Eds.) © 2015 Taylor & Francis Group, London, ISBN 978-1-138-02757-2
Investigation on stability problems as a second order theory problem for piers with practically infinite bending stiffness V. Karatzas & G. Karydis
Alkatech Consulting Engineers, Athens, Greece
E. Karatzas Roussou
Karatzas Consulting, Athens, Greece
T. Konstantakopoulos
Konstantakopoulos Consulting Engineers, Athens, Greece
ABSTRACT: Foundation failure is governed by excessive deflection. Therefore, it is evident that determining the deflection that occurs under certain loadings proves to be a handy tool in assessing a structure’s behavior. In this paper, the calculation method is shown of the Plimit load that a structure can undertake, taking into consideration its actual deformations, boundary conditions of the restraint supports (foundations), and the construction imperfections using spring rotational constants. Note that the Plimit load is less than the Pcritical joint load due to the existence of second order theory moments. The method presented defines the actual resistance for a structure with compression members whose slenderness is (λ) ≤ 25 after the old DIN 1045, commonly found in seismic regions i.e. strong piers and their respective foundations.
1 INTRODUCTION During design, while dimensioning a compression member engineers usually check the member against Euler’s critical load. This criterion always holds true when the slenderness ratio of the compression member is small (λ < 25) and its bending stiffness (EI ) is infinite, i.e. large shear walls or piers. Unfortunately, this check does not ensure that the structure is failure-proof. Hence, it was proposed to introduce an additional check that of the Pcritical joint load . However, the Pcritical joint load is a stability check directly related to the overall structural shape/form of the structural system. It does not ensure though, that the individual structural members and/or their restraints have the required resistance capacity in case a boundary limit is reached. In fact, individual structural members of a structural system may fail under a particular load that is less than the Pcritical joint load . Hence, it is of interest to determine the member’s and/or its restraint resistance capacity. In other words, the aim of this paper is to determine the actual limit load in relation to deflection. 2 STRUCTURAL ANALYSIS Considering Fig. 1 above, the following equations are derived:
where ψn = deflection angle of a member; Pn = member load; h = height; and c1 = spring rotational constant. 1279
Figure 1. Two member structural model.
Thus, for n number of members:
Solving further, the following equation is derived:
Due to very small rotational angles, the following is true:
From this homogeneous system of equations when the matrix denominator is equal to zero (N = 0), one can determine Pcritical joint load . Next, from equations 4, 5, 6 . . . one can determine the rotational angle ratio i.e. ψ1 /ψ2 , . . . , ψn−1 /ψn . In other words, one may compute the minimum failure load and its possible respective structure’s deformation mechanism figure. From these calculations one obtains the minimum deflection angle ratio for a Pcritical joint load , but not the actual absolute value of each deflection angle. Thus, it is classified as an eigenvalue problem. However, there are cases, where it is of importance to estimate the limit range of the magnitude of the deflection angles, in order to assess the structural safety limit and/or their possible failure mechanism. These are the cases where one needs to know the actual limit load. Hence, rearranging the initial equations it is obtained that:
Since the original equations contain circular functions (sine and cosine), these are generally solved by iteration. Thus, the actual magnitude/size of the angles is estimated theoretically like this: from equation (9) for an initial estimation value of ψ10 , angle ψ20 is computed. Then, using 1280
equation (10) ψ30 is computed. Similarly, ψ40 is found using equation (11). This procedure is repeated until the two values for the angle ψn0 obtained from either side of the equation are almost the same. In the case that the two values are not acceptably close, the procedure is carried out again until the outcomes match. This procedure may also be carried out with the use of a spreadsheet. The questions that rise are how does one estimate the initial value ψ10 to begin the iteration procedure and what is an acceptable deflection interval (ψ10 − ψ11 ) for two consecutive values. That is problem that the designer has to assess each time. 3 ESTIMATION OF THE DEFLECTION ANGLE RANGE FOR A CERTAIN STRUCTURAL LOADING 3.1 Discussion on equation (13) The system of equations (5) . . . (8), has more than just one apparent solution (ψ1,n = 0) that defines the eigenvalues of the system. But, only the critical eigenvalues are of interest to the designer. From the boundary conditions of the structure, i.e. allowable soil stress, one estimates the magnitude of the deflection angles for that particular loading condition. Thus, based on the structural loads, the bending moments acting upon the foundation of the structure are obtained. Then, from the known geotechnical relationships, one determines the allowable soil stress for the maximum allowable moment. The difference between these two moments M = Mmaxallowable − Mactual results in a range of additional moments (second order theory) and within this range lies an interval with the respective deflection angles. Therefore, two possible variable values for the deflection angles may be obtained. In the example presented below, a method for solving this type of problem is shown. The calculation of Plimit load is achieved using the following equations that have been derived from rearranging the equations presented afore. Hence,
Moreover, for n number of members:
where ψn = deflection angle of a member; Pn = member load; h = height; and c1 = spring rotational constant. 3.2 Investigation of equation 13 Equation (13) offers significant information for the assessment of structural behavior. Rewritting equation (13),
It is found that, the ratio of the deflection angles ψ2 /ψ1 depends upon the constant number 1, the ratio of spring rotational constants c1 /c2 , and the ratio of the total axial loads times the height of the 1281
Figure 2.
