New Austrian Tunneling Method (NATM) - Stability Analysis and Design by Walter Wittke, Berndt Pierau, Claus Erichsen

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Geotechnical Engineering in Research and Practice WBI-PRINT 5

Edited by Prof. Dr.-Ing. W. Wittke Beratende Ingenieure für Grundbau und Felsbau GmbH Consulting Engineers for Foundation Engineering and Construction in Rock Ltd.

New Austrian Tunneling Method (NATM) Stability Analysis and Design

Walter Wittke, Berndt Pierau, Claus Erichsen translated into English by: Jens Lüke and Johannes R. Kiehl

Translated from the German edition: Statik und Konstruktion der Spritzbetonbauweise. Geotechnik in Forschung und Praxis, WBI-PRINT 5, VGE-Verlag Glückauf GmbH, Essen 2002, ISBN 3-7739-1305-2

From the contents: >

Means of support

>

Geotechnical mapping and monitoring

>

Case Histories: ‚

Crown heading with open invert

‚

Crown heading with closed invert

‚

Sidewall adit heading

‚

Full-face heading

‚

Heading under the protection of pipe umbrellas

‚

Heading under the protection of jet grouting columns

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- III Preface Within the series "WBI-PRINT, geotechnical engineering in research and practice", volumes 4 to 7 are designed as a compendium of tunnel statics. This compendium started with the volume WBI-PRINT 4 "Stability analysis for tunnels, fundamentals", published in 1999 in German and in 2000 in English. The present volume WBI-PRINT 5 "New Austrian Tunneling Method stability analysis and design" covers, beside fundamentals of the New Austrian Tunneling Method (NATM), case histories of realized mined tunnels designed and constructed with participation of WBI. The selected case histories from the years 1985 to 2001 include crown headings with open and closed invert, sidewall adit headings, full-face headings and headings under the protection of pipe umbrellas and jet grouting columns. Analyses according to the finite element method have proved to be an indispensable tool for the design of tunnels. The stability analyses for all case histories presented were carried out using the program system FEST03. In order to enable this program system to be used by our professional colleagues as well, we have been offering it for sale for some little time now. WBI-PRINT 5 has been previously published 2002 in German as a paperback. Now the English translation is available online to provide a worldwide access to those who are interested in tunneling. It is also available on CD-ROM via WBI company. The next volume in the series WBI-Print is dedicated to the mechanized tunneling. This volume appears as WBI-PRINT 6 in German in December 2006. Special problems of tunnel statics will be covered in WBI-PRINT 7. I adress my special thanks to my two co-authors and directors at WBI, Dr.-Ing. B. Pierau and Dr.-Ing. C. Erichsen, who have been supporting my work substantially for many years. I am also obliged to Dr.-Ing. J. R. Kiehl for his editorial work. The translation into English was carried out by Dr.-Ing. J. Lüke as well as Dr.Ing. J. R. Kiehl. I convey my sincere thanks to them. Further thanks are due to our secretary and design office. Aachen, December 2006 Walter Wittke WBI-PRINT 5

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- IV Contents

Page

1.

Introduction

1

2.

Elements of the NATM

4

2.1

Shotcrete

4

2.1.1

Components and composition

4

2.1.2

Spraying methods

7

2.1.3

Early strength

11

2.1.4

Final strength

14

2.1.5

Deformability

15

2.1.6

Rebound

17

2.2

2.3

2.4

2.5

3.

Steel sets

17

2.2.1

Basic types

17

2.2.2

Load-carrying behavior

28

Anchors

30

2.3.1

Basic types

30

2.3.2

Load-carrying behavior

36

Advance support

37

2.4.1

Spiles

37

2.4.2

Pipe umbrellas

40

Geotechnical mapping and monitoring

48

2.5.1

Mapping

48

2.5.2

Monitoring

54

Crown heading with open invert

67

3.1

Glockenberg Tunnel near Koblenz, Germany

67

3.1.1

67

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- V Contents

3.2

3.3

Page 3.1.2

Structure

68

3.1.3

Ground and groundwater conditions

71

3.1.4

Excavation classes

73

3.1.5

Stability analyses

76

3.1.6

Crown heading and monitoring results

90

3.1.7

Conclusions

93

Gäubahn Tunnel in Stuttgart, Germany

94

3.2.1

Introduction

94

3.2.2

Structure

94

3.2.3

Ground and groundwater conditions

98

3.2.4

Excavation classes

99

3.2.5

Stability analyses for the design of the shotcrete support

102

3.2.6

Crown heading and monitoring results

108

3.2.7

Conclusions

110

Hellenberg Tunnel, Germany

111

3.3.1

Introduction

111

3.3.2

Structure

111

3.3.3

Ground and groundwater conditions

114

3.3.4

Excavation classes

116

3.3.5

Crown heading

119

3.3.6

Results of the crown face mapping

121

3.3.7

Stability analyses for the bench excavation

123

Construction and monitoring results

127

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- VI Contents

Page 3.3.9

4.

Conclusions

130

Crown heading with closed invert

131

4.1

Österfeld Tunnel in Stuttgart, Germany

131

4.1.1

Introduction

131

4.1.2

Structure

131

4.1.3

Ground and groundwater conditions

135

4.1.4

Fundamentals of the design

140

4.1.5

Stability analysis for the stages of construction

141

4.1.6

Excavation and support

147

4.1.7

Monitoring program and interpretation of the measuring results

151

Conclusions

155

4.1.8 4.2

4.3

Road tunnel "Elite" in Ramat Gan, Israel

156

4.2.1

Introduction

156

4.2.2

Structure

159

4.2.3

Ground and groundwater conditions

161

4.2.4

Design

164

4.2.5

Stability analyses

168

4.2.6

Construction

183

4.2.7

Monitoring

187

4.2.8

Conclusions

189

City railway tunnel to Botnang in Stuttgart, Germany

190

4.3.1

Introduction

190

4.3.2

Structure

190

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- VII Contents

5.

Page 4.3.3

Ground and groundwater conditions

190

4.3.4

Design

196

4.3.5

Stability analyses for the design of the shotcrete support

198

4.3.6

Construction

209

4.3.7

Monitoring

211

4.3.8

Conclusions

213

Sidewall adit heading 5.1

Road tunnel "Hahnerberger Straße" in Wuppertal, Germany

214

5.1.1

Introduction

214

5.1.2

Structure

215

5.1.3

Exploration

216

5.1.4

Design and construction

220

5.1.5

Stability analyses for the stages of construction

226

Stability analyses for the design of the interior lining

233

5.1.7

Monitoring

236

5.1.8

Conclusions

238

5.1.6

5.2

214

Limburg Tunnel, Germany

238

5.2.1

Introduction

238

5.2.2

Structure

241

5.2.3

Ground and groundwater conditions

243

5.2.4

Excavation and support

245

5.2.5

Sidewall adit excavation north

248

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- VIII Inhalt

Page 5.2.6

5.3

6.

Stability analyses for sidewall adit, excavation north

253

5.2.7

Monitoring results

260

5.2.8

Conclusions

262

Niedernhausen Tunnel, Germany

262

5.3.1

Introduction

262

5.3.2

Structure

264

5.3.3

Ground and groundwater conditions

266

5.3.4

Excavation and support

267

5.3.5

Three-dimensional stability analyses

270

5.3.6

Construction

284

5.3.7

Conclusions

290

Full-face heading 6.1

291

Urban railway tunnel underneath the Stuttgart airport runway, Germany

291

6.1.1

Introduction

291

6.1.2

Structure

292

6.1.3

Ground and groundwater conditions

294

6.1.4

Fundamentals of the design

300

6.1.5

Excavation and support

304

6.1.6

Stability analyses for the design of the shotcrete support

307

6.1.7

Monitoring

312

6.1.8

Interpretation of the monitoring results

316

6.1.9

Conclusions

322

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- IX Contents 6.2

Page Freeway tunnel "Berg Bock" near Suhl, Germany

323

6.2.1

Introduction

323

6.2.2

Structure

323

6.2.3

Ground and groundwater conditions

328

6.2.4

Excavation and support

330

6.2.5

Stability analyses for the stages of construction and design of the shotcrete support

334

Stability analyses for the design of the interior lining

340

6.2.7

Monitoring

348

6.2.8

Conclusions

350

6.2.6

7.

Heading under the protection of jet grouting columns 7.1

7.2

351

Road tunnel for the federal highway B 9 in Bonn-Bad Godesberg, Germany

351

7.1.1

Introduction

351

7.1.2

Structure

351

7.1.3

Ground and groundwater conditions

355

7.1.4

Design and construction

357

7.1.5

Stability analyses for the design of the shotcrete support

367

7.1.6

Monitoring

377

7.1.7

Conclusions

377

City railway tunnel "Killesberg-Messe" in Stuttgart, Germany

379

7.2.1

Introduction

379

7.2.2

Structure

380

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- X Contents

8.

Page 7.2.3

Ground and groundwater conditions

382

7.2.4

Excavation and support

385

7.2.5

Stability analyses

392

7.2.6

Monitoring

400

7.2.7

Conclusions

402

References

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- 1 1.

Introduction

The "New Austrian Tunneling Method" (NATM) was originally applied for tunnels in rock. Since the 1970ies, however, this tunneling method was carried out more and more also in soft rock with low overburden and in urban areas. Because of the outstanding importance of the shotcrete (sprayed concrete) for the application of this method the denotation "Sprayed Concrete Lining Method" or simply "Shotcrete Method" is mainly used in Germany. The NATM is a construction method, which is very adaptive according to changing subsoil conditions and changing shapes of crosssections. Interacting with the subsoil the primary function of the shotcrete membrane is to form an arch around the tunnel, which is capable to carry. With a favourable shape of the tunnel's crosssection and an adequate sequence of construction stages it is possible to avoid or at least to minimize bending moments and shearing forces in the shotcrete membrane. Thus, large underground openings can be supported by relatively thin shotcrete membranes. With an adequate design also the subsidence on the surface can be limited to relatively small values. Stability analyses, in which the interaction of the subsoil with the support are modeled in a realistic way, however, serve as a prerequisite for a successful tunnel heading using this method. The authors are convinced that this is possible only by numerical computation methods. Stability analyses, therefore, should be carried out generally using finite element codes. A powerful tool, which is suitable also for three-dimensional problems, is the finite element code FEST03 developed by WBI and documented in the volume WBI-PRINT 4 (Wittke, 2000). Since more than 20 years this program, which in this period of time has been improved and enlarged several times, serves as an valuable device for a safe and economic design of tunnels. The design of a tunnel according to the NATM is carried out stepwise with the following working steps, which are to be repeated several time, if required: -

Geotechnical investigations of the ground and groundwater conditions.

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- 2 -

Evaluation of the soil and rock mechanical parameters, based on test results as well as experience.

-

Stability analyses for both, the stability proof of the tunnel and the design of the shotcrete membrane as well as the interior concrete lining.

-

Design and assessment of excavation methods and support measures (excavation classes).

-

Supervision of stability by geotechnical mapping and monitoring during construction.

-

Back analysis of the results of measurements.

The authors of the given volume WBI-Print 5 since more than 20 years are experienced with the stability analysis and design of tunnels carried out by the NATM. With this volume it will be attempted to transmit this experience by case histories. In Chapter 2 an overview on the fundamentals of the NATM is given. Here also new developments such as non-alkaline shotcrete are treated. Moreover in Chapter 2 geotechnical mapping and monitoring, which are essential parts of this tunneling method, are dealt with. Advancing crown headings with open and closed invert are treated in the Chapters 3 and 4. In each chapter three case histories are presented. Advancing sidewall tunnel headings are subject of Chapter 5. Here also three case histories are documented. In Chapter 6 two more case histories are presented, in which a full-face excavation at least in sections were carried out. Two case histories for headings under the protection of jet grouting columns are dealt with in Chapter 7. The documentation of the case histories is, as a rule, arranged as follows: -

Description of the structure, WBI-PRINT 5

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- 3 -

-

characterization of the ground and the groundwater conditions,

-

design and excavation classes,

-

stability analyses and the design of the shotcrete membrane,

-

excavation methods and support measures carried out during construction,

-

geotechnical monitoring and interpretation of measurements,

-

conclusions.

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- 4 2.

Elements of the NATM

2.1

Shotcrete

2.1.1

Components and composition

Requirements In order to improve workers' protection, minimize environmental pollution (water and ground) and reduce the amount of eluates (alkalis, calcium hydroxides), shotcrete mixes may only be applied in general if they at least equivalent to conventional structural concrete mixes for support elements with respect to their physiological properties and their leaching behavior. The following requirements, among others, have to be met by the shotcrete: -

Low water permeability,

-

no use of alkali-containing additives,

-

a minimum strength of the green shotcrete, termed early strength (see Chapter 2.1.3).

The required early strength for the shotcrete can be achieved by either: -

The use of so-called spray bonding agents (SBM) or spray cements, which allow to dispense with setting activators, or

-

the use of alkali-free accelerating admixtures in powder or fluid form.

In special cases, e. g. with a high water discharge, spray bonding agents and alkali-free accelerating admixtures may also be applied in combination (ÖBV, 1998). Bonding agents According to DIN 18551, the following bonding agents may be used:

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Standard cements according to DIN 1164 (Parts 1 and 100, 1990),

-

spray bonding agents or spray cements certified by the supervisory authorities.

If spray bonding agents or spray cements have not been certified by the supervisory authorities, the suitability of the bonding agent for the production of shotcrete must be proved before construction by a testing certificate from an approved institute for materials testing. With respect to the leachability the amount of eluate must not be greater for this bonding agent than for standard cements. Proof of this must be provided by a testing certificate from a public health institute. On the basis of their rate of reaction, one distinguishes between two types of spray bonding agents (ÖBV, 1998): -

Spray bonding agent SBM-T: With a maximum processing time of less than one minute, this type of bonding agent can only be used for the production of shotcrete with dry aggregates (water content w ≤ 0.2 M.-% and according to the manufacturer's specification, respectively).

-

Spray bonding agent SBM-FT: With an admissible processing time of several minutes, this type of bonding agent can also be used for the production of shotcrete with wet aggregates (water content w generally 2 M.-% to 4 M.-%).

Admixtures With respect to the improvement of the shotcrete properties such as workability, stickiness, formation of dust, rebound, strength and tightness of the shotcrete fabric as well as reduction of the heat production, adding hydraulically active admixtures is useful (ÖBV, 1998). Fly ash is a proven admixture, but the use of other admixtures is also possible (e. g. silica dust, smelting sand, hydraulic lime). The total amount of added ground material and admixtures must not exceed 35 % of the bonding agent (ÖBV, 1998).

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- 6 Aggregates For shotcrete, concrete aggregates as specified in DIN 4226 (parts 1 and 2, 1983) must be used (DIN 18551, 1992). The maximum grain size must be selected between 4 and 16 mm (ÖBV, 1998). Additives Until a few years ago, alkali-containing accelerating admixtures were used as additives for shotcrete. This way it was possible to achieve a favorable development of early strength (see Chapter 2.1.3). These additives are strongly caustic and due to reasons of environmental protection they are not used anymore. Furthermore, they have a negative effect on the leaching behavior of the shotcrete. This has lead e. g. to drainages being clogged by encrustations and in some cases also to contaminations in the groundwater caused by the eluates. In addition, the shotcrete became porous and permeable to water with the leaching. This results in decreasing strength with progressing age. It is therefore state-of-the-art today to use spray bonding agents or spray cements without accelerating admixtures or with alkalifree accelerating admixtures, added as powder or in fluid form and certified by the supervising authorities. The suitability of the planned shotcrete recipe including the used additive must be proved before construction by laboratory testing of the setting behavior, the early strength and the strength development. Laboratory tests yield reference values, but they cannot capture all influences from the construction site and therefore cannot replace suitability testing on site (ÖBV, 1998). Furthermore, it has to be proven that the additives do not have a negative impact on the reinforcement and the remaining steel mounting parts. Composition According to ÖBV (1998), the mixes for dry-mix and wet-mix shotcrete are subdivided into: -

Dry mix (TM), moist mix, storable (FM-L), moist mix for immediate application (FM-S), wet mix (NM).

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- 7 These mixes are referred to as supply mixes. They differ in composition from the sprayed concrete due to the rebound occurring during spraying. The rebound is the share of the shotcrete mix which does not adhere to the surface of application during spraying and which must be disposed of. 2.1.2

Spraying methods

Dry-mix method For the dry-mix method, TM, FM-L and FM-S mixes can be used. The mix is conveyed intermittently to the spray nozzle via compressed air using a piston or rotary engine (thin stream transport). At the nozzle, it is wetted with water and sprayed onto the surface of application at a speed of 20 m/s to 30 m/s.

Fig. 2.1:

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Placing of the shotcrete by a manually guided spray nozzle (DB, 1985)

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- 8 -

Fig. 2.2:

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Placing of the shotcrete using a spray vehicle with a remote-controlled spray arm (Limburg Tunnel, new railway line Cologne – Rhine/Main) WBI GmbH, Henricistr. 50, 52072 Aachen, Germany

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- 9 Liquid accelerating admixture is added continuously to the supply water using dosing pumps. Accelerated admixture in powder form is added immediately before the mix transport using proportioners. Nowadays, the shotcrete is only rarely applied to the tunnel surface by a manually guided spray nozzle (Fig. 2.1). It is standard practice to apply the shotcrete using a spray vehicle with a remote-controlled spray arm (Fig. 2.2). Due to the high velocity of the shotcrete during placing, high rebound portions arise from the dry-mix method. A further problem is the resulting heavy formation of dust. The dry-mix method is therefore only permitted for special cases by the government safety organizations today. An alternative transportation technique for dry-mix shotcrete was developed by the Rombold und Gfröhrer Co. Here, the dry-mix shotcrete is subjected to compressed air in a pressure tank (silo) and conveyed via dust-encapsulated dosing screws continuously and dust-free with the air stream to the spray nozzle. As in the conventional dry-mix method, the water is added only just before the nozzle (Balbach and Ernsperger, 1986). As a bonding agent, spray cement with a swift development of strength (fast cement) and a high final strength is used (see Chapters 2.1.3, 2.1.4 and 4.1.4). The addition of an accelerating admixture is therefore not necessary. Advantages of the dry-mix method are the workability in small amounts and the transportability over long distances. Wet-mix method With the wet-mix method, the wet mix (NM) is conveyed by the spraying machine to the spray nozzle either by compressed air (thin stream transport) or hydraulically using piston pumps (thick stream transport). Like dry-mix shotcrete, wet-mix shotcrete is generally applied using a spray vehicle with a remote-controlled spray arm (Fig. 2.2). Manually guiding the spray nozzle is problematic because of the high weight of the wet-mix shotcrete. Less rebound, less formation of dust and a higher spraying performance are advantages of the wet-mix method over the dry-mix method. WBI-PRINT 5

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- 10 Processing and application Before the shotcrete is applied, loose rock must be removed from the excavation surface. The surfaces of application must be carefully cleaned with compressed air in order to achieve the best possible adhesion of the shotcrete. This particularly applies if the shotcrete lining is constructed in layers or if longer interruptions occur during the application of the shotcrete. An immediate sealing of the exposed rock surfaces with shotcrete of at least 3 cm thickness is intended to provide early support to the ground close to the excavation surface in order to largely avoid loosening and the resulting decrease in rock strength. The shotcrete must be applied in such a way that a homogeneous, dense shotcrete with a closed, even surface is achieved. Thick shotcrete linings are applied in two or more layers in order to avoid separation from the excavation surface. The shotcrete must be applied in such a way that spraying rebound and adherent spray dust into the shotcrete is avoided by all means. Rebound and dust must be removed before the next shotcrete layer is applied, and the shotcrete lining must always be constructed from the bottom to the top. The distance between the nozzle and the surface of application must be adapted to the delivery rate and the speed of application. It ranges between 0.5 and 2.0 m, depending on the air flow. The nozzle should be oriented at right angles to the surface of application, if possible. Exceeding or falling below the recommended nozzle distance as well as an inclined orientation of the nozzle relative to the surface of application generally lead to a reduced quality of the shotcrete and an increased amount of rebound. In case of steel insertions such as steel arches, steel girders, lagging plates, pipes, etc., spray shadows cannot be totally avoided, but they can be considerably reduced by proper nozzle control (ÖBV, 1998). Special care has to be taken when the connection is made to the existing shotcrete lining in the crown invert, the bench invert and the permanent invert. Existing rebound must be removed first. The reinforcement and the support arches should be completely wrapped up in shotcrete. It is important that the visible surfaces WBI-PRINT 5

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- 11 are constructed in a convex shape only, if possible, in order to achieve an arching effect. 2.1.3

Early strength

Shotcrete up to an age of 24 hours is termed green shotcrete. With respect to the requirements on strength development, green shotcrete is distinguished into the three early strength classes J1, J2 and J3 determined on the basis of years of experience (Fig. 2.3).

Fig. 2.3:

Early strength classes of green shotcrete (ÖBV, 1998)

Class J1 shotcrete is suitable for the application of thin layers on dry surfaces without specific statical requirements. It has the advantage of little dust development and rebound. If statical requirements exist with respect to the green shotcrete, e. g. for the exterior lining of a traffic tunnel, class J2 shotcrete is generally used. Class J3 shotcrete is required if rapWBI-PRINT 5

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- 12 idly developing loads from water pressure and/or rock pressure are to be expected. It is known from experience that dry-mix shotcretes allow to achieve the highest early strength values. Fig. 2.4 shows the comparison of the strength development of two wet-mix shotcretes, one with accelerating admixture containing alkali (1) and one with alkali-free accelerating admixture (2), and one dry-mix shotcrete with alkali-free spray bonding agent (3). The latter type reaches class J3. Using accelerating admixtures, either containing alkali or alkali-free, class J2 is reached. Here, the shotcrete with alkali-free accelerating admixture shows greater early strength.

Fig. 2.4:

Comparison of the strength development of shotcretes with accelerating admixtures, either containing alkali or alkali-free, and alkali-free spray bonding agent

Because of the disadvantages of the dry-mix method compared to the wet-mix method, which are discussed in Chapter 2.1.2, wet-mix shotcrete is often preferred over dry-mix shotcrete also when high WBI-PRINT 5

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- 13 demands are made on the early strength, notwithstanding that a higher early strength can be achieved with dry-mix shotcrete. An example for this is the shotcrete for the Schulwald Tunnel of the new railway line (NBS) Cologne – Rhine/Main of German Rail. Due to the predicted poor geological conditions, dry-mix shotcrete with spray cement as bonding agent was selected at first. After comprehensive preliminary testing, an early strength corresponding to class J3 was achieved with the recipe given in Fig. 2.5. Because of the high formation of dust during the application and the insufficient placement performance, it was later decided to change to wet-mix shotcrete with a liquid, alkali-free accelerator (BE U22). With this shotcrete, the recipe and early strength development of which are given in Fig. 2.6, a class J2 early strength was achieved which proved sufficient. A placement performance of up to 25 m3/h was obtained with this wet-mix shotcrete, which clearly exceeds the 14 m3/h achieved with the dry-mix method. Further, rebound values of less than 10 % were reached (Brötz et al., 2000).

Fig. 2.5:

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Recipe and early strength development of the dry-mix shotcrete for the Schulwald Tunnel of the new railway line Cologne – Rhine/Main (Brötz et al., 2000) WBI GmbH, Henricistr. 50, 52072 Aachen, Germany

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- 14 -

Fig. 2.6:

2.1.4

Recipe and early strength development of the wet-mix shotcrete for the Schulwald Tunnel of the new railway line Cologne – Rhine/Main (Brötz et al., 2000) Final strength

Besides its rapid strength development, shotcrete with alkali-free accelerating admixtures or spray bonding agents also possesses a high final strength. While accelerators containing alkali impede the hydration of the cement, this is not the case if alkali-free accelerating admixtures or spray bonding agents are used. A final strength of 30 to 40 N/mm2 can be obtained in practice (Brötz et al., 2000; Bauer, 2000; NATS, 1998). With shotcrete with accelerators containing alkali, even a final strength of 25 N/mm2 combined with a high early strength is difficult to achieve (NATS, 1998).

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- 15 2.1.5

Deformability

The deformation behavior of green shotcrete according to Hesser (2000) is essentially characterized by -

hardening with time and a creep ability decreasing with time,

-

overproportional, non-linear creep with increasing load.

The Young's moduli determined by Hesser (2000) in laboratory tests on dry-mix shotcrete test specimens of different ages show good agreement with the relations by Weber (1979) and by the Comité Euro-International du Béton (CEB, 1978) (Fig. 2.7a). Fig. 2.7b shows the development of Young's modulus in the first 24 hours. According to this, Young's modulus of the shotcrete amounts to approx. 15000 MN/m2 after 24 hours, with creep and shrinkage not being taken into account. Investigations by Manns et al. (1987) showed that the creep and shrinkage deformations of wet-mix shotcrete are larger than those of dry-mix shotcrete. In comparison to standard concrete, the creep and shrinkage deformations of shotcrete are generally clearly larger. In finite element stability analyses for tunnels, the development of Young's modulus of the shotcrete with time as well as creep and shrinkage are generally not taken into account. An equivalent modulus is instead assigned to the shotcrete to account for hardening during application of the load as well as creep and shrinkage. The interpretation of the displacements and stresses measured at different tunneling projects by means of back analyses has shown that a modulus of E = 15000 MN/m2 can be used as equivalent modulus of the shotcrete, corresponding to the 24-hour-value after CEB and Weber (see Fig. 2.7b). Specially in cases, where the shotcrete is already loaded at a young age due to short round lengths and early closing of the invert, values of 2000 to 7500 MN/m2 for the equivalent modulus have proven to be more realistic (see Chapters 6.1 and 7.1).

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Fig. 2.7:

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Development of Young's modulus for shotcrete: a) In the first 49 days; b) in the first 24 hours

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- 17 The development of the deformability of shotcrete with time including shrinkage and creep and its representation in numerical analyses are still a subject of further development. 2.1.6

Rebound

As mentioned above, the amount of rebound is higher for dry-mix shotcrete than for wet-mix shotcrete. Using dry-mix shotcrete with spray bonding agent, however, allows to reduce the rebound considerably as compared to conventional dry-mix shotcrete (Fig. 2.8).

Fig. 2.8:

Comparison of rebound between conventional dry-mix shotcrete, dry-mix shotcrete with spray bonding agent and wet-mix shotcrete (NATS, 1998)

The rebound further depends, among other things, on the water cement ratio, the aggregates and the cement type (Maidl, 1992). 2.2

Steel sets

2.2.1

Basic types

Steel sets are made with different profiles. Examples are shown in Fig. 2.9. One distinguishes between plain girders and lattice girders.

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Fig. 2.9:

Profiles for steel sets, survey (Heintzmann Ironworks Bochum, Germany)

Among the plain girders are e. g. GI profiles (mining I profiles), TH profiles, bell profiles and standard profiles. Their dimensions, weights, geometrical moments of inertia, section moduli and characteristic parameters are given in Fig. 2.10. Further plain girders are star profiles, the specifications of which are shown in Fig. 2.11.

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Fig. 2.10: WBI-PRINT 5

Specifications of GI-profiles, TH profiles, bell profiles and standard profiles (Maidl, 1984) WBI GmbH, Henricistr. 50, 52072 Aachen, Germany

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Fig. 2.11:

Specifications of star profiles (Heintzmann Ironworks Bochum, Germany)

Fig. 2.12 to 2.14 show examples of butt joints of HEB profiles, TH channel profiles and star profiles.

Fig. 2.12:

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HEB profile joints: a) Butt strap joints; b) slab joints WBI GmbH, Henricistr. 50, 52072 Aachen, Germany

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Fig. 2.13:

Flexible joint for TH gutter profiles

Fig. 2.14:

Star profile joint (Heintzmann Ironworks Bochum, Germany)

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- 22 For the same bending stiffness, lattice girders have less weight per meter of girder length than plain girders. They are therefore easier to handle. Lattice girders are distinguished into 3stringer girders (Fig. 2.15) and 4-stringer girders. Fig. 2.16 shows the specifications for 3-stringer lattice girders in which the rods are welded to the stiffening elements from the inside. The rods of the Pantex 3-stringer and 4-stringer PS-girders are welded to the stiffening elements from the outside (Fig. 2.17, 2.18 and 2.19).

Fig. 2.15:

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3-stringer lattice girder (Dernbach Tunnel, new railway line Cologne – Rhine/Main)

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Fig. 2.16:

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Specifications of 3-stringer lattice girders (Heintzmann Ironworks Bochum, Germany)

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Fig. 2.17:

Pantex 3-stringer PS-girder (Tunnel-Ausbau-Technik Ltd., Germany)

For the same height, plain girders like GI, TH and other profiles have a far greater normal and bending stiffness than lattice girders (Fig. 2.20). The normal stiffness of lattice girders is independent of their height, if the cross sectional area of the stringer rods As remains constant. In Fig. 2.20, As = 12.4 cm2 was assumed for the stringer rods (d1 = 20 mm, d2 = 28 mm, see Fig. 2.16).

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Fig. 2.18:

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Specifications of Pantex 3-stringer PS-girders (Tunnel-Ausbau-Technik Ltd., Germany)

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- 26 -

Fig. 2.19:

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Specifications of Pantex 4-stringer PS-girders (Tunnel-Ausbau-Technik Ltd., Germany) WBI GmbH, Henricistr. 50, 52072 Aachen, Germany

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- 27 -

Fig. 2.20:

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Bending and normal stiffness vs. height of steel sets

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- 28 2.2.2

Load-carrying behavior

Steel sets only become fully effective as a support if they form a closed ring. In this way they are often used as support in mining without additional support measures. In tunneling according to the NATM, the steel sets have further tasks as: -

Immediate support of the tunnel face area over the length of the foremost round,

-

template of the tunnel profile for the application of the shotcrete and the excavation of the next round,

-

support for spiles installed as advance support for the next round (see Fig. 2.15 and Chapter 2.4.1).

In tunneling according to the NATM, steel sets are only rarely installed as a closed ring after each round. Therefore, immediately after their installation, they only have very little bearing capacity. For the NATM, the load-carrying behavior of the steel sets is based on their bond with the shotcrete membrane. Immediately after a steel set is installed and covered with shotcrete, when the shotcrete still has a very low Young's modulus, it is mostly the steel set that carries the loads resulting from rock mass pressure. Since steel sets are usually installed closely behind the tunnel face, this loading at the beginning is generally small. With progressing hardening of the shotcrete with time, the normal strength and thus the bearing capacity of the shotcrete membrane increases. Finally, after the shotcrete has fully hardened, the normal stiffness of the steel sets can be neglected compared to the one of the shotcrete membrane. Fig. 2.21 illustrates this for the example of a 30 cm thick shotcrete membrane. It shows the ratio of the normal stiffnesses of the shotcrete membrane and the steel sets vs. the Young's modulus and the age of the shotcrete, respectively for two GI profiles and one 3-stringer lattice girder. A spacing of the steel sets of 1 m is assumed for all cases. It becomes evident that already at the age of a few hours the normal stiffness of the shotcrete membrane WBI-PRINT 5

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- 29 exceeds the one of the steel sets. After 24 hours, the normal stiffness of the shotcrete membrane amounts to 17 times of that of the lattice girders and 4 to 8 times the one of the GI profiles.

Fig. 2.21:

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Ratio of normal stiffnesses of shotcrete membrane and steel sets vs. Young's modulus and age of the shotcrete WBI GmbH, Henricistr. 50, 52072 Aachen, Germany

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- 30 The steel sets covered with shotcrete can also be accounted for as part of the reinforcement for the dimensioning of the shotcrete membrane (e. g. stringer rods of lattice girders). This requires however that the steel sets are completely covered with shotcrete. Especially if plain girders are used, however, the bond is reduced due to spray shadows. Therefore, in general steel sets are conservatively disregarded as reinforcement and in finite element analyses, the steel sets are generally not modeled. In Guideline 853 of German Rail (DB, 1999), the following criteria for the selection of steel sets are given: -

Plain girders yield more stable immediate support than lattice girders. This requires, however, friction-locked connections between the supports and the rock.

-

Lattice girders bond better with the shotcrete and lead generally to tighter shotcrete membranes than plain girders.

-

Spray shadows are generally smaller for lattice girders than for plain girders.

Details regarding the better bonding effect of lattice girders with shotcrete and the consequences for the strength and tightness of the shotcrete membrane can also be found in Eber et al. (1985). 2.3

Anchors

2.3.1

Basic types

Except for special cases, in tunneling mainly non-prestressed (untensioned) anchors, termed rock bolts, are used in boreholes. A detailed description of the terms and designations of rock bolts is given in the German standard DIN 21521 "Gebirgsanker für den Bergbau und den Tunnelbau" ("Rock bolts for mining and tunneling"). With respect to the load-carrying behavior (see Chapter 2.3.2), bond anchors, which are form-locked with the rock mass (Fig. 2.22a to c), are distinguished from anchors that are friction-locked with the rock mass (Fig. 2.22 d). In addition, there are anchors that are form-locked as well as friction-locked with the rock mass (Fig. 2.22e). WBI-PRINT 5

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Fig. 2.22:

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Anchors: Cement mortar anchor (SN-anchor); b) synthetic resin anchor; c) injection drill bolt (IBObolt); d) expansion shell bolt; e) injection bolt with expansion shell

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- 32 Among the form-locked anchors are the mortar anchors (e. g. SNanchors, IBO-rods), the resin anchors and the friction pipe anchors (e. g. Swellex anchors, split-set anchors). A frictionlocked bond with the rock mass results for the expansion shell bolt. In the case of the mortar anchor and the resin anchor, the bond between the anchor rod and the rock is effected by a setting agent (cement mortar or synthetic resin mortar) over a specific length of the borehole (Fig. 2.23). If the bond extends over the full length of the borehole, the terms full bond anchor or fully cemented anchor are used as well. Among the full bond anchors are also the friction pipe anchors, the anchor rod of which consists of a pipe folded in the longitudinal axis (Swellex-anchor) resp. slit open (split-set-anchor). In the borehole, this pipe is braced against the rock mass by pressing it against the borehole wall. Expansion shell bolts are rock bolts in the case of which the bottom end of the anchor rod is braced against the borehole wall using wedge-shaped or conical elements (expansion elements) (DIN 21521, Fig. 2.22d). The use of SN-anchors, resin anchors, friction pipe anchors and expansion shell bolts requires that the boreholes for the installation of the anchors are stable. Fig. 2.23 shows the working steps for the installation of a mortar anchor (SN-anchor). For unstable boreholes so-called injection drill bolts (IBO-bolts) are used (Fig. 2.22c and 2.24). Injection drill bolts consist of an anchor pipe which is made of high-strength steel with a continuously rolled-on thread and constitutes the drill bar. A drill bit is screwed on to the bottom end of the anchor pipe. After the borehole has been drilled to the desired depth, it remains with the anchor pipe in the borehole. The bond between the anchor pipe and the rock mass is accomplished by the injection of cement suspension via the anchor pipe through injection openings in the drill bit and the anchor pipe. The loosened rock mass surrounding the borehole is also injected and stabilized in the process (Fig. 2.24). Expansion anchors constructed as injection bolts (Fig. 2.22e) represent a combination of a friction-locked and form-locked connection between the anchor rod and the rock mass. The working steps for the installation of this anchor type are shown in Fig. 2.25.

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- 33 -

Fig. 2.23:

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Installation of a mortar anchor (SN-anchor)

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- 34 -

Fig. 2.24:

Installation of an injection drill bolt: a) Drilling; b) grouting

Fig. 2.25:

Installation of an expansion anchor constructed as an injection bolt (Ischebeck Titan Ltd.)

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- 35 In tunneling, non-prestressed (untensioned) anchors are generally installed as systematic anchoring for the excavation support in a raster determined on the basis of statical criteria. Anchor lengths between 3 and 6 m are common. Using sleeve connections, anchor lengths up to approx. 18 m are feasible. The anchors basically consist of an anchor rod and an anchor head with an anchor plate. Anchor manufacturers offer anchor rods made from steel and glass fiber reinforced synthetics with different strengths and crosssections (plain section, pipe cross-section). The advantages of glass fiber anchors over steel anchors lie mainly in the fact that they can be cut as well as bent. Glass fiber anchors are therefore often installed at locations where they have to be removed in the course of further excavation, e. g. for the support of the inner walls during sidewall adit heading or for the support of the tunnel face. Disadvantages of glass fiber anchors are the facts that they can carry only very small shear forces and that the transfer of point loads into the anchor at the anchor head is difficult to enable with the anchor design. The anchor heads should always be constructed in such a way that the anchor plates lie flat on the excavation surface or on the shotcrete and that the load transfer from the anchor plate to the anchor rod does not lead to bending or shear loading of the anchor rod. Therefore, mainly so-called sphere cap anchor plates are used that are spherically shaped around the hole. For the nuts screwed onto the anchor rods, so-called sphere cap nuts are used, which have a spherical surface in the contact area with the anchor plate. The anchor plates should have minimum dimensions of 150 x 150 x 10 mm. Special anchor head designs have been developed for the use of non-prestressed anchors in squeezing rock. If a certain anchor load is reached, these anchor heads yield, thus avoiding overstressing of the anchors. Since these anchor types are specialpurpose designs adjusted to the individual case, they will not be dealt with here in more detail. The admissible anchor force is the maximum force the rock bolt is permitted to be subjected to (maximum tensile force, maximum bond force or limit creep force) divided by a factor of safety. The maximum tensile force of the anchor rod is calculated according to WBI-PRINT 5

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- 36 DIN 21521 from the relevant cross-section in connection with the design strain limit and the design tension yield stress of the used material, respectively. The maximum bond force between the anchor rod and the rock mass must be determined for the individual case by pull tests according to DIN 21521. The limit creep force is the force which leads to the chosen creep rate according to DIN 4125 in the pull test. For glass fiber anchors, attention must be paid to the fact that the failure load of the anchor rods is generally far higher than the failure load of the anchor head parts. 2.3.2

Load-carrying behavior

In tunneling, rock bolts are installed as single anchors to support individual rock wedges susceptible to sliding, as surface anchoring to support the excavation surface (tunnel walls, tunnel face), and as systematic anchoring to improve the load-carrying capacity of the rock mass. Surface and systematic anchorings furthermore serve the purposes of supporting regions prone to collapse and of improving the load transfer between the steel sets and the shotcrete on the one hand and the rock mass on the other. For systematic anchorings mostly fully cemented anchors (SNanchors) or injection drill bolts (IBO-bolts) are used. Rock bolts must carry tensile and possibly shear forces. Failure of the anchors due to overstressing may occur. The failure mode of bond anchors depends on whether they are placed in rock or in soil. In rock, generally the anchor rod breaks before the bond fails. In soil, failure of the mortar over the bond length occurs first. Systematic investigations into the mechanism of operation and the load-carrying capacity of cemented steel anchors were first carried out by Bjurström (1974). These investigations led to formulae for the shear resistance of inclined and not inclined anchors. Further relations for the shear resistance were provided by Azuar and Panet (1980). Empirical equations for the shear and tensile resistance were developed by Dight (1983). Spang and Egger (1989) provide formulae to determine the shear resistance contribution T0 by which the shear resistance of a discontinuity increases as a consequence of anchoring. This contribution can be converted into an "anchor cohesion" WBI-PRINT 5

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- 37 -

cA =

T0 A

by which the cohesion of the rock atic anchoring. In (2.1) A is the systematic anchoring. Since glass little shear resistance, they are choring to increase the rock mass

(2.1)

mass increases due to a systemeffective area per anchor of a fiber anchors only have very not suited for a systematic anstrength.

Equation (2.1) makes it possible to model a systematic anchoring in FE-analyses by an increase in the cohesion of the rock mass. As an alternative, individual anchors can be modeled by truss elements, which, however, can only transfer axial forces. The combined tension and shear loading of cemented anchors can therefore not be captured by the common truss elements. Erichsen and Keddi (1990) and Keddi (1992) report on numerical analyses of the influence of anchor design and bond between rock bolt and rock mass on the load-carrying behavior of cemented anchors. 2.4

Advance support

2.4.1

Spiles

In loose rock, steel spiles are used as advance support of the workspace at the tunnel face in order to limit overbreak and to protect the miners against falling rock. The spiles are arranged in the roof area of the tunnel approx. parallel to the tunnel axis in the form of a fan. They are installed before the underlying round is excavated. The length of the spiles should be at least three times the round length in order to enable sufficient overlap. The spacing between the spiles should not exceed 30 cm. Fig. 2.26 shows exemplarily the advance support using spiles at the sidewall adit excavation of the Limburg Tunnel of the new railway line Cologne – Rhine/Main in longitudinal section and crosssection.

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Fig. 2.26:

Advance support using spiles (spile umbrella Limburg Tunnel, new railway line Cologne – Rhine/Main): a) Longitudinal section A-A; b) cross-section B-B

One distinguishes mortar spiles, driven spiles and injection drill spiles. WBI-PRINT 5

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- 39 Mortar spiles made from rebars are installed with lengths of 3 to 6 m. They are designed for a rock mass in which the drillholes are stable without a casing. The rebars (∅ 25 – 28 mm) are installed in mortar-filled drillholes. By the use of a accelerating admixture or by choosing an appropriate time of installation, respectively, it is ensured that the mortar achieves sufficient strength by the beginning of the following excavation works. Fig. 2.27 shows mortar spiles that were blasted free during excavation. In this case they do not fulfill their original purpose, which is to limit overbreak and to protect the workspace against falling rock. The blasting was not carried out smoothly with respect to the rock.

Fig. 2.27:

Mortar spiles with lattice girder, blasted free

Driven spiles consist of steel pipes 3 to 6 m long, which are driven into pre-drilled holes. The diameter of the steel pipes is slightly larger than the diameter of the drillholes. If the steel pipes are designed correspondingly and the rock mass is groutable, a later rock mass improvement using injections is possible. Fig. 2.28 shows driven steel pipe spiles installed as advance support in the Tunnel Deesener Wald of the new railway line Cologne – Rhine/Main. WBI-PRINT 5

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Fig. 2.28:

Driven steel pipe spiles (Tunnel Deesener Wald, new railway line Cologne – Rhine/Main)

Injection drill spiles (IBO-spiles) are installed with a length of 4 to 12 m. They are used in rock mass conditions where stable holes cannot be drilled and/or the rock is to be strengthened by grouting with cement suspension at the same time. Depending on the diameter of the anchor rod, the holes for the spiles are drilled with a diameter ranging from 42 to 76 mm. The spiles consist of steel pipes with a lost drill bit, which are used for flushing during drilling and for grouting the drillhole together with the surrounding rock mass after the planned depth is reached. Installation and grouting of injection drill spiles are carried out corresponding to Fig. 2.24. 2.4.2

Pipe umbrellas

If the spiles described in section 2.4.1 do not offer sufficient support, pipe umbrellas are used. They are applied mostly in coheWBI-PRINT 5

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- 41 sionless, loose ground if streets or structures are undercut with little cover. Just as spiles, pipe umbrellas are constructed over a certain part of the circumference of the excavation profile, preceding the excavation. Because of the dimensions of the pipes, pipe umbrellas are far stronger than spile umbrellas and extend further in advance of the excavation. Instead of pipe umbrellas, composite pile and jet grouting umbrellas are also constructed in connection with the NATM for the advance support of the workspace at the tunnel face. In the following, only pipe umbrellas will be covered. Examples for composite pile umbrellas and jet grouting column are described in Chapters 3.2 and 7. The jet grouting technique is covered in detail in Chapter 7. Two systems for the construction of pipe umbrellas are distinguished: -

Pipe umbrella with niches (Fig. 2.29), pipe umbrella without niches (Fig. 2.30).

If the pipe umbrella is constructed from niches, the excavation profile of the tunnel is widened in the course of the excavation, so that the drill points for the steel pipes of an umbrella are located outside of the shotcrete membrane (Fig. 2.29). Thus, the geometry of the steel sets must be adapted for each round and the niches must at least partially be filled with shotcrete before the sealing and the interior lining are installed. If the pipe umbrella is constructed without niches, the drill points for the steel pipes are located in the tunnel face of the standard excavation profile (Fig. 2.30). In the course of further excavation, the pipe sections lying within the cross-section of the shotcrete membrane must be cut off after each round. Pipe umbrellas are constructed with lengths of ca. 15 to 30 m with a heavy drill rig (e. g. drill carriage, Fig. 2.31). The outer diameter of the steel pipes used varies between 76 mm and 200 mm, while the wall thickness of the pipes ranges from 8 mm to 25 mm. The pipes serves as the casing while the boreholes are drilled, and remain in the soil or rock, respectively, as a structural element of the pipe umbrella. After the installation of the pipes, the annular gap between the pipes and the rock mass and as far as WBI-PRINT 5

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- 42 possible also the surrounding rock mass itself are grouted with cement suspension from out of the pipes in order to improve the support effect of the pipe umbrella. To this end, the pipes are provided with injection valves at a spacing of 0.5 to 1 m (Fig. 2.32). The pipes are subsequently closed with a cover and filled with suspension or mortar.