Piers and their respective foundations with their points of failure. Table 1. Computed results for spring rotational constants. Spring rotational constant c1 c2
11 m foundation
7 m foundation
9.29 × 1012
2.095 × 1012
pier 1.0125 × 1012
first structural member with the respective spring constant c2 regardless the number of structural members (or storys). Hence, during the design process it is quite easy to assess a future behavior, since it is quite easy to determine c1 , c2 , h1 , P1 , . . . , Pn . From these values, one may compute the deflection angle ratio ψ1 , ψ2 that prevails in the structural failure assessment. It is also found that when the rotational spring constant ratio is c1 /c2 ≤ 1 that means c2 > c1 then, a smaller deflection angle ratio results and hence, the convergence of the values for the homogeneous system ψ1 , ψ2 , . . . , ψn is achieved at a safer and faster rate. 4 EXAMPLE A paper (Drosos & Georgakos & Anastasopoulos & Gazetas, 2010), presents the behavior and the points of failure of two piers with respect to their foundations as tested in different experiments that took place in the National Technical University of Athens (see Fig. 2 below). Now, for this case study an effort is made to explain why the pier in the first case and the foundation on the other case failed the way they did using the term Plimit load discussed in this paper. Assume that the pier and its respective foundation act as a two member model, where the foundation is member 1 and the pier is member 2. Also, consider as a given that the superstructure’s loads are 12,000 kN and the pier’s loads are 2200 kN. First, the spring rotational constants are found using the formula below (Petersen, 1982):
where c = rotational spring constant; a = foundation length; b = foundation width; E = soil elastic modulus; i = foundation depth to foundation width variable; and k = foundation length to foundation width variable. Next, the respective Pcritical joint load values are determined using equations (5) and (6). Note that the actual computations are not presented (due to space limitation). It follows that equations (5) 1282
and (6) become:
And the following 2nd order equations result: 0.296 · h2 · P 2 − 0.142 · h · P + 0.11 · c2 = 0 for the large foundation and 0.296 · h2 · P 2 − 1.548 · h · P + 0.48 · c2 = 0 for the small foundation. Hence, for the large foundation: Pcritical joint load = 770,000 kN, ψ1 = 0.09945 ψ2 and ψ2 = 10.0548 ψ1 Similarly, for the small foundation: Pcritical joint load = 570,000 kN, ψ1 = 0.227 ψ2 and ψ2 = 4.403 ψ1 Next, the designer must check if the natural frequencies of the structure are close to the resonance frequency that produces additional dynamic loading. The natural frequence equation shown below is applied:
where ω = natural frequency; g = gravitational force; and h = height. Thus, the following results are obtained: ω = 6.06 sec−1 for the large foundation and −1 ω = 4.95 √ sec for the small foundation.√Hence, for a dynamic loading with a frequency greater than = 2 × 6.06 = 8.54 sec−1 and = 2 × 4.95 = 7.00sec−1 respectively, the structure is not be affected by dynamic loading due to resonance. Only static loads act upon the structure. Moreover, the response of the structure to dynamic loading results in oscillations with an amplitude, sdynamic , that are lesser in comparison to the amplified oscillations due static loading (sdynamic < sstatic ). 4.1 Evaluation of the deflection angle interval under a certain loading: Large foundation The largest deflection ψ1 of the system occurs when the axial force resultant will cause the soil stresses greater than the allowable ones. From boundary limits such as the allowable soil stresses one may estimate the possible size of the deflection angles for that particular structural loading. Assume an allowable soil stress of 250 kN/m2 . Since the resultant lies beyond the internal core of the foundation, but within the external core then solving further for the eccentricity e, results in: e∗ = a/2 − e = 11/2 − 3.44 = 2.07 m therefore, the allowable moment capacity for this foundation for that given soil stress is: M = (12000 + 2200) · 2.07 = 29394 mkN. The respective deflection angle α∞ for circular