Fig. 2.29:

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Pipe umbrella with niche (Niedernhausen Tunnel, new railway line Cologne – Rhine/Main): a) Longitudinal section; b) Detail A WBI GmbH, Henricistr. 50, 52072 Aachen, Germany

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- 43 -

Fig. 2.30:

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Pipe umbrella without niche (Dernbach Tunnel, new railway line Cologne – Rhine/Main): a) Longitudinal section; b) Detail A; c) Detail section 1-1 WBI GmbH, Henricistr. 50, 52072 Aachen, Germany

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- 44 -

Fig. 2.31:

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Pipe umbrella in the Dernbach Tunnel (new railway line Cologne – Rhine/Main): a) Pipe umbrella; b) installation of a pipe WBI GmbH, Henricistr. 50, 52072 Aachen, Germany

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- 45 -

Fig. 2.32:

Pipe with injection valve

The boreholes for the pipe umbrella are arranged over a certain portion of the circumference of the excavation profile at a spacing of 30 to 50 cm. They are drilled ascending at an angle of approx. 5° with respect to the tunnel axis. The pipes of two pipe umbrellas should overlap by at least 3 m in the longitudinal tunnel direction (Fig. 2.29 and 2.30). In soil and in weathered, soft rock the boreholes are generally drilled with an auger and air and/or water flushing (Fig. 2.33). In rock or for large pipe diameters, however, mostly down-hole hammers are used. Here, the drill bit and the pipe are either connected by a bayonet joint so that the casing is continuously pulled into the borehole with the advance of the drill pipe, or the pipes are pushed hydraulically by the drill rig. After the final depth is reached, the drill pipe and the drill bit are released from the pipe and pulled out.

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- 46 -

Fig. 2.33: WBI-PRINT 5

Auger for the installation of pipes

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Fig. 2.34:

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Construction of the pipe umbrella at the starting wall of the Dernbach Tunnel (new railway line Cologne – Rhine/Main) WBI GmbH, Henricistr. 50, 52072 Aachen, Germany

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- 48 Fig. 2.34 shows the construction of a pipe umbrella through the starting wall of the Dernbach Tunnel of the new railway line Cologne – Rhine/Main. The starting wall is supported by shotcrete and anchors. In this case, the shotcrete support of the starting wall was drilled through before the pipes for the pipe umbrella were installed. 2.5

Geotechnical mapping and monitoring

2.5.1

Mapping

Before the excavation of a tunnel, in general only surface exposures and boreholes are available to assess the ground. Their results must be extrapolated to the planned tunnel and its surrounding area. Because of the ensuing uncertainties in assessing the ground conditions, mapping of the temporary tunnel face during construction is absolutely necessary. They form an essential basis for adapting the support to the local conditions. The extent of a geotechnical mapping of the tunnel face depends essentially on the available time. A "detailed mapping" including comprehensive information on the intact rock, the fabric and the groundwater can usually only be carried out if the tunnel excavation is halted. With routine mappings during the excavation, it is in general only possible to record these properties randomly. In this case, however, it is important to recognize and document deviations or changes in the local conditions. In a detailed mapping, the following properties should be recorded as far as possible (Wittke, 1990): -

Rock types and rock boundaries (boundaries of layers),

-

degree of weathering of the intact rock,

-

orientation of the discontinuities (strike and dip angle),

-

location, spacing and trace lengths of discontinuities, i. e. of the intersections of the discontinuities with the excavated rock surface,

-

location and shape of visible discontinuities lying at the rock surface, WBI-PRINT 5

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-

opening widths of open or filled discontinuities,

-

fillings and coatings of discontinuities,

-

location, dimensions and properties of large discontinuities and faults,

-

location of seepage water and quantity of discharge.

Apart from these information, the field protocol should include further peculiarities, such as for example information on discontinuities which tracings are so closely spaced that they cannot be recorded individually for reasons of time alone. Half-quantitative descriptions such as "strongly jointed" or "strongly fractured" are sufficient in this context. These descriptions should be complemented with information on the magnitude of the spacing in cm or dm. Further, recordings of the unevenness and roughness of the discontinuities and of possible indications of weathering on the discontinuity surfaces should be made. The discontinuity fillings just as bedding parallel and foliation parallel discontinuities can be labeled by special signatures (Wittke, 1990). During mapping, samples for complementing laboratory tests must be taken, if indicated. The standard mapping gear consists of -

a geological compass to measure the orientation of discontinuities,

-

a measuring tape to record the location of the outcrops and the tracings of discontinuities,

-

a rock hammer to make discontinuities visible that are e. g. covered by a thin weathered layer, to loosen rock samples or to sound the rock (clear or dull sound),

-

a note pad in which a sketch of the mapping is entered. It is more appropriate, however, in most cases to use prepared forms for the mapping (Fig. 2.35).

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- 50 -

Fig. 2.35: WBI-PRINT 5

Example of a tunnel face mapping form WBI GmbH, Henricistr. 50, 52072 Aachen, Germany

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- 51 Fig. 2.36 shows an example of a detailed mapping of a tunnel face. For reasons of clarity, the information on the orientation and nature of the discontinuities is omitted here.

Fig. 2.36:

Detail mapping of a tunnel face (Hasenberg Tunnel, Stuttgart urban railway)

Examples for mappings during excavation are given in Chapters 3.1, 3.3 and 4.2 (see Fig. 3.20, 3.41 and 4.43). Discontinuity orientations are measured using the geological compass (Fig. 2.37). The dip direction αD is read from the compass circle, while the dip angle β is read from the vertical circle. The dip direction αD is the angle between the projection of the dip line of the discontinuity onto the horizontal plane and north. αD is correlated with the strike angle α by the relation given in Fig. 2.37 (Wittke, 1990).

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- 52 -

Fig. 2.37:

Measuring the orientation of a discontinuity using the geological compass (Wittke, 1990)

The orientation of a discontinuity, given by the angles αD and β or α and β, respectively, can be represented in a polar equal area net by a point, the so-called pole. A polar equal area net is an equal-area hemispheric projection of the lower half of the reference sphere. The term reference sphere denotes a sphere the equator of which is located in the horizontal plane. The pole is the projection of the point of intersection of the normal to the discontinuity through the center of the reference sphere onto the polar equal-area net (Wittke, 1990). Entering the discontinuity orientations measured during one or several mappings in the polar equal-area net leads to a so-called polar diagram. An example of a polar diagram prepared in this way and for the grouping of the measured discontinuities into several sets is shown in Fig. 3.42 (Chapter 3.3.6).

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- 53 For a great number of measured discontinuity orientations, or if the measured discontinuity orientations scatter considerably, a statistical evaluation of the discontinuity orientations may be appropriate. To this end, instead of the poles, areas representing the angular regions of equal relative frequency of the poles are entered in the polar diagram. These areas are assigned to pole densities. The pole density denotes the number of poles located on 1 % of the area of the polar equal area net, relative to the total number of measured discontinuity orientations, in percent (Wittke 1990). In this way, the areas with the most frequent orientations can be identified for the individual discontinuity sets. An example of the representation of discontinuity orientations by areas of equal pole density is shown in Fig. 5.39 (Chapter 5.2.6). The mapping results should be summarized for homogeneous areas in a geometric structural model. To this end, first the parameters describing the geometry of the discontinuity fabric are determined on the basis of a statistical evaluation of the measured fabric data. For each discontinuity set, the mean values of the measured dip and strike angles, of the spacing and, if possible, of the tracings of the discontinuities must be specified. However, since especially the spacing and tracings of discontinuities scatter considerably, in addition to the mean values the standard deviations of these parameters should be specified as well (Wittke, 1990). In addition to these data, a structural model should also include a description of the properties of individual discontinuities, which form the sets. This description should contain information on the unevenness and the roughness as well as on the degree of weathering of the discontinuity surfaces, which can be obtained e. g. from the mapping of profiles. Further, it should be noted whether the discontinuities are closed or open, and whether and to which degree they are filled. Inhomogeneities such as discontinuities with an extent reaching the magnitude of the dimensions of the tunnel must be described individually with respect to their location and orientation. Finally, the model should be illustrated in a block figure containing as many as possible of the referenced parameters (Fig. 2.38).

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- 54 -

Fig. 2.38:

2.5.2

Example of a structural model of a rock mass with three discontinuity sets (Wittke, 1990)

Monitoring

General

Just as geotechnical mapping, geotechnical measurements represent an essential element of the NATM. On the one hand, they serve to monitor -

the stability of the tunnel and of adjacent structures,

-

the deformations in the ground and the displacements on the surface,

-

the loading of the shotcrete membrane,

-

the vibrations during heading.

On the other hand, just as the mapping, they form one of the foundations for adapting the support measures to the local ground conditions. Finally, the interpretation of already existing measurement results can and should be used to verify and, if indicated, optimize the dimensioning of the temporary and permanent lining. Stability analyses on the basis of measurement results furthermore WBI-PRINT 5

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- 55 permit to reduce or remove the uncertainties associated with the assumption of the parameters (back analysis) and to capture the influences from actual construction. Geotechnical monitoring comprises: -

Positional surveying,

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leveling,

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convergency measurements,

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extensometer measurements,

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inclinometer measurements,

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stress measurements,

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anchor force measurements,

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vibration measurements,

-

water level and water pressure measurements.

Positional surveying and leveling

Displacements of points located at the ground surface and at the tunnel contour (Fig. 2.39) are measured geodetically by leveling and by using a tachymeter. To measure subsidence, measuring cross-sections with several leveling points are installed perpendicular to the tunnel axis (Fig. 2.39). In sloping locations the horizontal displacement components should be measured as well. The zero reading should be carried out when the tunnel face is still several tunnel diameters away from the respective measuring cross-section to capture the subsidence due to tunneling completely. If settlements are to be expected resulting from a lowering of the groundwater table due to tunneling, the zero reading should be carried out before the start of construction. For the measurement of the displacement of points on the tunnel contour, the measuring locations are installed closely (≤ 1 m) beWBI-PRINT 5

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- 56 hind the tunnel face. The measuring points are steel bolts cemented into the rock. Simple homing boards or triple prisms are mounted onto these bolts. The zero reading is carried out before the next round.

Fig. 2.39:

Surface leveling perpendicular to the tunnel axis and positional survey of points on the tunnel contour

With the tachymeter, not only the vertical but also the horizontal displacement components parallel and perpendicular to the tunnel axis are measured. The measuring points are surveyed threedimensionally from bench marks by determining the direction, distance and inclination from the bench mark to the measuring point for different points in time. The measurement data given in polar coordinates are converted to Cartesian coordinates. From the difWBI-PRINT 5

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- 57 ferences in coordinates from two measurements, the respective displacement vectors of the measuring points can be computed (Reik and Völter, 1996). Fig. 2.40 shows as an example the vertical displacements of measuring points on the tunnel contour during the crown heading of the southern tube of the Gäubahn Tunnel in Stuttgart at chainage 113 m (see Chapter 3.2). The upper part shows the vertical displacements of the measuring points versus time. In the lower part the advance of the tunnel face is shown versus time.

Fig. 2.40:

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Geodetically determined vertical displacements of measuring points on the tunnel contour during crown heading (Gäubahn Tunnel, southern tube, chainage 113 m)

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- 58 Convergency measurements

Convergency devices allow the measurement of changes in distance of points on the tunnel contour. Fig. 2.41 shows examples for the arrangement of convergency measurement sections for a crown heading and a sidewall adit heading. To measure the distance between two measuring points, the measuring points are connected by a tensioned measuring tape or a measuring wire and the convergency device (Fig. 2.42). The changes in distance can be determined as the difference of the measured lengths of consecutive measurements (Reik and Völter, 1996). Changes in length in the order of 0.1 mm can be registered with this technique. Since convergency measurements interfere with tunnel heading, measurements with a convergency device are rarely carried out any more in tunneling nowadays. Optical three-dimensional measurements using a tachymeter and triple prisms are preferred instead (Fig. 2.39).

Fig. 2.41:

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Convergency measurement cross-sections: a) Crown heading; b) sidewall adit heading

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- 59 -

Fig. 2.42:

Schematic drawing of a convergency measuring system (Reik and Völter, 1996)

Extensometer and inclinometer measurements

Displacements and relative displacements (extension or compression) in the ground are measured in boreholes with stationary gages or probes. Displacements parallel to the axis of a borehole are generally measured using extensometers (stationary devices) or sliding micrometers (borehole probes). Multiple extensometers are suitable for the determination of relative displacements of measuring points with larger distances. Fig. 2.43 shows the setup of a multiple-rod extensometer. Absolute displacements can be determined by tying the extensometer head (1 in Fig. 2.43) in to a bench mark by leveling, or if the deepest anchor is a fixed point. Details on displacement measurements with

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- 60 extensometers are given e. g. in Recommendation No. 15 of the German Geotechnical Society (DGGT) (Paul and Gartung, 1991).

Fig. 2.43:

Setup of a multiple-rod extensometer (Interfels, 2000)

With a sliding micrometer, rock displacements can be measured at short distances along the axis of a borehole. To this end, a plastic casing is cemented into a borehole ca. 100 mm in diameter. Measuring marks are fixed in this casing at a spacing of 1 m. Using an inserted probe, the changes in distance between the measurWBI-PRINT 5

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- 61 ing marks are recorded one after the other (German Rail, Guideline 853, DB 1999). Displacements perpendicular to the axis of a borehole are usually measured using inclinometer probes or stationary inclinometers. In the case of the inclinometer probe, a plastic pipe with four guiding grooves is installed in a borehole ca. 100 mm in diameter. With a measuring probe, the changes in inclination with respect to the borehole axis and thus the displacements perpendicular to the borehole axis can be determined in two directions at right angles to each other (Fig. 2.44).

Fig. 2.44:

Measuring principle of an inclinometer probe: a) Measurement; b) evaluation

Fig. 2.45 shows an example of the arrangement of the measuring points in the case of a combined extensometer and inclinometer measuring cross-section. Since during the undercrossing of the measuring cross-section by the tunnel none of the measuring points constitutes a fixed point, the measuring points on the ground sur-

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- 62 face are tied in to bench marks by levelings (L) and positional surveys to determine absolute displacements.

Fig. 2.45:

Measuring cross-section with extensometers and inclinometers (constructed from the ground surface)

If the vertical as well as the horizontal displacements in longitudinal and transverse direction are to be determined in one borehole instead of separate boreholes, a device can be used that allows to measure the displacements in three directions perpendicular to each other at the same time. Such a device is e. g. the Trivec probe, a combination of sliding micrometer and inclinometer probe (German Rail, Guideline 853, DB 1999). Just as for the positional surveys, the zero reading should be carried out as early as possible before the corresponding measuring cross-section is undercrossed. Stress measurements

Stress measurements are carried out e. g. to assess the loading of shotcrete membranes and interior linings. Typical arrangements of WBI-PRINT 5

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- 63 pressure cells in a shotcrete membrane are sketched in Fig. 2.46. Here, one measuring cell is arranged tangentially on the outside of the shotcrete to measure the radial stress (rock mass pressure). The other, radially arranged measuring cell, however, serves to measure the tangential stress resulting from the loading (concrete pressure).

Fig. 2.46:

Arrangement of pressure cells in a shotcrete membrane for the measurement of concrete and rock mass pressure (Wittke, 1990)

The stresses measured with the pressure cell arranged tangentially to the tunnel contours generally scatter significantly. On the one hand, this is due to the inhomogeneous stress distribution in the ground and the comparatively small dimensions of the measuring cells. On the other hand, the excavation process leads to local loosening along the tunnel contour. This exaggerates the nonuniform distribution of the radial stress, which is also referred to as rock mass stress or contact stress. Most of the times, therefore, radial stress measurements yield relevant results only in special cases (e. g. swelling rock). WBI-PRINT 5

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- 64 For radially arranged measuring cells, the scatter of the measurement results is generally significantly smaller. The tangential stresses in the shotcrete measured with these cells scatter less strongly, since even an inhomogeneous radial loading of the lining results in a relatively even distribution of the normal thrust and thus also of the tangential stresses. In addition, the dimensions of the measuring cells are generally in the same order of magnitude compared to the thickness of a shotcrete membrane. Finally, the absolute value of the tangential stresses is in general many times greater than that of the radial loading of a shotcrete membrane. Nevertheless, the measured tangential stresses may also be non-uniform, if for example the shotcrete membrane has a greatly varying thickness as a consequence of an uneven excavation profile (Wittke, 1990). Measurements of the tangential stresses in the lining and the radial stresses between lining and rock mass are carried out using hydraulic valve gages or oscillating chord gages. The valve gage of Glötzl Co. (Fig. 2.47) is a hydraulic pressure cell, in which a compressive stress develops due to the loading of the oil-filled flat jack. The measurement is effected by increasing the fluid pressure in a circuit separate from the flat jack, until the fluid circulates back due to a slight deformation of a membrane located between an inner and an outer chamber. The reflux leads to a pressure drop at a pressure gage positioned in the circuit. The maximum stress measured before the pressure drop at the pressure gage can therefore be equated to the loading of the pressure cell (Wittke, 1990). With oscillating chord gages, changes in length of a chord clamped freely oscillating in the gage are measured. These changes lead to a change in the natural frequency of the measuring chord excited to oscillate by a direct current impulse. The altered natural frequency measured while the oscillation of the chord fades out yields the chord strain. The change in stress is proportional to the measured change in length (Schuck and Fecker, 1997). Fig. 2.48 shows two oscillating chord gages of Geokon Co.

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- 65 -

Fig. 2.47:

Stress compensation measurement with pressure valve gage, system Glötzl

Fig. 2.48:

Oscillating chord gages, system Geokon (Geokon, 1993)

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- 66 If high stresses must be expected in the ground, the in-situ or primary stress state in the rock mass is determined. Aside from the already mentioned interpretation of displacement measurements using numerical analyses, which in general is only possible during construction, stresses can be measured by stress relief overcoring tests, also referred to as stress measurements by the overcoring method. In this method, which can only be applied to rock, the stress state in the rock mass is derived from the deformations of a drill core due to unloading. To this end, closed-form solutions for elastic isotropic and elastic anisotropic stress-strain behavior of the rock mass are applied (Kiehl and Pahl, 1990). Further, in-situ stresses can also be determined by the hydraulic fracturing method or the hard inclusion method in boreholes. Another possibility to measure stresses or stress changes are compensation measurements using flat jacks inserted in sawed or drilled slots. Anchor force measurements

The anchor forces of untensioned anchors are measured in special cases only. In the case of tendons with a free anchor length, force measuring cells equipped with strain gages or oscillating chord gages can be installed at the anchor head. Lately, the technique of tension measurement with integrated optical fiber sensors has been developed that can be used also with fully cemented anchors. Vibration measurements

In the case of a smooth blasting excavation, vibrations generally need not be considered with respect to the stability of tunnels. They may have an impact, however, on neighboring structures. Therefore, in these cases vibration velocity measurements are carried out to verify and ensure that the reference values according to DIN 4150 (Parts 2 and 3, 1999) are complied with. Compliance with these values is controlled primarily by limiting the maximum charge per ignition step in the case of a smooth blasting excavation (Wittke and Kiehl, 2001; DIN 4150, Part 1, 2001).

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- 67 3.

Crown heading with open invert

3.1

Glockenberg Tunnel near Koblenz, Germany

3.1.1

Introduction

In the course of the improvement of the federal highway 42 (B 42) between Bendorf and Lahnstein, Germany, the Glockenberg Tunnel was constructed on the side of the Rhine river opposite to the city of Koblenz. This tunnel enables a non-intersecting access from the B 42 to the Pfaffendorf Rhine bridge. The southern portal of the tunnel lies in the direct extension of the Pfaffendorf Bridge. For each direction, the tunnel includes one driving lane, one stopping lane and an emergency sidewalk. In the mountain, the tunnel axis performs a 180° turn with a radius of ca. 50 m and then branches out into two single-lane tunnel tubes. One of these tubes accesses the B 42 in the direction of Bendorf, while the other one enables exiting the B 42 coming from Lahnstein into the direction of Koblenz (Fig. 3.1).

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- 68 3.1.2

Structure

The 315 m long Glockenberg Tunnel was constructed by underground excavation starting at the southern portal 3 (Fig. 3.2). Up to location 220.5 m, the tunnel was excavated with the two-lane crosssection TC 3 shown in Fig. 3.3. From then on the cross-section was widened (TC 3A) up to chainage 248 m. In this way, the tunnel cross-section TC 3 is transformed into the two smaller tunnel tubes with the cross-sections TC 1 and TC 2. The smaller tunnel tubes each have a length of approx. 40 m and end at the portals 1 and 2 (Fig. 3.1).

Fig. 3.2:

Glockenberg Tunnel, portal 3, crown

For the two-lane standard profile TC 3, a mouth-shaped profile with a width of 17.5 m and a height of 12 m was carried out. The excavated cross-section amounts to 175 m2. The clearance is approx. 15 m wide and approx. 4.5 m high. The thickness of the shotcrete membrane is 25 cm, while the interior lining made from watertight concrete is 60 cm thick (Fig. 3.3). In the vault area, a radius of curvature of R = 8.43 m was selected for the interior lining. The sidewalls were constructed with a radius of R = 4.66 m. With a radius of R = 25.25 m, the inWBI-PRINT 5

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- 69 vert was only slightly rounded out. For the transition from the sidewalls to the invert, a radius of R = 2.66 m was selected (Fig. 3.3).

Fig. 3.3:

Glockenberg Tunnel, tunnel cross-section TC 3

Fig. 3.4 shows the vertical section through the tunnel axis in the area of tunnel cross-sections TC3 and TC3A as a development. The maximum overburden height amounts to approx. 50 m.

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- 70 -

Fig. 3.4: WBI-PRINT 5

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- 71 3.1.3

Ground and groundwater conditions

The ground along the alignment of the Glockenberg tunnel was explored using core drillings and test pits. One borehole was equipped as an observation well (Fig. 3.4). According to the exploration, an up to 9 m thick soil layer consisting of talus material, residual detritus, loess and loess loam is underlain by an alternating sequence of slate and fine- to medium-grained sandstones belonging stratigraphically to the Lower Devonian of the Emsian stage (Fig. 3.4). In the course of the Variscan orogenesis phase, the originally horizontal strata series were narrowed in NW-SE direction and bent to folds of different sizes. Within the large-scale folding structure, the alternating sequence of ribboned or flaser-like, locally secondarily silicated slates with sandstone banks up to several meters thick is in parts specially folded and sheared. The sandstones are mostly quartzitically, locally also calcareously and ferrugineously cemented. Quartz and calcite veins millimeters to centimeters thick often occur in the rock zones rich in sand. According to the results of the petrographic microscopy and X-ray diffractometry investigations, quartz, feldspar, mica and carbonates are the main rock-forming components. The tunnel area is characterized by a distinct tectonic deformation in the form of a "special folding". Shallow as well as steep dipping of the strata to the northwest and southeast, respectively, occurs here with changes taking place over very short distances. The rock is fractured by bedding-parallel discontinuities, the foliation and joints. As a result of the folding structure, the orientation of the discontinuities varies within the tunnel area. Fig. 3.5 shows the mean discontinuity orientations measured on oriented drill cores.

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- 72 -

Fig. 3.5:

Glockenberg Tunnel, site plan showing the orientation of the discontinuities and the location of the expected fault zones

As far as it can be assessed from the surface exposures, the bedding-parallel discontinuities must be assumed completely separated. Due to the exceeding of the shear strength on the beddingparallel discontinuities as a result of shear during folding, triturated material and slickenside lineations locally exist on these planes. The same applies to quartz and calcite segregations which deposited in the cracks that developed due to dilatancy during the folding process. Judging from surface exposures in the surroundings of the tunnel, the extension and the spacing of the foliation discontinuities depend distinctly on the appearance of the intact rock. In the case of greater thickness of the banks and mostly in the clayey rock WBI-PRINT 5

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- 73 zones, individual completely separated foliation parallel discontinuities exist. Locally also slickenside lineations and triturated material were found on the foliation parallel discontinuities, indicating tectonic loading. The joints contain many more rock bridges than the bedding- and foliation parallel discontinuities. Their degree of separation is correspondingly smaller. In some core drillings, locally an increased core fragmentation, a more frequent appearance of slickensides parallel to the bedding or to the foliation, numerous quartz and calcite veins as well as rather strongly disintegrated rock zones were encountered. This led to the assumption that a fault zone striking in NE-SW direction in the area of tunnel cross-section TC 3 may be present (Fig. 3.5). Another fault zone was assumed in the area of tunnel crosssection TC 2 on the basis of the exploration results (Fig. 3.5). Beneath the soil cover, which was encountered down to a depth of 9 m below ground surface during the exploration as mentioned above, the rock has only slightly been altered by weathering processes. An exception are more strongly fragmented rock zones, e. g. fault zones, in which weathering has clearly progressed. To investigate the hydrological conditions, the groundwater level in the observation well OW (see Fig. 3.4) was measured and permeability tests were carried out at different depths in the boreholes. The observation well was set up in the area in which the highest groundwater level was expected due to the location on the slope. The measurements showed that the groundwater is encountered here approx. 1 m above the tunnel roof (see Fig. 3.4). 3.1.4

Excavation classes

Because of the size of the excavated cross-section of approx. 175 m2 (see Fig. 3.3), it was necessary to carry out excavation and support separately for crown, bench and invert (Fig. 3.6).

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- 74 -

Fig. 3.6:

Excavation classes 4, 5 and 6A for tunnel crosssection TC 3 (final design)

As already mentioned, a uniform orientation of the discontinuities could not be assumed. In addition, due to the circular tunnel WBI-PRINT 5

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- 75 alignment, the orientation of the discontinuities relative to the tunnel axis changed continuously with the heading. Further, fault zones locally had to be expected (see Fig. 3.5). Therefore frequently varying rock conditions were assumed for the tender design and the final design. Correspondingly, the excavation classes 4, 5, 6A and 7A were specified in the final design in accordance with the recommendations of the working group "Tunneling" of the German Geotechnical Society (DGGT, 1995: Table 1). A crown heading with an open invert was planned. The excavation class as well as the trailing distances of the bench excavation and the closing of the invert lining (E and F in Fig. 3.6) were to be selected in the course of the crown heading on the basis of the results of mapping and monitoring during construction (see Chapter 3.1.6). In this context special emphasis was placed on the results of the stability analyses (see Chapter 3.1.5) and of the geotechnical mapping. The excavation methods, round lengths and support measures specified for the excavation classes 4, 5 and 6A for tunnel crosssection TC 3 in the final design are listed in Fig. 3.6. The excavation class 7A including advance support and support of the tunnel face was not carried out during construction. Heading was planned by smooth blasting and/or mechanical excavation by a tunnel excavator and hydraulic chisel. The unsupported round length during crown heading was selected as 2.0 m for excavation class 4 and 1.5 m for excavation classes 5 and 6A. In excavation class 6A an advance support of the crown with 4 to 5 m long mortar spiles was planned for every second round. Excavation class 5 included a sealing of the temporary crown face with shotcrete (Fig. 3.6). For the shotcrete membrane, shotcrete of grade B25 with a thickness of 25 cm and two layers of reinforcing mats Q295 were specified. In the crown area, a systematic anchoring with 7 SN-anchors (excavation classes 4 and 5) resp. 9 SN-anchors (excavation class 6A) per round was planned. In each round a lattice girder was to be installed in the crown area (Fig. 3.6). Due to reasons of construction management, a continuous crown heading was aimed for. If a trailing bench and invert should beWBI-PRINT 5

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- 76 come necessary for stability reasons, the maximum trailing distance E resp. F should amount to two tunnel diameters D (Fig. 3.6). Unsupported round lengths for bench and invert of 4 m for tion class 4 and 3 m for excavation classes 5 and 6A were In the bench area, a systematic anchoring with 12 anchors tion classes 4 and 6A) resp. 10 anchors (excavation class round was specified. In each round one lattice girder was extended to the bench invert (Fig. 3.6).

excavaplanned. (excava5) per to be

The anchors were designed to be 4 to 6 m long in the crown as well as in the bench for all excavation classes (Fig. 3.6). If fault zones had to be crossed, locally longer anchors were planned to be installed. 3.1.5

Stability analyses

In order to investigate the stability and to dimension the shotcrete membrane, two-dimensional FE-analyses were carried out using the program system FEST03 (Wittke, 2000). Fig. 3.1 shows the analysis cross-sections AC 1, AC 2 and AC 3 investigated for the tunnel cross-section TC 3. Because of the varying orientations of the discontinuities in the tunnel area and the changing direction of the tunnel axis due to the circular plan of the tunnel, four idealized structural models were established. All of them assume that the discontinuities strike parallel to the tunnel axis. They are thus on the conservative side with respect to the stability and the support measures ensuing from the analyses (see Fig. 3.5 and 3.7). Further, the dip angles of the discontinuities were varied over wide ranges (Fig. 3.7). In model A, the bedding (B) and the foliation (F) are assumed dipping each at 45° perpendicular to the tunnel axis in opposite directions. In addition, a vertical joint set (J) striking parallel to the tunnel is taken into account. In model B the foliation is assumed dipping at 30° while the bedding dips at 60° in the opposite direction. Here as well an additional vertical joint set exists. Model C includes a vertical bedding striking parallel to the tunnel and a horizontal foliation. Instead of the vertical bedding planes, vertical joints are assumed in model D. WBI-PRINT 5

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- 77 Aside from these different structural models, cases with a fault zone located in the area of the tunnel were investigated as well. Fig. 3.8 shows the assumed locations of these fault zones. Position L assumes a 5 m thick fault zone dipping at 70° close to the left side of the tunnel. With position M a vertical, 5 m thick fault zone running through the tunnel cross-section is taken into account. Position R assumes a 5 m thick fault zone dipping at 70° close to the right side of the tunnel.

Fig. 3.7:

Considered structural models: a) Model A; b) model B; c) model C; d) model D

In Fig. 3.9 and 3.10 the pseudo three-dimensional FE-mesh (slice with thickness of 1 m) used for analysis cross-section AC 3 is WBI-PRINT 5

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- 78 shown. The FE-meshes for analysis cross-sections AC 1 and AC 2 have a corresponding setup, except for the overburden height.

Fig. 3.8:

Accounting for fault zones

The selected computation section is 140 m wide in transverse direction of the tunnel (x-direction). The height amounts to 120 m. The FE-mesh consists of 3288 three-dimensional, isoparametric elements with 7777 nodes (Fig. 3.9). The boundary conditions consist of vertically sliding supports for the nodes on the vertical boundary planes and of horizontally sliding supports for the nodes on the lower boundary plane (z = 0). All nodes are fixed in the ydirection.

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- 79 -

Fig. 3.9:

Analysis cross-section AC 3, FE-mesh, boundary conditions, ground profile and parameters (reference case)

The setup of the ground layers is the same for all analysis crosssections. The rock mass is encountered below a 5 m thick surface soil layer. In some analyses a fault zone in one of the positions shown in Fig. 3.8 is modeled discretely. The parameters assumed in the stability analyses for the soil and the fault zone correspond to the specifications established during

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- 80 the tender design. The rock mechanical parameters were specified in accordance with the parties concerned.

Fig. 3.10:

Analysis cross-section AC 3, FE-mesh, detail

For the reference case, Young's modulus of the rock mass was assumed conservatively as E = 1000 MN/m2. In comparative analyses a value of 1500 MN/m2 was specified. Poisson's ratio was assumed as 0.33 (Fig. 3.9). The angles of friction on the bedding-parallel and foliation parallel discontinuities were assumed as ϕb = ϕf = 24°. For the friction angle on the joints a value of ϕj = 24° was selected (Fig. 3.9).

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- 81 For the bedding parallel discontinuities no cohesion was assumed in all cases. The same applies to the foliation parallel discontinuities for the reference case (Fig. 3.9). In comparative analyses a cohesion of cf = 100 kN/m2 resp. cf = 200 kN/m2 was specified. The cohesion on the joints was selected as cj = 150 kN/m2 in the reference case (Fig. 3.9). In the comparative analyses it was assumed as cj = 200 kN/m2 and cj = 300 kN/m2, respectively. The transfer of tensile stresses perpendicular to the discontinuities was excluded in all analysis cases. A value of 15000 MN/m2 was specified as statically effective Young's modulus of the shotcrete. This value is to include the hardening during the application of the load (see Chapter 2.1.5). As mentioned before, the thickness of the shotcrete membrane was selected as t = 25 cm (Fig. 3.10, see also Fig. 3.3 and 3.6). The analyses were based on elastic-viscoplastic stress-strain behavior (Wittke, 2000) of the ground and elastic stress-strain behavior of the shotcrete membrane. According to the findings of the exploration, locally a ground water table lying above the tunnel roof had to be expected. With the drainage during tunnel excavation, the ground water table is lowered to the level of the tunnel invert. There was no water pressure therefore to be considered in the stability analyses for the shotcrete membrane. Fig. 3.11 shows a schematic representation of the computation steps for the simulation of the construction stages "crown excavation" and "bench excavation" and the excavation of the full crosssection. The first computation step includes the computation of the state of stress and deformation resulting from the dead weight of the ground (in-situ state). In the computation steps preceding the simulation of the individual partial excavations (computation steps 2, 4 and 6), a preceding stress relief is accounted for in the respective cross-section area to be excavated (Wittke, 2000).

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Fig. 3.11: WBI-PRINT 5

Analysis cross-section AC 3, computation steps WBI GmbH, Henricistr. 50, 52072 Aachen, Germany

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- 83 To this end, Young's modulus in this area is reduced to the value Ered = av · E where E denotes Young's modulus of the ground and av < 1.0 is the so-called stress-relief factor. The preceding stress relief enables an approximate representation of the deformations and stress redistributions occurring in the rock preceding the tunnel excavation and the installation of the support. Due to the preceding stress relief, the shotcrete membrane installed in the following computation step is subjected to a smaller loading compared to a simultaneous simulation of excavation and installation of the shotcrete support without the intermediate step of the preceding stress relief. The stress relief factor is specified on the basis of experience gained from measurements on completed structures and comparisons with the results of three-dimensional analyses. In the present case, the preceding stress relief factor was chosen as av = 0.5 (Fig. 3.11). In addition to the installation of the shotcrete membrane, the installation of the SN-anchors in crown and bench for excavation class 6A (see Fig. 3.6) is simulated in computation steps 2 (crown excavation and support) and 4 (bench excavation and support). The anchors are modeled by 4 m long truss elements (Wittke, 2000; Fig. 3.11). In the following, the results of stability analyses for analysis cross-section AC 3 (see Fig. 3.1, 3.9 and 3.10) are shown exemplarily. Analysis cross-section AC 3 is located in the area of the maximum overburden of approx. 50 m (Fig. 3.9). In this area the occurring discontinuity orientations can be captured approximately with structural models B and C (see Fig. 3.7). The construction stage after the bench excavation (5th computation step) proves to be critical with respect to the stability of the tunnel, if structural model B and the parameters given in Fig. 3.9 are assumed. If the anchors are not taken into account, the displacements do not converge in the course of the viscoplastic iterative calculation in the 5th computation step, i. e. the stability of the tunnel cannot be proven by the analysis for this construction stage. Even if the anchors are taken into account, considerable viscoplastic displacements still result in this computation step, especially at the left bench lining toe. These displacements converge, however, in the course of the viscoplastic WBI-PRINT 5

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- 84 iterative calculation. Fig. 3.12 and 3.13 show the principal normal stresses, the areas where the shear strength on the discontinuities is exceeded and the displacements of the excavation profile computed for the bench excavation with the anchors taken into account. The computed roof subsidence amounts to approx. 40 mm, the maximum heave of the invert to approx. 65 mm (Fig. 3.13).

Fig. 3.12:

Analysis cross-section AC 3 (structural model B), principal normal stresses and exceeding of strength, bench excavation (5th computation step)

If structural model C is assumed, the stability of the tunnel can be proven in the analysis for all construction stages even if the systematic anchoring is not taken into account. Also, if a greater WBI-PRINT 5

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- 85 shear strength is assumed on the discontinuities, the computed displacements become considerably smaller.

Fig. 3.13:

Analysis cross-section AC 3 (structural model B), displacements, bench excavation (5th – 1st computation step)

Fig. 3.14 shows the bending moments and normal thrust in the shotcrete membrane calculated for the 5th computation step. For a safety factor of 1.35, the design results in a statically required reinforcement of up to 4.0 cm2/m. If the lattice girders are taken into account, this reinforcement is covered by the planned reinforcement (see Fig. 3.6).

Fig. 3.14:

Analysis cross-section AC 3 (structural model B), stress resultants in the shotcrete membrane, bench excavation (5th computation step)

In Fig. 3.15 the tensile anchor forces computed for the truss elements for the 7th computation step are shown. They reach a maximum value of approx. 130 kN. WBI-PRINT 5

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Fig. 3.15:

Analysis cross-section AC 3 (structural model B), tensile anchor forces, 7th computation step

The admissable anchor force of the installed anchors adm FA results from comparing the anchor force divided by the cross sectional area As with the tension yield point of the anchor steel βs taking into account a factor of safety of η = 1.75. SN-anchors made from BSt 500S (βs = 500 N/mm2) with a diameter of 25 mm were selected (see Fig. 3.6). This leads to the admissable anchor force: adm FA =

1 ⋅ βS ⋅ AS = 140 kN η

(3.1)

This value is not exceeded in the analysis (see Fig. 3.15). With the computation of the anchor forces it must be taken into account that the placement of the anchors is simulated simultaneously with the excavation and the installation of the shotcrete membrane (see Fig. 3.11). Thus the anchor loads are overestimated in the analyses, since in reality the anchors are placed after the installation of the shotcrete membrane. To neutralize this effect, a Young's modulus smaller than the one of steel was assumed for the anchors in the analyses.

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Fig. 3.16:

Analysis cross-section AC 3 (structural model B), fault zone to the left of the tunnel, principal normal stresses and exceeding of strength, bench excavation (5th computation step)

Due to the uncertainties involved with this approach regarding the computed anchor forces and stress resultants of the shotcrete membrane, a comparative analysis was carried out, in which the shear resistance of the anchors was converted into an equivalent cohesion on the discontinuities. This cohesion was assumed within the anchored area (see Chapter 2.3.2). The results of this analysis confirmed the reinforcement determined before and thus indirectly also the computed anchor forces.

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- 88 Fig. 3.16 to 3.19 show the results of an analysis based on the parameters given in Fig. 3.9 and 3.10, structural model B and in addition a 5 m thick fault zone dipping at 70° close to the left side of the tunnel (see Fig. 3.8, position L). For these unfavorable assumptions the stability of the construction stage after the bench excavation (5th computation step, see Fig. 3.11) can only be proved in the analysis, if in the area of the left half of the tunnel 10 to 15 m long IBO-bolts (see Chapter 2.3.1) reaching behind the fault zone into the undisturbed rock mass are simulated by truss elements. The principal normal stresses, the exceeding of strength and the displacements computed for this case for the 5th computation step are shown in Fig. 3.16 and 3.17. If the long anchors are simulated, the displacements of the excavation profile are of the same magnitude as for the corresponding case without a fault zone (see Fig. 3.13 and 3.17).

Fig. 3.17:

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Analysis cross-section AC 3 (structural model B), fault zone to the left of the tunnel, displacements, bench excavation (5th – 1st computation step) WBI GmbH, Henricistr. 50, 52072 Aachen, Germany

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- 89 -

Fig. 3.18:

Analysis cross-section AC 3 (structural model B), fault zone to the left of the tunnel, stress resultants of the shotcrete membrane, bench excavation (5th computation step)

Fig. 3.18 shows the bending moments and normal thrust in the shotcrete membrane determined for the 5th computation step. Compared to the corresponding case without a fault zone, greater maximum bending moments and smaller maximum normal thrusts occur in the shotcrete (see Fig. 3.14 and 3.18). For a safety factor of 1.35, the design results in a statically required reinforcement of up to 6.2 cm2/m. If the lattice girders are taken into account, this reinforcement is covered by the planned reinforcement as well.

Fig. 3.19:

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Analysis cross-section AC 3 (structural model B), fault zone to the left of the tunnel, tensile anchor forces, 7th computation step WBI GmbH, Henricistr. 50, 52072 Aachen, Germany

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- 90 In Fig. 3.19 the tensile anchor forces computed for the truss elements in the 7th computation step are given. Only the anchors running through the fault zone are loaded. The maximum computed anchor force amounts to 150 kN, which is only slightly more than the admissible anchor load according to (3.1) of the planned IBO-rods (∅ 25 mm) of 140 kN. 3.1.6

Crown heading and monitoring results

Geotechnical mappings of the crown face were carried out in regular intervals during the heading. Special emphasis was placed on recording the properties, extent and orientation of the discontinuities (Fig. 3.20). No fault zones were encountered in the tunnel cross-section. Further, in the area of tunnel cross-section TC 3 ca. 20 measuring cross-sections with gage bolts were installed to monitor the displacements caused by the heading. In tunnel cross-sections TC 1 and TC 2 further measuring cross-sections were installed. As already mentioned, the excavation classes were determined on the basis of the results of the stability analyses as well as the mappings and the displacement measurements. Fig. 3.21 shows exemplarily the maximum roof subsidence measured during crown heading in the area of tunnel cross-sections TC 3 and TC 3A. With the exception of one measurement in the portal area (δR = 16 mm), the measured values range from 1 to 7 mm. In the area of the three analysis cross-sections AC 1, AC 2 and AC 3, the measured roof subsidence can be compared with the results of the stability analyses. With this comparison it must be taken into account, however, that the displacements that occurred in the measuring cross-sections already before the zero reading cannot be measured. Since the elastic part of the displacements mostly occurred before the zero reading, the measured roof subsidence δR can approximately be compared with the viscoplastic part of the displacements computed for the roof δ Rvp . It becomes apparent that in the area of analysis cross-sections AC 2 and AC 3 the measured displacements are smaller than the viscoplastic parts of the displacements computed with no cohesion assumed on the foliationparallel discontinuities (cf = 0, see Fig. 3.9).

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Fig. 3.20: WBI-PRINT 5

Crown face, chainage 216.1 m, excavation class 5: a) Photograph; b) mapping WBI GmbH, Henricistr. 50, 52072 Aachen, Germany

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Fig. 3.21:

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Crown heading, comparison of measured and computed roof subsidence, tunnel cross-sections TC 3 and TC 3A WBI GmbH, Henricistr. 50, 52072 Aachen, Germany

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- 93 If a value of cf = 100 kN/m2 is chosen for the cohesion on the foliation parallel discontinuities, the measured roof subsidence corresponds to the computed viscoplastic part of the displacements. For analysis cross-section AC 1, cf = 0 yields good agreement of the measured roof subsidence with the computed viscoplastic part of the displacements (Fig. 3.21). The results of the geotechnical mapping of the tunnel face and the displacements measured during the crown heading allowed in consideration of the results of the stability analyses the tunnel to be driven with excavation classes 4, 5 and 6A (see Fig. 3.6). Excavation class 7A was not carried out. Further, it was possible to excavate the crown over the entire tunnel length without simultaneously trailing bench. 3.1.7

Conclusions

The Glockenberg Tunnel is located in an alternating sequence of slates and sandstones characterized by a tectonic deformation due to a "special folding". The strength and deformability of the rock mass is predominantly determined by the discontinuities (bedding parallel discontinuities, foliation parallel discontinuities and joints). As a consequence of the folding structure, the orientation of the discontinuities is not uniform. In addition, the strike directions of the discontinuities with respect to the tunnel axis vary due to the almost circular tunnel alignment. Fault zones had to be expected locally as well. Therefore, different structural models had to be assumed for the stability analyses. The comparison of the displacements measured during tunneling with the values computed for the different cases allowed to optimize the support measures and to apply economical excavation classes. It became apparent that the analyses represented an essential contribution to an economical construction of this tunnel.

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- 94 3.2

Gäubahn Tunnel in Stuttgart, Germany

3.2.1

Introduction

The Gäubahn Tunnel is part of the new construction of the federal highway B 14 between the Schattenring intersection and the Südheimer Square providing a northern bypass of the city district of Vaihingen in Stuttgart, Germany. It undercrosses a railway line (Gäubahn) and buildings adjacent to the Rudolf-Sophien institution. During the heading underneath the railway track railway traffic had to be maintained. Specific tunnel support measures were therefore planned for this section. 3.2.2

Structure

The Gäubahn Tunnel consists of two parallel ca. 300 m long tunnel tubes with two lanes each (Fig. 3.22). Approx. 270 m of each tunnel tube were driven by underground construction ascending from the eastern portal. The precuts in the portal areas were constructed by the cut-and-cover method (Fig. 3.22 and 3.23). The maximum overburden amounts to approx. 20 m. In the area of the undercrossing of the Gäubahn an overburden of approx. 4 m exists (Fig. 3.23). Fig. 3.24 shows the 11.5 m wide and 8.85 m high standard profile of a tunnel tube. The excavated cross-section amounts to approx. 93 m2. The shotcrete membrane has a concrete grade of B25 and a thickness of 25 to 30 cm. The interior lining consists of watertight grade B35 concrete with a thickness of 45 cm. The roof and the invert are shallowly rounded with radii of curvature of R = 8.15 m and R = 10.45 m, respectively. At the sidewalls the radius of curvature amounts to R = 5.45 m. The transitions from the sidewalls to the roof and to the invert, respectively, were designed with comparatively small radii of R = 3.06 m and R = 2.65 m, respectively (Fig. 3.24). The two tunnel tubes were constructed by advancing crown heading with trailing bench and invert excavation. The cross-section is correspondingly divided into crown and bench/invert (Fig. 3.24).

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Fig. 3.22: WBI-PRINT 5

Gäubahn Tunnel, site plan WBI GmbH, Henricistr. 50, 52072 Aachen, Germany

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Fig. 3.23:

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Gäubahn Tunnel, longitudinal section through the southern tube with ground profile and analysis cross-sections WBI GmbH, Henricistr. 50, 52072 Aachen, Germany

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- 97 -

Fig. 3.24:

Gäubahn Tunnel, standard profile

Fig. 3.25 is an aerial photograph of the eastern precut with the buildings of the Rudolf-Sophien institution after the excavation and support of both tunnel tubes. The picture shows the open-cut arch sections (cut-and-cover construction) in the transition from underground excavation to the portal blocks not yet built in the represented construction phase.

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Fig. 3.25:

3.2.3

Gäubahn Tunnel, eastern precut and Rudolf-Sophien institution

Ground and groundwater conditions

Below the ground surface a few meters thick overlying strata are present (Fig. 3.23). They consist mostly of rock weathered into sand and silt, respectively, and partially relocated. The rock mass underneath belongs stratigraphically to the Stubensandstone formation consisting of a sequence of sandstones and siltstones. The sandstone and siltstone layers are up to several meters thick. Locally, however, also narrowly spaced alternating sequences occur (Fig. 3.23). The sandstones as well as the siltstones are divided into banks by bedding planes. The thickness of these banks ranges between few centimeters and 1 to 2 m. The bedding planes are approximately horizontal. The joints are generally steeply dipping and deviate only little from the direction perpendicular to the bedding (here: the vertical direction). In general they extend over many meters horizon-

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- 99 tally. At least in some areas the joints are slightly opened according to the exploration results. In the siltstone layers slickensides were found as well in addition to the bedding planes and joints. These slickensides are randomly oriented and have mostly smooth surfaces. In the middle tunnel section two boreholes were equipped as observation wells. According to their readings, the groundwater table was located at the level of the two tunnel cross-sections (Fig. 3.23). In two observation wells located close to the tunnel portals no groundwater was encountered. 3.2.4

Excavation classes

The excavation of the two tunnel tubes was planned mostly as an advancing crown heading with open invert and trailing bench and invert excavation following excavation class 4A. The excavation sequence, excavation method, round lengths and support measures for excavation classes 4A-1, 4A-2 and 4A-3 (see DGGT, 1995: Table 1) are given in Fig. 3.26. These excavation classes differ with regard to in the unsupported round length, the lattice girder spacing and the number of anchors per round. The shotcrete membrane is reinforced inside and outside with steel fabric mats Q295. If necessary, the tunnel face was planned to be sealed with plain shotcrete. The trailing distance of bench (D) and invert (E) was to be determined depending on the monitoring results and the geotechnical conditions encountered during the heading as well as depending on the results of the stability analyses. The tunnel was excavated using a tunnel excavator and, in some areas, also by smooth blasting. Measurements carried out on buildings showed that the vibration velocities remained far under the reference values for the admissible structural vibrations given in DIN 4150, part 3 (DIN 4150-3, 1999).

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- 100 -

Fig. 3.26:

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Standard heading, excavation classes 4A-1, 4A-2 and 4A-3

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Fig. 3.27:

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Undercrossing of the Gäubahn, excavation classes 6A-K and 7A-K: a) Cross-section; b) longitudinal section (detail) WBI GmbH, Henricistr. 50, 52072 Aachen, Germany

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- 102 In the area of the undercrossing of the Gäubahn a crown heading with closed temporary invert with the excavation classes 6A-K and 7A-K, respectively, was planned (see DGGT, 1995: Table 1) in order to limit the tunneling-induced subsidence (Fig. 3.27). These excavation classes include an advancing support using so-called composite pile umbrellas. These consist of 8 m long piles with a diameter of 40 mm which were grouted in 116 mm boreholes using cement based suspension and are slightly inclined against the horizontal in the longitudinal section (Fig. 3.23 and 3.27). In each round one layer of composite piles (composite pile umbrella) was constructed. If required, spiles were planned to be installed in addition. In excavation class 7A-K, the tunnel face was in addition to be rounded out and supported with shotcrete (Fig. 3.27b). 3.2.5

Stability analyses for the design of the shotcrete support

For the design of the shotcrete membrane, two-dimensional analyses with the program system FEST03 (Wittke, 2000) were carried out. The investigation comprised the analysis cross-sections AC 1 to AC 4 shown in Fig. 3.23. Fig. 3.28 shows exemplarily the FE-mesh, the boundary conditions, the ground profile and the parameters used for analysis cross-section AC 2. Analysis cross-section AC 2 is located in the area of the undercrossing of the buildings adjacent to the Rudolf-Sophien institution. The overburden in this area amounts to approx. 20 m (see Fig. 3.23). The 52 m wide, 66 m high and 1 m thick computation section in the form of a slice is discretized by an FE-mesh with 2703 isoparametric elements and 13956 nodes. To simplify matters, only one tunnel tube is modeled. The plane of symmetry lies in the middle of the rock pillar between the two tunnel tubes. Thus the simultaneous excavation of both tunnel tubes is simulated in the analyses. In reality, the crown excavation of one tunnel tube precedes the other one. The simulation of a simultaneous heading of both tubes however is on the safe side with regard to the loading of the rock mass and the stability of the openings. For the nodes of the vertical boundary planes vertically movable supports are introduced as boundary conditions, whereas for the nodes on the lower boundary plane (z = 0) horizontally movable supports are specified (Fig. 3.28). All nodes are fixed in the y-direction.

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Fig. 3.28:

Analysis cross-section AC 2, FE-mesh, boundary conditions, ground profile and parameters

Four sandstone and siltstone horizons each of the alternating sequence of the Stubensandstone formation were modeled. The positioning of the siltstone horizon in the roof area is to be assessed unfavorable with respect to the stability of the tunnel, WBI-PRINT 5

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- 104 since the strength of the discontinuities is considered smaller in the siltstone than in the sandstone (see Fig. 3.28). At middle tunnel level a sandstone layer is modeled which is followed by a siltstone layer in the area of the tunnel invert. The groundwater table was assumed at tunnel invert level due to the drainage during construction. Below the groundwater table the rock mass is subjected to hydrostatic uplift. In the tender documents ranges for the soil and rock mechanical parameters are given. The stability analyses for the design of the shotcrete support are based on the most unfavorable values (Fig. 3.28).

Fig. 3.29:

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Analysis cross-section AC 2, FE-mesh, detail

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- 105 The intact rock strength of the sand- and siltstones was assumed infinitely high as a simplification. The rock matrix was assigned elastic stress-deformation behavior. This simplification is justified, because the strength on the discontinuities is markedly smaller than the intact rock strength. Two discontinuity sets were taken into account in the analyses. The horizontal bedding and a vertical joint set striking parallel to the tunnel axis with an unfavorable effect on tunnel stability. At and underneath the tunnel invert, the rock mass is unloaded during the tunnel excavation. Previous experience has shown that the modulus relevant for the unloading of the rock mass is higher than the loading modulus assumed as E = 300 MN/m2. Correspondingly, the elements in and underneath the invert area were assigned an unloading modulus of EU = 1000 MN/m2 (see Fig. 3.29). The statically effective Young's modulus of the shotcrete was specified as 15000 MN/m2 taking into account the hardening during the application of the load (Fig. 3.29). The excavation and support of the tunnel were simulated in five computation steps (Fig. 3.30). The first computation step comprises the determination of the state of stress and deformation resulting from the dead weight of the ground (in-situ state). With computation steps 2 and 3 the crown excavation and its support using shotcrete are simulated. To this end, Young's modulus of the elements in the crown was reduced in the 2nd computation step to the value Ered = av · E with av = 0.5. av is the so-called stress relief factor already mentioned in Chapter 3.1 (Wittke, 2000). The excavation and support of the crown follow in the 3rd computation step. In the 4th and 5th computation steps the excavation and support of the remaining cross-section are simulated correspondingly. Fig. 3.31 shows the displacements due to the crown excavation (3rd – 1st computation step) calculated for the excavation contour and the ground surface. The roof subsidence amounts to approx. 18 mm, the invert heave is approx. 8 mm and the maximum subsidence at the ground surface results to approx. 11 mm.

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Fig. 3.30:

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Analysis cross-section AC 2, computation steps

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Fig. 3.31:

Analysis cross-section AC 2, displacements, crown excavation (3rd – 1st computation step)

The computed bending moments and normal thrust in the shotcrete membrane are shown for the 3rd computation step in Fig. 3.32. The design of the shotcrete support for this construction stage yields that no reinforcement is required with respect to bending and normal thrust for a factor of safety of 1.35. Steel fabric mats Q295 were placed inside and outside as a minimum reinforcement (see Fig. 3.26).

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- 108 After the bench and invert excavation in the 5th computation step the displacements increase only slightly. For this construction stage as well, no reinforcement is statically required for a factor of safety of 1.35.

Fig. 3.32:

3.2.6

Analysis cross-section AC 2, stress resultants in the shotcrete membrane, crown heading (3rd computation step)

Crown heading and monitoring results

To monitor the stability and to verify the results of the stability analyses the heading was accompanied by a geotechnical monitoring program. The following measurements were carried out: -

Leveling and trigonometric measurements on the ground surface,

-

leveling on structures,

-

leveling on sleepers of the Gäubahn,

-

trigonometric measurements on overhead wire poles,

-

combined extensometer and inclinometer measurements in the area of the undercrossing of the Gäubahn and at the western portal,

-

leveling and convergency measurements in both tunnel tubes.

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Fig. 3.33:

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Subsidence of ground surface and structures, measured during crown heading WBI GmbH, Henricistr. 50, 52072 Aachen, Germany

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- 110 In both tunnel tubes the crown heading was far ahead of the bench and invert excavation. Fig. 3.33 shows exemplarily the ground surface subsidence due to the crown heading in the area of the undercrossing of the Gäubahn and the buildings adjacent to the RudolfSophien institution. In the area of the undercrossing of the buildings adjacent to the Rudolf-Sophien institution the crown heading was carried out with an open invert and small round lengths of 0.8 to 1.2 m following excavation class 4A-1 (see Fig. 3.26). The subsidence measured here ranges from 7 to 19 mm. In the area of the undercrossing of the Gäubahn the heading was changed over to a crown heading with closed invert under the protection of an advancing support using composite piles and spiles (see Fig. 3.27). The subsidence could thus be limited to 2 to 8 mm. This small subsidence did at no point affect the railway traffic. The measured ground surface subsidences are in good agreement with the ones determined in the stability analyses. In the area of the undercrossing of the buildings adjacent to the Rudolf-Sophien institution (analysis cross-section AC 2) a maximum surface subsidence of 11 mm was computed (see Fig. 3.31). The maximum surface subsidence computed for the undercrossing of the Gäubahn (analysis cross-section AC 3, see Fig. 3.23) amounts to 7 mm. 3.2.7

Conclusions

With a crown heading with open invert and small round lengths (excavation class 4A-1, see Fig. 3.26) it was possible to achieve small surface subsidence < 2 cm during tunneling in the alternating sequence of sandstone and siltstone horizons of the Stubensandstone formation in the region of Stuttgart. In order to avoid affecting the railway traffic during the undercrossing of the Gäubahn it was necessary in this area to limit the subsidence to even smaller values. This was achieved mainly by supporting the curved temporary crown invert with shotcrete. For safety reasons an additional advancing support with composite piles was carried out. With a very small surface subsidence ranging from 2 to 8 mm it was possible to maintain the railway traffic without any interference. In areas where the rock mass could not be excavated by a tunnel excavator smooth blasting was carried out. It was possible to prove by measurements that the vibration velocities fell well beWBI-PRINT 5

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- 111 low the reference values for the allowable structural vibrations given in DIN 4150, Part 3 (DIN 4150-3, 1999). The results of the geotechnical monitoring during construction are in good agreement with the surface subsidence computed in the stability analyses. It could thus be shown that FE-analyses are a good basis for estimating the subsidence due to tunneling, if the parameters are assessed appropriate and the excavation and the installation of the support are simulated realistically. 3.3

Hellenberg Tunnel, Germany

3.3.1

Introduction

The alignment of the new railway line Cologne-Rhine/Main runs through the Rhine schist mountains (Rheinisches Schiefergebirge) in NW-SE direction tightly bundled with the freeway (Autobahn) A3. Coming from Cologne, it crosses in particular the Siebengebirge, Westerwald and Taunus mountains. The alignment comprises more than 30 tunnels with a total length of approx. 47 km. The Hellenberg Tunnel is one of six tunnels driven in the Taunus mountains and lies to the south of Idstein. 3.3.2

Structure

The new railway line Cologne-Rhine/Main of German Rail (Deutsche Bahn AG) undercrosses the Hellenberg mountain in a 552 m long tunnel. The maximum overburden of the tunnel amounts to approx. 19 m. The tunnel sections in the portal areas were constructed by the cut-and-cover method. The central tunnel segment with a length of approx. 470 m was driven by underground construction ascending from southeast to northwest (Fig. 3.34). A mouth-shaped profile with an excavated width of 15.4 m and a height of 12.3 m was constructed for the double-tracked Hellenberg Tunnel (Fig. 3.35). This is the standard profile for doubleTracked tunnels of the new railway line Cologne-Rhine/Main. The excavated cross-section amounts to approx. 150 m2 (Fig. 3.35).

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Fig. 3.34: WBI-PRINT 5

Hellenberg Tunnel, site plan WBI GmbH, Henricistr. 50, 52072 Aachen, Germany

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- 113 -

Fig. 3.35:

New railway line Cologne-Rhine/Main, standard profile

In the vault area the radius of curvature of the shotcrete membrane amounts to R = 7.32 m. The invert is rounded with a radius of R = 16.5 m. For the transition area from the sidewalls to the

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- 114 invert radii of R = 4.38 m and R = 2.5 m, respectively, were selected (Fig. 3.35). Fig. 3.36 shows the southern portal at the start of underground excavation of the Hellenberg Tunnel.

Fig. 3.36: 3.3.3

Hellenberg Tunnel, southern portal

Ground and groundwater conditions

The ground along the alignment of the Hellenberg Tunnel was explored by test pits and core drillings. The boreholes were equipped as observation wells (Fig. 3.34 and 3.37). The Quaternary talus material, which is loamified in the upper region, extends to a depth of approx. 2 m. In the portal areas it is encountered down to a maximum depth of 5 m. Below the Quaternary cover Devonian rocks follow composed in the area of the Hellenberg Tunnel of phyllitic slate with sandstone and quartzite intercalations and embedded conglomerate lenses. These layers are referred to as Variegated Schist.

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Fig. 3.37: WBI-PRINT 5

Hellenberg Tunnel, longitudinal section with predicted excavation classes WBI GmbH, Henricistr. 50, 52072 Aachen, Germany

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- 116 The rock mass in the area of the Hellenberg Tunnel is characterized by a deep-reaching weathering extending to a depth of some 35 m. Unweathered rock mass is thus only encountered below the tunnel's invert. In the course of the exploration the drill cores and the test pits were geotechnically mapped. Further, television probing and laboratory and in-situ tests were carried out. From the northern tunnel portal the groundwater table rises to 12 m above the tunnel roof. In the following, it drops towards the southern tunnel portal to below the tunnel's invert. The groundwater table mapped in the tunnel's longitudinal section (Fig. 3.37) is based on the highest groundwater levels measured in the boreholes equipped as observation wells. 3.3.4

Excavation classes

Because of the size of the excavation profile it was necessary to subdivide excavation and support into crown, bench and invert (Fig. 3.35). An advancing crown excavation was carried out. A prediction of the distribution of the excavation classes based on the results of the exploration and of stability analyses and on experience is shown in Fig. 3.37. According to the drilling results, the strongly weathered rock extends in the portal areas to about the tunnel's invert. Accordingly, excavation classed with a temporary crown invert support were planned here (6A-K1 and 7A-K1, see DGGT, 1995: Table 1). In the central tunnel section it was expected that the crown heading can be carried out with an open invert. In parts excavation classes without advance support measures (4A-2 and 4A-3, see DGGT, 1995: Table 1) were predicted. For other sections it was assumed that the heading requires advance support (excavation classes 6A-1 and 6A-2, see DGGT, 1995: Table 1). The excavation methods, round lengths and support measures planned for excavation classes 6A-1 and 6A-2 are shown in Fig. 3.38 and 3.39. It was planned to carry out the excavation mainly by tunnel excavators and hydraulic chisels. In areas in which the tunnel was located in slightly weathered and slightly jointed rock, local loosening blasting was planned. WBI-PRINT 5

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Fig. 3.38:

Excavation classes 6A-1 and 6A-2

The unsupported round lengths during the crown excavation were specified in excavation class 6A-1 as 0.80 to 1.20 m and in excavation class 6A-2 as 1.21 to 1.60 m (Fig. 3.38 and 3.39). In both excavation classes an advance support of the crown with 3 to 4 m long mortar spiles was provided for (Fig. 3.38 and 3.39). Shotcrete of grade B25 was selected for the shotcrete membrane at a thickness of 25 cm and reinforced inside and outside by steel fabric mats Q221 (Fig. 3.38 and 3.39). In the sidewall area a systematic anchoring with at least twelve 4 m long SN-anchors per tunnel meter was planned (Fig. 3.38 and 3.39). WBI-PRINT 5

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Fig. 3.39:

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Support for excavation classes 6A-1 and 6A-2: a) Cross-section; b) longitudinal section WBI GmbH, Henricistr. 50, 52072 Aachen, Germany

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- 119 A possible installation of anchors in the roof area ought to be decided on site. For each round one steel set (lattice girder) was to be installed in the crown area (Fig. 3.38 and 3.39). The trailing distance of the bench (section D in Fig. 3.38) and the closing of the invert (section E in Fig. 3.38) was to be specified depending on the results of monitoring during heading, the encountered geotechnical conditions and the results of the stability analyses. The unsupported round lengths were specified in excavation class 6A-1 for the bench and the invert as 1.60 to 2.40 m and ≤ 3.60 m, respectively. In excavation class 6A-2 the corresponding round lengths amounted to 2.41 to 3.20 m and ≤ 4.80 m, respectively (Fig. 3.38). To improve the stability, the shotcrete membrane was to be widened to t = 60 cm in the area of the crown and bench support feet if necessary. It is shown calculational in Wittke et al. (1986), Wittke (1990) and Wittke (1998), however, that the stability of a crown heading with open invert can only slightly be improved by these kind of measures. 3.3.5

Crown heading

The excavation classes were finally specified during the crown heading on the basis of the results of stability analyses, of crown face mappings (see Chapter 3.3.6) and of the monitoring results. The crown was excavated over the entire tunnel length without a trailing bench excavation (unlimited length of section D in Fig. 3.38). In the area of the southern portal a crown heading with a temporary support of the invert was carried out over a length of approx. 80 m (excavation class 7A-K1, Fig. 3.40). At the northern portal a temporary support of the invert was only constructed in the beginning of excavation (excavation class 6A-K1, Fig. 3.40). In the remaining area a crown heading with open invert was carried out (excavation classes 6A-1 and 6A-2, Fig. 3.40).

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Fig. 3.40: WBI-PRINT 5

Hellenberg Tunnel, longitudinal section and excavation classes, as carried out WBI GmbH, Henricistr. 50, 52072 Aachen, Germany

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- 121 3.3.6

Results of the crown face mapping

The evaluation of the crown face mapping resulted in a modified elevation of the boundary between strongly and slightly weathered rock as compared to the exploration results (Fig. 3.37 and 3.40). Nevertheless, the predicted distribution of excavation classes agreed well with the construction (Fig. 3.37 and 3.40). During the tunnel face mapping the appearance, the extent and the orientation of the discontinuities were determined as well. Fig. 3.41 shows as an example the crown face mapping at chainage 345.4 m. The strike and dip angles of the discontinuities were measured using a geological compass (see Chapter 2.5.1). For reasons of clarity of the representation only some of the measured discontinuity orientations are mapped in Fig. 3.41.

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Crown face mapping, chainage 345.4 m WBI GmbH, Henricistr. 50, 52072 Aachen, Germany

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- 122 In Fig. 3.42 all discontinuity orientations measured from chainage 300 to 350 m are shown in a polar diagram (see Chapter 2.5.1). Accordingly the discontinuities can mainly be assigned to three sets.

Fig. 3.42:

Measured discontinuity orientations, chainage 300 to 350 m, polar diagram

The bedding planes and foliation discontinuities display the NE-SW strike typical for the Rhine schist mountains and dip at a moderate steep to steep angle (50 to 80°) towards northwestern directions. Two joint sets J1 and J2 further exist dipping mostly steeply (60 to 90°) and striking parallel to the tunnel axis (Fig. 3.42). Fig. 3.43 shows the bench at the southern portal. The photograph gives an impression of the strongly weathered rock in this area and of the discontinuity fabric.

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Fig. 3.43: 3.3.7

Bench at the southern portal

Stability analyses for the bench excavation

Within the tunnel sections driven according to excavation classes 6A-1 and 6A-2 an advancing crown excavation with open invert and trailing bench and invert excavation was planned (Fig. 3.38). The evaluation of the mapping carried out during the crown heading revealed, however, that the steep joints J1 and J2 striking parallel to acute-angled to the tunnel axis (see Fig. 3.42) appear frequently in some areas and have a large extent. Before the start of the bench excavation the stability of the tunnel in this construction stage was therefore investigated in FE-analyses with the program system FEST03 (Wittke, 2000). On the basis of the results of these analyses the length of the section E between bench and invert excavation (see Fig. 3.38) was to be specified. Fig. 3.44 shows the computation section, the FE-mesh, the ground profile and the parameters taken as a basis for the stability analyses (Wittke et al., 1999). The tunnel cross-section is located in the strongly to slightly weathered Variegated Schist. The overburden is 18 m high. Below the tunnel's invert the rock is unweathered. WBI-PRINT 5

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Fig. 3.44:

Computation section, FE-mesh, ground profile and parameters

As mentioned above, the foliation or bedding, respectively, strikes approx. perpendicularly to the tunnel axis and dips steeply (see Fig. 3.42). Since discontinuities of this orientation hardly influence the load transfer in transverse tunnel direction, the foliation/bedding was not taken into account in the analyses. The shear strength of the joints J1 and J2, which are relevant for the stability, was, however, accounted for. The shear strength was furthermore varied because of the different appearance of these discontinuities. Three cases were investigated, in which the friction angles on the joints ϕJ were assumed as 20°, 22.5° and 25° and the cohesion as cJ = 0. The anchoring of the rock provided for in all excavation classes was not taken into account in the analyses as a conservative assumption. The heading of the tunnel was simulated in four computation steps (Fig. 3.45). In the 1st computation step the in-situ state of stress and deformation resulting from the dead weight of the rock mass was computed. After that, the excavation of the crown (2nd WBI-PRINT 5

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- 125 computation step) and the bench (3rd computation step) was simulated, each time with simultaneous installation of the shotcrete membrane. In the 4th computation step the invert was excavated and supported using shotcrete.

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Computation steps WBI GmbH, Henricistr. 50, 52072 Aachen, Germany

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- 126 A thickness of t = 25 cm (see Fig. 3.35) and a Young's modulus of E = 15,000 MN/m2 were specified for the shotcrete membrane. Fig. 3.46 shows the development of the displacements computed for the bench support feet in the course of the viscoplastic iterative analysis in the 3rd computation step.

Fig. 3.46:

Viscoplastic displacements depending on the shear strength of the joint sets J1 and J2, 3rd computation step

For the case ϕJ = 25° the stability of the construction stage due to the bench excavation can be proven in the analysis. The horizontal and vertical displacements computed at the bench support feet converge in the course of the viscoplastic iterative analysis (see Fig. 3.46). The horizontal and vertical components of the viscoplastic displacements of the bench support feet amount to δ Hvp = 15.6 mm and δ Vvp = 9.2 mm. With decreasing friction angle ϕJ markedly larger viscoplastic displacements are computed (see Fig. 3.46). In the case ϕJ = 20° the displacements do not converge in the analysis (see Fig. 3.46). The

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- 127 stability of the construction stage as a result of the bench excavation can thus not be proven by the analysis for this case. Summarizing it may be stated that the stability of the tunnel during bench excavation depends to a large degree on the friction angle ϕJ on the joints. On the basis of the results of the stability analyses it was agreed upon for those sections in which the crown was excavated without invert support to set no limit for the length of the section E between bench excavation and the support of the invert (see Fig. 3.38). With the bench mapping during heading, however, great importance was attached to recording the appearance and extent of the joints. Whenever strongly jointed rock was encountered during bench excavation, the anchoring of the sidewalls was intensified. 3.3.8

Construction and monitoring results

Fig. 3.47 shows the sequence of the heading. Following the crown heading (1), the bench was excavated up to chainage approx. 80 m at a round length of 2.0 m. The invert trailed with a round length of 3.6 m and was supported after each round (2). In this area the strongly weathered rock mass extends into the tunnel cross-section (Fig. 3.47), and a crown invert support was installed. After that, the bench was excavated from chainage 80 m to 462 m with a round length of 2.4 m (3 in Fig. 3.47). As mentioned, the anchoring was locally intensified because of the heavy jointing of the rock mass. In the northern portal area the bench was excavated again with immediately trailing invert over a short tunnel section (4 in Fig. 3.47). The round lengths were the same as in the southern portal area. Finally, the invert was excavated backward and supported with round lengths of 3.6 m (5 in Fig. 3.47). The heading was accompanied by a geotechnical monitoring program including surface leveling, extensometer and inclinometer measurements from the ground surface and leveling and convergency measurements in the tunnel.

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Fig. 3.47: WBI-PRINT 5

Excavation, heading sequence WBI GmbH, Henricistr. 50, 52072 Aachen, Germany

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Fig. 3.48: WBI-PRINT 5

Measured vertical displacements at the bench support feet WBI GmbH, Henricistr. 50, 52072 Aachen, Germany

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- 130 Fig. 3.48 shows exemplarily the maximum subsidence at the bench support feet measured after bench excavation. Since the elastic part of the displacements had mostly occurred already before the zero reading, it could not be recorded by the measurements. The measurement results were therefore compared to the computed viscoplastic displacement parts shown in Fig. 3.46. It becomes apparent that the measured displacements are smaller than the viscoplastic vertical displacements computed assuming a friction angle of ϕJ = 25° (approx. 10 mm). It thus turns out that the construction stage following the bench excavation was stable as predicted. 3.3.9

Conclusions

Crown headings are carried out in excavation classes with and without invert support, depending on the rock conditions. Since the different excavation classes vary strongly in cost, great importance is attached to the appropriate specification. The example of the Hellenberg Tunnel shows how the excavation classes can be specified safely on the basis of the results of stability analyses, mapping during construction and monitoring. It becomes apparent that proofs of stability according to the FE-method can contribute essentially towards the prediction of and decision on the excavation classes.

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- 131 4.

Crown heading with closed invert

4.1

Österfeld Tunnel in Stuttgart, Germany

4.1.1

Introduction

The Österfeld Tunnel undercrosses residential areas as well as the railway tracks of the Gäubahn and the S-Bahn (urban railway) partly with low overburden. During the heading of the tunnel underneath the railway lines the railway traffic had to be maintained. In this area, therefore, special measures for the support of the tunnel were required. Special attention was required, because the road tunnel is located in mudstone layers of the Lias α, which are subjected to high horizontal stresses. 4.1.2

Structure

The Österfeld Tunnel, which has been excavated by means of the NATM is approx. 400 m long and part of the eastern by-pass around Vaihingen, a suburb of the city of Stuttgart. The new road connection with a total length of 1.9 km is joining the highway B 14 at the Vaihinger triangle (Fig. 4.1). The Österfeld Tunnel undercrosses the Paradiesstraße and the railway tracks of the Gäubahn and the S-Bahn (Fig. 4.2). Furthermore the new road crosses the Nesenbachtal by means of a 170 m long bridge. Following the southern end of this bridge, the road is running through the 780 m long Hengstäcker Tunnel. The subsequent road section is joining the Nord-Süd-Straße (Fig. 4.1). The new road connection was opened for traffic in September 1999. The northern portal of the Österfeld Tunnel is located at the end of the roadway Unterer Grund. Up to the Don-Carlos-Brücke the tunnel runs parallel to the tracks of the Gäubahn and the S-Bahn (Fig. 4.2). Along the first 100 m of this section at the base of the adjacent railway trench a 9 m high angular retaining wall is located. Moreover in the area of the tunnel some residential buildings are located. One of these houses is directly undercrossed by the tunnel. Behind the Don-Carlos-Brücke the tunnel undercrosses the four tracks of the above mentioned railway lines at an acute-angle. The tunnel ends at the northern flank of the Nesenbachtal (Fig. 4.2).

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Fig. 4.1:

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Map of the eastern by-pass in Stuttgart-Vaihingen

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Fig. 4.2:

Plan and longitudinal section of the Österfeld Tunnel

The overburden of the tunnel varies between 6 m and 14 m. The lowest overburden results at the undercrossing of the railway tracks (Fig. 4.2). WBI-PRINT 5

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Fig. 4.3:

Österfeld Tunnel, regular cross-section

For the tunnel a mouth-shaped cross-section was carried out with a total internal width of 11.2 m and a total height of 9.7 m. The excavated cross-section amounts to approx. 98 m2. The thickness of WBI-PRINT 5

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- 135 the shotcrete membrane is 25 to 30 cm and the thickness of the interior lining consisting of watertight reinforced concrete is 40 cm. For the traffic in each direction a traffic lane of 3.75 m width is available (Fig. 4.3). The radius of curvature at the area of the crown is R = 5.6 m. The invert (R = 10.4 m) as well as the temporary invert of the crown (R = 12.2 m) were carried out with larger radii. At the transitions from the crown to the temporary invert of the crown (R = 1 m) and from the sidewalls to the invert (R = 2.4 m) small radii were selected (Fig. 4.3). 4.1.3

Ground and groundwater conditions

During the design phase for the new by-pass road an extensive program for the exploration of the subsoil and groundwater conditions was carried out. This program includes core drillings, the investigation of soil, rock and water samples, water level observations, combined extensometer and inclinometer measurements as well as in-situ stress measurements. According to the results of this exploration program the Österfeld Tunnel is nearly completely located in the Psilonoten- and Angulatenlayers of the Lower Jurassic (Lias α1 and Lias α2, Fig. 4.4), consisting of mudstones with single limestone and lime-sandstone interbeds. The mostly solid mudstones are transversed by bedding parallel discontinuities with small spacings and are distinctly jointed. From the portal zones up to the central part of the tunnel the degree of weathering decreases and the strength of the intact rock as well as of the rock mass, respectively, increases. The generally very hard layers of limestone and lime-sandstone are characterized by two sets of vertical joints J1 and J2, which are oriented perpendicular to the bedding planes and enable the excavation with an excavator. The almost horizontal bedding B dips parallel to the tunnel axis towards the Nesenbachtal. The sets of discontinuities B, J1 und J2 are also present in the mudstones, however, with a considerable less extent and frequency as in the limestones and lime-sandstones. Concerning the rock mechanical parameters the mudstones, the limestones and the lime-sandstones are combined to one layer (Fig. 4.5). In the present case this is permissible, because the thicknesses of the limestones and the limesandstones are small in comparison to the thicknesses of the mud-

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- 136 stone layers as well as to the dimensions of the tunnel crosssection.

Fig. 4.4:

Stratigraphical and rock mechanical classification of the ground

Locally the invert of the tunnel intersects the mudstones of the Rät, which are located below the Lias α formation (Fig. 4.4 and 4.5).

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- 137 The transition to the unweathered Knollenmergel, which is located underneath the mudstones of the Rät, is formed by a disintegrated layer, the so-called reduction zone of the Knollenmergel (Fig. 4.4 and 4.5). In the reduction zone as well as in the unweathered Knollenmergel slickensides are present, which dip with 20° up to 40° and strike in all directions (Fig. 4.5).

Fig. 4.5:

Lias α, Rät and Knollenmergel, structural model (Wittke, 1990)

On top of the above described layers of the Lias α up to the ground overlying strata of reclaimed fill and weathered mudstone (clay) are located (Fig. 4.2). The thickness of these layers varies between 3 and 8 m. In the area of the railway trench the rock surface is situated only a few decimeters below the ground surface. The structural model for the Lias α, the Rät and the Knollenmergel, illustrated in Fig. 4.5, was developed within the scope of

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- 138 other projects in the city of Stuttgart with comparable ground conditions (Wittke, 1990). The soil mechanical parameters derived for the layers close to the surface are based on the results of laboratory investigations on samples taken from core drillings located in the area of the tunnel. Due to the grain-size distributions and moisture contents as well as data gained from experience a mean modulus of deformation of E = 25 MN/m2 as well as mean effective shear parameters of ϕ' = 20° and c' = 10 kN/m2 are assumed (Table 4.1). The elastic behaviour of the Lias α can be approx. described by a transversally isotropic stress-strain-law. Based on the volumetric distribution of the mudstones and the limestones and lime-sandstones the mean elastic constants were calculated according to Wittke (1990). In these calculations the mudstones were assumed to be transversally isotropic and the limestones and lime-sandstones were treated as isotropic elastic rocks. The elastic constants of the mudstones, limestones and lime-sandstones were derived from the results of dilatometer tests and laboratory tests on rock samples. Also the experiences gained from other projects in the area of Stuttgart were utilized for the estimation of these parameters. The general relationships for the elastic constants of alternating sequences consisting of transversally isotropic rocks are given in Salamon (1968). layer Fill and clay Mudstones with single layers of limestones and lime-sandstones Rät and reduction zone of the Knollenmergel

Knollenmergel

Table 4.1:

deformability strength E = 25 MN/m² ϕ' = 20°, c' = 10 kN/m² E1 = 1000 MN/m² Bedding B: E2 = 500 MN/m² ϕB = 20° cB = 0 Jointing J1, J2: ϕJ = 30°, cJ = 40 kN/m² E = 150 MN/m² Discontinuities in the Rät and slickensides in the reduction zone: ϕD = 17.5°, cD = 10 kN/m² E = 1000 MN/m² Slikensides: ϕS = 17.5°, cS = 10 kN/m²

Mean values of soil and rock mechanical parameters

The shear strength of the mudstones is dominated by the shear strength of the discontinuities, which is significantly smaller WBI-PRINT 5

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- 139 than the shear strength of the intact rock. For the shear strength of the bedding planes a friction angle of ϕB = 20° and a cohesion of cB = 0 are assumed. For the joints of sets J1 and J2 an angle of friction of ϕJ = 30° and a cohesion of cJ = 40 kN/m2 are estimated (Table 4.1). A tensile strength normal to the discontinuities is not taken into account. With regards to the rock mechanical properties the Rät and the reduction zone of the Knollenmergel are combined to a uniform layer. On the basis of experience for the modulus of deformation a value of E = 150 MN/m2 is assumed (Table 4.1). In laboratory tests on rock samples unconfined compressive strengths ranging from σu = 0.3 MN/m² to 5.0 MN/m2 were determined. Though these values are quite low, here also the strength on the discontinuities is decisive. The shear parameters of the discontinuities in the Rät and the slickensides at the reduction zone of the Knollenmergel, respectively, are estimated to be ϕD = 17.5° and cD = 10 kN/m2 (Table 4.1). A tensile strength normal to the discontinuities here also is not accounted for. The unconfined compressive strength of the unweathered Knollenmergel is somewhat higher than that of the reduction zone. Decisive for the strength of the unweathered Knollenmergel are however the slickensides with shear parameters of ϕs = 17.5° and cs = 10 kN/m2 (Table 4.1). The modulus of deformation of the unweathered Knollenmergel was not evaluated. From experience gained from other projects in the area of Stuttgart with comparable subsoil conditions (Wittke, 1990) for this parameter a value of E = 1000 MN/m2 is estimated (Table 4.1). According to the results of investigations, carried out in the past at different structures in the Lias α high horizontal in-situ stresses are to be expected (Grüter, 1988; Wittke, 1990; Wittke 1991). Thus in two exploratory boreholes in-situ stress measurements using the overcoring technique (Kiehl and Pahl, 1991) were carried out. These measurements resulted in horizontal in-situ stresses in an order of magnitude of ΔσH = 0.2 to 1.9 MN/m2 for the mudstone layers of the Lias α. These stresses have to be accounted for in addition to those resulting from the dead weight due to horizontally confined in-situ conditions. In the portal zones the ground-water level is located below the invert of the tunnel. In the remaining area the water table is loWBI-PRINT 5

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- 140 cated somewhat above the invert of the tunnel. In the central section of the tunnel it is levelled at the middle of the height of the cross-section of the tunnel (Fig. 4.2). 4.1.4

Fundamentals of the design

Because of the small overburden, the high horizontal stresses and the low shear strength of the bedding parallel discontinuities in the Lias α, in which the tunnel is located, special problems concerning the stability of the tunnel during construction arise. After the excavation of the tunnel the horizontal stresses have to be transmitted around the tunnel's cross-section. As a consequence stress concentrations at the roof and the invert of the tunnel occur, which may be considerably higher than in the horizontal stresses present in the undisturbed state of stress (in-situ state). If the tunnel remains unsupported over a greater section the mudstones would be highly stressed in horizontal direction. Because of the above mentioned comparable low strength shear failures on the bedding parallel discontinuities above the roof and beneath the invert of the tunnel would occur. Moreover a buckling of thin mudstone layers would be possible. To avoid failures and collapses under these difficult conditions a bolted shotcrete support has to be always installed immediately after excavation. Thus the stresses can be transmitted around the tunnel mainly through the shotcrete. To achieve a stable stage of construction the support must be adequately designed and must have an early bearing capacity. High requirements with regard to a high early strength of the shotcrete are to be fulfilled. Because of the large cross-section of the tunnel, which amounts to 98 m2 (Fig. 4.3), as well as for reasons of the construction process and for stability it was decided to subdivide the crosssection in crown, bench and invert. Because of the same reason and in order to minimize surface subsidence the crown was designed with a shotcrete membrane at the temporary invert. Also during crown excavation as well as bench and invert excavation, respectively, short round lengths and an early closing of the shotcrete support were foreseen. An alkali free shotcrete with a quick-setting spray cement was used according to the dry shotcrete mixture transport technique of the Rombold and Gfröhrer company , which is described in Balbach WBI-PRINT 5

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- 141 and Ernsperger (1996) (see Chapter 2.1.2). This shotcrete is characterized by a fast development of strength as well as a high ultimate strength. Although it is a shotcrete with a concrete grade of B25 corresponding to a C20/25 according to EUROCODE 2 (EC2), it was well known from experience gained from other projects that this shotcrete develops a considerably higher strength comparable to a concrete grade of B45 corresponding to a C 35/45 (EC2). As a conservative assumption the design of the shotcrete membrane, therefore, was based on a concrete grade of B35 corresponding to a C30/37 (EC2). The strength of the shotcrete during construction was continuously checked by concrete tests. 4.1.5

Stability analysis for the stages of construction

The stability analyses for the stages of construction of the Österfeld Tunnel were carried out at vertical slices according to the finite element method (Wittke, 2000). In the mudstone layers of the Lias α in addition to the stresses due to dead weight horizontal stresses ΔσH varying from 0.5 to 1.5 MPa - depending on the location of the considered computation section - were simulated. Analyses without consideration of increased horizontal stresses were carried out too. To account for the displacements which occur ahead of the temporary tunnel face and before the shotcrete membrane is installed in the two-dimensional analyses a stress relief was simulated by reducing the Young's modulus of the rock mass to be excavated (Wittke, 2000): Ered = av ⋅ E

with av ≤ 1

(4.1)

In the analyses the so-called stress relief factor av is varied between 1.0 (no stress relief) and 0.5. In the next step of analysis the excavation as well as the installation of the shotcrete membrane was simulated simultaneously. In the scope of the review of the design two- and threedimensional analyses using the computer code FEST03 (Wittke, 2000) were carried out. By means of these analyses the displacements monitored during excavation of the tunnel were back analyzed and the assumptions and thus the parameters taken as a basis for the stability analyses were checked. In the following the steps as

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- 142 well as the results of these analyses are explained by means of an example. In Fig. 4.6 the finite element mesh used for the three-dimensional analyses is illustrated (Hauck et al., 1998). The dimensions of the computation section are 100 m x 100 m x 40 m. The computation section is subdivided into 9074 isoparametric elements with 12720 nodes. The computation section is modeling the area of the Paradiesplatz, which is located approx. 120 m south of the portal "Unterer Grund" (Fig. 4.1 and 4.2). The setup of the finite element mesh enables the modeling of the stages of excavation, of the shotcrete membrane as well as of the subsoil profile and the railway trench. The analyses were carried out assuming an elasticviscoplastic stress-strain behaviour for the ground. For the mudstones of the Lias α with single layers of limestones and limesandstones as mentioned above a transversally isotropic stressstrain behaviour in the elastic domain as well as increased horizontal in-situ stresses were simulated. The soil and rock mechanical parameters as well as the three-dimensional finite element mesh are shown in Table 4.1 and Fig. 4.6.

Fig. 4.6:

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Computation section, finite element mesh and arrangement of extensometers and inclinometers

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- 143 -

Fig. 4.7:

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Steps of analysis for simulation of the heading of the tunnel WBI GmbH, Henricistr. 50, 52072 Aachen, Germany

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- 144 In the three-dimensional analysis the stages of construction arising during the heading of the tunnel were simulated in several steps, which approximately correspond to the driving and support of the tunnel in reality (Fig. 4.7). The applied so-called "step by step" method is explained in detail in Wittke (2000). In the two first steps of the analysis the in-situ state of stress due to dead weight and the additional horizontal stresses in the Lias α were calculated. For the simulation of the increased horizontal stresses ΔσH the nodes located at the boundary plane with the coordinate x = 100 m were displaced in x-direction. These displacements lead to horizontal stresses, which correspond to the stress ΔσH existing in the Lias α (Fig. 4.6). In the first step of analysis the whole computation section was horizontally loaded by these displacements. The unit weight γ was accounted for, however, only for the Lias α. In the second step of the analysis the soil layer underneath the surface as well as the Rät and the Knollenmergel, in which increased horizontal stresses are not existing, were substituted by materials, which have the same mechanical parameters as before, which however are not weightless any longer (γ > 0). Since the new materials were installed stress-free into the already deformed corresponding elements (Wittke, 2000) and because in the second step of the analysis the horizontal displacements of the boundary x = 100 m were not changed, the soil underneath the surface, the Rät and the Knollenmergel are loaded only by the dead weight and not subjected to increased horizontal stresses. The excavation of the railway trench was simulated within the third step of analysis. In the steps 4 up to 12 the crown excavation as well as the excavation of the bench and the invert following at a certain distance were simulated. In the computation case, which is illustrated in Fig. 4.7, unsupported round lengths of 3 m for the crown excavation and of 6 m for the excavation of the bench and the invert were simulated. These are larger than the real round lengths applied during construction which are mentioned below (Chapter 4.1.6). Hereby the development of strength of the shotcrete, which was not considered in the analysis, was roughly simulated. By modeling greater round lengths it was taken into account that the young shotcrete develops its complete bearing capacity only after a number of days (see Chapter 2.1). Therefore the distance between the load bearing shotcrete support and the tunnel face modeled in the analysis is larger as one round length in reality. The excavation of the bench and the invert was simuWBI-PRINT 5

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- 145 lated stepwise in a similar manner as the crown excavation (Fig. 4.7). By means of three-dimensional analyses it is possible to compute the stress redistributions in the area of the temporary tunnel face, which lead to stress concentrations in the rock mass ahead of the tunnel face, which is not yet excavated, as well as in the rock mass adjacent to the excavated cross-section and in the support already installed. To demonstrate the three-dimensional carrying behaviour the development of the calculated vertical displacement of a point at the roof during the heading of the tunnel is illustrated in Fig. 4.8. Up to the 7th step of analysis, in which the tunnel face passes the considered point, already 50 % of the final displacement due to excavation are evaluated as advancing displacement.