section foundations is obtained from following equation found in DIN 4019 Part 2 where e ≤ r/3 and solving results in α∞ = 0.004627. This is the maximum allowable system deflection for loads found using first order theory (a nondeflective system). However, due to second order theory (deflective system) the iteration should be confined to an interval between zero and maximum allowable deflection, otherwise surcharge pressure prevails and failure occurs. If one now wishes to narrow down this interval to shorten the iteration process, then the deflection angle that develops due to the actual must be evaluated. Assume axial loading with an eccentricity: i.e. 1.83 meters. This means that the axial loading lies near of edge of the internal core of the section and as such, a moment M equal to 26023 mkN is produced. It follows that, ψ1 = 0.004627 − 0.00028 = 0.004347 In other words, the deflection angle interval ranges from 0 to 0.004347 and with the iteration method, after 12 loops it is obtained that: ψ1 = 0.00002 and ψ2 = 0.00020342. The additional second order theory moments are: M1 = c1 ψ1 = 1858 mkN, M2 = 2060 mkN and ψ1 /ψ2 = 10.1711. Thus, the total moments are: M1 = 26023 + 1858 = 27881 mkN < 2994 mkN. Note, that they are less than the allowable moment found earlier and hence, there is no failure of the foundation as 1283
verified by the experiment. On the other hand, for the pier the actual moment was found to be 20818 mkN. The sum of the moments is 22878 mKN > 20818 mkN. Thus, the pier fails due to a surcharge of second order theory moments. 4.2 Evaluation of the deflection angle interval under a certain loading: Small foundation Acting accordingly, for comparison reasons assume the same type of loading on the structure and soil stress with a small foundation, inducing an eccentric axial loading then, e = a/6 = 7/6 = 1.1667 < 1.83 m < 2.3334 m. These axial loads are found within the external section’s core and so the force resultant lies at a distance of 1.67 m from the edge of the section. The moment produced is the same as before (26023 mkN). The soil stress was found to be 810 kN/m2 > 250 ⇒ unacceptable. Therefore, the foundation fails as verified by the experiment, despite the fact that the deflection angles may be quite small. The pier suffers from no damage whatsoever.
5 CONCLUSIONS It is concluded that special consideration must be given during the design of a structure to the evaluation of the deflection ratio. This paper investigates a method for solving second order theory stability problems with the help of spring rotational constants, for loads smaller than Pcritical joint load , that due to the existence of second order deflections structural failure may result. This method deals with actual loads that define the true load limits of a structure with compression members whose slenderness ratio (λ ≤25), i.e. large shear walls or piers of large span bridges, commonly found in seismic regions due to regulations’ requirements. From the investigation of equation (13) it is deduced that, in seismic regions where large axial accelerations may occur, i.e. significant fluctuations of axial loads, structures fail at a greater extend. That is because as observed from equation (13) a reduction in the axial loading results in an increase of the ψ2 /ψ1 ratio in relation to the increased lateral loading. Hence, it goes without say that the ψ2 /ψ1 ratio is the principle factor for evaluating structural failures. A case study was used to present the application of this proposed methodology for dealing with stability problems. The experimental outcomes obtained from the National Technical University of Athens were verified by the analysis presented in this paper. Namely, in the case of a large foundation, the pier failed at the base due to a large second order moment that created a respective large deflection, although Pcritical joint load was not reached. There was no failure of the soil (only small stresses developed). On the contrary, for the case with the small foundation, the soil crushed due to the formation of large stresses, while smaller moments formed at the base of the foundation and yet again, Pcritical joint load was not reached.
REFERENCES Drosos, V. & Georgarakos, P. & Anastasopoulos, I. & Gazetas, G. 2010. Experimental Validation of Bridge Pier Seismic Design Employing Soil Ductility Karatazas, Elisabeth & Karatzas, Velvet & Karydis, George & Konsttigatantakopoulos, Theodore. IBSBI 2014. Investigation on Stability Problems as a Second Order Theory Problem for Piers with Practically Infinite Stiffness, Greece. Karatzas, Velvet & Karatzas, Elisabeth, fib 2010. Instability Problems – Investigation of P critical joint load under moment loads Karatzas, Velvet & Karatzas, Elisabeth & Karydis, George, STESSA 2012. Investigation of structures whose slenderness ratio is λ ≤25 and is based on P critical joint and the eigenvalue ratio Petersen, Christian. 1982. Statik und Stabilität der Baukonstruktionen, Vieweg: pp. 910–912
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BIBLIOGRAPHY Drosos, V. & Georgarakos, P. & Anastasopoulos, I. & Gazetas G, 2010. Experimental Validation of Bridge Pier Seismic Design Employing Soil Ductility, Volos Greece. Karatazas, Elisabeth & Karatzas, Velvet & Karydis, George & Konsttigatantakopoulos, Theodore. IBSBI 2014. Investigation on Stability Problems as a Second Order Theory Problem for Piers with Practically Infinite Stiffness, Greece. Karatzas, Velvet & Karatzas, Elisabeth, fib 2010. Instability Problems – Investigation of P critical joint load under moment loads, USA Karatzas, Velvet & Karatzas, Elisabeth & Karydis, George, STESSA 2012. Investigation of structures whose slenderness ratio is λ ≤25 and is based on P critical joint and the eigenvalue ratio Petersen, Christian. 1982. Statik und Stabilität der Baukonstruktionen, Wiesbaden: Vieweg.
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Suspension cables bridge and arches L.M. Laginha Instituto de Engenharia de São Paulo, Divisão de Estruturas, Brasil
ABSTRACT: The behavior of Suspension Bridges without stiffness girder under action of a moving load is analyzed theoretically and with experimentation. The traditional assumption of the inextensibility of the suspension cable is replaced by the Equality of the Areas, explaining what really happens with this formulation. The length of the parabolic cable is calculated in an original way, only by his forces, solving the problem with the compatibility of deformations. The behavior of the arches is analyzed together with the cables, generalizing the application of the formulation Equality of the Areas.
1 INTRODUCTION This paper is a summary of the Master’s Thesis of the author supervised by Prof. Hideki Ishitani at the Polytechnic School of the University of São Paulo in 1997. We define the reference configuration as corresponding to the balance of the permanent load w, and after the cable receive accidental load P, the displaced configuration. The balance of the suspension cable in the reference configuration (Fig. 1) in the generic x section:
results in the cable equation in the reference configuration:
The above equation represents the reference setting, that is, the function of the position of the dead load and the horizontal component Hw from Tw , the tensile force. By hypothesis, the cable’s own weight was considered evenly distributed horizontally and added to the own weight of the board, resulting the dead load w.
Figure 1. The reference configuration.
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Figure 2. The offset configuration where kl is the abscissa of equilibrium of the load P.
This is the study of the suspension cable only submitted to a concentrated load P on the general equilibrium abscissa kl (Fig. 2 – offset configuration). For this, we use two Cartesian systems oriented vertically from a line linking the anchors down, and horizontally, the anchors to the concentrated load, respectively. 2 EQUALITY OF THE AREAS Assuming zero the work of the permanent load w to be applied to accidental concentrated load P (Fig. 2), it follows the classical expression for the HP value. In fact, consider null the work of the dead load w along the cable, implies that the integral of displacement v = yw − y is zero. Mathematically:
This implies that the Aw area, enclosed by the reference configuration, and the area A, encompassed by the displaced configuration, are equal. Therefore, matching Aw = A we obtain the classical value of literature:
The Equality of the Areas has the same meaning as the vertical immobility of the center of gravity of the dead load. In Figure 5 are shown plots of vertical displacement of the cable to a concentrated load obtained from experimental forms. 3 EXPERIMENTAL ANALYSIS Two experiments were conducted in the Laboratory of Structures and Structural Materials of the Polytechnic School of the University of São Paulo. A cable with 1.5 mm nominal diameter formed by 19 galvanized steel wires of 0.3 mm was charged with 19 weights of 0.625 kgf = 6.25 N. Resulted in a load w uniformly distributed horizontally 0.10 kgf/cm = 1 N/cm, simulating the dead load 1288
Figure 3.
Layout and experimental structure with dial mm hundredth. LEM-EPUSP, 1994.
Figure 4. 1994.
Layout and experimental structure with electromagnetic transducers “HP DC-DT”. LEM-EPUSP,
of the board. The range of 125 cm was divided into ten segments, resulting in 9 stations for the measurement of displacements in the model. The adopted fw construction sag was 12.5 cm, φ = P/W and W = wl. With the hypothesis of Equality of the Areas, there are points whose abscissas have no vertical displacement (Fig. 5) and depend only on the equilibrium position x = kl of the load P. Fixed 1289
Figure 5. The theoretical fixed points in the central graph, and the experimental vertical displacements v.