Fig. 4.8:

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Vertical displacements of a selected point at the roof in the course of analysis WBI GmbH, Henricistr. 50, 52072 Aachen, Germany

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- 146 Fig. 4.9 illustrates the calculated principal normal stresses in the rock mass resulting from excavation and support of the crown (12th step of analysis) computed for a cross-section perpendicular to the axis of the tunnel, and located in the area of the unsupported crown (cross-section C-C in Fig. 4.7). For the considered computation case an increased horizontal stress of ΔσH = 1 MN/m2 existing in the Lias α was assumed. As a result of computation an arch is formed in the rock mass around the crown. As a consequence of the increased horizontal stresses beneath the crown's invert near the contour of the excavation stresses up to 1.6 MN/m2 are computed. Above and underneath the unsupported cross-section of the crown as well as at the foot of the slope of the railway trench and also in the Knollenmergel the strengths on the discontinuities are exceeded (Fig. 4.9).

Fig. 4.9:

Principal normal stresses and subsoil areas in which strength is exceeded resulting from the crown heading , 12th step of analysis (section C-C in Fig. 4.7)

In a corresponding illustration Fig. 4.10 shows the displacements due to the crown heading. Fig. 4.10 represents the differences between the nodal point displacements computed for steps 12 and 3 of the analysis for cross-section B-B (Fig. 4.7). The displacements oriented towards the excavated opening amount to approx. 6 mm. For the foot of the slope of the railway trench displacements of approx. 10 mm are computed (Fig. 4.10).

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- 147 The comparison of the results of the two- and three-dimensional analyses shows that for the two-dimensional analyses a stress relief factor of av, which ranges from 0.5 to 0,8, is to be taken into account to achieve displacements corresponding to the results of the three-dimensional analyses. According to the results of the comparative analyses the shotcrete membrane can be designed with a thickness of 25 cm and a reinforcement consisting of an inner and outer steel fabric met Q 295 considering a safety factor of 1.35.

Fig. 4.10:

4.1.6

Displacements due to the crown heading, 12th - 3rd step of analysis (section B-B in Fig. 4.7)

Excavation and support

For the excavation of the tunnel with a total cross-section of 98 m2 (Fig. 4.3) a tunnel excavator was used. The heading was subdivided into a crown excavation and an excavation of bench and invert following the crown at some distance (Fig. 4.11). The distance from the excavation of the bench and the invert to the crown was chosen to at least 50 m. The round lengths for the crown's excavation were chosen between 80 cm and 1.2 m. For the excavation of the bench and the invert round lengths ranges from 1.6 to 3.0 m (Fig. 4.11). Thus during the crown's excavation as well as the excavation of the bench and the invert an early closure of the shotcrete membrane was realized. This measure has proven to limit the subsidence especially during the undercrossing of the buildings and the railway to a low level.

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- 148 -

Fig. 4.11:

Undercrossing of the Gäu- and S-Bahn (urban railway), longitudinal section

The regular support of the tunnel consists of a shotcrete membrane reinforced by two layers of steel fabric mats and with thicknesses varying from 25 cm to 30 cm. The temporary invert was supported by a 20 cm to 25 cm thick shotcrete membrane. Also steel sets and a systematic bolting around the crown and the bench are part of the support (Fig. 4.3 and 4.12). In Fig. 4.13 details of the support at the foot of the crown are illustrated. Due to the low overburden and the high frequency of the discontinuities near the ground surface in the area of the two portals grouted spiles were used as advancing support. The intensively jointed mudstones located immediately above the Oolithenbank (Fig. 4.4) turned out to be caving to a major degree. In case of a unfavorable location of this rock layer at the tunnel roof, therefore, also in greater distance to the portals the installation of grouted spiles were required. WBI-PRINT 5

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- 149 -

Fig. 4.12:

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Tunnel cross-section with support measures: a) Regular cross-section; b) undercrossing of the railway tracks WBI GmbH, Henricistr. 50, 52072 Aachen, Germany

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- 150 -

Fig. 4.13:

Design of support in the area of the crown's foot: a) Crown excavation; b) excavation of the bench and the invert

Fig. 4.14:

Crown excavation under a pipe umbrella

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- 151 In the area of the railway trench the excavation of the crown was carried out underneath the protection of a total of ten advancing pipe umbrellas (Fig. 4.11, 4.12 and 4.14). The steel pipes in each case were installed from niches and have a diameter of 140 mm and a length of 14.50 m. The overlap between two successive pipe umbrellas was selected to 3 m. The lengths of the niches, which were expanded continuously up to a maximum depth of 75 cm, is 6.5 m (Fig. 4.11). Before the excavation of the bench and the invert in the area of the niches was carried out, the niches were filled with shotcrete. The pipes consisting of 2 m long pieces were installed using a solid eccentric bit with an overcut of approx. 1 cm. Using prefabricated openings for injection (Fig. 4.12) as well as a double packer the annulus was grouted by means of cement based suspension. From the recorded grout volumes it could be concluded that by this procedure only the annulus between the pipe and the borehole wall was filled with grout. An appreciable grouting of the rock mass located between the pipes was not achieved. Finally, in a separate working operation, the steel pipes were filled with suspension (Hauck et al., 1998). The average completion time for a pipe umbrella was 5 days. During the corresponding interruption of the crown heading the bench and the invert excavation was carried out which was started after the installation of the first pipe umbrella because of their higher rate of advance. In this way an optimum rate of advance could be achieved (Hauck et al., 1998). For the time interval between the excavation and the installation of the shotcrete membrane the bearing behaviour of the steel pipe umbrella in the longitudinal direction of the tunnel is activated. In other words the space between the tunnel face and the load bearing shotcrete support in this stage is bridged by the pipes. As a consequence caving and loosening of the rock mass in this area as well as resulting subsidence are largely avoided. 4.1.7

Monitoring program and interpretation of the measuring results

During the heading of the tunnel an extensive monitoring program with special emphasis on the area of the undercrossing of the railway was carried out. Before the excavation started four main WBI-PRINT 5

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- 152 measuring cross-sections (MC2, MC5, MC9 and MC11, Fig. 4.2) with vertical combined extensometer and inclinometer measuring equipments on both sides of the tunnel (Fig. 4.6) were installed. On the surface above and aside of the tunnel, at the buildings, at the "Don-Carlos-Brücke" as well as at the sleepers of the railway tracks points for levelling were installed. The measuring program was complemented by underground convergency and displacement measurements within various tunnel cross-sections, which were carried out parallel to the tunnel driving. Eventual subsidence of the railway tracks was monitored by optical measurements carried out from fixed points located outside of the area of the railway tracks, by means of installation of reflectors, which were fixed at the sleepers. The measuring results show small subsidence of the ground surface with a maximum of 2.5 cm. During the undercrossing of the house Paradiesstraße no. 69 the maximum vertical displacement of the building could be limited to 12 mm. The differential settlements of the building were so small that the admissible angular rotations of the building were not reached and thus no visible damages of the building occurred. In the area of the railway tracks also no inadmissible subsidence or differential settlements could be observed. The results of the monitoring were evaluated and interpreted by means of finite element analyses (Hauck et al., 1998). Exemplarily the measuring results of the measuring cross-section 5 (MC5), which is situated in the area of the "Paradiesplatz" (Fig. 4.2) will be discussed. It reflects the situation, in which the tunnel is running immediately adjacent to the slope of the railway trench and the roof of the tunnel is approx. located at the elevation of the railway trench. The horizontal displacements measured by inclinometers in four boreholes during the crown's excavation above the tunnel are oriented towards the railway trench and amount up to 9 mm (Fig. 4.15a). In the height of the tunnel's cross-section the measured horizontal displacements are oriented on both sides towards the excavated opening and amount 6 to 7 mm. These displacements lead to horizontal convergencies of the side walls of the tunnel. WBI-PRINT 5

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- 153 -

Fig. 4.15:

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Measured and calculated horizontal displacements (MC5): a) Crown excavation; b) full excavation of the tunnel's cross-section WBI GmbH, Henricistr. 50, 52072 Aachen, Germany

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- 154 -

Fig. 4.16:

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Measured and calculated vertical displacements (MC5): a) Crown excavation; b) full excavation of the tunnel's cross-section WBI GmbH, Henricistr. 50, 52072 Aachen, Germany

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- 155 Above the roof of the tunnel a change of direction of the horizontal displacements occurs (drilling E7). This change can be observed also at the foot of the slope and was indicated by the tilting of the masts of the contact wire in the direction of the railway tracks. The following excavation of the bench and the invert of the tunnel has lead to an increase of the horizontal displacements from 50 up to 100 % (Fig. 4.15b). The measurement of the vertical displacements resulting from the crown excavation in the measuring cross-section 5 show a subsidence at the ground surface above the tunnel as well as in the rock mass above and adjacent to the tunnel. On the other hand in the area of the railway trench small heavings were measured (Fig. 4.16a). During the following excavation of the bench and the invert no significant changes of the vertical displacements were measured (Fig. 4.16b). The interpretation of the results of the measurements by means of finite element analyses leads to the result that the measured displacements can only be understood, if increased horizontal stresses in the rock mass are taken into account. The rock mechanical parameters on which the stability analyses are based on as well as the increased horizontal in-situ stresses in the order of magnitude of ΔσH ≈ 1 MN/m2 could be veryfied by the comparison of measured and calculated displacements (Fig. 4.15 and 4.16). 4.1.8

Conclusions

The design and construction of the eastern by-pass of StuttgartVaihingen can be considered as a challenge. Based on the experience gained in connection with large tunneling projects for road as well as for urban railway traffic carried out in the city of Stuttgart the complex and partly new tasks, which are related to the heading of the Österfeld tunnel, could be solved rather excellently. With regard to the stability and the displacements resulting from tunnel driving special attention was required because of the small overburden as well as the increased horizontal stresses and low WBI-PRINT 5

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- 156 shear strengths on the bedding parallel discontinuities of the mudstones of the Lias α, in which the tunnel is located. These problems were solved by the following measures: -

Subdivision of the excavation of the tunnel's cross-section in crown, bench and invert,

-

installation of a shotcrete membrane and a systematic bolting immediately after excavation,

-

short round lengths and an early closing of the shotcrete support during the crown's excavation as well as the excavation of the bench and the invert,

-

reinforced shotcrete support of the temporary invert of the crown,

-

use of a shotcrete with a quick-setting cement with a high early and ultimate strength,

-

carrying out of steel pipe umbrellas in the area of the undercrossing of the railway tracks with low overburden.

By these measures the excavation of the tunnel could be carried out with very little subsidence at the ground surface. At no stage of construction the railway traffic was affected. Moreover it has been found that the results of the threedimensional analyses using the finite element method, which were carried out within the scope of this project, have made an essential contribution for prognosis as well as for the design of the measures for excavation and support. 4.2

Road tunnel "Elite" in Ramat Gan, Israel

4.2.1

Introduction

The two-lane road tunnel "Elite" was headed in Ramat Gan, a city in the Tel Aviv area. The tunnel was started from a underground parking lot located adjacent to the 260 m high Gate Tower (Fig. 4.17), the highest building in the Middle East. The tunnel under-

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- 157 crosses eight-lane Jabotinsky Street to Tel Aviv in the area of an intersection (Fig. 4.18).

Fig. 4.17: WBI-PRINT 5

Gate Tower, Ramat Gan (Israel) WBI GmbH, Henricistr. 50, 52072 Aachen, Germany

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- 158 -

Fig. 4.18:

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Elite Tunnel, site plan with drill points

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- 159 4.2.2

Structure

The axis of the approx. 110 m long Elite Tunnel runs firstly along a circular arc with a radius of approx. 120 m and then changes into a straight line (Fig. 4.18). The tunnel clearance is approx. 4.9 m high and approx. 10 m wide (Fig. 4.19). The tunnel is located in a calcareous sand with low cohesion (Kurkar), which contains local lenses of cohesionless fine sands (see Chapter 4.2.3). The overburden ranges from 3 to 4.5 m.

Fig. 4.19:

Elite Tunnel, tender design, cross-section

The tender design included the advance installation of steel pipes 90 to 110 cm in diameter above and adjacent to the tunnel over the entire tunnel length using a microtunneling machine (Fig. 4.19 and 4.20). Under the protection of this umbrella of steel pipes filled with concrete, the tunnel was subsequently to be excavated. The approximately rectangular reinforced concrete tunnel cross-section (Fig. 4.19) was to be constructed in blocks in the process with the excavation being interrupted for each block (Fig. 4.20). In cooperation with Walter construction company (Walter Bau AG), WBI prepared a contractor's design proposal described in Chapter 4.2.4. This contractor's design proposal is based on the NATM and was later carried out. WBI-PRINT 5

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- 160 -

Fig. 4.20:

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Elite Tunnel, tender design, construction stages: a) Installation of block 1; b) excavation; c) installation of block 2 WBI GmbH, Henricistr. 50, 52072 Aachen, Germany

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- 161 4.2.3

Ground and groundwater conditions

In the course of the exploration for other construction projects, eight boreholes were sunk in the tunnel area. Fig. 4.18 shows the drill points of seven of these boreholes. In Fig. 4.21 the drill logs of four boreholes are projected onto a longitudinal section through the tunnel axis. According to these borehole logs, fill or clayey sands and clays exist down to a depth of 2 m. Below, medium dense to dense calcareously bonded, partially cemented sands with a fraction of gravel are encountered (Fig. 4.22). In these slightly cohesive sands, termed "Kurkar", locally cohesionless fine sands are intercalated, as mentioned above.

Fig. 4.21:

Elite Tunnel, longitudinal section with drill logs

To assess the relative density and the bulk modulus, Standard Penetration Tests (SPT) were carried out in each borehole. Fig. 4.23 shows exemplarily the drill log and the SPT results for borehole B13 located closest to the tunnel alignment (see Fig. 4.18). WBI-PRINT 5

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- 162 -

Fig. 4.22: WBI-PRINT 5

View of the temporary tunnel face located in the "Kurkar" formation WBI GmbH, Henricistr. 50, 52072 Aachen, Germany

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- 163 -

Fig. 4.23:

Borehole B13, drill log and Standard Penetration Test (SPT), results

According to this, Kurkar is encountered at the level of the tunnel cross-section, with the exception of a thin soil layer of uniWBI-PRINT 5

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- 164 formly graded sand with fines. On the basis of the blow counts per 30 cm of penetration depth determined with the SPTs ranging between N30 = 16 and N30 = 46, the Kurkar and the uniformly graded sand can be classified as medium dense to very dense (D = 0.5 to 0.7) according to DIN 4094, appendix 1 (1990). An estimate of the bulk moduli according to DIN 4094, appendix 1 (1990) on the basis of the determined values for N30 leads to moduli ranging from approx. 40 to approx. 100 MN/m2. The groundwater table was found at 0 to 1 m a.s.l. in the exploration boreholes, which is equal to 6 to 7 m below the tunnel's invert (Fig. 4.23). 4.2.4

Design

Fig. 4.24 shows the tunnel cross-section according to the contractor's design proposal prepared by WBI together with Walter Bau AG. The excavation contour circumscribes the clearance with a height of approx. 8.2 m and a width of approx. 12.1 m. Because of the low overburden, the roof was designed shallow with a radius of curvature of 11.75 m. For the sidewalls and the invert large radii were selected as well with 11.55 m and 14.06 m, respectively. At the transitions from the roof to the sidewalls and from the sidewalls to the invert the selected radii are comparatively small with 2.65 m and 2.12 m. The excavated cross-section amounts to approx. 85 m2. For construction management reasons and for reasons of the stability of the tunnel face, the cross-section is subdivided into the crown with temporary invert support and the trailing bench and invert (Fig. 4.24 and 4.25). The height of the crown amounts to 5.6 m. The transitions from the sidewalls to the temporary invert have radii of 1.7 m. The temporary invert of the crown is rounded with a radius of 18.38 m (Fig. 4.24). A thickness of 25 cm is selected for the shotcrete membrane in the vault. In the area of the bench, the invert and the temporary crown invert the shotcrete membrane is planned with a thickness of 20 cm. The thickness of the reinforced concrete interior lining amounts to 40 cm (Fig. 4.24).

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- 165 -

Fig. 4.24:

Elite Tunnel, contractor's design proposal, crosssection

To protect the work space and to relieve the area of the temporary tunnel face, the tunnel is planned to be excavated under the protection of pipe umbrellas consisting of 12 m long steel pipes with a diameter of ca 17 cm (Fig. 4.25 and 4.26). The pipes are spaced at approx. 40 cm and have a wall thickness of 7 mm. Four rebars are entered into each pipe to increase the section modulus. Further, the pipes are filled with B25 concrete (Fig. 4.26). The steel pipes overlap by 3 m (Fig. 4.25). As mentioned above, the WBI-PRINT 5

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- 166 pipe umbrellas are designed to bridge the work space and should therefore be able to carry the entire load resulting from overburden and traffic. Their design is therefore based on these loads (see Chapter 4.2.5).

Fig. 4.25:

Elite Tunnel, contractor's design proposal, excavation and support, longitudinal section

If the sand has a cohesion or an apparent cohesion, the gaps between the pipes can be bridged by the arching effect. If dry, loose sand or cohesionless fill appears in the roof area, the gaps between the pipes must be supported, e. g. by the installation of additional pipes (Fig. 4.26). Alternatively, the gaps may be bridged by steel plates welded to the pipes, or they may be supported using spiles (see Chapter 4.2.6). The contractor's design proposal includes a steep temporary tunnel face inclined at approx. 65° and supported by reinforced shotcrete and tunnel face anchors (SN-anchors) spaced at 2 to 3 m. The tunnel face anchors shall be 12 m long, just as the pipes (Fig. 4.25). By installing the anchors in parallel with the construction of the pipe umbrellas, the interruptions of the heading necessary due to the construction of the pipe umbrellas should be minimized. WBI-PRINT 5

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- 167 -

Fig. 4.26:

Elite Tunnel, contractor's design proposal, crosssection with pipe umbrella and detail

Round lengths of approx. 1 m were specified for the crown heading (Fig. 4.25). After each round, the shotcrete support including the closing of the temporary invert are to be completely installed before the excavation continues. This way the unsupported span in the crown never amounts to more than 1 m. Round lengths of approx. WBI-PRINT 5

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- 168 2 m were specified for the excavation of bench and invert (Fig. 4.25). Here as well the shotcrete support are to be completely installed after each round including its closing at the invert, so that the unsupported length amounts to 2 m at the most. The contractor's design proposal has essentially two advantages over the tendering design (see Fig. 4.19 and 4.20). The heading and the installation of the reinforced concrete interior lining are carried out in two separate working steps, which leads to a considerable gain of time and thus to cost savings. Further, the contractor's design proposal does not require the use of a microtunneling machine. 4.2.5

Stability analyses

Design of the shotcrete support For the dimensioning of the shotcrete support two-dimensional analyses were carried out using the program system FEST03 (Wittke, 2000). Fig. 4.27 shows the computation section, the FE-mesh, the boundary conditions, the ground profile and the parameters the analyses were based on. The computation section consists of a 48 m wide, 45 m high and 1 m thick slice. The FE-mesh was divided into 630 isoparametric elements with a total of 3958 nodes. For the nodes on the bottom boundary (z = 0) and on the lateral boundaries (x = 0 and x = 48 m) sliding supports were selected as boundary conditions. On the two planes perpendicular to the tunnel axis, equal displacements were prescribed for the nodes with equal x- and z-coordinates (Wittke, 2000). All nodes were assumed fixed in y-direction. The traffic load acting on the ground surface was accounted for by a surface loading (pt = 23 kN/m2). The overburden amounts to 4.5 m (Fig. 4.27). The ground was subdivided into two soil layers. Down to a depth of 8.5 m a medium dense sand was assumed with a Young's modulus of E = 100 MN/m2 and a Poisson's ratio of ν = 0.35, corresponding to a bulk modulus of Es = 160 MN/m2. Below that, a dense sand with E = 250 MN/m2 and ν = 0.35 corresponding to Es = 400 MN/m2 was specified. To be on the safe side, no cohesion was assumed for any of the two sands (c' = 0). An angle of friction of ϕ' = 30° WBI-PRINT 5

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- 169 was specified. For the shotcrete membrane, a Young's modulus of E = 15000 MN/m2 was assumed (Fig. 4.27).

Fig. 4.27:

Computation section, FE-mesh, boundary conditions, ground profile and parameters for two-dimensional analyses

Since the ground profile and the cross-section of the tunnel are symmetrical to the tunnel axis, only one half of the tunnel crosssection was modeled (Fig. 4.27). Fig. 4.28 illustrates the computation steps. The 1st computation step comprises the determination of the state of stress and deformation resulting from the dead weight of the ground and the traffic load pt. In the 2nd computation step the excavation and the shotcrete support of the crown are modeled. The 3rd computation WBI-PRINT 5

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- 170 step represents the excavation and shotcrete support of bench and invert.

Fig. 4.28:

Two-dimensional analysis, computation steps

It should be pointed out that the simulation of excavation and support in one computation step leads to a overestimation of the loading of the shotcrete membrane, since the displacements preceding the excavation, which have already occurred before the support is installed, are not taken into account in the analysis. This in turn results in the tunneling-induced ground surface subsidence being underestimated. Fig. 4.29 shows the displacements of the ground surface and the tunnel contour (crown) computed for the 2nd computation step relative to the 1st computation step. The largest displacements result at the roof with 25 mm and at the ground surface with a maximum of 22 mm. These values change only marginally in the 3rd computation step. Fig. 4.30 depicts the bending moments and normal thrust in the shotcrete membrane determined for the 2nd and 3rd computation step. Because of the small radius in the case of the crown heading the WBI-PRINT 5

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- 171 bending loading of the shotcrete support is larger in the 2nd computation step than in the 3rd computation step. The largest moments result on the sidewalls. In the 3rd computation step the largest bending moment occurs at the transition from the roof to the sidewalls. Compressive normal thrust are computed for the entire tunnel circumference.

Fig. 4.29:

Displacements, 2nd – 1st computation step

In Fig. 4.31 the statically required amounts of reinforcement are given for the design of the shotcrete membrane according to DIN 1045 (1988) for a safety factor of η = 1.45. This factor of safety includes the safety factors of ηt = 1.6 for the traffic load (pt) and ηγ = 1.4 for the overburden weight (γ · Ho) which were predetermined by the constructor:

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- 172 -

η =

Fig. 4.30:

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ηt ⋅ pt + η γ ⋅ γ ⋅ H o pt + γ ⋅ H o

(4.2)

≈ 1.45

Stress resultants in the shotcrete membrane: a) 2nd computation step; b) 3rd computation step

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- 173 -

Fig. 4.31:

Statically required reinforcement of the shotcrete membrane: a) 2nd computation step; b) 3rd computation step

For a concrete grade of B25, a steel grade of BSt 500/550 and a surface distance of the reinforcement of t1 = 3 cm, steel crosssections of ≤ 2.3 cm2/m on the inside and ≤ 5.2 cm2/m on the outside are evaluated as maximum statically required reinforcement. WBI-PRINT 5

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- 174 These can be covered by steel fabric mats Q295 on the inside and on the outside and by supplementary reinforcement within the upper sidewall area (Fig. 4.31). In the bench and invert area, the inside steel fabric mat Q295 can be omitted. Fig. 4.32 shows the design and the reinforcement of the shotcrete membrane in the area of the crown's foot.

Fig. 4.32:

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Design and reinforcement of the shotcrete membrane in the area of the crown's foot

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- 175 Stability of the temporary tunnel face To investigate the stability of the temporary tunnel face, threedimensional analyses were carried out using the program system FEST03 (Wittke, 2000). Fig. 4.33 shows the computation section, the FE-mesh, the boundary conditions, the ground profile and the parameters these analyses were based upon. The computation section is 40 m wide, 40 m high and 62 m long. The FE-mesh was divided into 5848 isoparametric elements with a total of 14945 nodes.

Fig. 4.33:

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Computation section, FE-mesh, boundary conditions, ground profile and parameters for three-dimensional analyses WBI GmbH, Henricistr. 50, 52072 Aachen, Germany

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- 176 Sliding supports were selected as boundary conditions for the nodes on the bottom boundary (z = 0). The nodes on the planes x = 0 and x = 40 m were assumed fixed in x-direction, and the nodes on the planes y = 0 and y = 62 m were fixed in y-direction. The traffic load acting on the ground surface was increased by the safety factor ηt = 1.6 and applied as a uniform surface load. An overburden of 4 m was specified (Fig. 4.33). The ground profile and the parameters corresponded to the assumptions made for the two-dimensional analyses. The weight of the soil (γ = 20 kN/m3) and thus also the overburden pressure were increased by the safety factor ηγ = 1.4. The tunnel face was assumed inclined at 80°. The tunnel face anchors were considered by a cohesion in the respective area (Fig. 4.33). Five cases were investigated: c' = 0 (no tunnel face anchors), c' = 25 kN/m2, c' = 50 kN/m2, c' = 75 kN/m2 and c' = 100 kN/m2, corresponding to an anchor arrangement with a raster spacing between 1 m x 1 m and 2.1 m x 2.1 m.

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Three-dimensional analyses, computation steps WBI GmbH, Henricistr. 50, 52072 Aachen, Germany

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- 177 The three-dimensional analyses were carried out in two computation steps as a simplification (Fig. 4.34). Following the analysis of the primary state in the 1st computation step, the excavation and simultaneous shotcrete support of the crown and the installation of the pipe umbrella were simulated in the 2nd computation step. The unsupported area adjacent to the tunnel face was assumed to be 1 m deep. The shotcrete support of the tunnel face was not taken into account. It should be pointed out that with these analyses as well the tunneling-induced displacements of the soil are underestimated, since the support exerted by the shotcrete membrane is overestimated with the computation sequence outlined in Fig. 4.34. For a realistic computation of the displacements, one of the two procedures for the simulation of a three-dimensional tunnel heading described in Wittke (2000) (step-by-step method or iterative method) would be necessary. The analyses, however, had the only purpose of assessing the stability of the temporary tunnel face.

Fig. 4.35:

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Principal normal stresses computed after completion of the viscoplastic iterative analysis and elements with exceeded strength (c' = 25 kN/m2), 2nd computation step (vertical section through the tunnel axis) WBI GmbH, Henricistr. 50, 52072 Aachen, Germany

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- 178 As a consequence of the excavation, the strength of the soil is locally exceeded in the area of the tunnel face. Fig. 4.35 shows the principal normal stresses computed for the 2nd computation step and the elements with exceeded strength, specifically marked, in a vertical section through the tunnel axis for the case of a cohesion of the anchored area of c' = 25 kN/m2. A corresponding representation of the computed displacements is given in Fig. 4.36. The displacements result from elastic and inelastic deformations of the soil. The latter are computed in a viscoplastic iterative analysis (Wittke, 2000). Fig. 4.35 and 4.36 show the results of the 2nd computation step after completion of the viscoplastic iterative calculation.

Fig. 4.36:

Displacements computed after completion of the viscoplastic iterative analysis (c' = 25 kN/m2), 2nd – 1st computation step (vertical section through the tunnel axis)

The criterion for the proof of stability of the tunnel face is the convergency of the nodal displacements in the course of the viscoplastic iterative analysis. Fig. 4.37 shows the development of the WBI-PRINT 5

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- 179 displacements computed for a node on the tunnel face in the course of the viscoplastic iterative analysis in the 2nd computation step for the five investigated cases. For the case without tunnel face anchors (c' = 0), the convergency of the displacement of this node cannot be proven by the analysis. In all other cases (c' ≥ 25 kN/m2) the displacement converges in the computations. For c' = 25 kN/m2 a displacement of only 7 mm results (Fig. 4.36 and 4.37). The cohesion of 25 kN/m2 corresponds to an anchor raster of 2.1 x 2.1 m.

Fig. 4.37:

Displacement of a node on the tunnel face in the course of the viscoplastic iterative analysis, 2nd computation step

Design of the pipe umbrella As mentioned above, the pipe umbrella must be able to carry the loads from traffic and from the weight of the overburden. As for the proof of stability of the tunnel face, the traffic load was increased by the factor of safety ηt = 1.6 and the overburden pressure of the soil by the factor of safety ηγ = 1.4 for the design of the pipe umbrella, as requested by the client.

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- 180 As explained in Chapter 4.2.4, the round length for the crown heading amounts to 1 m and the shotcrete support is to be installed and temporarily closed at the invert after each round. For the design of the pipe umbrella it is conservatively assumed that its maximum free span amounts to approx. 2 m. It is taken into account here that the green shotcrete does not have any bearing capacity yet directly after its application (see Chapter 2.1.3).

Fig. 4.38:

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Design of the pipe umbrella, statical systems and loading WBI GmbH, Henricistr. 50, 52072 Aachen, Germany

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- 181 At 80°, the inclination of the tunnel face is assumed somewhat steeper than in the design (see Fig. 4.25). This results in a distance between the tunnel face and the closed support of approx. 2 m at the roof and approx. 1 m at the temporary crown invert (Fig. 4.38).

Fig. 4.39:

Design of the pipe umbrella, calculation of the section modulus of the pipes

The beam on two supports and the beam fixed at both ends are considered as statical systems for the design of the pipe umbrella. For the reasons given above, the maximum span of the beam is assumed as 2 m. For the fixed beam, 0.5 m each on both ends of the beam are added to the length and assumed to be fixed. From the superposition of the traffic load and the overburden pressure, taking into account the spacing of the pipes (L = 418 mm), the loading of the beam results to q = 65.7 kN/m (Fig. 4.38). WBI-PRINT 5

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- 182 In Fig. 4.39 the calculation of the section modulus W of the pipes, which is required for the stress proof is specified. For the proof of safety, the computed stresses are compared to the yield stress of the pipes made from steel of the grade St37 (βy = σadm = 240 N/mm2). This is permissible, since the assumed loads were provided with factors of safety. The beam fixed at both ends is decisive for the design with a computed tension of 192.2 N/mm2 (Fig. 4.40).

Fig. 4.40:

Design of the pipe umbrella, stresses and deflection

The deflection of the pipe umbrella is estimated at approx. 3 mm (Fig. 4.40).

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- 183 4.2.6

Construction

The Elite Tunnel was the first tunnel in Israel to be constructed by the NATM. Particularly high demands were therefore made on the technical construction supervision provided by WBI. The demands focused mainly on the works for the excavation and the installation of the shotcrete membrane, which had to be continuously supervised. Differing from the design the pipes only had a length of 10.4 m. Since the overlap of the pipe umbrellas amounts to 3 m, it was possible to excavate for 7.4 m under one umbrella before the pipes for the next umbrella had to be installed (Fig. 4.41).

Fig. 4.41:

Elite Tunnel, excavation and support, longitudinal section

At the roof and above locally fill or layers with cohesionless, fine grained, loose sands were encountered. Here, the gaps between the steel pipes had to be supported (Fig. 4.42). In the area of the first two pipe umbrellas the gaps were supported by steel WBI-PRINT 5

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- 184 plates welded to the pipes (Fig. 4.42a). In the area of pipe umbrellas 3 to 14 cemented rebars (spiles) were installed between the steel pipes (Fig. 4.42b). The drillings for the spiles served at the same time to explore the ground in advance. In the area of the last two pipe umbrellas, additional steel pipes, filled with B25 concrete but not reinforced, were installed (Fig. 4.42c).

Fig. 4.42:

Support of the gaps between the pipes in case of locally occurring dry, loose sand or fill: a) Pipe umbrellas 1 and 2; b) pipe umbrellas 3 to 14; c) pipe umbrellas 15 and 16

Due to reasons of construction, a vertical tunnel face was carried out. For stability reasons the crown generally had to be excavated in several steps and supported immediately with reinforced shotcrete (t = 5 to 10 cm) (Fig. 4.41). Fig. 4.43 depicts a geotechnical mapping of the crown face at chainage 17. In Fig. 4.44, the supported crown is shown at chainage 36. It can be seen here that in the middle of the crown a sup-

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- 185 port core was carried out (see Fig. 4.41). Further pictures from crown heading construction are shown in Fig. 4.45 to 4.47.

Fig. 4.43:

Geotechnical mapping of the crown face, chainage 17

Fig. 4.44:

View of the supported crown face, chainage 36

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- 186 -

Fig. 4.45:

Start at the northern portal

Fig. 4.46:

Construction of a pipe umbrella

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- 187 -

Fig. 4.47:

Crown excavation in parts

The excavation of bench and invert was only started after the crown of the entire tunnel had been excavated (see Fig. 4.41). 4.2.7

Monitoring

To measure the subsidence due to the tunneling, 13 leveling points were installed at the ground surface above the tunnel roof, and eight measuring cross-sections with roof bolts were installed in the tunnel. The measurements served to monitor the stability of the tunnel. Fig. 4.48 shows exemplarily the subsidence of the ground surface and the tunnel roof measured during the crown heading with the temporary face located at chainage 39. The subsidence of the ground surface amounts to approx. 20 to 50 mm, the measurable subsidence of the roof accounts for approx. 15 to 20 mm. Fig. 4.48 further shows the ground surface subsidence measured after completion of the crown heading. It amounts to approx. 20 mm in the area of the northern portal and to approx. 40 to 60 mm in the remaining tunnel section. WBI-PRINT 5

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- 188 -

Fig. 4.48: WBI-PRINT 5

Surface and roof subsidence measured during crown heading WBI GmbH, Henricistr. 50, 52072 Aachen, Germany

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- 189 For the comparison of the measured subsidence with the results of the stability analyses it must be noted that the subsidence preceding the excavation, which has already occurred before the shotcrete support is installed, was not captured in the analyses (see Chapter 4.2.5). The subsidence that has occurred at the ground surface before the shotcrete support is installed amounts to approx. 20 mm in the example of Fig. 4.48. By adding this subsidence to the calculated subsidence of approx. 22 mm (see Fig. 4.29) and to the estimated deflection of the pipe umbrella of approx. 3 mm (see Fig. 4.40), a total subsidence of approx. 45 mm can be derived from the analysis results which agrees well with the measured subsidence (see Fig. 4.48). 4.2.8

Conclusions

The Elite Tunnel in Ramat Gan crosses through a medium dense to dense, calcareously bonded, slightly cohesive sand termed "Kurkar". Loose, cohesionless sands are locally embedded. The overburden amounts to 3 to 4 m. Since the tunnel undercrosses an eight-lane road, the subsidence had to be kept small. The tunnel was excavated by the NATM under the protection of a pipe umbrella as a crown heading with a closed support at the temporary invert and trailing bench and invert excavation. The crown was excavated in several parts with a vertical tunnel face, a support core and short round lengths and supported by reinforced shotcrete and tunnel face anchors. The shotcrete membrane was installed and closed at the invert at a distance of approx. 1 m to the tunnel face (see Fig. 4.41). Following this procedure the tunnel was excavated in a stable way. The ground surface subsidence amounted to approx. 4 to 6 cm. The results of the FE-analyses contributed essentially towards the design, the statics and the specification of the excavation and support measures.

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- 190 -

4.3

City railway tunnel to Botnang, in Stuttgart, Germany

4.3.1

Introduction

In the early 1990's the Stuttgart city railway line U9 was improved up to Botnang terminal. As a part of this project the "Herder Street" and "Lindpaintner Street" stops were connected by a 550 m long, double-tracked tunnel (Fig. 4.49a). The tunnel undercrosses the Botnang saddle in a wide turn as well as the Gäubahn and some built-up areas (Fig. 4.49). 4.3.2

Structure

Between the Botnang portal and chainage km 4+392 the two-tracked standard profile was constructed over a length of 379 m (Fig. 4.49b) with a height of approx. 8.5 m and a width of approx. 10.5 m (Fig. 4.51). From chainage km 4+392 to the Herder Street portal at chainage km 4+239.5 the height of the cross-section increases from approx. 8.5 to approx. 14 m (enlarged profile, Fig. 4.49b). The maximum cross-section at the Herder Street portal has a height of approx. 14 m and a width of approx. 12 m (Fig. 4.50). The maximum overburden amounts to some 47 m (Fig. 4.49b). The maximum cross-section at chainage km 4 + 239.5 is shown in Fig. 4.50. The shotcrete membrane is 30 to 35 cm thick, the thickness of the reinforced concrete interior lining amounts to 60 cm. In the area of the sidewalls and the invert, with R = 13.16 m and R = 14.5 m, respectively, comparatively large radii were selected for the rounding of both linings. In the roof area a smaller radius of R = 4.6 m was designed for statical reasons. At the transitions from the sidewalls to the invert smaller radii of R = 1.1 m were selected. The excavated cross-section amounts to approx. 142 m2. 4.3.3

Ground and groundwater conditions

To explore the ground and groundwater conditions core drillings were sunk along the tunnel alignment and equipped as observation wells (Fig. 4.49b).

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- 191 -

Fig. 4.49:

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City railway tunnel to Botnang: a) Site plan; b) longitudinal section with ground profile WBI GmbH, Henricistr. 50, 52072 Aachen, Germany

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- 192 -

Fig. 4.50:

City railway tunnel to Botnang, maximum crosssection (km 4+239.5)

The following formations exist in the area of the tunnel alignment from top to bottom (Fig. 4.49b): -

Fill,

-

residual loam, talus deposits,

-

Kieselsandstone layer and Lehrberg layers (claystones and marlstones), WBI-PRINT 5

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- 193 -

-

Untere Bunte Mergel (marlstones),

-

Dunkle Mergel and Schilfsandstone layer (marlstone, sandstones and claystones),

-

Gypsum Keuper.

Fig. 4.51:

Excavation and support, standard profile, excavation class 7A, cross-section

Fill of larger thickness occurs at the ground surface and at the portal areas. It mainly consists of firm to stiff sandy silt with embedded solid and hard rock fragments and to a minor degree also building rubble. Fill with a thickness of up to 17 m was found at the settlement-sensitive Gäubahn embankment, which the tunnel undercrosses over a tunnel length of 100 m (Fig. 4.49b).

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- 194 The flanks of the Botnang saddle are covered with talus material, the thickness of which usually amounts to 1 to 2 m. Up to 4 m of thickness are reached in shallowly dipping terrain where the talus material changes into residual loam. The latter consists of soft to stiff silts containing heavily varying amounts of solid and hard rock fragments. In the area of the Botnang Tunnel portal an inclinometer was installed already before the start of construction. The measurements gave no indication of slope movements. Layers of partially plastic leaching silts appear in the marls, which are belonging to the Keuper formation. The Schilfsandstone layer is composed of a series of gray-green, gray and brown, mostly solid and hard marlstone and sandstone banks that are weathered close to the surface. The upper part of the profile is dominated by marlstones with a varying however high fine sand content. Gray-black, sand-free claystones are intercalated in some areas as well. The thickness of this layer amounts to approx. 25 m. The lower part of the Schilfsandstone layer, located within the tunnel's cross section, consists mainly of hard sandstone banks with clay flasers. The fine- to medium-grained sandstones are clay-bonded or cemented by immediate siliceous grain bonding. Quantitative mineralogical investigations yielded quartz contents ranging from 30 to 40 %. The thickness of the layers varies between 10 and 100 cm. The bedding parallel discontinuities are occasionally marked by soft or firm clay layers, on which the banks tend to separation. Sandstones of the Schilfsandstone layer have a medium to wide joint spacing, while joints in clay- and marlstones are narrow- to medium-spaced. Two vertical joint sets exist, which intersect at an angle of 60 to 70°. Their acute angle includes the N-Sdirection. Thin sandstone banks are fractured into plates, while thick banks are fractured into columns rather. The joints are mostly rough, slightly undulating and closed. Open joints and clayey coatings occur mostly in the hillside area (slope dilatation) as well as close to the surface. The Gypsum Keuper consists of an alternating sequence of soft to firm silts, stiff claystone layers and hard claystone and dolomite layers. The Gypsum Keuper layers encountered are almost completely

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- 195 leached. In several boreholes single, strongly leached gypsum nodules as well as thin gypsum layers were found. The soil and rock mechanical parameters listed in Table 4.2 are based on the results of borehole expansion tests using the borehole jack model "Stuttgart", on laboratory tests on samples taken from the exploration boreholes as well as on experience gained from other projects located in comparable ground conditions. The stability analyses (see Chapter 4.3.5) were based on these parameters.

Layer

Deformability

Strength

Fill in the Gäubahn area

E = 10 MN/m2 ν = 0.35

ϕ' = 27.5° c' = 5 kN/m2

Talus material/ residual loam

E = 5 MN/m2 ν = 0.4

ϕ' = 27.5° c' = 5 kN/m2

Kieselsandstone

E = 2000 MN/m2 ν = 0.2

Joints J: ϕJ = 40°, cJ = 20 kN/m2

Lehrberg layers, E = 300 MN/m2 Untere Bunte Mergel, ν = 0.3 Dunkel Mergel

ϕ = 30° c' = 130 kN/m2

Schilfsandstone, portal areas

ϕ = 35° c' = 30 kN/m2

E = 150 MN/m2 ν = 0.3 2

Bedding B: ϕS = 40°, cS = 50 kN/m2 Joints J: ϕK = 40°, cK = 50 kN/m2

Schilfsandstone

E = 500-1000 MN/m ν = 0.2

Gypsum Keuper, weathered

E = 20 MN/m2 ν = 0.35

ϕ = 27.5° c' = 25 kN/m2

Gypsum Keuper, unweathered

E = 300-500 MN/m2 ν = 0.3

ϕ = 27.5° c' = 25 kN/m2

Table 4.2:

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Soil and rock mechanical parameters

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- 196 According to the results of the piezometer measurements, the undisturbed groundwater table is located above or at the tunnel roof almost over the entire tunnel length. In the area of the Herder Street portal it falls to about the middle of the tunnel's crosssection (Fig. 4.49b). 4.3.4

Design

From the Herder Street portal to chainage km 4 + 310 the doubletracked standard profile was planned to be excavated first as a temporary stage. After that the final profile was to be excavated from chainage km 4 + 310 to the Botnang portal. The final profile is equal to the enlarged profile from chainage km 4 + 310 to chainage km 4 + 392 and to the double-tracked standard profile starting at chainage km 4 + 392. After the cut-through the crosssection was planned to be enlarged in the invert to the final profile from chainage km 4 + 310 backwards to Herder Street (Fig. 4.49b). Fig. 4.51 shows the standard profile divided into crown, bench and invert and the excavation and support measures planned for the standard excavation procedure. The excavation class was designated as 7A according to the recommendations of the working group "Tunneling" of the German Geotechnical Society (DGGT, 1995: Table 1). Crown and bench were each excavated with an immediate installation of the support. The crown face may only be ahead of the bench face by a maximum of approx. 3 m (Fig. 4.52, phase I in the left portion). Shotcrete with a thickness of 25 to 30 cm and reinforced by two layers of steel fabric mats Q257, support arches spaced at 0.8 to 1.1 m and a systematic anchoring using SN-anchors were planned (Fig. 4.51). Mortar spiles should be used as advancing support. The support should be closed at the invert no more than 2.4 m behind the invert excavation and no more than some 16 m behind the excavation at the roof (Fig. 4.2, phase I in the left portion). The reinforced shotcrete should be placed at the invert with a thickness of 20 to 25 cm (Fig. 4.51). The tunnel face was planned to be excavated steeply and in steps, with a support core in the crown area. Immediately after excavation the tunnel face was to be supported using 5 to 10 cm thick shotcrete, possibly reinforced by a steel fabric mat Q257 (Fig. WBI-PRINT 5

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- 197 4.52, left portion). During the excavation the shotcrete support should be opened in sections, and the exposed partial areas should immediately be sealed again by shotcrete after the excavation.