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Figure 6.
Point is defined as a point of reference yw setting where it crosses setting the shifted y (yk or yj ). These abscissas, result v = 0, does not correspond more to the same physical point on the wire. The abscissas that locate the Fixed Points obey the following law:
4 NEW FORMULATION OF THE SUSPENSION CABLE LENGTH 4.1 The length formula deduction To determine the suspension cable length, we adopt the notation: T – traction force on the cable support; R – vertical component of T on support; H – horizontal component of T , invariant in x and w – vertical load evenly distributed horizontally. Below, we will calculate the length of the cable only submitted to w dead load, which is why we will omit the indices w (T indicates Tw , H indicates Hw , V indicates Vw , etc.):
this is the exact parabolic cable length formula, highlighting the reaction forces in the funicular polygon. 4.2 The elastic deformation of the cable After applying the concentrated load P, each cable element ds, will bring with it a ds end deformation, due to the final traction force T installed on the cable. Thus, the total stretch in the cable s is: calculating, results in:
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Figure 7. Comparison of Central Displacement with the Equality of Areas, Elastic, Elastic with EA/2 and Inextensible.
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where sw is the total elongation of the cable in the reference configuration. As sP is the deformation due to the load P, we have: sP = s − sw . Applying the above deducted formulations and Aw = wl 3 /(12Hw ) results:
where sP is the deformation increased due solely to the application of the load P. The elastic elongation was calculated by the expression (7). It was solved by the program “Mathematica”, and obtained the value of HP = βHw . – The problem of compatibility of the axial deformation of the cable is resolved equaling the length difference sP = s − sw ; – The effect of varying the cable EAc axial stiffness was verified by comparing the values obtained with EAc and EAc /2; – The vertical displacements was calculated with the hypothesis of Equality of Areas; – Finally, determined the displacement obtained under the assumption of inextensibility with: s = sP − sw = 0 through the Mathematica program. Adopted for the cable: E = 1.6 E6 kgf/cm2 Ac = 0.013 cm2 and EAc ∼ = 20.000 kgf. 5 CONCLUSIONS 5.1 The term Equality of the Areas exactly means the area encompassed by the reference configuration and the displaced configuration, respectively, or also the immobility of the center of gravity of w, the dead load. It should be noted, that this is not the cable inextensibility condition, usually adopted by the literature; 5.2 Analyzing the curves shown in Figure 8, it can be said: the vertical displacements of the suspension cable are mainly due to the change of cable equilibrium line, with little influence of the variation of axial stiffness EAc . We can conclude that the analyzed phenomenon shows little sensitivity to changes in the value of the axial stiffness of the cable; 5.3 The formulation of the Equality of Areas can be applied to solve the Biarticulated Arc. So HP is obtained, equation (4), ignoring the compatibility of the axial deformation of the cable or arch. The values of “Leveraged Moments by Displacement” vH in the cable are identical to the values of the bending moments M in the analogue biarticulated arc. Cables
Arches
Equation of displacement: (9) MP = yw HP + vH where v is the vertical displacement of the cable. Integrating the equation along the span l: ∫ MP dx = ∫ yw HP dx + ∫ vHdx Apply Equality of Areas for the displacements: ∫ vHdx = 0
Equation of flexural moments: MP = yw HP + M (10) where M are the flexural moments on the arch. Integrating the equation along the span l: ∫ MP dx = ∫ yw HP dx + ∫ Mdx Apply Equality of Areas for the moments: ∫ Mdx = 0