Fig. 4.52:

Excavation and support, longitudinal section

Fig. 4.52 (phase II, right portion) and 4.53 show the excavation and support measures planned for the later lowering of the invert in the area with enlarged cross-section. In the upper part of the cross-section (crown, bench I and invert I), excavation and support was to be carried out as for the standard heading, but with a thicker shotcrete membrane (t = 30 to 35 cm) and systematic anchoring (see Fig. 4.51 and 4.53). A reinforced shotcrete membrane with a thickness of 30 to 35 cm should also be installed in the lower part of the cross-section (bench II and invert II). The support should be closed at the invert no more than 3.3 m behind the excavation of invert II and no more than 9.9 m behind the excavation of bench II (Fig. 4.52, right portion). The support of the tunnel face and the advancing support should be constructed as for the standard heading.

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- 198 -

Fig. 4.53:

4.3.5

Excavation and support, maximum cross-section, lowering of the invert, cross-section

Stability analyses for the design of the shotcrete support

For the design of the shotcrete support, two-dimensional FEanalyses on vertical slices were carried out with the program system FEST03 (Wittke, 2000). Fig. 4.49b shows the locations of the analysis cross-sections investigated in the final design analyses. The analyses were based on the parameters given in Table 4.2. In the following, analysis cross-section 2 will be exemplarily treated. It is located at chainage km 4+296 in the area of the undercrossing of the Gäubahn (Fig. 4.49b). In Fig. 4.54 the computation section, the FE-mesh, the boundary conditions, the ground WBI-PRINT 5

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- 199 profile and the parameters are shown. The computation section consists of a 63.9 m high, 45.7 m wide and 1 m thick slice of rock mass. The FE-mesh was divided into 1104 isoparametric elements with a total of 1320 nodes.

Fig. 4.54:

Computation section, FE-mesh, boundary conditions, ground profile and parameters, analysis crosssection 2 (km 4+296)

For the nodes on the lower boundary plane (z = 0) and on the lateral boundary planes (x = 0 and x = 45.7 m), sliding supports were introduced as boundary conditions (Fig. 4.54). For the two planes perpendicular to the tunnel axis (y = 0, y = 1 m), equal displacements were assumed as boundary conditions for nodes with equal xWBI-PRINT 5

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- 200 and z-coordinates (Wittke, 2000). All nodes were assumed fixed in y-direction. In the area of analysis cross-section 2, the tunnel has approx. 23 m of overburden. The upper part of the cross-section is located in the Schilfsandstone, the lower in the weathered Gypsum Keuper. Above the Schilfsandstone, hillside loam and the fill of the embankment constructed for the Gäubahn are modeled, with thicknesses of approx. 3 m and 15 m, respectively. The weathered Gypsum Keuper is underlain several meters below the tunnel's invert by the unweathered Gypsum Keuper (Fig. 4.54). The vertical section through the tunnel axis constitutes a plane of symmetry with respect to the geometry of the tunnel crosssection and to the ground profile. Therefore, only one half of the tunnel was modeled in the analysis (Fig. 4.54). Fig. 4.55 and 4.56 show the six computation steps used to simulate the excavation and the support during tunneling according to the design (reference case). In the 1st computation step, the state of stress and deformation resulting from the dead weight of the ground (in-situ state) is determined. In the 2nd computation step the excavation of the crown and its support using shotcrete are simulated. The 3rd computation step comprises the excavation and shotcrete support of bench I. In the 4th computation step excavation of the invert I and the closing of the temporary support at the invert are simulated. Since the first two construction stages (computation steps 2 and 3) were simulated with an open invert, this analysis sequence accounts for a late closing of the invert as specified in the design with a distance of approx. 16 m to the tunnel face (see Fig. 4.52). The connection of the temporary invert support to the sidewall was simulated at first without any curvature corresponding to the design (Fig. 4.55 and 4.56). In the 5th and 6th computation step, the excavation of bench II and the excavation of invert II with simultaneous installation of the shotcrete support also at the invert are simulated. Fig. 4.57 presents the principal normal stresses determined for the 6th computation step (complete excavation), as well as those areas in which the strength has been exceeded. It can be seen that the plastic zones are limited to the area of the tunnel contour.

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- 201 -

Fig. 4.55:

Analysis cross-section 2, reference case, computation steps 1 to 3

Fig. 4.56:

Analysis cross-section 2, reference case, computation steps 4 to 6

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- 202 -

Fig. 4.57:

Analysis cross-section 2, reference case, principal normal stresses and elements with exceeded strength, 6th computation step

Fig. 4.58 depicts the displacements computed for the 6th tion step and thus the displacements in the stage after vation of the total cross-section. The analysis results paratively large vertical displacements of 12 cm at the 7.4 cm on the ground surface.

computathe excain comroof and

Fig. 4.59 shows the stress resultants computed for the stage after the excavation and support of invert I (4th computation step) and of the total cross-section (invert II, 6th computation step). In the 4th computation step (excavation of invert I) extremely high bending moments result in the area of the connections of the temporary invert support to the sidewalls (Fig. 4.59a). For this loading, the shotcrete membrane cannot be reasonably designed. In the 6th computation step as well, comparatively large bending moments occur in the area of the sidewalls (Fig. 4.59b). The design of the shotcrete membrane for a concrete grade of B25, a shotcrete WBI-PRINT 5

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- 203 thickness of t = 35 cm and a safety factor during construction of 1.35 yields that in the lower sidewall areas a higher amount of reinforcement of the shotcrete membrane compared to the original design is required (Fig. 4.60).

Fig. 4.58:

Analysis cross-section 2, reference case, displacements, 6th – 1st computation step

Subsidence of the computed magnitude (Fig. 4.58) could also not be permitted for the undercrossing of the Gäubahn. On the basis of these analysis results and of the displacements measured during the heading (see Chapter 4.3.7), it was decided in agreement with all parties concerned to close the support at the invert earlier in order to keep the subsidence smaller. The computation steps for the analyses simulating an early support closing at the invert are shown in Fig. 4.61 and 4.62. The excavation of bench I and invert I as well as the installation of the temporary invert support were simulated in one computation step (3rd computation step, Fig. 4.61). In order to reduce the loading of the shotcrete membrane in the area of the connection of the temporary invert with the sidewalls, a rounded connection was modeled (Fig. 4.61).

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- 204 -

Fig. 4.59:

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Analysis cross-section 2, reference case, stress resultants in the shotcrete membrane: a) 4th computation step; b) 6th computation step

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- 205 -

Fig. 4.60:

Analysis cross-section 2, reference case, statically required inside reinforcement of the shotcrete membrane

To reduce the high bending loading of the shotcrete membrane in the sidewall areas after the excavation of the complete crosssection (H ≅ 14 m, see Fig. 4.59 and 4.60), an anchoring of the shotcrete membrane in the sidewall areas by untensioned anchors was accounted for in the 4th computation step in addition to the excavation of bench II. The anchoring in the sidewall areas was modeled by truss elements. Five untensioned anchors per tunnel meter were assumed in the analysis with a cross sectional area of 4.5 cm2 each. By the 5th computation step the excavation and support of the invert was simulated (Fig. 4.62).

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- 206 -

Fig. 4.61:

Analysis cross-section 2, case 2, early closing of the invert and rounding of the temporary crown invert, computation steps 1 to 3

Fig. 4.62:

Analysis cross-section 2, case 3, early closing of the invert and support of the sidewalls using anchors, computation steps 4 and 5

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- 207 The results of this analysis show first, that the computed displacements decrease considerably due to the simulation of an early closing of the invert and of the anchoring (see Fig. 4.58 and 4.63). The subsidence of the ground surface amounts to only 27 mm as opposed to 74 mm in the reference case. The horizontal displacements of the sidewalls decrease from 70 mm in the reference case to 18 mm.

Fig. 4.63:

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Analysis cross-section 2, case 3, displacements, 5th – 1st computation step

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- 208 The rounding of the cross-section in the area of the invert leads to a marked reduction of the bending moment (see Fig. 4.59a and 4.64). For a concrete grade of B25, a shotcrete support thickness of t = 35 cm and a factor of safety of 1.35 the analysis yields that supplementary reinforcement is required in the lower sidewall area in addition to the planned steel fabric mats Q257. Further, in the area of the greatest change in bending moment corresponding to the greatest shear force a shear reinforcement is necessary. It can be covered by diagonally bent-up reinforcement.

Fig. 4.64:

Analysis cross-section 2, case 2, stress resultants in the shotcrete support, 3rd computation step

In Fig. 4.65a the tensile anchor forces determined in the 5th computation step (see Fig. 4.62) are given. Values of up to 100 kN (10 t) per anchor are computed. The bending moments in the shotcrete membrane computed for the stage after the excavation of the complete cross-section (5th computation step) with the anchoring taken into account are given in Fig. 4.65b. Compared to the reference case (without anchoring) a strong reduction of the bending loading becomes evident (see Fig. 4.59b and 4.65b). The dimensioning yields that no supplementary reinforcement is required in the sidewalls if steel fabric mats Q257 are installed. At the transition from the sidewalls to the

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- 209 temporary invert, however, supplementary reinforcement in addition to the mats is required.

Fig. 4.65:

4.3.6

Analysis cross-section 2, case 3, 5th computation step: a) Tensile anchor forces; b) bending moments

Construction

Fig. 4.66 shows the excavation and support measures carried out within the area of the enlarged cross-section during the heading (Beiche and Kagerer, 1993). The shotcrete membrane was constructed with a thickness of 35 cm and rounded in the transition zones from the sidewalls to the temporary crown invert. The reinforcement included inside and outside reinforcement mats Q257 and supplementary reinforcement in the transition zones. Further, according to the design a systematic anchoring of the vault and the sidewalls was carried out (Fig. 4.66). In the central sidewall areas (invert I, bench II) the anchoring was intensified.

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- 210 -

Fig. 4.66:

Excavation and support within the area of enlarged cross-section, construction

In addition, advancing injections were carried out to improve the ground ahead the tunnel face and above the crown, and injection drill spiles with mortar filling were installed. The rounds were carried out with lengths of ≤ 80 cm. The temporary crown invert was supported in the beginning at a distance of 9 to 14 m to the tunnel face. In the area of the undercrossing of the WBI-PRINT 5

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- 211 Gäubahn and in the immediately following part of the heading, the support was closed at the invert at a maximum of 6 m behind the tunnel face (Beiche and Kagerer, 1993). 4.3.7

Monitoring

To measure the tunneling-induced displacements, measuring crosssections with extensometers and inclinometers were installed in the area of the undercrossing of the Gäubahn and of Zamenhof Street, among other locations. In the tunnel, measuring crosssections for levelings as well as stress measurement crosssections with rock mass pressure and concrete pressure measuring cells were installed. In addition, levelings were carried out on the grout surface, on the tracks of the Gäubahn and on structures (Fig. 4.67).

Fig. 4.67:

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Monitoring program

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- 212 The subsidence of the ground surface due to the crown heading amounted to approx. 25 mm at the beginning of the excavation and reached 51 mm at the edge of the Gäubahn embankment (Fig. 4.68). Up to this time the temporary crown invert was supported 9 to 14 m behind the tunnel face. To reduce the subsidence, especially during the undercrossing of the Gäubahn, the support was closed at the invert in the further course of the heading at the minimum possible distance of 6 m to the tunnel face, as mentioned above. It was thus possible to reduce the subsidence, which then amounted to only 34 mm in the area of the tracks of the Gäubahn (Fig. 4.68). The magnitude of this value is in reasonable agreement with the subsidence of 27 mm computed under the assumption of an early closing of the invert for the analysis cross-section 2 (see Fig. 4.63).

Fig. 4.68:

Measured ground surface subsidence, longitudinal section through the tunnel axis

In the further course of the heading the tunneling-induced subsidence decreased to very low values due to the more favorable geotechnical conditions (Beiche and Kagerer, 1993).

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- 213 4.3.8

Conclusions

During the construction of the Stuttgart city railway tunnel to Botnang, the tracks of the Gäubahn and built-up areas had to be undercrossed (see Fig. 4.49). Since in the area of the undercrossing of the Gäubahn the ground had a high deformability and a low strength, special measures had to be taken to limit the ground surface subsidence due to tunneling and to avoid interference with railway operations and damage to the buildings. A crown, bench and invert heading with closed support at the invert and a following lowering of the invert was chosen. With an early closing of the invert the tunneling-induced subsidence could be limited to admissible values. It was possible to reduce the loading at the transitions from the sidewalls to the temporary invert decisively by rounding the temporary crown invert. The loading of the high sidewalls after the enlargement of the crosssection could be clearly reduced by a systematic anchoring (see Fig. 4.65). With these measures it was possible to limit the ground surface subsidence during the undercrossing of the Gäubahn to an admissible value of some 3 cm. Railway operations were not interfered with and damage to the structures undercut and to the railway facilities did not occur. The the the the

FE-analyses contributed essentially to the specification of excavation and support measures, such as the early closing of invert, the curvature of the temporary crown invert as well as systematic anchoring of the sidewalls.

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- 214 5.

Sidewall adit heading

5.1

Road tunnel "Hahnerberger Straße" in Wuppertal, Germany

5.1.1

Introduction

To create a high-capacity and therefore non-intersecting east-west connection between the freeways A46 and A1, it was planned to newly construct and improve the state highway L418 running through the city of Wuppertal, Germany, as a four-lane road. Within the scope of this construction project, the Hahnerberger Straße, a street running approximately in north-south direction, was to be undercrossed in the city district of Hahnerberg at an angle of 60° by underground construction (Fig. 5.1). This task proved to be quite difficult, since the tunnel has a low overburden and a large excavated cross-section. Also, a limitation to the tunnelinginduced ground surface subsidence had to be kept due to the buildings at the ground surface in the area of the tunnel alignment.

Fig. 5.1:

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Road tunnel Hahnerberger Straße, site plan and exploration

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- 215 5.1.2

Structure

The approx. 130 m long tunnel must provide room in each direction for two main driving lanes and one exit or approach lane, respectively (Fig. 5.2 and 5.3). The widths of the required clearances range between 13.5 m and 16.2 m. The height of the clearances amounts to 4.90 m. The tunnel has a total width of approx. 37 m and a total height of approx. 12 m (Fig. 5.3). To illustrate these very large dimensions, the cross-section of a double-tracked tunnel for the new high-speed railway lines of German Rail (Deutsche Bahn AG) is shown in Fig. 5.3.

Fig. 5.2:

Road tunnel Hahnerberger Straße, longitudinal section

Fig. 5.3:

Road tunnel Hahnerberger Straße, cross-section

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- 216 The tunnel has an overburden height of approx. 10 m (Fig. 5.2). To avoid damage to the buildings above the tunnel and to the water and gas mains running along the Hahnerberger Straße, the subsidence due to tunneling was to be limited to a magnitude of 2 cm. 5.1.3

Exploration

Within the framework of route planning and in order to assess the feasibility of underground construction, an investigation program was carried out to explore the ground conditions and to determine the rock mechanical parameters. Test pits were excavated and core drillings were sunk with depths ranging between 10 and 34 m (Fig. 5.1). According to the exploration results, below an about 2 to 4 m thick layer of top soil and loam with cobbles, a narrowly bedded alternating sequence of sandstones and claystones of the Middle Devonian Brandenberg layers is found. The sandstones and claystones are weathered close to the surface and intensely jointed. The orientation of the bedding planes and joints in the rock mass was measured on oriented drill cores and in test pits using a geological compass. Fig. 5.4 shows the idealization of the discontinuity fabric of the rock by a structural model (see Chapter 2.5.1). The rock mass is separated by bedding planes (B), which dip shallowly at approx. 30° and persist widely, and by three steeply dipping main joint sets (J1 to J3). According to the drilling results, the spacing of the bedding planes ranges between several centimeters and a few decimeters. The joint spacing amounts to a few decimeters on average. The bedding planes are partially filled with clay and mixedgrained soils at a thickness of up to 10 to 40 cm. This leads to a greater deformability perpendicular to the bedding than parallel to it. After Wittke (1990), transversely isotropic deformation behavior can be assumed in the elastic stress domain for an alternating sequence of this kind. This kind of anisotropy can be described by 5 independent elastic constants: Two Young's moduli E1 and E2, one shear modulus G2 and two Poisson's ratios ν1 and ν2 (Fig. 5.4). The modulus E1 relevant for loading parallel to the bedding was derived from the results of the pressuremeter tests and from experi-

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- 217 ence gained from other projects with comparable ground conditions as E1 = 1000 MN/m2 (Fig. 5.4).

Fig. 5.4:

Structural model and rock mechanical parameters

The modulus E2, relevant for loading perpendicular to the bedding which is smaller than E1, was determined according to Wittke (1990) from the following relation: E2 =

1 α β + EIR EBF

=

1 0.9 0.1 + 1000 70

≈ 400 MN/m 2

(5.1)

The symbols in (5.1) denote: α

=

0.9:

β

=

0.1:

EIR =

1000 MN/m²:

EBF = 70 MN/m2:

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Fraction of the sandstone in the alternating sequence. Fraction of the bedding plane filling in the alternating sequence. Young's modulus of the jointed sandstone with the bedding plane filling not taken into account. Bulk modulus of the bedding plane filling.

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- 218 α and β were derived from a statistical evaluation of the mapping results of the drill cores and test pits. Taking the joints in the sandstone into account, the modulus E1 derived from the pressuremeter test results was chosen to EIR as represented above. The bulk modulus of the bedding plane filling EBF was estimated on the basis of the results of soil mechanical laboratory tests. The shear modulus G2, relevant for shear loading parallel to the bedding and thus strongly dependent on the mechanical properties of the bedding plane filling, was estimated at G2 = 325 MN/m2. Poisson's ratio ν1 corresponds approximately to Poisson's ratio of the sandstone. In unconfined compression tests on intact rock specimens a value of νIR = 0.25 resulted on average for the latter. This value was taken as a basis for the analyses. According to Wittke (1990), Poisson's ratio ν2 can be computed as follows: ν2 =

EBF ⋅ νIR α ⋅ EBF + β ⋅ EIR

=

70 ⋅ 0.25 ≈ 0.1 0.9 ⋅ 70 ⋅ + 0.1 ⋅ 1000

(5.2)

Laboratory tests resulted in very high values for the shear parameters of the intact rock. The failure behavior of the rock is thus essentially determined by the shear strength along the discontinuities, which was modeled by the Mohr-Coulomb failure criterion. For the shear strength parallel to the bedding, the bedding plane filling is relevant. On the basis of the grain-size distribution and water content and of experience, a friction angle of ϕB = 25° and no cohesion were assumed (Fig. 5.4). Unlike the bedding planes, the joints are mostly undulating and contain sandy, rusty coatings. Close to the surface, however, they are also partially filled with clayey, sandy silt. It is essential for the assessment of their shear strength that they mostly only extend through one layer and thus extend considerably less far than the bedding. Therefore, a cohesion of cJ ≈ 0 to 0.02 MN/m2 and a friction angle of ϕJ ≈ 30 to 35° were assumed for the joints (J1 to J3) (Fig. 5.4). A tensile strength normal to the discontinuities was not accounted for.

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- 219 The water permeability of the rock is relatively low perpendicularly to the bedding, because the bedding plane filling has a sealing effect. The groundwater table is located approx. 2 m above the tunnel roof. To further explore the rock mass conditions and to test the planned construction method, an adit was driven in the middle between the two tunnel tubes as an advance construction measure (Fig. 5.5). In the course of construction, a reinforced concrete buttress was installed in the adit. The additional ground exposure by the adit confirmed the findings about the ground which had been gained in the first exploration phase.

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Central adit for additional exploration and for testing of construction techniques WBI GmbH, Henricistr. 50, 52072 Aachen, Germany

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- 220 The adit was driven by means of smooth and true to profile blasting. To keep the blasting-induced vibrations low, the round lengths and the charges per ignition step were limited. Especially close to the buildings, the excavation profile had to be subdivided for the same reasons (I and II in Fig. 5.5). This experience could be used for the further planning and tendering of the tunnel. During the heading displacements were measured in the adit and at the ground surface and interpreted using FE-analyses based on the rock mechanical parameters given above (back analysis). The measured and the computed displacements were in substantial agreement. The derived parameters were thus confirmed and remained unchanged in the further course of the design (Modemann and Wittke, 1988). 5.1.4

Design and construction

In the conditions present here, only the NATM is suitable for an underground construction of the tunnel. In view of the width of the tunnel structure of approx. 37 m and of the overburden of approx. 10 m, which is very low by comparison, it was necessary to provide for one tunnel tube for each direction. For reasons of the alignment only a width of approx. 1 to 2 m was available for the buttress between the two tunnel tubes. The option to leave a rock pillar between the tubes was thus ruled out. Therefore, as mentioned above already, a 1.2 m thick reinforced concrete buttress with a mushroom-shaped widening of the head and a base enlargement was constructed in the exploration adit driven in advance (Fig. 5.6 and 5.7). The two tunnel tubes thus formed still have a very large span compared to the overburden. To enable at least a slight an arching effect in the remaining rock mass above the tunnel tubes, the tunnel profiles were designed with a very small camber above the prescribed clearance. The rounding of the inverts served to carry the water pressure on the interior lining better. Only a slight rounding was necessary here in spite of the wide span, because due to the small height of the clearance sufficient space was available to strengthen the invert of the interior lining. Since the groundwater table could be lowered during the heading, the shotcrete support did not have to be dimensioned for water pressure. The interior lining, however, was to be constructed waWBI-PRINT 5

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- 221 tertight. By request of the client, a PVC sealing was planned between the shotcrete membrane and the interior lining (Fig. 5.6). As a consequence of the shape of the cross-section, each tunnel tube was provided with a separate sealing, which did not include the central buttress. The arrangement and installation of the sealing was considerably simplified this way.

Fig. 5.6:

Design

The reinforced concrete buttress was constructed in segments of 9 m each. Connecting reinforcement was provided for at the head and at the base for the shotcrete support of the two tunnel tubes to be constructed later (Fig. 5.7). Further, a sound load transfer between the concrete buttress and the rock mass above was effected by means of a contact injection. After that, the tunnel tubes were excavated in parts of the cross-section for reasons of stability, limitation of the heading-induced subsidence and vibrations. The first step was the excavation and support of a sidewall adit in the northern tube. The cross-section of the sidewall adit was again subdivided by a crown heading with trailing invert (Fig. 5.8 and 5.9). As illustrated in Fig. 5.8, the outside sidewall was supported by a lattice girder and a 30 cm thick shotcrete membrane as well as by SN-anchors. The side of the sidewall adit facing the tunnel tube, on the other hand, was supported by only 15 cm of shotcrete. Glass fiber anchors were further installed on this side. This part of the support was removed again in the further course of the works. WBI-PRINT 5

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- 222 -

Fig. 5.7: WBI-PRINT 5

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- 223 -

Fig. 5.8:

Sidewall adit excavation and support

Fig. 5.9:

Sidewall adit and central adit

In the third step, the crown of the northern tube was excavated and supported in sections (Fig. 5.10 and 5.11). The lattice girders and the reinforcement of the shotcrete membrane were connected WBI-PRINT 5

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- 224 with the corresponding support elements of the sidewall adit in the process. The connection of the support to the central buttress was already mentioned above (see Fig. 5.7). The length of the SNanchors was increased in the vault area to improve the arching effect in the rock above the tunnel roof.

Fig. 5.10:

Crown excavation and support

Fig. 5.11:

Crown heading

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- 225 The excavation of the bench and the invert constituted the last steps of the excavation sequence of the northern tube (Fig. 5.12 and 5.13). The invert was supported by 30 cm of shotcrete. Lattice girders and an invert anchoring could be dispensed with.

Fig. 5.12:

Bench and invert excavation with closed invert support

Fig. 5.13:

Bench and invert excavation

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- 226 The stepwise excavation and installation of the support measures and the short round lengths of approx. 1 m, specified for stability reasons and to limit the blasting vibrations, lead to a comparatively low heading performance. This had the advantage that the shotcrete membrane was only fully loaded when the shotcrete had mostly reached its final strength. The same applies to the SNanchors. To keep the heading-induced subsidence as small as possible, the interior lining of the northern tube was to be installed before the southern tube was excavated. For reasons of construction management the contractor was allowed, however, to carry out the excavation and support of the southern sidewall adit in parallel with the concreting works in the northern tube. The sequence of construction in the southern tube was analogous to the northern tube (Modemann and Wittke, 1988). 5.1.5

Stability analyses for the stages of construction

A rock slice with a width of 65 m and a height of 37 m was specified as computation section for the stability analyses for the construction stages (Fig. 5.14). The FE-mesh was subdivided into 440 three-dimensional isoparametric elements with 1140 nodes. Sliding supports were specified as boundary conditions for the nodes of the lower boundary plane (z = 0). Since the rock mass is anisotropic and the bedding is neither oriented perpendicularly nor parallel to the tunnel axis (see Fig. 5.4 and 5.14), displacements in x- and y-direction already occur due to the dead weight of the rock mass. Therefore the nodes on the vertical boundary planes of the computation section must not be fixed perpendicularly to the respective boundary plane. The displacements in x- and y-direction resulting from the dead weight of the rock mass were determined by an advance analysis using a column-like computation section. These displacements were introduced as boundary conditions for the nodes on the vertical lateral boundary planes (x = 0 and x = 65 m) (Wittke, 2000). For the two planes normal to the tunnel axis equal displacements were assumed as boundary conditions for opposite nodes with equal x- and z-coordinates (Wittke, 2000). WBI-PRINT 5

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- 227 -

Fig. 5.14:

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- 228 -

Fig. 5.15:

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- 229 The stability analyses were carried out with the program system FEST03 (Wittke, 2000) in 4 computation steps: In the 1st computation step, the stresses and deformations in the rock mass resulting from the dead weight of the rock (in-situ state) are determined. In the 2nd computation step, the central adit as well as the northern sidewall adit are simulated to be excavated and supported. The 3rd computation step further accounts for the installation of the reinforced concrete buttress in the central adit and the crown excavation in the northern tunnel tube. Then, in the 4th computation step, the stresses and deformations occurring after the complete excavation and support of the northern tunnel tube are computed. The principal normal stresses determined for the 2nd computation step show clearly how the loads resulting from the overburden weight are diverted around the central adit and the sidewall adit (Fig. 5.15a). A mutual influence of the two excavations is not apparent. In the 4th computation step the loads resulting from the overburden weight are diverted around the entire excavation. Above and below the tunnel unloaded areas develop, whereas stress concentrations occur in the rock mass beside the northern tunnel tube as well as above and below the central buttress (Fig. 5.15b and 5.16).

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- 230 According to the results of the analyses, the dead weight of the overburden is supported to a large degree by the rock mass beside the tunnel and by the central buttress. The resulting stress distribution in the concrete buttress is shown in Fig. 5.17. It becomes apparent that a horizontal tension loading results for the mushroom-shaped buttress head, whereas the shaft is loaded by normal thrust and bending.

Fig. 5.17:

Stresses in the central buttress, 4th computation step

An appreciable bending loading of the shotcrete membrane only occurs at the connection to the central buttress (Fig. 5.18). A connecting reinforcement is provided for to carry this bending loading (Fig. 5.18). The maximum computed displacements due to the excavation of the northern tunnel tube amount to 22 mm at the tunnel's roof and to 8 mm at the ground surface (Fig. 5.19).

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- 231 -

Fig. 5.18:

Connection of the shotcrete membrane to the central buttress: a) Bending moments, 4th computation step; b) connecting reinforcement

Fig. 5.19:

Displacements, 4th – 1st computation step

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- 232 An interpretation of the loading of the shotcrete membrane and the central buttress shows that due to the arching in the rock mass above the tunnel (see Fig. 5.15) only a portion of the overburden weight is carried by the shotcrete membrane. To determine whether the shotcrete membrane and the central buttress can carry the overburden weight alone, an analysis in which no arching can develop in the rock mass above the tunnel was carried out within the framework of safety considerations. For this purpose, above and beside the tunnel tube a vertical discontinuity set was assumed in the rock mass, striking parallel to the tunnel axis and having no shear strength (ϕ = 0, c = 0). Shear stresses thus cannot be transferred across these discontinuities. The principal normal stresses and stress trajectories determined for this case show clearly that the weight of the overburden must be completely carried by the shotcrete membrane and the central buttress (Fig. 5.20).

Fig. 5.20:

Principal normal stresses and stress trajectories without arching in the rock mass, 4th computation step

A comparison of the normal thrust in the shotcrete membrane computed for this case with the ones of the previously investigated WBI-PRINT 5

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- 233 case, in which arching was possible in the rock mass above the tunnel, shows that the normal thrust increases considerably (Fig. 5.21). Also for this case, however, it was possible to dimension the shotcrete membrane taking into account the inside and outside steel fabric mats Q221.

Fig. 5.21:

5.1.6

Normal thrust in the shotcrete membrane with and without arching in the rock mass above the tunnel, 4th computation step

Stability analyses for the design of the interior lining

A number of stability analyses with varying assumptions was carried out for the design of the interior lining of the two tunnel tubes. Among other things it was assumed that the shotcrete membrane becomes ineffective over time.

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- 234 -

Fig. 5.22:

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Simulation of the water pressure acting on the interior lining

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- 235 A corresponding assumption was made in some analyses for the central buttress as well, since it is not protected from the groundwater by the PVC sealing. Furthermore, cases with and without arching in the rock mass were investigated. The result of merely one analysis shall be presented here, in which the interior linings are only loaded by their dead weight and by the water pressure. This case is relevant as long as the shotcrete membrane remains intact and carries the rock mass pressure. The water pressure is applied to the linings in the form of equivalent nodal forces (Fig. 5.22, detail I). The bedding of the interior lining is simulated by truss elements arranged between corresponding nodes of the shotcrete membrane and the interior lining and by the behavior of the shotcrete membrane and the rock mass, which is assumed elastic (Fig. 5.22, detail II). The truss elements possess a very high stiffness. They can only transfer compressive stresses, but not tensile or shear stresses. In this way the lack of a shear bond between the interior lining and the shotcrete membrane due to the PVC sealing is simulated. Since the rock mass pressure is not to be taken into account for the load case "dead weight and water pressure", the rock mass is assumed weightless. The computed bending moment distribution represented in Fig. 5.23 shows clearly the bending loading of the inverts and the central buttress resulting from the water pressure. The corresponding distribution of the normal thrust is shown in Fig. 5.24.

Fig. 5.23:

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Bending moments in the interior lining, load case "dead weight and water pressure"

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- 236 -

Fig. 5.24:

5.1.7

Normal thrust in the interior lining, load case "dead weight and water pressure"

Monitoring

A measuring program was planned to monitor the stability as well as the subsidence occurring at the ground surface and at the buildings. It included leveling and convergency measurements in the tunnel. In addition, extensometer measurements as well as levelings at the ground surface and at buildings were carried out during construction. Further, the groundwater levels in observation wells located beside the tunnel (see Fig. 5.1) were read continuously, and the blasting-induced vibrations were measured at the buildings. Of course, evidence was perpetuated at the buildings before the start and after the end of construction. The subsidence measured at a building located right above the tunnel is shown exemplarily in Fig. 5.25 for different stages of construction. It can be seen that the subsidence of the building at the end of construction is distinctly less than 2 cm, and that the building has subsided almost evenly. In the course of the different construction stages only small differential subsidence of the buildings occurred as well. No damage was found on this or the surrounding buildings (Modemann and Wittke, 1988).

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Fig. 5.25: WBI-PRINT 5

Measured subsidence of a building located above the tunnel WBI GmbH, Henricistr. 50, 52072 Aachen, Germany

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- 238 5.1.8

The road span and keep the brations

Conclusions

tunnel Hahnerberger Straße is a tunnel with a very large low overburden, for which the requirement was made to heading-induced subsidence at the ground surface and vias small as possible.

This task was solved with the following measures: -

Construction of two tunnel tubes with a reinforced concrete buttress in the middle providing additional support of the rock mass,

-

subdividing the cross-section of both tunnel tubes into several parts,

-

limiting the round length to approx. 1 m,

-

rounding the cross-sections of the two tunnel tubes in the vault area to enable the formation of a vault in the shotcrete membrane, the interior lining and the rock mass,

-

designing the shotcrete membrane and the reinforced concrete buttress in such a way that they can carry the complete rock mass pressure,

-

installation of the interior lining after the first tube was excavated and before the excavation of the second tube started,

-

smooth blasting.

Essential for the success of the construction project were furthermore the appropriate characterization of the ground and the proofs of stability established using the FE-program developed by the authors and their co-workers. 5.2

Limburg Tunnel, Germany

5.2.1

Introduction

The new railway line Cologne-Rhine/Main undercrosses a business area of the city of Limburg in a approx. 2.4 km long tunnel between chainage km 109+680 and km 112+075. WBI-PRINT 5

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- 239 -

Fig. 5.26: WBI-PRINT 5

Limburg Tunnel, site plan WBI GmbH, Henricistr. 50, 52072 Aachen, Germany

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- 240 -

Fig. 5.27: WBI-PRINT 5

Limburg Tunnel, longitudinal section with ground profile WBI GmbH, Henricistr. 50, 52072 Aachen, Germany

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- 241 In addition to business and industrial structures, the tunnel undercrosses among other things the freeway (Autobahn) A3 and the highway B 49 (Fig. 5.26 and 5.27). With overburden heights from 10 to 30 m it crosses through varying and geotechnically difficult ground in the form of predominantly weathered to decomposed slate as well as Tertiary and Quaternary clay, silt and sand (Fig. 5.27). For reasons of stability the excavation profile had to be subdivided during tunnel heading. In most parts a sidewall adit excavation with additional tunnel face support measures was carried out. 5.2.2

Structure

In the portal areas and in a short middle section (intermediate starting point), the Limburg Tunnel was constructed by the cutand-cover method. The sections in between, approx. 1190 m and 1090 m in length, were excavated by underground construction (Fig. 5.27). Seen from the direction of Cologne, the tunnel undercrosses first the sports grounds "Im Finken" with an overburden of approx. 16 m. Next, it undercrosses the freeway A3 and the traffic lanes of the freeway exit Limburg North at an acute angle with an overburden varying between 10 and 20 m. Between chainage km 110+750 and km 110+810, the highway B 49 and its underpass of freeway A3 are undercrossed with a small roof cover of approx. 4 m. Afterwards, the tunnel crosses under parts of the Massa shopping center and under the county road K 472 with an overburden of approx. 15 to 19 m. At chainage km 111+680 the tunnel undercrosses a high bay warehouse and between chainage km 111+800 and km 111+870 a warehouse of Tetra-Pak Co. (Fig. 5.26). Fig. 5.28 shows the geometry of the standard profile and of the sidewall adits. The double-tracked tunnel was constructed with a mouth-shaped profile with a width of 15.2 m and a height of 12.4 m. In the vault area a radius of curvature of R = 7.28 m was selected. The transition from the sidewalls to the invert was constructed with a radius of R = 4.43 m. For statical reasons the invert was deeply rounded with R = 11.63 m. The inside walls of the sidewall adits had a radius of curvature of R = 8.0 m. The roof and the invert of the sidewall adits were rounded with small radii of R = 0.4 m (Fig. 5.28). WBI-PRINT 5

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- 242 -

Fig. 5.28:

Limburg Tunnel, standard profile

The excavated cross-section amounts to approx. 150 m2. The sidewall adits as well as the remaining cross-section (core) were subdivided for the excavation into crown, bench and invert (Fig. 5.28). The shotcrete membrane has a thickness of 35 cm. The thickness of the inside shotcrete membranes of the sidewall adits amounts to 30 cm. The interior lining is 40 cm thick (Fig. 5.28). Fig. 5.29 shows the start of underground excavation at the northern portal.

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- 243 -

Fig. 5.29:

5.2.3

Limburg Tunnel, start of underground excavation at the northern portal

Ground and groundwater conditions

The ground at the alignment of the Limburg Tunnel was explored by core drillings, some of which were equipped as observation wells. In the area of Tertiary clays This formation lated 1 to 9 m

the northern portal the tunnel is located in the of the western sunken block of the Limburg rift. consists predominantly of silty clay with intercathick sand and gravel layers (Fig. 5.27).

At about chainage km 109+915 follows the middle horst block of the Limburg rift, consisting at first of slate decomposed to sandy and clayey silt. At the core of the uplifted block, solid Devonian slate is encountered. The slate has been strongly loaded in terms of fracture tectonics. It is strongly weathered in the upper zone up to a depth of approx. 45 m. At approx. chainage km 110+250, the tunnel leaves the uplifted block and crosses through the central sunken block of the Limburg

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- 244 rift up to about chainage km 111+640. As in the western sunken block, predominantly Tertiary clays are encountered here. At about chainage km 111+640, in the area of the high bay warehouse of Tetra Pak Co., the tunnel reaches the uplifted block of the Greifenberg horst. On the first approx. 70 m, the tunnel is located partially in decomposed keratophyre tuff and partially in strongly weathered, in some areas also completely decomposed slate, into which the tunnel enters completely in the following section. From about chainage km 111+910 to the southern portal, the tunnel crosses through slightly weathered to decomposed slate (Fig. 5.27). The foliation represents the dominant discontinuity set in the Devonian slate. It strikes from southwest to northeast and dips mostly at 40 to 70° in southeastern direction. Because of the tight (isoclinal) folding, however, foliation planes dipping steeply to the northwest occur as well. In connection with two steeply dipping joint sets, an approximately orthogonal discontinuity fabric exists in most cases. To determine the deformability of the ground, 9 dilatometer tests and 40 borehole jack tests were carried out in exploration boreholes located in the area of the Limburg Tunnel. According to these tests, a modulus of deformation for initial loading of 200 to 1000 MN/m2 can be assumed for slightly weathered slate. In decomposed slate, however, comparatively small moduli of deformation for initial loading of 5 to 80 MN/m2 were determined. Even smaller values of 5 to 60 MN/m2 resulted for the soil. For 63 samples from the clayey and silty surface layers as well as from the decomposed rock the unconfined compressive strength was determined. The unconfined compressive strength of the clay and silt of the surface layers scatters over the relatively wide range of 50 to 1900 kN/m2. The results of the tests on decomposed rock, however, are in the range of 100 to 500 kN/m2. The strength of the slightly to strongly weathered or decomposed slate is substantially determined by the low strengths on the discontinuities (see Chapter 5.2.6). Between the northern portal and the eastern boundary of the middle horst block, the groundwater table follows the course of the ground surface at a depth of approx. 5 to 10 m (Fig. 5.27). In the WBI-PRINT 5

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- 245 area of the central sunken block of the Limburg rift, the groundwater table is located at a depth of about 25 to 45 m at the level of the two receiving streams, the Elbach and the Lahn. To the east of the central sunken block, at the Greifenberg horst, the groundwater table rises again to approx. 5 to 10 m below the ground surface (Fig. 5.27). The seasonal variations in the water table can amount to several meters. The groundwater table shown in the longitudinal section (Fig. 5.27) is based on the highest groundwater levels measured in the boreholes equipped as observation wells. The groundwater is assumed to flow roughly from northeast to southwest. The water permeability of the Devonian rock and the Tertiary layers is estimated at 10-7 to 10-5 m/s. 5.2.4

Excavation and support

In the areas of the Greifenberg horst and the middle horst block, in which the tunnel cross-section is located in slightly weathered slate, a crown excavation with closed invert was planned according to excavation classes 5A-K, 6A-K or 7A-K, depending on the degree of weathering. In the other much longer sections the tunnel is located in soil or in decomposed slate. Here, a sidewall adit excavation was planned according to excavation classes 4A-U, 5A-U, 6AU or 7A-U (see DGGT, 1995: Table 1). Fig. 5.30 shows the sequence of excavation and the support for excavation class 7A-U-0, which was carried out for the most part. Excavation class 7A-U-0 is characterized by short round lengths for crown and bench (A = 0.6 m to 0.8 m), tunnel face support with plain shotcrete (t ≥ 7 cm), advance support with spiles and early closing of the invert (C ≤ 3.2 m). The tunnel profile is supported by a reinforced shotcrete membrane with two layers of steel fabric mats Q285 and by steel sets spaced at ea = 0.6 to 0.8 m. A systematic anchoring of the sidewall adits with SN-anchors was planned on the outside and as required also on the inside. Just as the excavation of the advancing sidewall adits, the excavation of the core is characterized by a short round length at the crown, tunnel face support with plain shotcrete and advance support with spiles (Fig. 5.31). The round lengths at bench and invert amounted to B = 2.8 to 3.2 m. The support was closed at the

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- 246 invert at a maximum of approx. 18 m behind the crown excavation (Fig. 5.31).

Fig. 5.30:

Excavation and support of the sidewall adits, excavation class 7A-U-0

The rock mass and the soil could be excavated mechanically using a tunnel excavator. The excavation was carried out temporarily at four locations simultaneously: -

Excavation north (starting from the northern portal), WBI-PRINT 5

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- 247 -

-

excavation south (starting from the southern portal),

-

excavations center north and center south (starting from the intermediate starting point, see Fig. 5.27).

Fig. 5.31:

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Excavation and support of the core, excavation class 7A-U-0

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- 248 5.2.5

Sidewall adit excavation north

The sidewall adit excavation north (Fig. 5.32 to 5.35) started at first with the excavation class 7A-U-0 with the support being closed at the invert after 4 rounds (C = 4A ≤ 3.2 m, see Fig. 5.30).

Fig. 5.32:

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Excavation north, right sidewall adit

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- 249 -

Fig. 5.33:

Excavation of the crown of the sidewall adit

After 160 m, as the heading reached the decomposed slates, which were fractured to small sizes, water entered more intensely through joints and foliation parallel discontinuities within the area of the tunnel face. This lead to stability problems at the vertical tunnel face and thus impeded the heading. Starting at chainage 193 m the excavation sequence was therefore modified. The step between crown and bench excavation was extended and the bench face was steepened. The support was closed at the invert at a distance of 6 rounds behind the crown excavation (Fig. WBI-PRINT 5

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- 250 5.36). In this manner a tunnel face more shallowly inclined on average was achieved. In addition, the safety of the staff during closing of the invert was increased.

Fig. 5.34:

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Excavation of the bench of the sidewall adit

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- 251 -

Fig. 5.35:

Invert of the sidewall adit

Fig. 5.36:

Modified sequence of excavation and closing of invert support after 6 rounds, sidewall adits, excavation north, chainage 193 to 267 m

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- 252 As stability problems continued at the tunnel face, an inclined crown and bench face supported with plain shotcrete was constructed starting at chainage 267 m. In addition, a reinforced shotcrete support was installed at the invert of the temporary bench of the sidewall adits to achieve a rapid stabilization of the excavation. Furthermore, the final closing of the support at the invert was carried out separately for the crown and bench heading. After the excavation and support of the crown and the bench of the sidewall adits had been completed ≤ 20 m in advance, the heading was interrupted and the tunnel face was sealed with plain shotcrete. Subsequently the excavation and support of the invert was carried out (Fig. 5.37). To drain off the water inflow through joints and foliation parallel discontinuities in the tunnel face area, a drainage was laid in the invert of the bench of the sidewall adits and a shaft sump reaching down to below the tunnel's invert was constructed every 10 to 20 m.

Fig. 5.37:

Inclined tunnel face and support of the invert of the temporary bench of the sidewall adits, excavation north, chainage 267 to 336 m

At chainage 336 m the heading was changed back again to excavation class 7A-U-0, since the strength of the rock mass increased and the jointing as well as the water inflow decreased. Fig. 5.38 is a photograph of the core excavation at excavation north.

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- 253 -

Fig. 5.38: 5.2.6

Excavation north, core excavation

Stability analyses for sidewall adit excavation north

The problems with the tunnel face stability that had occurred during the sidewall adit excavation north in the decomposed slate of the middle horst block (see Fig. 5.27) were attributed to the low shear strengths on the discontinuities of the strongly weathered to decomposed slate. The foliation and bedding parallel discontinuities F the orientations of which were measured during tunnel face mapping, strike approximately perpendicularly to the tunnel axis and dip mostly steeply with dip angles between 40° and 70° towards the tunnel face. In addition, two joint sets J1 and J2 exist, which strike in parallel and perpendicularly to the tunnel axis and dip steeply as well (Fig. 5.39). The water pressure acting in the foliation and bedding discontinuities, respectively, and in the joints had a further unfavorable effect on the stability of the tunnel face. Three-dimensional stability analyses were therefore carried out using the program system FEST03 (Wittke, 2000) for the sidewall adit excavation in the area of the middle horst block to investiWBI-PRINT 5

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- 254 gate the influence of the strength on the discontinuities and of the seepage pressure on the stability of the tunnel face.