5.4 With respect to stiffness, the heavier the deck bridge, i.e., the greater the span l or w dead load, the cable will be more rigid. The vertical displacements are directly proportional to fw construction sag in the formulation of the Equality of the Areas. In this approach, the vertical displacements not depend directly on the span l (l, the span, is hidden in the load factor ϕ = P/W , since W = wl), as can be seen in the following formula for the displacement of the central point:
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5.5 The formulation for the length of the parabolic cable, according to its reaction forces (6), does not need the geometry of the cable. REFERENCES Ammann, O.H. 1923. Possibilities of the Modern Suspension Bridge for Moderate Spans. Engineering NewsRecord, 21/06. Cullimore, M.S. 1986. The Clifton Suspension Bridge-Preservation for Utilisation. IABSE ProceedingsAugust: 100-86. Ferry Borges, J.; Arantes e Oliveira, E.; Silva Lima, C. 1966. International Symposium on Suspension Bridges, LNEC. Franco, M. 1983. Apostila PEF-602/FAU-USP. Gravina, P.B.J. 1951. Teoria e Cálculo das Pontes Pênseis, EPUSP. Hardesty, S.; Wessman, H.E. 1939. Preliminary Design of Suspension Bridges, Transactions, ASCE, Paper N◦ 2029. Honshu – Shikoku Bridge Authority The Akashi Kaikyo Bridge. Horton, T. 1983. Superspan – The Golden Gate Bridge, Squarebooks. Hudson, Ralph G. 1977. Manual do Engenheiro, Livros Técnicos e Científicos. Johnson, J.B.; Bryan, C.W.; Turneaure F.E. 1911. The Theory and Practice of Modern Framed Structures, John Wiley & Sons, 9a Edição. Jones, V. & Howells, J. 2008. Manual of Bridge Engineering, ICE. Judd, B.J., Wheen, R.J. 1978. Nonlinear Cable Behaviour, Proceedings ASCE, Vol. 104, No ST3, March, p. 567–575. Kerensky, O.A. 1959. The Maitland Lecture, ISE. Laginha, L.M. 1997. O Equilíbrio do Cabo Pênsil, Dissertação de Mestrado EPUSP. Laginha, L.M. 2006. Nova Formulação para o Comprimento do Cabo Parabólico, Revista Engenharia, N◦ 576. Langendonck, T. 1936. Prova de Carga das Pontes de São Vicente e de Jacarehy, Boletim 16, IPT. Norris, C. Head; Wilbur, J. Benson; Utku, S. 1976. Elementary Structural Analisys, McGraw-Hill Kogakusha Ltd., 3a Edição. O’Connor, C. 1976. Pontes – Superestruturas, Vol.2, LTC/EDUSP. Peters, T.F. 1987. Transitions in Engineering, Birkhäuser Verlag Pippard, A.J.S. 1947. The Experimental Study of Structures, Edward Arnold p. 5. Pippard, A.J.S. & Baker, J. 1968. The Analysis of Engineering Structures, Edward Arnold. 4a Edição. Pugsley, A.G. 1949. Some Experimental Work on Model Suspension Bridges, The Structural Engineer, Aug., Vol. 27, N◦ 8, p. 327–347. Pugsley, A.G. 1952. The Gravity Stiffness of a Suspension Bridge Cable, Quarterly Journ. Mech. and Applied Math., Vol. 5, p. 384–394. Pugsley, A.G. 1968. The Theory of Suspension Bridges, Edward Arnold, 2a Edição. Ramm & Wagner. 1967. Praktische Baustatik, Stuttgart – 4a Edição. Steinman, D.B. 1929. A Practical Treatise on Suspension Bridges, John Wiley & Sons, 2a Edição. Steinman, D.B. 1918. A Generalized Deflection Theory For Suspension Bridges, Transactions ASCE, No, 1935, p. 1133–1170. Steinman, D.B., Grove, W.G. 1926. The Eye-Bar Cable Suspension Bridge at Florianópolis, Brazil, Transactions ASCE, N◦ 1662, Jan., p. 267–393. Steinman, D.B. & Watson, S.R. 1941. Bridges and Their Builders, G. P. Putnam’s Sons, Nova York. Timoshenko, S.P. 1983. History of Strength of Materials, Dover. Timoshenko, S.P. & Young, D.H. 1945. Mecânica Técnica – Estática, Editora Gertum Carneiro, p.123. Timoshenko, S.P. & Young, D.H. 1965. Theory of Structures, McGraw-Hill Kogakusha. 2a Edição. Vogel, R. 1983. Building Brooklyn Bridge, The Design and Construction: 1867–1883, Smithsonian Institution, Civil Engineering, May. Widmer, U. 1979. IABSE Bulletin, B11. Williams Jr., J.W. 1977. J.B. Eads and his St. Louis Bridge, Civil Engineering, Oct.