Fig. 5.39:

Discontinuity orientations measured during sidewall adit excavation north, polar diagram

Fig. 5.40 shows the three-dimensional computation section, the FEmesh and the parameters the analyses were based upon. One sidewall adit was modeled. The overburden amounts to 23 m. In order to investigate the influence of the shear strengths on the discontinuities on the tunnel face stability three threeWBI-PRINT 5

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- 255 dimensional analyses were carried out (cases A, B and C, Table 5.1).

Fig. 5.40:

Computation section, FE-mesh and parameters Case

Discontinuities in the slate

ϕF [°] A

Table 5.1:

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ϕJ1 [°]

ϕJ2 [°]

no discontinuities

B

20

20

20

C

10

10

10

Limburg Tunnel, three-dimensional analyses regarding the tunnel face stability WBI GmbH, Henricistr. 50, 52072 Aachen, Germany

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- 256 -

Fig. 5.41:

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Computation steps for the simulation of the sidewall adit heading WBI GmbH, Henricistr. 50, 52072 Aachen, Germany

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- 257 -

Fig. 5.42:

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Seepage flow analysis, equipotential lines and seepage forces WBI GmbH, Henricistr. 50, 52072 Aachen, Germany

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- 258 In case A discontinuities were not taken into account. In cases B and C the foliation/bedding F, dipping at 70° towards the tunnel face, was simulated, together with the joint set J1, vertical and striking parallel to the tunnel axis, and the joint set J2, dipping at 60° opposite to the direction of heading (see Fig. 5.40). In case B the friction angle on the discontinuities was chosen as ϕF/J = 20° and in case C as ϕF/J = 10° (Table 5.1). No cohesion was assumed on the discontinuities. Young's modulus of the slate was specified as E = 50 MN/m2 in all cases (Fig. 5.40). The heading of a sidewall adit with crown, bench and invert excavation was simulated in 10 computation steps, which are outlined in Fig. 5.41. In the 11th computation step the seepage forces due to the water seeping through the rock mass were applied (Fig. 5.41), which had been determined in a three-dimensional seepage flow analysis using the program system HYD03 (Wittke 2000). This seepage flow analysis results in the distribution of piezometric heads h, from which the seepage forces can be calculated. The analysis was based on an undisturbed groundwater table located 16 m above the roof of the sidewall adit. Fig. 5.42 shows the location of the groundwater table lowered due to the tunnel excavation, the computed equipotential lines (h = const.), as well as, qualitatively, the direction and magnitude of the seepage forces FS oriented perpendicularly to the equipotential lines. The slate was simplificatively assigned a homogeneous and isotropic permeability with a permeability coefficient of kf = 10-6 m/s. The distribution of the piezometric heads and the seepage forces is, however, independent of the selected permeability coefficient, since a homogeneous and isotropic ground was assumed for the analysis (Wittke, 2000). Fig. 5.43 shows the development of the computed horizontal displacement δH of a point on the unsupported tunnel face in the course of the viscoplastic iterative analysis for computation steps 10 and 11. Case A results in a stable tunnel face with a maximum horizontal displacement of δH ≈ 3 cm. In case B the displacements are larger than in case A, with max. δH ≈ 6 cm in the 10th computation step and max δH ≈ 10 cm in the 11th computation step (with seepage pressure). The tunnel face is stable, however, since the displacements converge in the course of the viscoplastic iterative analysis. Case C WBI-PRINT 5

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- 259 results in a maximum displacement of δH ≈ 22 cm for the 10th computation step and of δH ≈ 35 cm for the 11th computation step. Although the displacement δH converges in this analysis as well, the tunnel face cannot be regarded stable anymore due to the magnitude of δH. For a horizontal displacement of 35 cm it must be assumed that the rock loosens up and disintegrates in the tunnel face area.

Fig. 5.43:

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Horizontal displacement of the tunnel face depending on the shear strength on the discontinuities WBI GmbH, Henricistr. 50, 52072 Aachen, Germany

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- 260 In summary it can be stated on the basis of the FE-analysis results that the stability of the tunnel face of the sidewall adit is substantially influenced by the discontinuity fabric. For shear parameters of cF = cJ1 = cJ2 = 0 and ϕF = ϕJ1 = ϕJ2 = 10°, which had been estimated from the results of the geotechnical mapping during heading, the tunnel face must be considered unstable. The analysis results further show that for low shear strengths on the discontinuities the seepage pressure acting on the tunnel face has a strong influence on the magnitude of the horizontal displacements. If follows that by an efficient drainage of the rock in advance of the sidewall adit excavation the problems during the heading could have been reduced decisively. 5.2.7

Monitoring results

The heading of the Limburg Tunnel was accompanied by a geotechnical monitoring program including surface leveling, extensometer and inclinometer measurements as well as leveling and convergency measurements in the tunnel. Fig. 5.44 shows exemplarily the vertical displacements of the sidewall adit roofs δSL and δSR and of the tunnel roof δR measured after the excavation of the entire cross-section in excavation north from chainage 160 to 375 m (km 109+900 to km 110+115). The measured vertical displacements of the roofs of the two sidewall adits are larger than the one of the roof of the total crosssection, because with the latter only a part of the displacements occurring during the core excavation can be captured. The measured displacements of the sidewall adit roofs include a part of the subsidence that occurred during the sidewall adit heading and in addition the subsidence resulting from the core excavation. The maximum subsidence of δSL = 45 mm was measured at chainage 300 m (Fig. 5.44). In this area the sidewall adit was excavated with an inclined tunnel face and with a supported invert of the temporary bench (see Fig. 5.37). From chainage approx. 340 m on the measured subsidence decreased to a few millimeters. In this area the sidewall adit excavation was changed over to excavation class 7A-U-0 (see Fig. 5.30). Starting with chainage approx. 360 m the roof was located in slightly weathered to unweathered slate. Here, the measured roof subsidence decreased to ≤ 3 mm (Fig. 5.44).

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- 261 -

Fig. 5.44:

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Measured vertical displacements, chainage 160 to 375 m

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- 262 5.2.8

Conclusions

With the Limburg Tunnel of the new railway line Cologne/Rhine-Main stability problems often occurred at the tunnel face during the sidewall adit excavation north in the area of the middle horst block. These problems impeded the heading and required intensified support measures. In this section the tunnel cross-section was located in strongly weathered slate which is strongly jointed to fractured to small sizes. The groundwater table was encountered approx. 15 m above the roof. By three-dimensional FE-analyses the influence of the shear strength on the discontinuities of the slate and of the seepage pressure in the rock mass on the stability of the tunnel face was investigated. The results show that the tunnel face stability is substantially determined by the discontinuity fabric. It could further be proven that the seepage pressure resulting from the seepage flow in connection with the low strength on the discontinuities caused the stability problems during the sidewall adit excavation north. The impediments during the heading and the intensified support measures could have been considerably reduced by an efficient drainage of the rock mass in advance of the heading. 5.3

Niedernhausen Tunnel, Germany

5.3.1

Introduction

The Niedernhausen Tunnel lies in the southern part of the new railway line Cologne-Rhine/Main in the area of the town of Niedernhausen, south of Idstein, Germany. Over a length of approx. 350 m following the northern portal the tunnel is located in the completely weathered and decomposed slates of the Schwall layers with an overburden of about 17 to 50 m. Further, the groundwater table lies at roof level or above in this area and the freeway A3 had to be undercrossed (Fig. 5.45). Under these conditions the tunnel face was not stable without supplementary support measures. Moreover, the tunnelinginduced subsidence of the ground surface in the area of the undercrossing of the freeway A3 had to be kept small. The high demands and the difficult ground conditions made it necessary to drain the rock in advance and to construct the tunnel in this area in partial excavations with additional tunnel face support measures. WBI-PRINT 5

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- 263 -

Fig. 5.45:

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Niedernhausen Tunnel: a) Site plan; b) longitudinal section with ground profile WBI GmbH, Henricistr. 50, 52072 Aachen, Germany

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- 264 5.3.2

Structure

The two-tracked Niedernhausen Tunnel is 2765 m long (Fig. 5.45). From chainage km 140+867 to km 141+499 as well as in the southern portal area the tunnel was constructed by the cut-and-cover method. The overburden in this area amounts to less than 10 m. Over the remaining 2101 m the tunnel was excavated by underground construction. The maximum overburden height is 95 m (Fig. 5.45b).

Fig. 5.46:

Niedernhausen Tunnel, standard profile with sidewall adits

Between chainage km 141+600 and km 141+700 the tunnel undercrosses the federal highway A3 at an acute angle with an overburden of 20 to 30 m. The tunnel roof lies in the hillside loam in this area. The remainder of the cross-section is located in the already mentioned deeply weathered and decomposed Schwall layers (Fig. 5.45b). A sidewall adit heading was planned in this area. WBI-PRINT 5

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- 265 Fig. 5.46 shows the 15.7 m wide and 12.6 m high standard profile with the geometry of the sidewall adits. The excavated crosssection amounts to approx. 150 m2. The vault area was constructed with a radius of curvature of R = 7.25 m. In the sidewalls a radius of R = 9.05 m was selected. The transitions from the sidewalls to the invert were constructed with a radius of R = 2.8 m. The invert was rounded with R = 14.15 m. The inside walls of the sidewall adits had radii of curvature of R = 6.375 m and R = 10.925 m, respectively. The roof and the invert of the sidewall adits were constructed with a small radius of R = 1.0 m. The temporary crown inverts of the sidewall adits were rounded with R = 6.25 m (Fig. 5.46). The shotcrete membrane was carried out unusually strong with a thickness of 35 to 45 cm. The inside shotcrete membrane of the sidewall adits were constructed 30 cm thick. The thickness of the interior lining amounts to 40 cm (Fig. 5.46). The sidewall adits were subdivided into crown and bench for the heading. The remaining cross-section (core) was subdivided into crown, bench and invert (Fig. 5.46). Fig. 5.47 shows the starting wall of the northern heading at the northern portal.

Fig. 5.47: WBI-PRINT 5

Niedernhausen Tunnel, starting wall of the northern heading WBI GmbH, Henricistr. 50, 52072 Aachen, Germany

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- 266 5.3.3

Ground and groundwater conditions

The ground at the alignment of the Niedernhausen Tunnel was explored by core drillings. Some of these boreholes were equipped as observation wells. The Quaternary hillside loam and talus material extends to a depth of approx. 40 m in the northern tunnel section. In the southern tunnel area, however, the Quaternary surface layer is only 12 m thick at the most (Fig. 5.45b). Below the Quaternary cover Devonian rock sequences follow (Fig. 5.45b) consisting of silty and finely sandy slate (Schwall layers), micaceous sandstone and quartzite with slate intercalations (Hermeskeil layers), quartzitic sandstone and quartzite with embedded sandy slate (Taunus quartzite) and phyllitic slate with sandstone and quartzite layers (Variegated Schist). The slate of the Schwall layers is characterized by a far-reaching weathering which can extend to a depth of about 150 m. It corresponds in strength to a cohesive soil. Young's modulus, however, is roughly that of a well-compacted gravel sand. The discontinuity fabric has remained intact in spite of the weathering (Fig. 5.48).

Fig. 5.48: WBI-PRINT 5

Deeply weathered, decomposed slate of the Schwall layers WBI GmbH, Henricistr. 50, 52072 Aachen, Germany

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- 267 The bedding planes and foliation discontinuities strike in NE-SW direction, which is typical for the Rhine schist mountains, and dip mostly steeply at 60 to 90° in varying directions. In combination with two discontinuity sets usually an orthogonal discontinuity fabric exists. To the south of the Taunus ridge upthrust, in the middle and the southern part of the tunnel, the rock mass is mostly unweathered. Individual steep bedding-parallel fault zones here exist (Fig. 5.45b). From the northern tunnel portal to about km 142+500 the groundwater table follows roughly the course of the ground surface at a depth of 5 to 25 m. In the southern tunnel section it drops to approx. 50 m below the ground surface due to a drinking water supply facility. This level is still up to approx. 40 m above the tunnel roof, however (Fig. 5.45b). The seasonal variations in the groundwater table amount to up to 10 m. The groundwater table shown in the longitudinal section (Fig. 5.45b) is based on the highest groundwater levels measured in the boreholes equipped as observation wells. 5.3.4

Excavation and support

In the approx. 350 m long section in which the tunnel's crosssection is partially or completely located in the weathered Schwall layers, a sidewall adit heading was planned according to excavation classes 7A-U-0 and 7A-U-1, respectively (Fig. 5.49, see DGGT, 1995: Table 1). The two excavation classes differ with regard to the unsupported round lengths, the spacing of the steel sets and the number of anchors per m2 (Fig. 5.49). In both excavation classes the shotcrete membrane is reinforced by two layers of steel fabric mats Q295. As mentioned above, the thickness of the shotcrete membrane of 35 to 45 cm is unusually strong. The inside shotcrete membrane of the sidewall adits was to be carried out with a thickness of at least 30 cm. Further, a tunnel face support with plain shotcrete (t ≥ 7 cm) and an advancing support with spiles were planned (Fig. 5.49).

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- 268 -

Fig. 5.49: WBI-PRINT 5

Excavation and support, excavation classes 7A-U-0 and 7A-U-1 WBI GmbH, Henricistr. 50, 52072 Aachen, Germany

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- 269 -

Fig. 5.50: WBI-PRINT 5

Pipe umbrella for the undercrossing of freeway A3 WBI GmbH, Henricistr. 50, 52072 Aachen, Germany

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- 270 In the area of the undercrossing of the freeway the workspace for the tunnel excavation was to be additionally supported by pipe umbrellas made of 14.5 m long pipes with an inclination of approx. 10 % and an overlap of > 3 m. The pipe umbrellas were to be constructed from niches (Fig. 5.50). 5.3.5

Three-dimensional stability analyses

Calibration of the analysis model (analysis cross-section AC 1)

In the course of the final design for the tunnel driven by sidewall adit excavation three-dimensional stability analyses were carried out with the program system FEST03 (Wittke, 2000). A total of four cases were investigated (cases A, B, C and D, see Table 5.2). Temporary support of c the sidewall [kN/m2] adit invert 5 no

Slate, decomposed Ho Case [m]

E [MN/m2]

ϕ [°]

Tunnel face support

Unloading modulus

-

-

A

17

20

25

B

25

"

"

"

yes

crown: p = 0.04 MN/m2

EU = 3E

C

"

"

"

"

no

"

"

D

"

40

10

yes

"

"

Table 5.2:

27.5

Niedernhausen Tunnel, three-dimensional analyses of the sidewall adit heading

To calibrate the analysis model and to verify the soil and rock mechanical parameters the analyses were to be based upon, the vertical and horizontal ground displacements measured at measuring cross-section MC 2 (Fig. 5.51) during the sidewall adit heading were back-analyzed first. MC 2 (analysis cross-section AC 1) is located at km 141+580 approx. 20 m in front of the undercrossing of freeway A3 (Fig. 5.51). To record the vertical and horizontal displacements resulting from the heading, leveling points were set up at the ground surface and 3 multiple extensometers as well as 2 inclinometers were installed beside and above the tunnel. Zero readings were taken sufficiently ahead of the excavation. The arWBI-PRINT 5

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- 271 rangement of these measuring devices is shown in Fig. 5.52 together with the tunnel's cross-section.

Fig. 5.51:

Longitudinal section, location of measuring crosssection MC 2 and analysis cross-sections AC 1 and AC 2

Fig. 5.53 shows the three-dimensional FE-mesh for the backanalysis of the measured displacements together with its main dimensions, the boundary conditions and the modeled stratigraphy. To keep the computing time within justifiable bounds, symmetric ground conditions as well as a horizontal ground surface and horizontal layers were assumed. It was thus sufficient to model one half of the tunnel's cross-section and the surrounding ground.

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- 272 -

Fig. 5.52:

MC 2 (km 141+580), measuring devices

According to the assumed symmetry the two sidewall adits are simulated as being excavated parallel in time in the analyses. In reality the western sidewall adit heading ran ahead of the eastern one by 10 to 20 m (g in Fig. 5.49). The influence of the simplified simulation in the analysis on the final result is small, however. The overburden in the area of measuring cross-section MC 2 amounts approx. 17 m. The boundary between the soil and the strongly weathered rock mass was modeled in the FE-mesh at about half the height of the tunnel's cross-section. The characteristic parameters of the soil include a Young's modulus of 10 MN/m2 and shear parameters of ϕ' = 25° and c' = 7.5 kN/m2. For the weathered rock mass, the corresponding values are E = 20 MN/m2 to 40 MN/m2, ϕ = 25° to 27.5° and c = 5 kN/m2 to 10 kN/m2 (Fig. 5.53). Thus both layers differ essentially in their deformability and only negligibly with respect to their shear strength.

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- 273 -

Fig. 5.53:

Computation section, FE-mesh, boundary conditions, ground profile and parameters for the threedimensional analyses (AC 1 and AC 2)

The outside shotcrete membrane was constructed 45 cm thick in the course of the sidewall adit heading, and it is modeled accordingly (Fig. 5.54). The shotcrete membrane of the inside walls of the sidewall adits is 30 cm thick. The shotcrete support of the temporary crown invert of the sidewall adits shown in Fig. 5.54 was not installed in the area of MC 2. It was therefore not modeled in analysis case A for the back-analysis of the measured displacements (see Table 5.2).

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- 274 -

Fig. 5.54:

FE-mesh, detail

The computation sequence of the analyses comprising a total of 36 computation steps is shown schematically in Fig. 5.55. The analysis of the in-situ state (1st computation step) is followed in the 2nd computation step by the excavation of a 23 m long crown section of the sidewall adits together with an invert section trailing by 2 m. The 2nd computation step is the starting stage for the actual heading simulation carried out in computation steps 4 to 36 according to the iteration technique described in detail in Wittke (2000).

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- 275 -

Fig. 5.55:

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Computation steps for the simulation of the sidewall adit heading WBI GmbH, Henricistr. 50, 52072 Aachen, Germany

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- 276 The basic idea of this iterative analysis is that the computation section moves with the heading or with the simulation of each round, respectively. With increasing number of iterations the constraints are reduced which develop in the starting stage due to the restraint of the displacements in longitudinal tunnel direction. In this way the computed displacements approach the actual displacements of the excavation surface at the time of the installation of the shotcrete membrane. In this manner the loading of the shotcrete membrane and the deformations due to the heading can be realistically determined.

Fig. 5.56:

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Sidewall adit heading, comparison of measured and computed vertical displacements

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- 277 According to excavation class 7A-U-1 applied in construction (see Fig. 5.49) a round length of 1 m is simulated. The distance between the shotcrete support and the temporary tunnel face amounts to 1 m in the crown and 2 m in the invert. The closing of the invert of the sidewall adits is thus completed in each case at a distance of 4 m from the crown face (Fig. 5.55). The tunnel face support using shotcrete is not taken into account. The vertical and horizontal displacements in the rock mass and on the ground surface computed in case A for the sidewall adit heading are compared with the measurement results in Fig. 5.56 and 5.57. The computed displacements were taken from the plane located 14 m behind the crown face in the 36th computation step. This distance represents approximately the average of the distances of MC 2 to the crown faces of the two sidewall adit headings at the time of the measurement (see Fig. 5.56 and 5.57).

Fig. 5.57: WBI-PRINT 5

Sidewall adit heading, comparison of measured and computed horizontal displacements WBI GmbH, Henricistr. 50, 52072 Aachen, Germany

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- 278 A comparatively good agreement results between the measured and the computed displacements (see Fig. 5.56 and 5.57). It is only in the computed vertical displacements that heave show at the level of the tunnel's invert and below which was not measured. According to experience this is because in the FE-analysis the same modulus was specified for the areas unloaded by the excavation as for the loaded areas above and beside the tunnel. The modulus relevant for unloading is generally markedly higher than the one for loading. In the stability analyses for the sidewall adit heading in the area of the undercrossing of the freeway (cases B, C and D, see Table 5.2), which are described in the following, an unloading modulus EU equal to three times the loading modulus was therefore specified below the tunnel's invert level (see Fig. 5.54). Three-dimensional analyses of the sidewall adit heading (analysis cross-section AC 2)

In the area of the undercrossing of the freeway (analysis crosssection AC 2) the overburden height varies between 23 and 27 m (see Fig. 5.51). The stability analyses for this area were correspondingly based on an average overburden of 25 m (see Fig. 5.53). The investigation included cases B, C and D (see Table 5.2). Differing from case A, an unloading modulus EU = 3 E was specified below the invert in these analyses, as already mentioned (see Fig. 5.54), together with a crown face support with p = 0.04 MN/m2 (see Fig. 5.55 and Table 5.2). In this way the supporting effect of horizontal cemented anchors installed in advance of the crown face was modeled. In cases B and D a shotcrete support of the temporary crown invert (t = 10 cm, see Fig. 5.54) is accounted for. In case C the temporary crown invert support is not taken into account. In case D the specified shear strength of the decomposed slate is higher than in cases A, B and C (see Table 5.2). In Fig. 5.58 the stress resultants are shown as a function of the distance from the tunnel face and from the boundary of the mesh for the example of the normal thrust and the moment in the roof. The broken lines in Fig. 5.58 show the extrapolated course of the stress resultants which would ensue without the influence of the boundaries. The stress resultants in the section located 16.5 m behind the tunnel face (dimensioning section) can be regarded as

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- 279 decisive and taken as a basis for the design of the shotcrete membrane.

Fig. 5.58:

Sidewall adit, stress resultants vs. distance from the crown face, case B, 36th computation step

Fig. 5.59 shows the stress resultants M, N and S in the dimensioning section for case B. Very large bending moments M occur in the areas of the small radii of curvature of 1.0 m at the roof and at the transition from the inside wall of the sidewall adit to the invert (see Fig. 5.46). It was assumed for the design of the shotcrete membrane that the computed bending moments would not develop in reality to their full extent since the shotcrete membrane is WBI-PRINT 5

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- 280 installed in several working steps. The construction joints in the areas mentioned above enable the formation of links which result in a reduction of the bending moments but not of the normal thrust. In agreement with the parties concerned the calculated bending moments were reduced for the design to 65 %.

Fig. 5.59:

Sidewall adit, stress resultants in the shotcrete membrane, dimensioning section, case B, 36th computation step

Fig. 5.60 shows the statically required reinforcement crosssections for bending and normal thrust determined for cases B, C and D. In addition to the reduction of the moments already mentioned, the design was based on a B25 concrete grade, a BSt 500 steel grade and a distance of the reinforcement from the edge of t1 = 4 cm. In the roof (sections 1 and 2) and at the transition from the outside wall of the sidewall adit to the invert (section 3) statically required reinforcement cross-sections are computed for case B which are not covered by the planned reinforcement (Q295 inside and outside) and require supplementary reinforcement. In the remaining area (section 4), no reinforcement is required for statical reasons. A minimum reinforcement of Q295 inside and outside suffices here. For case C, in which a support of the temporary crown invert is not accounted for, even larger statically required reinforcement cross-sections result in sections 1 to 3 than for case B (see Fig. 5.60). For case D, in which a higher Young's modulus and a higher shear strength are specified for the decomposed slate, statically required reinforcement cross-sections

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- 281 larger than the minimum reinforcement only ensue in the roof (sections 1 and 2, Fig. 5.60).

Fig. 5.60:

Sidewall adit, required reinforcement cross-sections for bending and normal thrust

As a consequence of the results of these analyses, during the undercrossing of the freeway A3 by sidewall adit excavation supplementary reinforcement was installed in the roof area and the temporary invert of the crown of the sidewall adits was supported with shotcrete. The tunnel face was additionally supported by tunnel face anchors (see Chapter 5.3.6). Further, an advancing pipe umbrella was installed for the excavation of the core (see Fig. 5.50). Three-dimensional analyses of a crown heading with closed invert and tunnel face anchoring

Because only a low heading performance could be achieved with the sidewall adit heading due to the extensive support measures, WBI investigated in further analyses whether in the section following the undercrossing of freeway A3 also a crown heading with closed invert and tunnel face anchoring could be carried out in a stable way for the same ground conditions and higher overburden (see Fig. WBI-PRINT 5

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- 282 5.45b; Wittke and Pierau, 2000; Wittke and Sternath, 2000). For this purpose three-dimensional FE-analyses were carried out for the computation section shown in Fig. 5.61 with the program system FEST03 (Wittke, 2000).

Fig. 5.61:

Computation section, FE-mesh, boundary conditions and parameters for three-dimensional analyses of the crown heading with closed invert and tunnel face anchoring

The analyses were based on an overburden of 50 m. Because of the higher overburden compared to the undercrossing of freeway A3, a higher Young's modulus and a higher shear strength were assumed for the decomposed slate with E = 100 MN/m² and EU = 300 MN/m², respectively, ϕ = 27.5° and c = 20 kN/m2 (see Table 5.2 and Fig. 5.61). WBI-PRINT 5

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- 283 -

Fig. 5.62:

Crown heading with closed invert and tunnel face anchoring: a) Stress resultants in the shotcrete membrane; b) required reinforcement

The computation sequence for the simulation of the crown heading was analogous to the analyses simulating the sidewall adit heading (see Fig. 5.55). The analysis results show that the stability of the crown heading can be proved if the tunnel face is supported by advance anchoring and the support is closed soon at the invert. As in the analyses described before, the tunnel face anchors were accounted for by a support of the tunnel face with p = 0.04 MN/m2 (see Fig. 5.61). It can be proven that the subsidence due to the heading can be kept small in this way. The computed loading of the shotcrete membrane WBI-PRINT 5

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- 284 of the crown due to bending and normal thrust (Fig. 5.62a) reveals a high bending compression loading at the transition to the temporary invert. The reason for this is the small radius of curvature of the shotcrete membrane of only 2 m. For reasons inherent to the construction process it is often attempted to keep the radius of curvature as small as possible. For statical reasons, on the other hand, the radius of curvature should not be less than 2 m. Supplementary reinforcement cannot be dispensed with in this case, however, although the membrane is rounded according to this requirement (Fig. 5.62b). 5.3.6

Construction

The Niedernhausen Tunnel was excavated between the northern tunnel portal at chainage km 141+499 and chainage km 141+771.5 by sidewall adit heading. The mapping during tunneling showed that the layer boundary talus material / slate is located in the crown area of the tunnel's cross-section (Fig. 5.63).

Fig. 5.63: WBI-PRINT 5

Construction, northern section WBI GmbH, Henricistr. 50, 52072 Aachen, Germany

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- 285 The excavation work proved very difficult mainly because of the high groundwater level. As the closing of the invert was delayed due to the inflow of water, large subsidence resulted at the ground surface (Wittke and Sternath, 2000). To stabilize the tunnel face it was necessary to drain the ground in advance and also to support the tunnel face using shotcrete and anchors (Fig. 5.64 and 5.65). The heading performance was correspondingly low with approx. 1 m/day in the sidewall adits.

Fig. 5.64: WBI-PRINT 5

Tunnel face support of the sidewall adit, crown

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- 286 -

Fig. 5.65:

Tunnel face support of the sidewall adit, crown and bench

Fig. 5.66 depicts the support measures during the sidewall adit heading in cross- and longitudinal section. It shows that excavation class 7A-U-1 was modified as follows with respect to the original design (see Fig. 5.49): -

Supplementary reinforcement in the sidewall adit roofs,

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- 287 -

support of the temporary crown inverts of the sidewall adits,

-

tunnel face anchoring in the crown area of the sidewall adits,

-

closed inverts in the sidewall adits at ≤ 8 m behind the crown excavation.

Fig. 5.66:

Construction, sidewall adit heading: a) Crosssection; b) longitudinal section

Pipe umbrellas (see Fig. 5.50) were only carried out in the core area of the tunnel's cross-section during the undercrossing of the freeway A3.

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- 288 For the advance drainage of the ground, at first up to 50 m long horizontal vacuum wells were constructed from the starting wall (see Fig. 5.47). In addition, vacuum lances were installed from the tunnel face during the heading. The effectiveness of the vacuum lances remained limited, however. The outflow of water in the tunnel face area could only be reduced but not be prevented in this way. The ensuing mud formation interfered considerably with the excavation. As a consequence, the groundwater was wells drilled from the ground surface down to below the tunnel's invert. In vance it was possible to increase the afterwards.

lowered using deep vacuum in advance of the excavation the ground drained in adheading performance markedly

Fig. 5.67 is a photograph of the northern heading during the excavation of the core.

Fig. 5.67:

Excavation of the core

From the south the tunnel was driven up to chainage km 141+771.5 by crown heading. In the section where the tunnel cross-section was located in the decomposed slate the crown was excavated up to WBI-PRINT 5

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- 289 the cut-through with a closed invert and tunnel face anchoring (Fig. 5.63). The basis for this were the results of the FEanalyses described in Chapter 5.3.5. Fig. 5.68 depicts the support measures carried out during the crown heading in cross- and longitudinal section. The round lengths and the excavation sequence are shown in the longitudinal section as well.

Fig. 5.68:

WBI-PRINT 5

Construction, crown heading with closed invert and tunnel face anchoring: a) Cross-section; b) longitudinal section WBI GmbH, Henricistr. 50, 52072 Aachen, Germany

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- 290 The computed displacements were well confirmed by the results of surface leveling and convergency measurements in the tunnel. The heading performance in the crown amounted to almost 2 m/day (Wittke and Pierau, 2000). 5.3.7

Conclusions

The Niedernhausen Tunnel was excavated over a length of approx. 350 m in the completely weathered and decomposed slates of the groundwater-bearing Schwall layers, which have a low strength and a high deformability. In this ground the freeway A3 had to be undercrossed with an overburden of 20 m to 30 m. A sidewall adit heading with short round lengths was carried out. To stabilize the tunnel face the rock had to be drained in advance and the tunnel face had to be supported by shotcrete and anchors. The results of three-dimensional FE-analyses showed that additional measures were required to support the work space at the tunnel face, such as e. g. the support of the temporary crown invert of the two sidewall adits and the installation of spiles. Because of the extensive support measures a very low heading performance could only be achieved. It could be proven by further three-dimensional analyses that in sections with higher overburden also a crown heading with closed invert and systematic tunnel face anchoring could be carried out in a stable way in these unfavorable ground conditions. As a consequence, the Niedernhausen Tunnel was successfully excavated by crown heading outside of the sphere of influence of freeway A3. Almost twice the heading performance of the sidewall adit heading was achieved here. The experience gained with the excavation of the Niedernhausen Tunnel should also lead to economic solutions for future tunnel structures in comparable ground conditions. It should be possible in many cases to replace an expensive sidewall adit excavation by a crown heading with closed invert and tunnel face anchoring.

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- 291 6.

Full-face heading

6.1

Urban railway tunnel underneath the Stuttgart airport runway, Germany

6.1.1

Introduction

The urban railway of Stuttgart, Germany, was extended in 2001 by a section starting at the Airport station and running underneath the airport area to the city of Filderstadt-Bernhausen. The Airport station (construction lot 72) and a continuation of limited extent towards Bernhausen (lot 92) were constructed by the cut-and-cover method. A 2.15 km long tunnel section driven by underground construction (lot 601) follows. It undercrosses among other areas the apron and the runway of the airport. The approx. 500 m long tunnel section of the Filderstadt-Bernhausen station (lot 602) was constructed by the cut-and-cover method (Fig. 6.1).

Fig. 6.1:

Undercrossing of Stuttgart airport, site plan

The present Chapter describes the tunnel section of lot 601 driven by underground construction, with an emphasis on the undercrossing of the runway of the airport. In this section the absolute and WBI-PRINT 5

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- 292 differential ground surface subsidence due to the tunneling had to be kept as small as possible. This task proved to be quite demanding, because settlement-sensitive, soft valley deposits and fill are locally encountered in the runway area and the groundwater table lies above the tunnel roof. Lowering the groundwater table during tunneling could therefore not be permitted, because of the risk of large settlements. The shotcrete support had thus to be constructed with a low water permeability and designed to withstand the water pressure. For the design of the shotcrete membrane it had further to be taken into account that high horizontal stresses exist especially in the mudstones of the Lias α formation, in which the major part of the mined tunnel section is located (see Chapter 4.1). 6.1.2

Structure

The course of the alignment and the ground profile are shown in Fig. 6.2. Following lot 92, the alignment descends in the direction of Bernhausen up to the area in front of the runway.

Fig. 6.2:

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Undercrossing of Stuttgart airport, longitudinal section with ground profile WBI GmbH, Henricistr. 50, 52072 Aachen, Germany

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- 293 The tunnel then runs horizontally over a length of approx. 1400 m (lot 601). The following gradient extends into the cut-and-cover section (lot 602). In the area of the station the alignment then runs approximately horizontally again.

Fig. 6.3:

Single-tracked tunnel tube, standard profile

Two single-tracked tunnel tubes are planned for the mined tunnel section (lot 601), only one of which has been built for the time being, however. The tunnel tube was constructed with a circular profile, among other reasons also in order to be able to design the shotcrete membrane for the full water pressure. The 30 to 35 cm thick shotcrete membrane was made from alkali-free dry-mix shotcrete with a low water permeability using spray cement as a bonding agent (see Chapter 2.1.2). The excavated diameter is apWBI-PRINT 5

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- 294 prox. 8.7 m, the inside diameter 7.0 m. The excavated crosssection of one tunnel tube amounts to approx. 60 m2 (Fig. 6.3). In the area of the airport runway the tunnel roof is located approx. 21 m below the ground surface (Fig. 6.4). Stuttgart Airport Co. (Flughafen Stuttgart GmbH, FSG) demanded that the tunnelinginduced subsidence in this area has to be limited to 15 mm and the differential subsidence at the ground surface to 1 o/oo.

Fig. 6.4:

6.1.3

Undercrossing of the runway of Stuttgart airport, longitudinal section with ground profile Ground and groundwater conditions

The ground profile in the area of the mined tunnel of lot 601 is similar to the one in the area of the Österfeld Tunnel (see Chapter 4.1). The stratigraphic sequence includes the layers of the Knollenmergel, the Rät, the Lias α and the Quarternary (see Fig. 4.4).

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- 295 According to geotechnical criteria the ground can be subdivided as follows from bottom up (Fig. 4.4, 6.2 and 6.4): -

Knollenmergel and Rät,

-

Lias α, predominantly mudstone with single layers of limesandstone,

-

Lias α, alternating sequence of mudstone and lime-sandstone,

-

overlying strata consisting of Lias α residual clay and Filder loam.

To explore the ground and the groundwater conditions core drillings were sunk along the tunnel alignment. Some of these boreholes were equipped as observation wells.

Fig. 6.5:

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Photograph of the tunnel face showing the transition from the alternating sequence of mudstone and limesandstone to the layers consisting predominantly of mudstone

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- 296 -

Fig. 6.6:

Fill and valley deposits in the runway area of Stuttgart airport

According to the exploration results, the mined tunnel of lot 601 is located over its entire length in the rock layers of the Lias α WBI-PRINT 5

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- 297 formation (Fig. 6.2 and 6.4). In the area of the airport runway it lies completely within the layers consisting predominantly of mudstone. Adjacent to the cut-and-cover tunnel sections, i. e. in the areas with descending or climbing alignment, the tunnel crosses through the alternating sequence of mudstone and banks of limesandstone (Fig. 6.2). Fig. 6.5 is a photograph of the temporary tunnel face, in which the lowest banks of lime-sandstone of the alternating sequence, which are also referred to as "main sandstone", are clearly recognizable. The runway of the airport is founded on cohesive layers (Fig. 6.4) consisting mainly of Filder loam and Lias α residual clay. To the west of the urban railway alignment a lake was formerly located in the area of today's runway, flowing out into a creek running towards the east. In the course of the construction of Stuttgart airport, the area of the lake and the creek was filled up. The existing partially soft and settlement-sensitive valley deposits remained under the fill in the process. The fill and the valley deposits are locally several meters thick (Fig. 6.6). Fig. 6.7 shows the structural model (see Chapter 2.5.1) derived for the ground in the area of the tunnel section driven by underground construction. The discontinuity fabric is characterized by an orthogonal system of horizontal bedding parallel discontinuities and steep to vertically dipping joints. Unlike to the banks of lime-sandstone, the bedding parallel discontinuities and joints in the mudstone layers are mostly closed or filled with clay and only vaguely recognizable. In the mudstones of the Rät and the Knollenmergel slickensides exist, dipping at 20 to 40° and striking in all directions (see Chapter 4.1.3). The soil and rock mechanical parameters given in Table 6.1 were specified on the basis of the results of laboratory and in-situ tests as well as experience gained from projects in comparable ground conditions (see Chapter 4.1). For the rock layers of the Lias α formation encountered in the area of the mined tunnel, a transversely isotropic elastic stress-strain behavior, described by 5 independent elastic constants (Wittke, 2000), was assumed for loading below the strength. A further characteristic of the Lias α layers are the low shear strengths on the bedding parallel discontinuities and the joints.

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- 298 -

Fig. 6.7:

Structural model of the ground in the area of the mined tunnel section (lot 601)

The parameters given in Table 6.1 should be interpreted as characteristic parameters according to DIN 4020 (1990) and were taken as a basis for the stability analyses of the mined tunnel.

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- 299 -

Layer

Deformability E = 15 MN/m2 ν = 0.4

c' = 25 kN/m

E1 = 1500 MN/m2 E2 = 750 MN/m2 G2 = 200 MN/m

Permeability

ϕ' = 25°

Overlying strata

Alternating sequence of mudstone and limesandstone

Strength

2

ν1 = 0.25 ν2 = 0.2

kf ≤ 10-8 m/s

2

Bedding B: ϕB = 20°

Horizontally:

cB = 80 kN/m2

kfH = 5 ⋅ 10-5 m/s

Joints J1, J2:

Vertically:

ϕJ = 35°

kfV = 10-6 m/s

cJ = 40 kN/m2

E1 = 1000 MN/m2 E2 = 500 MN/m2

Mudstone with single layers of G2 = 200 MN/m2 lime-sandstone ν1 = 0.25 ν2 = 0.2 Rät and leached zone of the Knollenmergel

Knollenmergel, unweathered

Table 6.1:

E = 150 MN/m2 ν = 0.3

E = 1000 MN/m2 ν = 0.25

Bedding B: ϕB = 20°, cB = 0

kf = 10-7 m/s

Joints J1, J2: ϕJ = 35°, cJ = 0 Discontinuities: ϕD = 17.5°

kf = 5 ⋅ 10-7 m/s

cD = 10 kN/m2 Slickensides: ϕS = 17.5°

kf ≤ 10-8 m/s

cS = 10 kN/m2

Characteristic soil and rock mechanical parameters

The results of in-situ tests and measurements on different structures in the area of Stuttgart have shown that increased horizontal in-situ stresses exist in the Lias α (Grüter, 1988; Wittke, 1990; Wittke, 1991). According to these results, in addition to the horizontal stresses resulting from the dead weight, horizontal stresses of ΔσH = 1 to 2 MN/m2 exist in the unweathered mudstone layers and horizontal stresses of ΔσH = 0.5 to 1.0 MN/m2 exist in the alternating sequence (Böttcher et al., 1998). The magnitude of these stresses was confirmed by the results of the stress measurements by the overcoring method (Kiehl and Pahl, 1990) carried out in the course of the project presented here. Horizontal in-situ WBI-PRINT 5

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- 300 stresses of ΔσH = 0.5 to 1.7 MN/m2 were derived from the results of these stress measurements. The permeability tests carried out as part of the exploration showed that the mudstone layers have a much lower water permeability than the lime-sandstone banks. Accordingly, the alternating sequence is inhomogeneous with respect to its permeability. However, since the tunnel diameter is large compared to the thicknesses and the spacing of the layers of the alternating sequence, the alternating sequence can be overall considered approximately homogeneous, if the different permeabilities of the layers is taken into account by introducing an anisotropy (Wittke, 2000). The horizontal permeability of kfH = 5 · 10-5 m/s is determined here by the banks of lime-sandstone, whereas the vertical permeability of kfV = 10-6 m/s is due to the mudstone (Table 6.1). The overlying strata, the Lias α layers consisting predominantly of mudstone, the Rät and the Knollenmergel have a much smaller water permeability than the alternating sequence (Table 6.1). In the northern part of the airport the groundwater of the alternating sequence is mostly artesian. The water table is encountered within the overlying strata (see Fig. 6.2 and 6.4). In the adjacent section up to the southern limit of the airport the groundwater table lies within the alternating sequence. A further section with locally artesian groundwater follows (see Fig. 6.2). The mined tunnel section of lot 601 is thus located almost over its entire length completely underneath the groundwater table. The maximum height of the water table above the tunnel's invert is reached in the area of the airport runway with approx. 26 to 27 m (see Fig. 6.4). 6.1.4

Fundamentals of the design

During the cut-and-cover construction of the first 400 m of the tunnel (lot 92) the groundwater was lowered to below the construction pit's invert using drawdown wells. Measurements of the groundwater level in the airport and specially in the runway area showed that the drawdown cone had a range of approx. 450 m (Fig. 6.8). To avoid subsidence in the airport area, it became necessary to recharge the groundwater using injection wells (Erichsen and Tegelkamp, 1998). WBI-PRINT 5

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- 301 -

Fig. 6.8:

Advance construction (lot 92), site plan with drawdown cone

Especially the soft valley deposits in the runway area (see Fig. 6.6) are very sensitive to settlements. Especially here, but also in other areas, a lowering of the groundwater table as a consequence of the tunnel heading would lead to subsidence due to loss of the hydrostatic uplift. The FSG therefore demanded that no groundwater lowering must occur during the underground tunneling. Three-dimensional, transient seepage flow analyses were therefore carried out by WBI in the early stages of the project to investigate the influence of the tunnel heading on the groundwater condiWBI-PRINT 5

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- 302 tions. The program system used for this purpose was developed by WBI (Erichsen, 1994). It is described in detail in Wittke (2000). The assumptions made and the results of the analyses are described and explained in detail in Wittke (2000) as well, and also in Wittke-Gattermann and Wittke (1997). The permeability of the layers consisting predominantly of mudstone (see Chapter 6.1.3), in which the tunnel is located in the area of the undercrossing of the airport runway, as well as the permeability of the shotcrete membrane of the tunnel were varied in the analyses.

Fig. 6.9:

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Analyzed drawdown curves of the groundwater table due to the heading, steady state, vertical section through the tunnel axis

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- 303 Fig. 6.9 shows the drawdown of the groundwater table determined for the steady state in a vertical section through the tunnel axis. For a low permeability of the mudstone layers of kf = 10-7 m/s and an impermeable shotcrete membrane, the groundwater drawdown of < 5 cm can be neglected. An increase in the permeability of the mudstone to kf = 10-6 m/s already leads to a groundwater drawdown of 0.5 m. An increased permeability of the shotcrete membrane yields a considerably increased groundwater drawdown even for a value of kf of 10-7 m/s for the mudstone. The steady state is always reached within a period of time which is short in relation to the construction time (Wittke-Gattermann and Wittke, 1997; Wittke, 2000; Tegelkamp et al., 2000). Corresponding to the results of the groundwater modeling analyses and to the demand by FSG that groundwater lowering must not occur during the construction of the tunnel, the tunnel had to be supported during the heading by a shotcrete membrane with a low water permeability. The membrane had to be designed to withstand the water pressure. An alkali-free dry-mix shotcrete with spray cement as bonding agent was used. A statically favorable circular profile was selected for the tunnel's cross-section (see Fig. 6.3). The water pressure taken into account as well as the loads resulting from the rock mass pressure, which are essentially determined by the increased horizontal in-situ stresses in the mudstone layers, lead to a required thickness of the reinforced shotcrete membrane of 30 to 35 cm. An advancing crown excavation was not a reasonable option under the given conditions, because on the one hand an open invert over great lengths could not be permitted since this would lead to a groundwater lowering, and on the other hand a watertight support of the temporary crown invert could not be designed to withstand the water pressure with economically justifiable expense. A fullface heading with a stepped tunnel face was therefore provided for in order to achieve as soon as possible a circular cross-section with closed invert. This was important for statical reasons as well as under the aspect of watertightness, since this design enables water to enter during the tunneling only through the temporary tunnel face and thus only through a comparatively small cross sectional area. This design was also taken as a basis for the three-dimensional seepage flow analyses.

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- 304 6.1.5

Excavation and support

Fig. 6.10 specified nantly of crown and

shows the sequence of excavation and means of support for the tunnel heading in the layers consisting predomimudstone. The full-face excavation was subdivided into bench/invert excavation.

Fig. 6.10:

Standard heading in the mudstone, excavation and support

The tunnel cross-section was excavated using tunnel excavators, which were additionally equipped with a heavy hydraulic chisel. Thicker banks of lime-sandstone were loosened by blasting. The round lengths ranged between 1 and 1.5 m in the crown and between 2 and 3 m in the bench/invert (Fig. 6.10). The support was closed at the invert after 4 to 6 m.

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- 305 In addition to the reinforced shotcrete support of the excavation profile, a systematic anchoring with 3 m long SN-anchors was carried out locally. Steel sets were placed at a spacing of 1 m (Fig. 6.10). The advancing support using spiles, included in the design for more unfavorable rock mass sections, could be completely dispensed with. In the deep tunnel section, in which the tunnel is located completely within the low-permeability layers consisting predominantly of mudstone, no further measures apart from the lowpermeability shotcrete membrane were taken to maintain the groundwater table. In the northern tunnel section, in which the ascending tunnel cuts into the strongly water-bearing layers of the alternating sequence, an advance sealing of water-bearing discontinuities was required to prevent a strong inflow of water through the tunnel face. To this end cement grouting was carried out through boreholes drilled from the tunnel face with an average length of 15 m (Fig. 6.11). Cement based suspensions with water cement ratios of 0.7 to 2 were used for the grouting.

Fig. 6.11:

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Sealing by advance cement grouting of the alternating sequence

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- 306 The advance grouting was carried out in sections in three working steps: 1.

Sealing of the temporary tunnel face with shotcrete.

2.

Construction of a grouting umbrella above the tunnel roof.

3.

Construction of a transverse bulkhead in the area of the alternating sequence to seal off the area to be excavated against groundwater flow in longitudinal tunnel direction.

The grouting boreholes were sealed against the borehole head with fabric packers (geotextiles) filled with cement based suspension. During the grouting works the heading had to be interrupted for approx. 2 weeks each time. By the advance sealing of the alternating sequence stronger inflow of water through the open tunnel face could be prevented to a large extent. Only when the grouting boreholes were drilled (Fig. 6.12), some inflow rates occurred for a short time due to the construction process (Tegelkamp et al., 2000).

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- 307 6.1.6

Stability analyses for the design of the shotcrete support

For the dimensioning of the shotcrete support, two- and threedimensional FE-analyses were carried out using the program system FEST03 (Wittke, 2000). These analyses were based on the characteristic parameters given in Table 6.1. In the Lias α increased horizontal in-situ stresses were taken into account. For the layers consisting predominantly of mudstone ΔσH = 1.5 MN/m2 was specified. The alternating sequence was assigned a value of ΔσH = 0.5 MN/m2. Further analyses were carried out with no additional horizontal stresses assumed, as well as with horizontal stresses in the mudstone increased to ΔσH = 2 MN/m2. In Fig. 6.13, the location of the analysis cross-sections investigated in the design analyses is given in the geological longitudinal section.

Fig. 6.13:

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Location of the analysis cross-sections

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- 308 Fig. 6.14 shows exemplarily the computation section, the FE-mesh, the boundary conditions and the ground profile of a threedimensional analysis for the area of the airport runway (analysis cross-section 4A, see Fig. 6.13) with an overburden of 21 m.

Fig. 6.14:

Analysis cross-section 4A, FE-mesh, boundary conditions and ground profile for three-dimensional analyses

In Fig. 6.15 the computation steps chosen for the simulation of the tunnel heading by the "step-by-step" method (Wittke, 2000) are illustrated.

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- 309 -

Fig. 6.15:

Analysis cross-section 4A, computation steps

In the first two computation steps the in-situ state of stress is determined, taking into account the dead weight of the rock mass and the increased horizontal stresses in the Lias α. To simulate the different horizontal stresses ΔσH in the alternating sequence and in the mudstone, the nodes lying on the boundary planes WBI-PRINT 5

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- 310 x = 40 m and y = 65 m (see Fig. 6.14) were assigned horizontal displacements in x- and y-direction corresponding to the respective horizontal stresses ΔσH. To prevent shear stresses from being transferred across the boundary between the alternating sequence and the mudstone due to the different horizontal displacements, an interface layer is arranged between the two layers. In the 1st computation step, the dead weight is only taken into account for the alternating sequence and the mudstone. The overlying strata, the Rät and the Knollenmergel are assumed to be weightless. In the 2nd computation step the latter two layers are replaced by materials with the same mechanical parameters, however with their dead weight. Since the new materials are installed stress-free in the already deformed corresponding elements (Wittke, 2000) and the horizontal displacements remain unchanged in the 2nd computation step, only the stresses due to dead weight but not the increased horizontal stress ΔσH are effective in these layers. In the 3rd computation step, the excavation and shotcrete support of the crown and the bench and invert, trailing by 2 m, are simulated. Computation steps 4 to 16 include the simulation of the heading according to the "step-by-step" method. In each computation step, the crown and the trailing bench and invert are both advanced by 2 m (Fig. 6.15). Fig. 6.16 and 6.17 show the stress resultants in the shotcrete membrane determined in this analysis. The representation of the maximum stress resultants along the tunnel in Fig. 6.16 illustrates that the loading of the shotcrete membrane continuously develops with increasing distance from the tunnel face, until the maximum values are reached at a distance of approx. twice the tunnel's diameter. This cross-section is designated the dimensioning section, because the stress resultants determined here have to be considered decisive for the design of the shotcrete membrane (see Chapter 5.3). The stress resultants M, N and S in the dimensioning section are shown in Fig. 6.17. As expected, great compressive normal thrust exist. Because of the high horizontal stresses the maximum values result in the roof and invert areas. The bending moments are relatively small due to the favorable geometry of the cross-section (circle), and the shear forces follow corresponding to the moment distribution.

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- 311 -

Fig. 6.16:

Analysis cross-section 4A, stress resultants vs. distance from the tunnel face, 16th computation step

Fig. 6.17:

Analysis cross-section 4A, stress resultants in the shotcrete membrane, dimensioning section, 16th computation step

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- 312 In addition to the three-dimensional analyses, two-dimensional analyses ones were carried out to investigate the influence of a variation of the parameters and of the horizontal in-situ stresses in the Lias α on the loading of the shotcrete membrane. To model the influence of the displacements occurring in the tunnel face area before the shotcrete membrane is installed, either the computed stress resultants in the shotcrete membrane were reduced in the two-dimensional analyses, or a preceding stress relief was simulated (Wittke, 2000). For the section in which the tunnel's cross-section is located entirely within the alternating sequence, a calibration of the results of two-dimensional analyses on the basis of the results of a corresponding three-dimensional analysis resulted in a stress relief factor according to (4.1) (see Chapter 4.1) of av = 0.35. The load case water pressure acting on the shotcrete membrane was investigated in separate two-dimensional FE-analyses. Because the water pressure builds up only at a certain distance from the tunnel face, it suffices to superpose the stress resultants ensueing from the rock mass pressure and the water pressure and to design the shotcrete membrane on the basis of the stress resultants thus obtained. According to the results of the stability analyses, the shotcrete membrane of the standard tunnel sections could be designed with a thickness of 35 cm for a factor of safety of 1.7 without requiring more than the minimum reinforcement (inside and outside steel fabric mats Q285). The specified limits for the tunneling-induced subsidence and differential subsidence were not exceeded either. 6.1.7

Monitoring

The construction was accompanied by an extensive monitoring program aimed at compliance with all requirements as well as control and optimization of the heading works. This program was coordinated in mutual agreement with the FSG. The monitoring program included the monitoring of the groundwater level as well as displacement measurements at the ground surface and in the tunnel. Further, stress measurements were carried out in the shotcrete membrane. For the monitoring of the groundwater level, the system of observation wells already existing on the WBI-PRINT 5

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- 313 airport from previous construction and supplemented during the exploration works for the urban railway tunnel could be used. Fig. 6.18 shows the location of the measuring cross-sections and observation wells as well as the heading location in November 1999. From mid-February to mid-March 2000 the airport runway was undercrossed by the excavation south coming from the northern starting shaft. The excavation north emanating from the southern starting shaft cut through to the excavation south on April 22, 2000 (Fig. 6.18). The tunnel section excavated between November 1999 and April 2000 is specifically marked in Fig. 6.18. The cutthrough from lot 601 to the already existing lot 92, which have been constructed before by the cut-and-cover method, was carried out on June 5, 2000.

Fig. 6.18:

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Stuttgart urban railway (lot 601), monitoring program

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- 314 In Fig. 6.19 the hydrographs of the observation wells in the area of the airport runway are shown for the time between November 1999 and June 2000. For the winter months a general rise of the groundwater level is apparent at all observation wells, which declines again in the spring of 2000. A reaction of the groundwater level to the tunnel heading cannot be recognized. The measured variations in water level can be attributed to the natural course of the groundwater flow and are not related to the tunneling.

Fig. 6.19:

Observation well hydrographs during the undercrossing of the runway of Stuttgart airport

A temporary drop of the groundwater table by a maximum of 2 to 3 m however occurred in the boreholes located in the area of the northern apron beside the rescue adit (Fig. 6.18). This decrease correlates in time with the grouting works carried out in the tunnel section located within the alternating sequence (see Fig. 6.11). With distances of up to 100 m to the observation wells and in view of the limited discharge in the tunnel through the grout-

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- 315 ing boreholes of approx. 1 l/s, these observations demonstrate the sensitivity of the aquifer in the alternating sequence. Overall it must be stated, however, that with the advance grouting a large-scale, long-term groundwater lowering could be avoided. Even in the northern apron area, subsidence due to interference with the groundwater did not occur.

Fig. 6.20:

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Results of the displacement measurements in MC 6

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- 316 In Fig. 6.20, the results of the displacement measurements are illustrated exemplarily for measuring cross-section MC 6 (see Fig. 6.18). Here, a tunneling-induced subsidence of 4 mm at the most was measured at the ground surface above the tunnel roof. Along the entire tunnel, predominantly a subsidence of between 4 and 7 mm was determined above the tunnel axis. Because of these comparatively low values, which were markedly smaller than the admissible subsidence of 15 mm, the airport facilities as well as the buildings of Bernhausen, neighboring to the south, could be undercrossed without damage. Comparatively large horizontal displacements in the ground of up to 9 mm were measured at the level of the tunnel (see Fig. 6.20). This displacement distribution is typical for this construction project and has to be attributed to the increased horizontal insitu stresses in the Lias α. 6.1.8

Interpretation of the monitoring results

During the construction the displacements measured in the tunnel as well as the tangential stresses measured in the shotcrete membrane were compared to the values computed in the design analyses, which are based on the characteristic parameters (see Table 6.1). The resulting differences can be attributed to the conservative assumptions made in the design analyses, mainly on the high horizontal stresses ΔσH assumed in the Lias α (see Fig. 6.14). To be able to back-analyze the measured results, parameter studies were carried out. The following parameters, which have a large influence on the loading of the shotcrete membrane, were varied: -

Horizontal stresses in the mudstone (ΔσH),

-

deformability of the mudstone layers (E1, E2),

-

shear strength on the bedding parallel discontinuities in the mudstone (ϕB),

-

deformability of the shotcrete membrane (ESC).

The best agreement with the monitoring results was achieved in a comparative analysis, in which a Young's modulus of the shotcrete support of ESC = 2000 MN/m2 was selected as opposed to the WBI-PRINT 5

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- 317 Young's modulus taken as a basis for the design analyses (ESC = 15000 MN/m2), and in which the horizontal stress in the mudstone was reduced from 1.5 MN/m2 (design analysis) to 1.0 MN/m2 (Table 6.2).

Design analyses

Comparative analysis

Shotcrete

ESC = 15000 MN/m2

ESC = 2000 MN/m2

Mudstone with single layers of lime-sandstone

ΔσH = 1.5 MN/m2

ΔσH = 1.0 MN/m2

Table 6.2:

Comparison analysis for the interpretation of monitoring results: Differences to the construction design analyses

The low modulus of ESC = 2000 MN/m2 accounts for the deformability development as well as the creep properties of the shotcrete. Because of the short round lengths and the early closing of the support ring, the shotcrete is loaded in the present case at a very young age, in which it still possesses a low strength and a high deformability as well as a high creep potential (see Chapter 2.1). The value of ESC = 15000 MN/m2 used in the design analyses therefore represents a conservative assumption, by which the loading of the shotcrete membrane is overestimated. A good agreement with the measured values can however only be achieved if the water pressure and the seepage forces are accounted for that act on the ground and the shotcrete membrane during construction. Since the water permeability of the alternating sequence is high compared to the one of the mudstone, there is no significant decrease of piezometric heads within the alternating sequence. Almost the entire decrease of piezometric heads thus occurs in the mudstone. Fig. 6.21 shows qualitatively the flow net with its equipotential lines and streamlines which represents the groundwater flow towards the tunnel. Thus, as an approximation, the entire water pressure pw resulting from the difference in elevation between the groundwater table and the top boundary of the mudstone layer acts on the mudstone layer. The loading of the shotcrete membrane due to the groundwater flow towards the tunnel can thereWBI-PRINT 5

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- 318 fore be replaced as an approximation by a surface load pW acting on the mudstone layer (Fig. 6.21).

Fig. 6.21:

Qualitative distribution of equipotential lines and streamlines in the mudstone and replacement of the water pressure by an equivalent load

Fig. 6.22 illustrates the determination of the stresses and deformations resulting from the flow towards the tunnel in two computation steps. In the 1st computation step the surcharge from the water pressure pw is applied onto the mudstone layer. In the 2nd computation step, the excavation and the installation of the shotcrete membrane are simulated. The rock mass and the shotcrete are assumed weightless here. The deformations and the loading of the shotcrete membrane can therefore be determined approximately by superposing the deformations and stresses computed for the load cases "rock mass pressure" and "water pressure on the mudstone layer".

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- 319 -

Fig. 6.22:

Computation steps to determine the stresses and deformations resulting from the flow towards the tunnel

Fig. 6.23 shows the comparison between the measured and the computed values of the displacements of the tunnel contour and the tangential stresses in the shotcrete. The comparison between measured and computed displacements is based on so-called "representative displacements" determined as the mean values of the displacements measured in different crosssections. It can be seen that the representative displacements can be captured by the analyses if the water pressure is taken into account (Fig. 6.23). Around the circumference of the shotcrete membrane differing tangential stresses were measured. The very low stresses measured in the roof area (see Fig. 6.23, bottom left), are not considered representative. In the other areas the measured tangential stresses are captured by the analysis. During the heading of the rescue adit as well, geotechnical monitoring was carried out (see Fig. 6.18). The representative displacements of the tunnel contour and the tangential stresses in the shotcrete measured in the rescue adit are pictured in Fig. 6.24 (left). The displacements are somewhat smaller than the representative displacements in the urban railway tunnel. The measWBI-PRINT 5

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- 320 ured tangential stresses are markedly higher in the roof than the tangential stresses shown for in the shotcrete membrane of the urban railway tunnel.

Fig. 6.23:

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Comparison of the measured displacements and tangential stresses in the runway area with the corresponding values computed with and without consideration of the water pressure

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- 321 The monitoring results obtained in the rescue adit can be backanalyzed with the same parameters as the monitoring results in the urban railway tunnel. In Fig. 6.24 the monitoring results are compared with the values computed with and without consideration of the water pressure. If the water pressure is taken into account, the displacements as well as the tangential stresses agree well.

Fig. 6.24:

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Comparison of the measured displacements and tangential stresses in the rescue adit with the corresponding values computed with and without consideration of the water pressure WBI GmbH, Henricistr. 50, 52072 Aachen, Germany

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- 322 6.1.9

Conclusions

Stuttgart airport was to be undercrossed by a tunnel with low to medium overburden. The water table is located above the tunnel roof, and settlement-sensitive layers are locally encountered at the ground surface. Because of the demands with respect to the admissible subsidence, groundwater lowering had to be avoided during tunneling. Accordingly, the shotcrete membrane had to be dimensioned for the water pressure. It had to be further taken into account for the design of the shotcrete membrane that the Lias α, in which the tunnel cross-section is located, shows increased horizontal in-situ stresses and low strengths on the bedding parallel discontinuities. This task was solved by the following measures: -

Full-face excavation with a stepped tunnel face and an early closing of the support ring,

-

construction of a low-permeability shotcrete membrane with high strength,

-

specification of a circular profile which is statically favorable for the design of the shotcrete membrane,

-

sealing of water-bearing discontinuities in the alternating sequence of mudstone and lime-sandstone by advance grouting.

By these measures, the tunnel could be excavated with very small subsidence of the ground surface. There was no interference with air traffic at any time. It further showed that the transient seepage flow analyses and two- and three-dimensional stability analyses carried out in the course of this project with the program systems HYDOPO and FEST03 (Wittke, 2000) represented an essential contribution towards the design, the statics and the specification of the excavation and support measures.

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- 323 6.2

Freeway tunnel "Berg Bock" near Suhl, Germany

6.2.1

Introduction

In the course of the new freeway (Autobahn) A71 connecting Erfurt and Schweinfurt (Kleffner, 2000), the "Berg Bock" freeway tunnel was excavated between the exit Suhl/North and the intersection Suhl (Fig. 6.25). The two tubes each 2700 m in length from north to south pass through the Suhl granite, the base sediments and porphyrite as well as, after passing the southern edge fault, the layers of the Lower Triassic sandstone (Fig. 6.26).

Fig. 6.25:

Tunnel Berg Bock, site plan

The two tunnel tubes were driven mainly by the full-face excavation method within a construction time of approx. 10 month. With four tunnel faces, maximum performances of approx. 30 m/d were attained. 6.2.2

Structure

Two tunnel tubes, eastern tube and western tube each 2700 m in length were constructed (Fig. 6.27). The spacing of axis of both tubes ranges between approx. 23 m at the northern portal and approx. 30 m at the southern portal. Each tube comprises two traffic lanes each 3.75 m wide and two emergency sidewalks (Fig. 6.28). The excavated cross-section of the tunnel ranges between 80 m2 and 100 m2 (Fig. 6.28 and 6.29). The alignment dips continuously towards the southern portal with 1.1 %. The maximum overburden amounts to approx. 190 m (Fig. 6.26).

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- 324 -

Fig. 6.26: WBI-PRINT 5

Tunnel Berg Bock, geological longitudinal section with analysis cross-sections WBI GmbH, Henricistr. 50, 52072 Aachen, Germany

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- 325 -

Fig. 6.27:

Safety conception

Fig. 6.28:

Tunnel cross-section with closed invert in weathered rock mass, portal zones

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- 326 -

Fig. 6.29:

Tunnel cross-section with open invert in stable rock mass

The safety conception is based on the German standard for the equipment and the operation of road tunnels (RABT, 1994). This standard was currently revised on the basis of the evaluation of several fire accidents in tunnels happened in recent years. Thus, additional, supplementary demands on the safety conception had to be fulfilled at short notice in the final planning for the tunnel Berg Bock (Schmidtmann and Erichsen, 2001). The safety conception comprises nine connection tunnels between the tubes at distances of ≤ 300 m (Fig. 6.27). The connection tunnels are equipped with emergency call niches and fire protection locks. Three breakdown bays are furthermore situated in each tunnel tube with distances of ≤ 600 m. Three connection tunnels located close to the breakdown bays are passable for rescue vehicles. Besides that, emergency call niches and niches provided with WBI-PRINT 5

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- 327 fire extinguishing devices are located in each tube at distances of ≤ 150 m. Fire emergency lights are installed on one side of each tube at distances of 24 m (Table 6.3). Installation connection tunnels breakdown bays emergency call niches hydrant niches electrical niches niches for drainage flushing shafts on both sides of the lanes road sign displays jet fans fire emergency lighting Table 6.3:

Distance

Quantity

≤ 300 m ≤ 600 m ≤ 150 m ≤ 150 m approx. 180 m 50 - 80 m

9 2 2 2 2 2

approx. 300 m approx. 300 m 24 m

2 x 9 2 x 9 2 x 110

x 3 x 19 x 19 x 14 x 104

Installations for operational safety

In the areas of the portals weathered rock of the Lower Triassic sandstone and completely weathered granite, respectively, were encountered (Fig. 6.26). In these sections the tunnel tubes were carried out with an approx. 11.6 m wide and approx. 9.9 m high mouth-shaped cross-section with a closed invert and an excavated cross-section of approx. 100 m². The thickness of the shotcrete membrane (concrete grade B 25 corresponding to C 20/25) is 25 cm. The interior lining (concrete grade B 35 corresponding to C 30/37) is 40 cm thick (Fig. 6.28). In the remaining sections, located in stable rock mass, a crosssection of approx. 11.4 m width and approx. 8.4 m height with an open invert and an excavated cross-section of approx. 80 m² was carried out. The tunnel walls mainly were supported by fiber reinforced shotcrete with a concrete grade of B 45 corresponding to C 35/45 and a thickness of t = 15 cm. Locally reinforced shotcrete with a concrete grade of B 25 and a thickness of t = 25 cm was installed. The shotcrete membrane was carried out with radii of R = 5.9 m, R = 4.3 m and R = 8.4 m. The 30 cm thick interior lining with a concrete grade of B 35 was founded on 50 cm high concrete shoulders with the same grade (Fig. 6.29). Between the interior lining and the shotcrete membrane a non-woven synthetic and as a sealing a 2 mm thick foil were installed (Fig.

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- 328 6.28 and 6.29). In the tunnel sections with closed invert between the interior lining and the shotcrete membrane a separation foil was installed at the invert (Fig. 6.28). Thus, in this area between interior lining and shotcrete membrane no tensile and shear forces can be transferred. Both tubes were carried out as fully drained road tunnels. The seepage water was drained off by two lateral drainage pipes in the area of the lower sidewalls and the gradient of the tunnel of 1.1 % (Fig. 6.28 and 6.29). Flushing shafts for the washing of the drainage ducts are provided at distances of 50 to 80 m on both sides of the lanes. Fig. 6.30 shows the northern portal of the Berg Bock tunnel.

Fig. 6.30: 6.2.3

Tunnel Berg Bock, northern portal

Ground and groundwater conditions

The ground profile is shown in Fig. 6.26 in a geotechnical longitudinal section. The ground conditions and the overburden height are approximately the same for both tunnel tubes.

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- 329 In the starting area at the northern portal, the granite is predominantly decomposed. The tunnel cross-section is alternatingly located here in hard, mostly strongly jointed granite and in completely decomposed granite. In the further course of the tunnel, unweathered, mostly very hard granite was encountered. The rock is streaked with a multitude of veins of different thickness. The rock mass is compact, with a joint spacing of more than 1 m, to narrowly jointed, and in some areas it is traversed by joints with large extent.

Fig. 6.31:

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View of the working face in the Lower Triassic sandstone WBI GmbH, Henricistr. 50, 52072 Aachen, Germany

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- 330 The granite is followed by the base sediments (sand-, mud- and siltstone). In the area of the highest cover, the tunnel is located in the porphyrite, which is predominantly hard to very hard and slightly to narrowly jointed. After that, the tunnel crosses the southern edge fault, which consists of water-bearing, strongly decomposed and mylonized zones with a thickness of a few decimeters. In the last tunnel section, the tunnel is located in the Lower Triassic sandstone (Fig. 6.26). The layers of the Lower Triassic sandstone and the base sediments consist of an alternating sequence of sandstone and mudstone. The widely persistent bedding parallel discontinuities are mostly horizontal in the Lower Triassic sandstone (Fig. 6.31) and predominantly steeply inclined in the base sediments. The joints normal to the bedding usually end at the bedding parallel discontinuties. The groundwater table is located up to 180 m above the tunnel roof. 6.2.4

Excavation and support

Heading in completely weathered granite In the area of the northern portal (Fig. 6.26), the following supporting measures were carried out: -

Preceding drainage borings,

-

preceding pipe umbrella,

-

crown heading with closed invert,

-

tunnel face support core.

The round lengths ranged between 0.75 and 1.0 m. A heading performance of approx. 1 m/d was achieved in the area of the completely weathered granite.

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- 331 Heading in granite and porphyrite Stable rock mass jointed to varying degrees was encountered in the unweathered granite and porphyrite. The rock mass conditions allowed a full-face blasting excavation of the tunnels in these areas (Fig. 6.32). Mainly excavation class A0 was applied with a round length of up to 3.5 m (Fig. 6.33). In part, an even greater round length was chosen during construction. By the use of steel fiber shotcrete (t = 15 cm) with a fiber content of 40 kg/m3 for the support of the excavation contour, the expenses for the support were kept low (Fig. 6.33). Anchors were installed as required depending on the jointing.

Fig. 6.32:

Loading of blastholes during the full-face excavation in granite

With this optimized heading scheme, approx. 2 to 3 rounds could be achieved per day and tunnel face. The heading performance thus amounted to up to 10 m/d per tunnel face. For the 4 tunnel faces, maximum performances of ≥ 30 m/d were reached.

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- 332 -

Fig. 6.33:

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Excavation and support in granite and porphyrite, excavation class A0 (full-face excavation)

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- 333 Heading in Lower Triassic sandstone and base sediments In these layers, a crown heading with closed invert and trailing bench excavation was carried out. In order to support the tunnel face and the working area against the dropping of so-called "coffin lids", if the bedding was approximately horizontal, preceding spiles were installed. In areas with the bedding dipping moderately steeply towards the tunnel, a support core was left standing to support the tunnel face. A performance of approx. 4 to 5 m per day and tunnel face was achieved with this heading technique. Heading schedule Both tunnel tubes were excavated within some 10 months each, corresponding to an average heading performance of approx. 9 m per day and tube.

Fig. 6.34: WBI-PRINT 5

Heading progress of the eastern tube WBI GmbH, Henricistr. 50, 52072 Aachen, Germany

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- 334 Fig. 6.34 shows exemplarily the heading progress over time for the eastern tube (see Fig. 6.27). The western tube was excavated in parallel with the eastern tube by so-called opposite heading. The eastern tube was successfully cut through on January 19, 2001. The cut-through of the western tube and thus the heading of the entire tunnel was celebrated on February 2, 2001. 6.2.5

Stability analyses for the stages of construction and design of the shotcrete support

To analyze the stability during construction and to design the shotcrete support, two-dimensional FE-analyses were carried out using the program system FEST03 (Wittke, 2000). A total of eight analysis cross-sections shown in Fig. 6.26 (AC 1 to AC 6, AC 2a and AC 6a) were investigated. Fig. 6.35 shows exemplarily the FE-mesh, the boundary conditions, the ground profile and the parameters for analysis cross-section AC 2, which was used to analyze the stability of the tunnel tubes in the granite. The overburden amounts to 130 m (see Fig. 6.26). Due to symmetry, only one tunnel tube is modeled as a simplification. The plane of symmetry lies at the center of the rock pillar between the two tunnel tubes. Such a model is to be considered conservative with respect to the loading of the shotcrete membrane, because this way a simultaneous excavation of both tunnel tubes is simulated. The specified computation section consists of a 1 m thick slice with a width of 75 m (x-direction). The height amounts to 157 m. The FE-mesh consists of 2135 isoparametric elements with 13004 nodes. As boundary conditions, vertically sliding supports are introduced for the nodes of the vertical boundary planes (x = 0 and x = 75 m). For the nodes of the lower boundary plane (z = 0) horizontally sliding supports are specified (Fig. 6.35). All nodes are assumed fixed in y-direction. The tunnel cross-section is entirely located in the granite. Below the ground surface, two 11 m thick layers of decomposed granite and slightly weathered granite, respectively, are simulated. The assumed mechanical parameters for these layers are also given in Fig. 6.35. The loading modulus of the unweathered granite was assumed as EL = 15000 MN/m2. Underneath the tunnel's invert an unWBI-PRINT 5

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- 335 loading modulus was specified in the analyses which at 30,000 MN/m2 was twice as high as the loading modulus (Fig. 6.36).

Fig. 6.35:

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Analysis cross-section AC 2 (shotcrete support), FEmesh, boundary conditions, ground profile and parameters WBI GmbH, Henricistr. 50, 52072 Aachen, Germany

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- 336 -

Fig. 6.36:

Analysis cross-section AC 2 (shotcrete support), FEmesh, detail

The orientations of the joints which are present in the granite were not clearly determined. Thus randomly distributed joint orientations were assumed. The rock mass therefore was modeled with

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- 337 an isotropic strength with shear parameters of ϕJ = 45° and cJ = 50 kN/m2 (Fig. 6.35).

Fig. 6.37:

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Analysis cross-section AC 2 (shotcrete support), computation steps

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- 338 -

Fig. 6.38:

Analysis cross-section AC 2, principal normal stresses and rock mass areas in which strength is exceeded, 3rd computation step

For the shotcrete, a statically effective Young's modulus of 15000 MN/m2 was assumed taking into account the hardening during the application of the load (Fig. 6.36). In Fig. 6.37 the computation steps are outlined. In the 1st computation step, the state of stress and deformation resulting from WBI-PRINT 5

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- 339 the dead weight of the ground is determined (primary state). In computation steps 2 and 4, a preceding stress relief is each simulated in those areas of the cross-section, of which the excavation and the shotcrete support are simulated in computation steps 3 and 5, respectively. The stress relief factor according to (4.1) is specified as av = 0.5. Fig. 6.38 shows the principal normal stresses in the rock mass around the excavation after the full-face excavation and the installation of the shotcrete support at the end of the 3rd computation step. The stress redistribution that occurred with the excavation as well as the areas of exceeded strength can be recognized. Although these areas extend around the entire circumference of the excavation, a pronounced arching is apparent. Due to the low deformability of the rock mass (E = 15000 MN/m2, see Fig. 6.35), the shotcrete membrane is only marginally loaded by the rock mass and thus takes on a slightly stabilizing and assisting function only. In Fig. 6.39 the heading-induced displacements computed for the full-face excavation are shown. The total computed roof subsidence amounts to 3.3 mm (3rd – 1st computation step, Fig. 6.39a). In the 2nd computation step (preceding stress relief) the displacements preceding the heading are determined. The shotcrete membrane is therefore only loaded in the 3rd computation step. The displacement of the excavation profile resulting from this loading (3rd – 2nd computation step) is shown in Fig. 6.39b. Thus the computed displacement of the shotcrete membrane amounts to 2.1 mm at the roof.

Fig. 6.39:

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Analysis cross-section AC 2, displacements due to full-face excavation: a) 3rd – 1st computation step; b) 3rd – 2nd computation step WBI GmbH, Henricistr. 50, 52072 Aachen, Germany

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- 340 Fig. 6.40 shows the stress resultants in the shotcrete membrane for the 3rd computation step. An approximate membrane state of stress is computed. The normal thrust is ranging between 400 and 1100 kN/m corresponds to only one tenth of the force resulting from the overburden. This confirms that the shotcrete membrane is only subjected to slight loading because of the high Young's modulus of the rock mass and the arching effect. The dimensioning yields that no reinforcement is statically required for the shotcrete membrane.

Fig. 6.40:

Analysis cross-section AC 2, stress resultants in the shotcrete membrane, 3rd computation step

The excavation of the shoulders (4th and 5th computation step) does not lead to significant changes relative to the 3rd computation step. 6.2.6

Stability analyses for the design of the interior lining

Investigated load cases and load combinations Two-dimensional FE-analyses were carried out for the design of the interior lining as well. Fig. 6.41 shows exemplarily the FE-mesh, the boundary conditions and the parameters specified for the design of the interior lining for analysis cross-section AC 2 (see Fig. 6.26). The computation section consists of a 1 m thick, 75 m wide and 162 m high slice subdivided into 2397 isoparametric elements with 14090 nodes.

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- 341 -

Fig. 6.41:

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Analysis cross-section AC 2 (interior lining), FEmesh, boundary conditions and parameters

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- 342 -

Fig. 6.42:

Analysis cross-section AC 2 (interior lining), FE-mesh, detail

Unlike the stability analysis for the design of the shotcrete support, the decomposed granite and weathered granite (see Fig. 6.35) were not modeled here, since these layers are insignificant for WBI-PRINT 5

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- 343 the loading of the interior lining. A loading modulus of EL = 10000 MN/m2 was assumed for the unweathered granite (Fig. 6.41 and 6.42). This value was derived from a comparison of the displacements measured during heading and the analysis results ("back-analysis", see Chapter 6.2.7). The remaining parameters, the overburden and the specified boundary conditions correspond to those of the stability analysis for the design of the shotcrete support. In Fig. 6.42, a detail of the FE-mesh is shown. The interior lining is modeled with a thickness of 30 cm , the shoulders with a thickness of 50 cm. The seepage water drainage is not discretized. For the design of the interior lining it is assumed that the shotcrete will be decomposed in the course of time and lose its bearing capacity. The assumed parameters for the decomposed shotcrete are given in Fig. 6.42. Since the interior lining is only subjected to significant loads after having reached its final strength, the calculation value for Young's modulus of 34000 MN/m2 commonly used for concrete of grade B35 is assumed (DIN 1045, 1988). The following load cases and load combinations, respectively, were investigated for the design of the interior lining: -

Dead weight of the interior lining (DW)

-

dead weight as before and rock mass pressure (DW + RP)

The seepage water is to be lowered to the invert's level by the lateral seepage water drainages (see Fig. 6.28 and 6.29). Thus there is not any water pressure acting on the interior lining, and no seepage pressure is applied to the rock mass above and closely beside the tunnel's cross-section. Such loadings are therefore not accounted for in the stability analyses for the interior lining. The computation steps of the stability analyses for the design of the interior lining are shown in Fig. 6.43. In the 1st computation step, the primary state in the undisturbed rock mass is computed. A preceding stress relief is simulated in the 2nd computation step with av = 0.5 according to (4.1). WBI-PRINT 5

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- 344 -

Fig. 6.43:

Analysis cross-section AC 2 (interior lining), computation steps

The 3rd computation step includes the full-face excavation of the cross-section and the simultaneous installation of the shotcrete support, which carries the rock mass pressure with a Young's modulus of 15000 MN/m2. WBI-PRINT 5

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- 345 In the 4th computation step, the installation of the interior lining and thus load case DW is simulated. The shotcrete support is still sustainable in this state and therefore able to continue to carry the rock mass pressure. To account for the sealing between the shotcrete membrane and the interior lining, no shear and tensile forces can be transferred in computation steps 4 and 5. This is simulated by insertion of a thin row of elements between the shotcrete membrane and interior lining elements. This row of elements is assigned a stiffness of approx. zero in the 4th and 5th computation step. The opposing nodes of this element row are linked by truss elements (see Fig. 6.42), which can transfer compressive forces, but not tensile forces or shear. In the 5th computation step, the shotcrete is assumed decomposed. As a result, the shotcrete membrane loses its bearing capacity and the interior lining must carry the rock mass pressure in addition to its dead weight. The stress resultants of the interior lining due to dead weight (4th computation step) are shown in Fig. 6.44. If a B35 concrete grade, a lining thickness of 30 cm in the vault and 50 cm at the shoulders, a cover of the reinforcement of d1 = 5.5 cm and safety factors according to DIN 1045 (1988) are assumed, it follows that circumferential reinforcement is not statically required. Due to the small shear forces in load case DW also a shear reinforcement is not needed.

Fig. 6.44:

Analysis cross-section AC 2, stress resultants in the interior lining, load case DW (4th computation step)

In the 5th computation step (load combination DW + RP) the shotcrete membrane is assumed decomposed. Stress redistributions result for this computation step compared to the preceding one. The ground carries a portion of the load previously supported by the WBI-PRINT 5

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- 346 intact shotcrete. On the other hand, a portion of the rock mass pressure carried by the shotcrete membrane before is taken on by the interior lining. This is apparent from the increase of the stress resultants from the 4th to the 5th computation step (Fig. 6.44 and 6.45). Particularly the normal thrust in the vault and the shear force in the shoulders increase markedly as compared to the 4th computation step.

Fig. 6.45:

Analysis cross-section AC 2, stress resultants in the interior lining, load combination DW+RP (5th computation step)

Vault For the tunnel section, which is located in granite and porphyrite, the design of the interior lining yields that reinforcement is not statically required (Fig. 6.46). The interior lining was therefore constructed with plain concrete in this section. Only in the area of the special cross-sections as the breakdown bays, the cross-connections, the niches and the blockouts, a constructive reinforcement was installed (Fig. 6.47). In the portal zones, in the weathered rock mass, in the base sediments and in the Lower Triassic sandstone a reinforcement was statically required, however. In total, it was possible to construct the Berg Bock Tunnel over approx. 50 % of its length with an interior lining made of plain concrete.

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- 347 -

Fig. 6.46:

Analysis cross-section AC 2, statically required reinforcement of the interior lining.

Fig. 6.47:

Not reinforced interior lining, reinforcement in the area of a niche

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- 348 Shoulders For the shoulders, the load combination dead weight and rock mass pressure results in a required shear reinforcement of 7.7 cm2/m2 for analysis cross-section AC 2 (Fig. 6.46). The proof of limitation of crack width according to DIN 1045 (1988) leads to a required reinforcement for the shoulders of 12.88 cm2/m in both, longitudinal and transverse direction. This amount of reinforcement is covered by top and bottom rebars Ø 10 mm spaced at s = 10 cm, to be placed in longitudinal and transverse direction. To cover the required shear reinforcement, steel fabric mats were bent to stirrup cages (Fig. 6.48). The shoulders were reinforced over the entire tunnel length.

Fig. 6.48:

6.2.7

Analysis cross-section AC 2, reinforcement of the shoulders

Monitoring

The heading of the Berg Bock Tunnel was accompanied by a geotechnical monitoring program. In the longitudinal section of Fig. 6.49 the range of the roof subsidence measured after the heading of the entire tunnel was completed is exemplarily shown. WBI-PRINT 5

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- 349 -

Fig. 6.49:

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Range of measured roof subsidence δR, longitudinal section WBI GmbH, Henricistr. 50, 52072 Aachen, Germany

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- 350 The largest subsidence was measured in the portal areas with values between 10 and 50 mm. In those tunnel sections where the cross-section is located in granite or porphyrite, a roof subsidence between 2 and 5 mm was measured. In the base sediments and in the Lower Triassic sandstone the roof subsidence ranges from 5 to 15 mm. As an approximation, the measurements captured only those displacements that occurred after the installation of the shotcrete membrane. Therefore, the measurement results in the granite area must be compared to the computed displacements shown in Fig. 6.39b (3rd – 2nd computation step). A roof subsidence of approx. 2 mm was computed. This analysis is based on a loading modulus of the unweathered granite of EL = 15000 MN/m2 (see Fig. 6.35). A comparative analysis with EL = 10000 MN/m2 yields a roof subsidence of approx. 4 mm. It is thus possible to reproduce the roof subsidence measured in the granite and the porphyrite well in the analyses using a loading modulus of approx. 10000 MN/m2 (see Fig. 6.49). This value was therefore taken as a basis for the stability analyses of the interior lining. 6.2.8

Conclusions

The Berg Bock Tunnel is situated in granite and porphyrite over a length of 2 × 2000 m corresponding to 75 % of its total length. In these sections, the tunnel was headed by full-face excavation using the drill and blast method with comparatively great round lengths and limited support measures. Due to the low deformability and the high strength of the rock mass, the ground was able to carry approx. 90 % of the overburden load, and the means of support only had a slightly assisting function. On this basis it was possible to optimize the heading conception and to excavate both tunnel tubes in a very short time. With the construction of the interior lining using plain concrete over approx. 50 % of the total tunnel length, the costs for the interior lining could be kept low as well. An important tool for the optimization of the heading and the means of support were the FE-analyses. Their results were confirmed by the experience made and the monitoring during the excavation.

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- 351 7.

Heading under the protection of jet grouting columns

7.1

Road tunnel for the federal highway B 9 in Bonn-Bad Godesberg, Germany

7.1.1

Introduction

In the city of Bonn–Bad Godesberg, Germany, the federal highway B 9 was relocated into a tunnel over a length of approx. 1200 m (Fig. 7.1). The tunnel undercrosses buildings as well as tracks of the German Rail (Deutsche Bahn AG) and the city railway. The tunnel cross-section is located in gravel sand. The groundwater table lies above the tunnel's invert. To guarantee the stability and to limit the subsidence due to tunneling, the tunnel was headed by the NATM under the protection of advancing jet grouting columns (DIN EN 12716, 2001). 7.1.2

Structure

Two tunnel tubes with two lanes and a width of approx. 11 m each were excavated over a length of approx. 1000 m (Fig. 7.1 and 7.2). The overburden of the tunnel tubes amounts to approx. 6 to 8 m (Fig. 7.2). Approximately in the middle of this section the tunnels undercross the ICE/IC (Intercity Express/Intercity) line Cologne-Koblenz of German Rail as well as a city railway tunnel. The latter is located at the construction pit Moltke square (Fig. 7.1). This tunnel section is followed by the construction pit Van-Groote square (Fig. 7.1). In the approx. 200 m long tunnel section south of the construction pit Van-Groote square the four lanes run through a tunnel tube which is divided into three sections (Fig. 7.1b and 7.3). This tunnel tube has a total width of approx. 30 m and an overburden of approx. 7 m (Fig. 7.3). A mouth-shaped profile was selected for the cross-section of the two two-lane tunnel tubes. Fig. 7.4 shows the geometry of the 11.2 m wide and 10.1 m high standard profile. The excavated crosssection amounts to approx. 92 m². The shotcrete membrane is 25 cm thick. The cross-section was excavated in two parts, as detailed in Chapter 7.1.4. The temporary invert of the partial excavation

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- 352 was rounded with R = 13.3 m and supported by a 20 cm thick shotcrete membrane. The reinforced concrete interior lining was constructed 40 cm thick (Fig. 7.4).

Fig. 7.1:

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Road tunnel Bonn–Bad Godesberg: a) Site plan; b) schematic representation of the structure

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- 353 -

Fig. 7.2:

Two-lane tunnel tubes, cross-section

Relatively large radii of curvature were selected for the shotcrete support in the roof and sidewall areas with R = 4.726 m and R = 6.426 m. As a consequence the loading of the shotcrete membrane by bending moments and shear forces is small in these areas, as shown below. The transitions from the sidewalls to the temporary invert and the permanent invert, respectively, were constructed with relatively small radii (Fig. 7.4). This leads to bending moments and shear forces in the shotcrete membrane in these areas. As shown below, at a radius of 1.2 m supplementary reinforcement is required at

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- 354 the transitions from the sidewalls to the temporary invert in addition to the planned reinforcement with inside and outside steel fabric mats Q188. At the transitions from the sidewalls to the permanent invert, however, no additional reinforcement is required at a radius of 1.8 m (see Chapter 7.1.5).

Fig. 7.3:

Tunnel tube divided into three sections, crosssection

The invert was slightly rounded with a radius of curvature of R = 15.4 m because of its loading by the water pressure (Fig. 7.4). A greater curvature would lead to a smaller amount of reinforcement in the interior lining, but also to additional excavation and would therefore not be economical.

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- 355 -

Fig. 7.4: 7.1.3

Two-lane tunnel tube, standard profile Ground and groundwater conditions

The tunnels were headed in the mostly sandy and gravelly soil layers of the gravel deposits of the lower terraces typical for the WBI-PRINT 5

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- 356 Rhine valley in the area of Bonn. Below the tunnel invert silt lenses are sporadically embedded in the sand and gravel layers which will be termed gravel sand in the following (Fig. 7.5).

Fig. 7.5:

Road tunnel Bonn-Bad Godesberg, longitudinal section with ground profile

The gravel sand is covered by an up to 7 m thick silt layer extending to the ground surface. Below the gravel sand is the Devonian base rock consisting of mudstone and sandstone layers (Fig. 7.5). Fig. 7.6 shows the grading ranges of the encountered gravel sand and the silt, together with the parameters derived from the subsoil exploration results. The gravel sand has a high porosity and permeability. It constitutes an aquifer connected to the Rhine river. The groundwater table is therefore influenced by the water levels of the Rhine. On average the groundwater is encountered approx. 3 m above the tunnel's invert (Fig. 7.5).

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- 357 -

Fig. 7.6:

7.1.4

Grain-size distribution of the soils and soil mechanical parameters Design and construction

The sequence of excavation and the support measures for the heading of the two two-lane tunnels are shown in Fig. 7.7 and 7.8 in cross- and longitudinal section. First the part of the tunnel cross-section located above the groundwater table was excavated. For statical reasons the temporary invert was rounded and supported by shotcrete (c in Fig. 7.7). Regularly spaced gravity wells were drilled from the temporary invert. With these wells, the groundwater table was lowered to the final tunnel invert (d in Fig. 7.7). Protected by this groundwater drawdown the tunnels were excavated in stages down to the invert and supported (e in Fig. 7.7). WBI-PRINT 5

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- 358 -

Fig. 7.7:

Heading of the two-lane tunnels, excavation and support, cross-section

The partial excavation above the groundwater table was subdivided into crown, bench and invert and carried out with a stepped tunnel face. The round lengths of the partial excavations amounted to 1 m each. To limit the subsidence the temporary invert was closed 6 to 8 m behind the roof excavation. This lead to an inclination of the tunnel face of 60° (Fig. 7.8). Because the tunnel face inclination

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- 359 exceeds the angle of friction of the gravel sand (see Fig. 7.6), the tunnel face was not stable if the apparent cohesion was not taken into account. Since the latter quickly vanishes with the soil drying up, only small sections could be excavated in one step. These sections were immediately sealed with reinforced shotcrete.