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Author index
Adão da Fonseca, A. 37, 215, 549 Aketa, O. 417 Akyazı, U. 351 Albuquerque, C.M.C. 1091 Almeida, M. 301 Almeida, P. 607 Alvárez, J.L. 209 Amara, A.B. 215 Amaro, N. 317 Angelmaier, V. 231 Apaydin, N.M. 1169 Argüelles, G.R. 173, 737 Arici, M. 1137 Astaneh-Asl, A. 1169 Babu, G.S. 263 Baldomir, A. 491 Bani´c, D. 443 Barata, V. 459 Barcik, W. 257 Barros, R.C. 607, 1181 Barthélémy, J.-F. 871 Bastien-Masse, M. 1001 Bastos, R. 1181 Baumhauer, A. 409 Bednarski, C.M. 549 Beer, C. 641 Begg, D. 1107 Bella, G. 215 Benko, V. 975 Bento, J. 799 Biliszczuk, J. 257, 277 Bishop, D. 789 Bittencourt, T.N. 831 Blanco, L. 167, 381 Bondonet, G. 871 Borderon, A.E.C. 759 Bordin, F. 879 Bouška, P. 649 Brandtsegg, A.S. 365 Brekke, D.E. 729 Brito, V. 285, 301, 513 Brock, C.T. 887 Brühwiler, E. 1001
Buhin, M.M. 443 Bæksted, A.F. 83 Caetano, E. 771 Calado, L. 895 Calçada, R. 1091, 1099, 1143 Calvo, M. 1137 Cañueto, M.M. 337, 599, 1057 Caprani, C.C. 1029 Cardoso, A.M.L. 1197 Carvalho, A. 1001 Carvalho, D. 559 Carvalho, M.T.M. 955 Casper, J. 409 Castro, L.G. 1181 Castrodale, R.W. 985 Cauvin, B. 871 Cavadas, F. 807 ˇ Cech, J. 649 Chakrabarti, S. 847 Chen, A. 93 Chernyshova, L.A. 1115 Chung, J.Y. 1217 Coelho, H. 559, 583, 655 Costa, B.J.A. 807 Crémona, C. 895 Criado, S. 167 Cruz, J.M.S. 721 Cruz, J.P. 459 Cunha, A. 771 Curran, P. 179 Custódio, J. 1121 da Conceição Neto, A.P. 831 da Silva Araújo, S. 955 Dahl, K.B. 729 Daraban, M. 527 de Argote, J.I.D. 209 de Barros, F.B. 159 de Boer, A. 1107 de Freitas Simões, J. 159 de Jesus, A.M.P. 1091 de Miranda, M. 143 1295
Defaucheux, L. 863 Desprets, H. 863 Diniz, S.M.C. 1073 Dobashi, H. 781 Domínguez, J.M. 167 Düzbasan, S. 351 Eilzer, W. 201, 231 El-Belbol, S. 887 Escamilla, M. 167 Failla, I. 1189 Farook, H.M. 263 Fernández, S. 187, 389, 583 Ferreira, A. 507, 1151 Ferreira, J.R. 1073 Ferreira, P.S. 1163 Figueiras, J. 799, 807 Figueiredo, H. 239 Filho, H.R. 955 Fillo, L’. 975 Foremniak, S. 575 Fossbakken, S. 365 Fu, M. 309, 1175 Fuente, S. 167, 187, 381 Ganesh, K. 331, 345, 357 García, M.B. 337, 599, 1057 Gil, M.A. 167, 381 Gil, S. 721 Gorkos, P. 879 Granata, M.F. 1137 Guarascio, M. 215 Güngör, N. 435, 521 Gupta, A. 567 Gupta, J. 591 Gutsch, A.W. 841, 1009 Haixue, L. 1065 Hajar, Z. 863 Halvoník, J. 975 Hart, J.A. 467, 475 Haugerud, S.A. 365 He, X. 309 Heggade, V.N. 673
Hendy, C.R. 789, 847, 855, 887 Hernández, S. 491 Hipólito, A. 317 Hofstadler, C. 751 Hołowaty, J. 271, 293 Holtberget, S.H. 373 Humpf, K. 201, 231 Hussain, N. 17 Iglesias, C. 583 Imam, M. 927, 937, 967 Ishitani, H. 1197 Islami, K. 823 Jang, S.H. 247, 1217 Johannesen, S.M. 127 Joly, E. 135 Jung, R. 201 Jungwirth, J. 409 Jurado, C. 1247 Kalný, M. 993 Kamleithner, M. 641 Kanaji, H. 417 Karatzas, V. 1279 Karydis, G. 1279 Kashefi, K. 1239 Kasuga, A. 67 Kattenstedt, S. 1203 Khairussaleh, N.A.M. 937 Kilic, S.A. 1169 Kim, S.B. 1217 Kittoli, E. 467, 475 Kolísko, J. 649 Kollegger, J. 575 Konstantakopoulos, T. 1279 Körfgen, B. 1169 Kosugi, T. 781 Kralj, S. 443 Kulakowski, M. 879 Kummer, M. 751 Kye, M. 707 Lacort, A.G. 1271 Laginha, L.M. 1287 Lasa, I. 1065 Laube, M. 1009 Li, G.P. 1255 Li, X. 1175 Lima, B. 507, 1151 Liu, C. 919 Liu, Y. 1037
Löker, E. 1263 Long, A.E. 567 Long, Q.Q. 1019 Lopes, F. 507 López, F.J.M. 337, 599, 1057 Lu, D.G. 1037 Lu, Y. 309, 1211 Ma, R. 93 Magalhães, F. 771 Makita, T. 815 Malindretou-Vika, M. 451 Malveiro, J. 1143 Manjure, P.Y. 1049 Mansperger, T. 201 Manterola, J. 9, 167, 187, 209, 381, 389, 395 Marenzi, L. 625 Márquez, M. 483, 499 Martín, B. 381 Martínez, A. 167, 209, 381, 389, 395 Mathur, A.K. 591 Matute, L. 151 Maurer, R. 1203 McPolin, D. 567 Melo, L.R.T. 831 Mendonça, T. 285, 301, 513 Meng, N. 743, 823, 1081 Mihal, M. 641 Miklashevich, L.V. 223, 1115 Millanes, F. 151, 239, 1129 Minoretti, A. 365 Miquilena, I. 715 Moine, P. 135 Moir, G.D. 247, 1217 Monteiro, M. 285, 513 Moor, G. 743, 1081 Moutinho, C. 771 Mundell, C. 789 Muñoz-Rojas, J. 187, 389, 583 Murali, P. 357 Nakamura, Y. 781 Nanukuttan, S. 567 Navarro, J.A. 167, 187, 209, 381 Nebreda, J. 1129 Nestegård, A. 365 Neves, A.S. 799, 1099 Neves, M.S. 721 1296
Ni, Y.S. 919 Nicholls, A.D.J. 887 Nie, Z. 1175 Niu, X.J. 1019 Nogueira, T. 317 Nolte, T. 1009 Nunes, F.P.S. 325 Nunes, S. 1001 Ökte, E. 1263 Olamigoke, O.A. 927, 967 Onoufriou, T. 947 Onysyk, J. 257, 277 Orcesi, A. 895 Ortega, M. 151, 239 Osborne, G. 167 Oshiro, R.A. 1197 Oyamada, R.N. 1197 Öztürk, B.D. 1263 Pacheco, P. 103, 507, 559, 583, 655 Palacio, C.D.U. 955 Paolacci, F. 1037 Parejo, J.M.G. 337, 599, 1057 Parke, G.A.R. 927, 937, 967 Paulík, P. 975 Pavi, S. 879 Pecker, A. 135 Peng, G.F. 1019 Pereira, P. 459 Pimentel, M. 1001 Pipenbaher, M. 427 Póvoas, A.A. 615 Prabucki, P. 257, 277 Preuer, A. 641 Raatschen, H.J. 1169 R˘ac˘anel, I.R. 527, 1223, 1231 Radic, J. 427 Raina, V.K. 533 Recupero, A. 1137, 1189 Rees, J. 1217 Reiso, M. 365 Resende, A. 559, 583, 655 Ribeiro, A.B. 1121 Ribeiro, C.F. 51 Ribeiro, D. 1143 Ribeiro, F. 1001 Ricciardi, G. 1189 Robb, D. 567 Rojas, M. 715 Romo, J. 401
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1297
Valenzuela, M.A. 483, 499 Vallejo, I. 483, 499 Vázquez, J.I.C. 663 Veie, J. 373 Venkatram, P.G. 345 Veronez, M. 879 Villoria, B. 127 Virlogeux, M. 863 Virtuoso, F. 1163 Vítek, J.L. 993 Vokáˇc, M. 649 Votsis, R.A. 947 Walther, M. 841 Watanabe, H. 417 Waterfall, P. 855 Weiher, H. 903 Weiss, W. 575 Wheeler, W.K. 193 Whitmore, D. 1065 Wielgosz, J.B. 127 Winkler, J. 855 Wo´zny, P. 257 Xu, D. 919 Yagi, N. 815 Yakıt, E. 351 Yang, J. 1019 Zeybek, F. 435 Zhang, X. 93 Zhou, C. 309 Zhou, J.Y. 1029 Zhu, P. 1211 Zucconi, M. 215