Fig. 7.8:

Heading of the two-lane tunnels, excavation and support, longitudinal section

The excavation contour was supported using reinforced shotcrete and steel sets (Fig. 7.7 to 7.9). As already mentioned, the tunnel tubes were excavated under the protection of advancing jet grouting columns forming a jet grouting vault (Fig. 7.7 and 7.8). This jet grouting vault transfers loads in transverse and longitudinal tunnel directions (Fig. 7.10). Thus the green shotcrete close to the tunnel face as well as the tunnel face area were less strongly loaded. Furthermore, with the jet grouting columns, the subsidence is limited, col-

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- 360 lapses are avoided, and the safety of the tunneling staff is thus ensured as well.

Fig. 7.9:

View of the temporary tunnel face

The jet grouting columns are approximately horizontal, 15 m long and have a design diameter of 63 cm. Two successive jet grouting vaults overlap by 3 m (Fig. 7.8 and 7.11).

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- 361 -

Fig. 7.10:

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Load transfer by the jet grouting vault and the shotcrete membrane: a) Cross-section; b) longitudinal section WBI GmbH, Henricistr. 50, 52072 Aachen, Germany

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- 362 -

Fig. 7.11:

Heading of the two-lane tunnels, excavation and support, plan view (section I-I, see Fig. 7.8)

In order to comply with the tunnel clearance, the jet grouting columns were not constructed horizontally but rather at a slight outward slant (Fig. 7.8 and 7.11). The excavation had thus to be widened in a trumpet-shaped way (Fig. 7.8, 7.11 and 7.12). An additional stabilization of the tunnel face was achieved by the support core shown in Fig. 7.11 and 7.12. Beside and above this support core was enough space for the jet grout drill carriage (Fig. 7.11 and 7.13). The tunnel face was supported in sections using reinforced shotcrete (Fig. 7.8 and 7.11). The jet grouting columns were constructed by the single-phase method (DIN EN 12716, 2001), according to which in a borehole cement based suspension is injected under high pressure (approx. 500 bar) into the soil via a rod and a nozzle at its end. WBI-PRINT 5

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- 363 -

Fig. 7.12:

Tunnel face with support core and niche for the construction of the jet grouting columns

Fig. 7.13:

Jet grout drill carriage in operation

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- 364 The nozzle rotates with the rod which is slowly pulled out of the borehole. In this way a column develops due to the mixing of the suspension with the ground. After hardening of the cement, this column possesses a high strength in comparison with the undisturbed soil. The surplus mixture of suspension and soil exits as backflow through the annular gap between borehole and rod. Fig. 7.13 shows the jet grout drill carriage in operation. To optimize the production parameters six test columns were constructed and dug out. Column diameters ranging from 60 to 90 cm were obtained. The production parameters are listed in Table 7.1. The parameters in the lines marked with arrows in Table 7.1 were selected for the construction of the jet grouting columns. Column diameters between 60 and 70 cm were achieved with these parameters (Wittke et. al, 2000). Retracting Injection Water/ Cement rate pressure cement ratio quantity [cm/min] [bar] [-] [kg/m] Î 30 500 1.05 251 " " 1.0 260 Î 27 500 1.05 283 24 " " 313 " " 1.0 325 20 " " 510 Î Parameters selected for the construction of the columns Table 7.1:

Column diameter [cm] 60 – 70 " 60 – 70 " " 80 - 90 jet grouting

Production parameters of six test columns

A comparatively small unconfined compressive strength of σD = 0.75 MN/m2 was demanded for the jet grouting columns and accounted for in the stability analyses (see Chapter 7.1.5). This strength was achieved already after a short time and the idle time causing high cost and long construction times could be minimized. Fig. 7.14 shows the unconfined compressive strength measured on drill cores taken and on samples from the backflow as a function of the sample age (Wittke et al., 2000). The diagram also includes the unconfined compressive strength determined on drill cores and backflow samples of jet grouting columns constructed during the heading of the city railway tunnel "Killesberg-Messe" in Stuttgart

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- 365 (see Chapter 7.2). According to this diagram, values of σD = 0.75 MN/m2 are achieved already after one day.

Fig. 7.14:

Unconfined compressive strength measured on backflow samples and drill cores versus sample age

The construction of the jet grouting columns included comprehensive quality management measures. The success of these measures became apparent during the excavation. No defects were found in the jet grouting vaults. To determine the influence of the vibrations due to tunneling as well as railway and road traffic on the stability of the tunnel face, the vibration velocity was measured during tunneling. The occurring vibrations turned out to be small. As an optional position it was planned to additionally stabilize the tunnel face usWBI-PRINT 5

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- 366 ing jet grouting columns (Fig. 7.15). To reduce the strength a portion of the cement in the suspension would have been replaced with bentonite to facilitate the later demolition of these columns. The construction of tunnel face columns did not become necessary, however.

Fig. 7.15:

Stabilization of the tunnel face using jet grouting columns (not carried out)

In the area where the tunnel tube is divided into three sections, the side tubes were excavated first under the protection of jet grouting columns and supported. The central tube was only excavated after the interior lining had been installed in both side tubes. A crown excavation with trailing bench and invert was carried out here. The shotcrete membrane of the central tube hereby is supported by the interior linings of both side tubes (Fig. 7.16).

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- 367 -

Fig. 7.16:

7.1.5

Excavation sequence/construction stages for the tunnel tube divided into three sections

Stability analyses for the design of the shotcrete support

To design the shotcrete support two- and three-dimensional FEanalyses were carried out with the program system FEST03 (Wittke, 2000). Fig. 7.17 shows the location of the 10 analysis crosssections (AC 1 to AC 10) in the geological cross-section which are investigated in the design analyses. Fig. 7.18 shows exemplarily the computation section, the FE-mesh, the boundary conditions, the ground profile and the parameters for a three-dimensional analysis.

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- 368 -

Fig. 7.17:

Location of the 10 analysis cross-sections

In the following the results of a two-dimensional analysis for analysis cross-section 2 are presented. Analysis cross-section 2 is located in the section of the two two-lane tunnel tubes (see Fig. 7.17). Fig. 7.19 shows the computation section, the FE-mesh, the boundary conditions, the ground profile and the parameters this analysis was based upon. The computation section consists of a 25 m wide, 30.5 m high and 1 m thick slice of the ground. The FE-mesh was subdivided into 669 three-dimensional isoparametric elements with a total of 797 nodes.

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- 369 -

Fig. 7.18:

Three-dimensional computation section, FE-mesh, boundary conditions, ground profile and parameters

For the nodes on the lower boundary (z = 0), horizontally sliding supports were assumed as boundary conditions. Vertically sliding supports were introduced for the nodes located on the vertical lateral boundary planes (x = 0 and x = 25 m) (Fig. 7.19). All nodes were assumed fixed in y-direction. The loading due to build-

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- 370 ings that acts on the analysis cross-section was represented by a surface load (p = 60 kN/m2).

Fig. 7.19:

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Analysis cross-section 2, FE-mesh, boundary conditions, ground profile and parameters for twodimensional analyses WBI GmbH, Henricistr. 50, 52072 Aachen, Germany

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- 371 In the stability analyses for analysis cross-section 2 the heading of only one tunnel tube was investigated. Because the distance of the two tunnel tubes amounts to more than one tunnel diameter in this area, the two tubes influence each other only to a small degree. Since symmetry exists with respect to the tunnel axes, the ground profile and the cross-sectional shape of the tunnel tubes, only one half of a tunnel tube was modeled. The vertical section through the tunnel axis constitutes the plane of symmetry (Fig. 7.19).

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Analysis cross-section 2, FE-mesh, detail WBI GmbH, Henricistr. 50, 52072 Aachen, Germany

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- 372 The shotcrete membrane with a thickness of d = 25 cm was modeled by one element layer. Three element layers with a total thickness of 60 cm were selected for the simulation of the jet grouting vault (Fig. 7.20). The parameters chosen for the undisturbed soil and the soil stabilized with jet grouting columns were determined on the basis of the exploration results by the parties concerned during technical discussions. They are shown in Fig. 7.19. The deformability and the strength of the soil stabilized by the jet grouting columns develop with time (see Fig. 7.14). In agreement with the parties concerned, values of E = 500 MN/m2, ϕ' = 35° and c' = 200 kN/m2 were specified for Young's modulus and the shear strength parameters of the jet grouting vault. These shear strength parameters correspond to an unconfined compressive strength of σD = 0.75 MN/m2. According to Fig. 7.14, these values are attained after a few days already. Young's modulus assumed for the shotcrete was varied in the stability analyses. In the following the computation sequence and the results of an analysis are presented, in which a modulus of E = 7500 MN/m2 was selected for the shotcrete (Fig. 7.19). This relatively small value reflects the development of strength and deformability and the creep properties of the shotcrete (see Chapter 2.1). In the case presented here the shotcrete is loaded at a very young age due to the early closing of the shotcrete support approx. 6 to 8 m behind the crown excavation (see Fig. 7.8). Fig. 7.21 shows the eight computation steps applied to simulate the excavation and support of the tunnel. In the 1st computation step, the state of stress and deformation resulting from the dead weight of the soil and the loading due to buildings (in-situ state) is determined. In the 2nd computation step the installation of the jet grouting vault is simulated. A preceding stress relief of the soil in the area of the crown excavation is modeled in the 3rd computation step (Wittke, 2000). The reduced Young's modulus of the gravel sand of Ered = 45 MN/m2 corresponds to a stress relief factor of aV = 0.6 according to (4.1). The 4th computation step represents the crown excavation and its support using shotcrete. In the 5th and 6th computation steps, the preceding stress relief of the soil in the area of the bench excavation as well as the bench excavation with temporary invert support using shotcrete are simulated.

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- 373 -

Fig. 7.21:

Analysis cross-section 2, computation steps

After the preceding stress relief of the soil in the area of the invert excavation in the 7th computation step, the drawdown of the WBI-PRINT 5

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- 374 groundwater table to the invert level and the excavation and shotcrete support of the invert are simulated in the 8th computation step. In Fig. 7.22 the nodal displacements computed for the 8th computation step related to the in-situ state (1st computation step) are shown in horizontal sections above the tunnel roof. The subsidence of the ground surface above the tunnel roof amounts to 33 mm.

Fig. 7.22:

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Analysis cross-section 2, displacements, 8th – 1st computation step

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- 375 -

Fig. 7.23:

Analysis cross-section 2, bending moments in the shotcrete membrane: a) 6th computation step; b) 8th computation step

Fig. 7.24:

Analysis cross-section 2, statically required outside reinforcement of the shotcrete membrane: a) 6th computation step; b) 8th computation step

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- 376 -

Fig. 7.25:

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Design of the support in the area of the bench's foot

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- 377 Fig. 7.23 shows a comparison of the bending moments in the shotcrete lining computed for the 6th and the 8th computation step. Due to the small radius of curvature of R = 1.2 m in the area of the connection of the temporary invert to the sidewalls (see Fig. 7.4), comparatively large bending moments occur at this location in the 6th computation step as previously mentioned (Fig. 7.23a). In addition to the planned steel fabric mats Q188 supplementary outside reinforcement becomes necessary in this area (Fig. 7.24a). In the 8th computation step the radius amounts to R = 1.8 m in the area of the connection of the permanent invert to the sidewalls (see Fig. 7.4). Smaller bending moments are computed for this step (Fig. 7.23 a and b). They can be carried by the shotcrete membrane without any reinforcement for an assumed safety factor of 1.35 (Fig. 7.24b). Fig. 7.25 shows the design of the support in the area of the bench's foot. 7.1.6

Monitoring

The subsidence due to the heading was measured by leveling using a closely spaced raster of measurement cross-sections. In addition, subsidence and convergency measurements were carried out in the tunnel tubes. Fig. 7.26 shows the maximum values of the ground surface subsidence measured during the heading between the two construction pits Moltke square and Van-Groote square. The largest subsidence of approx. 6.5 cm occurred close to the construction pit Moltke square in an area where the tunnel tubes were headed by sidewall adit excavation without jet grouting columns. In the other areas the maximum subsidence of the ground surface ranged between approx. 15 mm and approx. 45 mm (Fig. 7.26). The maximum ground surface subsidence of 33 mm computed for analysis cross-section 2 is in good agreement with the measured values (see Fig. 7.22 and 7.26). 7.1.7

Conclusions

With the relocation of the federal highway B 9 in Bonn-Bad Godesberg into a tunnel the whole tunnel cross-section was located in cohesionless gravel sand. Buildings and railway facilities had to be undercrossed with little overburden. Therefore, the subsidence WBI-PRINT 5

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- 378 of the ground surface due to the heading had to be limited to small values.

Fig. 7.26:

Maximum values of the measured ground surface subsidence, longitudinal section between the two construction pits Moltke square and Van-Groote square

The tunnel was excavated by the NATM under the protection of advancing jet grouting columns. With this measure a part of the overburden load could be transferred in lateral and longitudinal tunnel direction (see Fig. 7.10). Thus the green shotcrete in the tunnel face area and the tunnel face itself was less strongly loaded. Furthermore, with the jet grouting columns, the subsidence was limited, collapses were avoided, and the safety of the tunneling staff was thus ensured as well. The tunnel was excavated with a steeply inclined stepped tunnel face, short round lengths and a fast closing of the invert support (see Fig. 7.8). To guarantee the stability of the tunnel face, an immediate tunnel face support in sections using reinforced shotcrete was necessary. WBI-PRINT 5

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- 379 With these measures, a stable excavation of the tunnel was feasible, and the subsidence of the ground surface could be limited to 2 to 4 cm (see Fig. 7.26). No damage occurred on buildings or railway facilities. The FE stability analyses carried out for this project represented an essential contribution towards the design, the statics and the design of the excavation and support measures. 7.2

City railway tunnel "Killesberg-Messe" in Stuttgart, Germany

7.2.1

Introduction

Between July 1990 and April 1991 the "Killesberg-Messe" city railway tunnel was constructed in Stuttgart, Germany. A 64 m long section of the tunnel alignment is located immediately adjacent to the State Academy of Art and Design (Academy of Art, Fig. 7.27 and 7.28). In addition, the tunnel was driven through a quarry fill in this area. To keep the subsidence due to tunneling small, the tunnel was headed in this section under the protection of an advance support constructed by jet grouting columns (EN 12716, 2001).

Fig. 7.27:

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Killesberg-Messe city railway tunnel, site plan

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- 380 -

Fig. 7.28:

7.2.2

Killesberg-Messe city railway tunnel, longitudinal section with ground profile, excavation and support measures

Structure

With the "Killesberg-Messe" city railway line in Stuttgart, the so-called Trade Fair Line, a fast and convenient rail transit connection was constructed between the central junction Stuttgart Central Station and the Killesberg heights with the hill park. Over a length of approx. 360 m up to the Killesberg-Messe station the city railway line runs in a tunnel (Fig. 7.27). The overburden of the tunnel first rises to approx. 16 m, then decreases and amounts to approx. 6 m at the beginning of the underground Killesberg-Messe station (Fig. 7.28). From km 0+500 to km 0+710 the double-track standard profile was excavated. With a width of approx. 10.5 m, the excavated crosssection amounts to approx. 70 m² in this section. In the further course up to the Killesberg-Messe station the tunnel widens to a width of approx. 17.8 m. The excavated cross-section at the beginning of the station amounts to approx. 174 m2 (Fig. 7.27 to 7.29). WBI-PRINT 5

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- 381 -

Fig. 7.29:

Enlarged cross-section at the beginning of the Killesberg-Messe station (km 0+860)

A mouth-shaped profile was selected for the tunnel cross-section. The geometry of the enlarged cross-section at the beginning of the Killesberg-Messe station (largest cross-section) is shown in Fig. 7.29. The shotcrete membrane has a thickness of t = 35 cm, the inside sidewalls of the two sidewall adits had a 25 cm thick shotcrete membrane. If required the shotcrete membrane of the sidewall adits was planned to be closed at the invert with t = 20 cm. The interior lining was constructed 80 cm thick with watertight concrete of grade B35. In the roof and sidewall areas of the largest cross-section radii of curvature of R = 9.8 m and R = 5.4 m, respectively, were selected for the shotcrete support. The transitions from the sidewalls to the invert were constructed with the comparatively small WBI-PRINT 5

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- 382 radii R = 3.8 m and R = 1.9 m. Because of the water pressure loading of the interior lining the invert was slightly rounded with a radius of curvature of R = 15.8 m (Fig. 7.29). 7.2.3

Ground and groundwater conditions

The plateau of the Killesberg heights is formed by stone, which was mined in numerous quarries in the quarries were later closed and backfilled 80 to 90 quarry fill (Fig. 7.30). Below the Schilfsandstone the Gypsum Keuper are encountered (Fig. 7.28).

the Schilfsandpast. These years ago with the layers of

At the portal (km 0+500) the city railway tunnel cuts into talus material and the upper layers of the Gypsum Keuper (Estherien layers). Starting at km 0+650 the Schilfsandstone enters into the tunnel profile from the roof. From km 0+700 to the end of the tunnel at km 0+860 the roof and the upper part of the sidewalls are located in the quarry fill. The lower part of the sidewalls and the tunnel invert cut into the Schilfsandstone. The boundary to the Gypsum Keuper is located in this area at the level of the tunnel invert or slightly below (Fig. 7.28). The ground conditions were mainly derived from the results of core drillings. In addition, test pits were excavated and dynamic probing were carried out. The evaluation of old aerial photographs and seismic measurements served to localize the quarry edges. In front of the Academy of Art a test shaft 4 m in diameter was sunk. Here samples for soil mechanical laboratory tests were taken and plate loading tests were carried out to determine the deformability. The quarry fill is very heterogeneously composed. It consists of sandstone blocks of different sizes with edges up to 80 cm long. Partly the pieces of rock lie in a sandy silt matrix, partly they constitute a pure sandstone deposit without any filling material (Fig. 7.30). According to the geological report the porosity ranges between 3 and 35 %. By means of density measurements on large-scale samples a content of 10 % of large voids or cavities, respectively, was estimated. The grain size distributions determined for five samples are shown in Fig. 7.31. According to this, the fill consists of smoothly graded gravel/sand (GW), gravel/silt (GU) and sand/silt mixtures (SU).

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- 383 -

Fig. 7.30:

Quarry fill

Fig. 7.31:

Grain size distribution of samples from the quarry fill

The undisturbed Schilfsandstone consists mostly of hard sandstones with clay flasers and a marked horizontal bedding. The banks are

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- 384 between 10 and 50 cm thick. The sandstone is mostly vertically jointed with a medium to wide joint spacing. The Gypsum Keuper layers in the tunnel invert area belong to the White and Grey Estherien layers. Anhydrite or gypsum deposits were not encountered in the course of the ground exploration down to approx. one tunnel diameter below the invert. Swelling phenomena due to the tunneling were therefore not to be expected in the ground. No leaching cavities were drilled into either in the area specified above. The soil and rock mechanical parameters of the different ground layers determined or estimated from the exploration results are listed in Table 7.2.

layer

intact rock discontinuity sets unit Young's Poisbedding jointing modulus son's - weight ϕ c ratio E γ cB ϕJ cJ [°] [kN/m2] ϕB 2 3 2 ν [MN/m ] [kN/m ] [°] [kN/m ] [°] [kN/m2]

talus material

7

0,40

21,5

25

25

-

-

-

-

quarry fill

6

0,25

19

30

0

-

-

-

-

2000

0,20

25

40

3000

40

30

40

0

100

0,33

23

30

50

30

20

30

20

Schilfsandstone Gypsum Keuper

Table 7.2:

Soil and rock mechanical parameters

Five core drillings were equipped as observation wells. The measured water levels show that the ground water table lies approximately on the level of the quarry base and thus above the tunnel's invert. The boundary between the permeable Schilfsandstone and the Gypsum Keuper with its low permeability forms a preferred groundwater horizon. Above the quarry edge seepage water has accumulated. An amount of 1 l/s of water was pumped out of the test shaft mentioned above over a period of several days.

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- 385 7.2.4

Excavation and support

Between km 0+556 and km 0+808 the city railway tunnel was driven as a crown heading with trailing bench and invert excavation (Fig. 7.32a).

Fig. 7.32:

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Excavation and support: a) Crown heading, crosssection; b) crown heading, longitudinal section; c) sidewall adit heading, cross-section

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- 386 Crown and bench were excavated with a stepped tunnel face and a support core in the crown area and with round lengths of 2.4 to 3.3 m. To ensure the stability of the temporary tunnel face, it was supported in sections with reinforced shotcrete. The invert was excavated 11 to 15 m behind the bench. The distance between the crown face and the closing of the support at the invert amounted to between 16 and 22 m, depending on the ground conditions encountered (Fig. 7.32b). Because of the large cross-section at the end of the widening segment the tunnel was driven in the further course up to KillesbergMesse station (km 0+808 to 0+860) by sidewall adit excavation (Fig. 7.28, 7.29 and 7.32c). The excavation profile was supported with reinforced shotcrete, steel sets and a systematic anchoring using SN-anchors, the length and spacing of which were determined as required. The foundations of the Academy of Art are located at close distance from the tunnel (Fig. 7.33a). In this section of the alignment the tunnel roof and the upper sidewalls are located in the quarry fill. Because of the high deformability and low strength of the fill, the close distance of the foundations of the main building of the Academy of Art to the tunnel and the large tunnel cross-section, it was feared that tunneling-induced subsidence would lead to damages to the main building of the Academy of Art. To limit the subsidence, the tunnel was excavated in the area of the quarry fill under the protection of an advance support by jet grouting columns. To this end, from km 0+710 to km 0+785 seven jet grouting vaults were constructed in sections by the single phase method described in Chapter 7.1 (Fig. 7.33b). The jet grouting columns were constructed with a length of 11.0 to 12.75 m, measured from the tunnel face. At the end of each excavation section the columns extended 3 m beyond the tunnel face. With 1 m of nongrouted borehole length, this results in an overlap of the columns of 2 m (Fig. 7.34b). The drillings were directed with a 6 degree outward slant with respect to the tunnel axis. Thus, at the end of the enlarged excavation section there was enough clearance to construct the columns for the following section (Fig. 7.34b).

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- 387 -

Fig. 7.33:

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- 388 -

Fig. 7.34:

Advancing grouting of the quarry fill and jet grouting columns: a) Cross-section; b) longitudinal section

The nature of the ground in the area of the jet grouting columns to be constructed required particular measures. To prevent the ce-

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- 389 ment based suspension from seeping into large voids or cavities in the quarry fill, and in order to ensure that the columns could be constructed according to plan with respect to diameter, continuity, strength and position, the quarry fill was grouted in advance to fill existing cavities (Fig. 7.34). To this end drillings 114 mm in diameter with air flushing were sunk, equipped with PVC sleeve pipes and grouted with packers in steps of 1 m from the bottom up with a cement-bentonite grout (250 kg cement and 40 kg bentonite for 1000 l) which is stable with respect to sedimentation (Table 7.3). The purpose of this was to achieve a void-free matrix filling potential cavities in the quarry fill which would not impede the construction of the jet grouting columns and the tunnel excavation. 7 sections Per section 64 m Boreholes Drilling/sleeve pipe Grouted volume Cement Covered volume of soil Achieved grouting volume Table 7.3:

number m m3 t m3 %

152 1950 361 99 2280 16

Amount of grouting of the quarry fill

Water/cement (PZ 35 F) ratio of grout Pump pressure Pump capacity Retracting rate Rotational speed Grout flow rate Cement quantity Table 7.4:

14 - 25 283 52 14 325 16

bar l/min m/min min-1 l/min kg/m

0,8 400 95 0,41 20 233 205

Operational parameters of the jet grouting

Only a few hours after the completion of the grouting of the quarry fill the construction of the jet grouting columns started. With a planned minimum diameter of the columns of 0.5 m, the drillings were placed at a distance of 0.35 m to the tunnel contour. To confine temporary decrease in strength in the soil locally, the columns were constructed first with a spacing of 1.4 m, WBI-PRINT 5

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- 390 then halfway between two columns each, and finally in the gaps between two adjacent columns. This way a vault supported by the Schilfsandstone was constructed made up of intersecting or, in the area more widely fanned out, touching columns. The operational parameters of the jet grouting and the amount of work done are given in Tables 7.4 and 7.5 (Beiche et al., 1991). 7 vaults 64 m

Per segment Columns Drilling Grout Cement

number m m3 kg

28 - 44 308 - 561 103 91000

268 3280 723 637000

Table 7.5:

Amount of jet grouting work performed

Fig. 7.35:

Special construction equipment SR 510

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- 391 Fig. 7.14 shows the unconfined compressive strength measured on backflow samples and drill cores as a function of the sample age (Wittke et al., 2000). The heading generally recommenced 12 to 15 hours after the completion of a jet grouting vault. At this time the columns were between 12 hours and 2 days old. An essential technical requirement for advance support by jet grouting is the ability to construct the single structural elements – the soil-concrete columns – in a self-contained continuous operation. Suitable technical equipment must therefore above all possess a feeding length at least equal to the length of a jet grouting column. For economic reasons it is important that the drill mount can be quickly positioned as desired over the entire excavation profile. Fig. 7.35 shows the construction equipment SR 510 used for the Killesberg-Messe city railway tunnel in operation. The boreholes for the grouting of the quarry fill were drilled with the same equipment (Beiche et al., 1991). Jet grouting vault

1

Borehole length

11,00 m

Heading segment length

8,00 m

Days

10

Systematic advance filling

Number of boreholes Construction time Number of columns

2

3

4

5

11,00 12,00 12,75 12,75 8,00 20

9,00 30

9,75

9,75

40

6

7

12,75

12,75

9,75

9,75

50

60

70

14 (s = 0,8 m)

20

22

23

24

24

25

28 (s = 0,33 m)

34

38

40

42

42

44

Construction time Heading

Table 7.6:

Construction of the jet grouting vaults

The grouting of the quarry fill and the construction of the jet grouting columns had to be worked into the heading scheme. Since the columns could not be constructed in parallel with the heading, the established heading operation in ten-days periods was generWBI-PRINT 5

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- 392 ally changed over for the area with jet grouting columns to a continuous operation on demand for reasons of time (Beiche et al., 1991). The length of the heading section protected by jet grouting columns was 64 m in total. This section was constructed in 68 days. 24 days thereof account for the advance grouting of the quarry fill, 18.5 days for the construction of the columns and 25.5 days for the heading (Table 7.6). 7.2.5

Stability analyses

For the design of the shotcrete support and the interior lining two-dimensional FE-analyses were carried out using the program system FEST03 (Wittke, 2000). Eight analysis cross-sections were investigated in total, differing with respect to -

the geometry of the tunnel cross-section,

-

the ground conditions and overburden height,

-

the construction stages and/or

-

the support installations.

Two of these analysis cross-sections are located in the area of the Academy of Art. In the following a stability analysis for the analysis cross-section km 0+785 (see Fig. 7.27 and 7.28) is exemplarily presented (see Beiche et al., 1991). Fig. 7.36 shows the computation section, the FE-mesh, the boundary conditions, the ground profile and the parameters for this analysis cross-section. The upper part of the tunnel cross-section lies in the quarry fill. The edge of the former quarry is located approx. 2 m away from the tunnel sidewall opposite to the Academy of Art. The lower part of the tunnel's cross-section is situated in the Schilfsandstone. The boundary to the Gypsum Keuper is encountered approx. 1.2 m underneath the tunnel's invert (Fig. 7.36).

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Fig. 7.36:

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Analysis cross-section km 0+785, FE-mesh, boundary conditions, ground profile and parameters

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- 394 The computation section consists of a 1 m thick, 40 m high and 66 m wide slice including the tunnel's cross-section as well as the surrounding ground. It is divided into 934 isoparametric elements with a total of 2206 nodes. Vertically sliding supports are selected as boundary conditions for the nodes on the vertical lateral boundary planes (x = 0 and x = 66 m), while horizontally sliding supports are chosen for the nodes on the lower boundary plane (z = 0) (Fig. 7.36). All nodes are fixed in y-direction. The loading resulting from the Academy of Art building to the right of the tunnel is simulated by a surface load. The surface load is applied to the corresponding nodes of the FE-mesh in the form of point loads (Fig. 7.36). The shotcrete membrane (t = 25 cm) is simulated by one row of elements, the interior lining (t = 50 cm) by three and the jet grouting vault (t = 50 cm) by two element rows. The effect of a separating non-woven material preventing the transfer of tensile and shear stresses between shotcrete support and interior lining is simulated by a thin row of elements with no stiffness assigned and by radially arranged truss elements of high stiffness (pendulum rods), which only allow the transfer of normal compressive stresses (Fig. 7.37). The soil and rock mechanical parameters of the different layers as well as the parameters assigned to the shotcrete, the reinforced concrete of the interior lining and the jet grouting vault are given in Fig. 7.36. It was not possible to carry out advance tests to determine the mechanical parameters of the quarry fill improved by jet grouting. These parameters therefore had to be estimated for the stability analyses. The deformability and the strength of the soil improved by jet grouting develop with time (see Fig. 7.14). Since tunneling recommenced already approx. 15 hours after the completion of each jet grouting vault, Young's modulus and the strength of the improved soil are still low when the cross-section is excavated. Young's modulus and the shear parameters of the jet grouting vault were therefore assumed comparatively low and thus conservative in the analyses with values of E = 500 MN/m2, ϕ = 35° and c = 100 kN/m2. These shear parameters correspond to an unconfined compressive WBI-PRINT 5

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- 395 strength of σu = 0.5 MN/m2. According to Fig. 7.14 this value is reached after one day already.

Fig. 7.37:

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Analysis cross-section km 0+785, FE-mesh, detail with shotcrete membrane and interior lining

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- 396 A three-dimensional analysis was not considered necessary in this case, since a three-dimensional arching cannot develop in the quarry fill due to the low strength. Although the jet grouting vault constructed in advance reaches down to the quarry base and is thus supported by the Schilfsandstone, it cannot transfer any significant loading during crown excavation. The reason is that the time span between the construction of the jet grouting vault and the crown excavation is very short. The vault has therefore only reached a small fraction of its final strength at this stage. It is only after the bench excavation that the shotcrete support reaches down to the Schilfsandstone and a setting process has taken place in the vault. Both means of support are then ready to carry loads. It was therefore determined that the crown heading must not be ahead of the bench by more than twice of the lattice girder spacing (2.4 to 3.3 m) (see Fig. 7.32b). Fig. 7.38 shows the seven computation steps simulating the in-situ state and the construction stages, which are the installation of the jet grouting vault, crown, bench and invert excavation, installation of the interior lining and rise of the groundwater table to roof level. Fig. 7.39 to 7.41 show the analysis results for the 4th computation step (crown and bench excavation). Above the right half of the tunnel practically no arching develops in the fill due to its low strength and large deformability. This can be recognized in Fig. 7.39a from the fact that the major principal normal stresses are almost vertically oriented. Above the left half of the tunnel in the area of the vertical quarry wall the major principal normal stress deviates from the vertical. The vertical stresses are lower in the fill and higher in the Schilfsandstone than the overburden weight. This result is due to the different Young's moduli of the Schilfsandstone and the fill. The quarry fill is thus "hung" on the edge already in the stage before the tunnel excavation and the construction of the jet grouting vault. This effect should be less pronounced in reality than in the analysis, since the fill was placed in layers rather than in one step as simulated in the analysis.

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Fig. 7.38: WBI-PRINT 5

Analysis cross-section km 0+785, computation steps

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Fig. 7.39:

Analysis cross-section km 0+785: a) Principal normal stresses, 4th computation step; b) displacements, 4th – 1st computation step

Fig. 7.40:

Analysis cross-section km 0+785, 4th computation step: a) Vertical stresses in horizontal sections; b) horizontal stresses in vertical sections

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- 399 The computed roof subsidence amounts to approx. 2.5 cm. Heave occurs at the bench base (Fig. 7.39b). An arch develops in the jet grouting vault above the crown. Due to its bond with the shotcrete membrane the latter is strongly loaded as well and stress concentrations result in the area of the bench base. To illustrate the load transfer described above, the horizontal and vertical stresses are shown in sections in Fig. 7.40. Stress concentrations in vertical as well as in horizontal direction are apparent at the base of the bench and of the jet grouting vault. The loading of the shotcrete membrane exceeds the one of the jet grouting vault. This is due to Young's modulus of the shotcrete being markedly higher than the one of the jet grouting columns (see Fig. 7.36). Fig. 7.41 shows the computed bending moments and normal thrust in the shotcrete membrane. Large bending moments together with a comparatively small normal compressive thrust occur in the roof and the lower sidewall areas.

Fig. 7.41:

Analysis cross-section km 0+785, bending moments and normal thrust in the shotcrete membrane, 4th computation step

These results change only slightly with the invert excavation and the immediate closing of the support (5th computation step). The design of the interior lining was based on the following load cases: -

Dead weight of the interior lining,

-

dead weight of the interior lining and rock mass pressure,

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- 400 -

dead weight of the interior lining, water pressure (groundwater table at roof level) and rock mass pressure,

-

dead weight of the interior lining and water pressure (groundwater table at roof level).

In the load cases with consideration of the rock mass pressure (Fig. 7.38, 6th and 7th computation step) it is assumed that the jet grouting columns and the shotcrete membrane are decomposed and are thus not effective any more. The rock mass pressure generally leads to a great normal compressive thrust in the interior lining with favorable effects on the dimensioning for bending and normal thrust. In the load cases without rock mass pressure it is assumed that the support effect of the jet grouting vault and the shotcrete membrane remains intact. The surrounding rock mass and the shotcrete membrane are assumed weightless. A bedding of the interior lining is given, however, since for the shotcrete as well as for the bedrock Young's moduli listed in Fig. 7.36 are effective. The load case dead weight and water pressure is generally decisive for the bending reinforcement of the interior lining at the invert and the sidewalls. The statically required cross sectional areas of reinforcement for the interior lining range between 0 and 10.5 cm2/m. 7.2.6

Monitoring

To supplement and verify the results of the stability analyses, the tunnel stability and the ground surface subsidence were monitored by comprehensive measurements during construction. One extensometer measuring cross-section each was positioned in front of and behind the Academy of Art building. Surface measuring points every 8 m on the tunnel alignment, bolts on the outside and inside of the buildings and elevation measuring points on the ceilings facing the tunnel completed the measuring points on and above ground surface. In addition, convergency measuring crosssections were installed every 10 m in the tunnel, with roof bolts in the middle between them.

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- 401 Fig. 7.42 shows the subsidence of a selected surface measuring point in the area of the 4th jet grouting vault as a function of the heading location. It can be seen that the total subsidence of approx. 30 mm can be divided into three parts. Approx. one third of the subsidence is preceding subsidence caused by the approaching heading. Another third occurs during the construction of the jet grouting vault while the heading is stopped. At this stage, the tunnel face is located 3 m away from the measuring point. The last third of the subsidence occurs as a trailing impact after the undercrossing of the measuring point (Beiche et al., 1991).

Fig. 7.42:

Subsidence of surface measuring point 111 vs. heading location

In Fig. 7.43 measured and computed subsidences in the area of the Academy of Art are compared. In the FE-analysis subsidences in the order of up to 5 mm were computed for the area of the Academy of Art. An expert's report on the Academy's actual state and sensitivity to settlements lead to the result that subsidences of this magnitude would not cause any damages. The computed subsidence was confirmed by the monitoring results. A subsidence trough from the front to the back side of the building was hardly recognizable. The subsidences measured in the tunnel and on the ground surface WBI-PRINT 5

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- 402 above the tunnel could be reproduced by the FE-analysis as well (Fig. 7.43, Beiche et al., 1991).

Fig. 7.43:

7.2.7

Comparison of measured and computed subsidence in the area of the Academy of Art

Conclusions

The Killesberg-Messe city railway tunnel crosses partly through a quarry fill. In this area the tunnel alignment is located immediately adjacent to the Academy of Art, which is founded on the quarry fill and sensitive to settlements. To limit the headinginduced subsidence, the tunnel was driven in this area under the protection of an advance support, constructed by the jet grouting method. To do this it was necessary to fill the large voids and cavities existing in the quarry fill in advance by cement grouting. This way during jet grouting the cement based suspension was prevented from seeping into the cavities. The tunnel was driven by crown heading with trailing bench and invert excavation. It was constructed with a steep tunnel face, short round lengths and a fast invert support closing (see Fig.

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- 403 7.32b). To guarantee its stability, the temporary tunnel face was supported in sections using reinforced shotcrete. With these measures it was possible to limit the tunneling-induced subsidence at the Academy of Art building to a maximum of 4 mm (see Fig. 7.43). The FE-analyses carried out have been an important tool for the design, the statics and the specification of the excavation and support measures. The comparison of the FE-analysis results with the measured displacements in the ground showed good agreement (see Fig. 7.43).

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- 404 8.

References

Azuar, J. J.; Panet, M.: Shear behaviour of passive steel in rock mass. Rock bolting, Revue Industrie Minérale, St. Etienne, 1980. Balbach, W.; Ernsperger, H.: Entwicklung und Praxiserfahrungen mit einer neuen Verfahrenstechnik zur Herstellung und Verarbeitung von umweltfreundlichem Spritzbeton. BMI/96, 43 - 47, 1996. Beiche, H.; Erichsen, C.; Kagerer, W.; Schilcher, E.; Wooge, M.: Praktische Lösungen bei der vorauseilenden Schirminjektion unter Einsatz von Hochdruckinjektionen (HDI). Taschenbuch für den Tunnelbau 1992. Verlag Glückauf, Essen, 173 - 211, 1991. Beiche, H.; Kagerer, W.: Begrenzung der Setzungen beim innerstädtischen Tunnelbau. Proc. 10th Nat. Rock Mech. Symp., Aachen 1992, Special Edition Geotechnik, 49 - 56, 1993. Bauer, M.: Liquid non-alkaline accelerators of the latest generation. Tunnel 6/2000, 30 - 38, 2000. Bjurström, S.: Shear strength of hard rock joints reinforced by grouted untensioned bolts. Proc. 3rd ISRM Congress, Denver 1974, Vol. II B, 1194 - 1199, 1974. Böttcher, K.-H.; Schmid, I.; Erichsen, C.: S-Bahn Stuttgart Streckenverlängerung nach Filderstadt-Bernhausen. Die Untertunnelung des Flughafens. Vorträge der Baugrundtagung 1998 in Stuttgart, DGGT, 67 - 74, 1998. Brötz, K.; Löschnig, P.; Müller, F.: Shotcrete in the Schulwald Tunnel. Tunnel 6/2000, 22 - 29, 2000. CEB: Model code for concrete structures. Comité Euro-International du Béton, Paris, 1978. Deuse, T.; Mann, K.; Rüßmann, F.: New spray cement for processing naturally moist aggregates for the dry spraying method. Tunnel 6/1998, 37 - 41, 1998. DB (German Rail): Guideline 853, "Eisenbahntunnel planen, bauen und instandhalten", 1999.

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- 405 DB (German Rail): Der Tunnel; Verbindungsbahn der S-Bahn Stuttgart; Dokumentation ihrer Entstehung, 1985. DGGT (German Geotechnical Society): Empfehlungen des Arbeitskreises "Tunnelbau" (ETB). Verlag Ernst & Sohn, Berlin, 1995. Dight, P. M.: Improvements to the stability of rock slopes in open pit mines. Ph.D. Thesis, Monash University, Australia, 1983. DIN 1045: Beton und Stahlbeton; Bemessung und Ausführung, 1988. DIN 1164, Part 1: Portland-, Eisenportland-, Hochofen- und Traßzement; Begriffe, Bestandteile, Anforderungen, Lieferung, 1990. DIN 1164, Part 100: Zement; Portlandölschieferzement; Anforderungen, Prüfungen, Überwachung, 1990. DIN 4020: Geotechnische Untersuchungen für bautechnische Zwecke, 1990. DIN 4094 (Beiblatt 1): Erkundung durch Sondierungen; Anwendungshilfen, Erklärungen, 1990. DIN 4150, Part 1: Erschütterungen im Bauwesen; Vorermittlung von Schwingungsgrößen, 2001. DIN 4150, Teil 2: Erschütterungen im Bauwesen; Einwirkungen auf Menschen in Gebäuden, 1999. DIN 4150, Part 3: Erschütterungen im Bauwesen; Einwirkungen auf bauliche Anlagen, 1999. DIN 4226, Part 1: Zuschlag für Beton; Zuschlag mit dichtem Gefüge; Begriffe, Bezeichnungen und Anforderungen, 1983. DIN 4226, Part 2: Zuschlag für Beton; Zuschlag mit porigem Gefüge (Leichtzuschlag); Begriffe, Bezeichnungen, Anforderungen, 1983. DIN 21521, Part 1: Gebirgsanker für den Bergbau und den Tunnelbau; Begriffe, 1990.

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- 406 DIN 21521, Part 2: Gebirgsanker für den Bergbau und den Tunnelbau; Allgemeine Anforderungen für Gebirgsanker aus Stahl, Prüfungen, Prüfverfahren, 1993. DIN EN 12716: Ausführung von besonderen geotechnischen (Spezialtiefbau); Düsenstrahlverfahren, 2001.

Arbeiten

DIN 18551: Spritzbeton; Herstellung und Güteüberwachung, 1992. Eber, A.; Betzle, M.; Baumann, T.: Untersuchungen zum Einsatz von Gitterträgern im Tunnelbau. Bauingenieur 60, 137 - 141, 1985. EC 2: Eurocode 2 – Design of concrete structures. Part 1: General rules and rules for buildings, 1991. Erichsen, C.: Grundwassermodell für räumliche, instationäre Strömungen in doppelt porösen Medien. Veröffentlichungen des Instituts für Grundbau, Bodenmechanik, Felsmechanik und Verkehrswasserbau der RWTH Aachen, Vol. 26, 247 - 261, 1994. Erichsen, C.; Keddi, W.: Das Tragverhalten vermörtelter Anker und Entwicklung eines neuen Ankertyps. Proc. 9th Nat. Rock Mech. Symp., Aachen 1990, Special Edition Geotechnik, 10 - 21, 1991. Erichsen, C.; Tegelkamp, M.: S-Bahn Stuttgart - Streckenverlängerung vom Flughafen nach Filderstadt-Bernhausen. Die Untertunnelung des Flughafens. Taschenbuch für den Tunnelbau 1999, Verlag Glückauf, Essen, 245 - 262, 1998. Exostra International BV: Brochure, 2000. Forschungsgesellschaft für Straßen- und Verkehrswesen, Arbeitsgruppe Verkehrsführung und Verkehrssicherung: Richtlinien für die Ausstattung und den Betrieb von Straßentunneln (RABT), Guidelines, 1994. Geocon Inc.: VW Concrete Stress Cells. Instruction Manual, 1993. Grüter, R.: Erkundung des Spannungszustandes in den Schichten des Schwarzjura durch einen Großversuch bei der S-Bahn zum Flughafen Stuttgart/Echterdingen. Proc. 8th Nat. Rock Mech. Symp., Aachen 1988, Special Edition Geotechnik, 99 - 110, 1988.

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- 407 Hauck, C.-D.; Erichsen, C.; Tegelkamp, M.: Die Ostumfahrung von Stuttgart-Vaihingen: Städtischer Tunnelbau unter komplexen Bedingungen der Technik und Umwelt. Vorträge der STUVA-Tagung '97 in Berlin, Forschung + Praxis, Vol. 37, 113 - 120, 1998. Hesser, J.: Zum Einfluß unterschiedlicher Spritzbetonqualitäten auf das Tragverhalten tiefliegender Gesteinsstrecken. Laborative Untersuchungen und numerische Analysen. Institut für Aufbereitung und Deponietechnik der TU Clausthal, Vol. 10, 2000. Interfels GmbH: Brochure, 2000. Keddi, W.: Numerische Untersuchungen zum Tragverhalten vermörtelter Felsdübel in klüftigem Fels. Veröffentlichungen des Instituts für Grundbau, Bodenmechanik, Felsmechanik und Verkehrswasserbau der RWTH Aachen, Vol. 23, 1992. Kiehl, J. R.; Pahl, A.: Empfehlung Nr. 14 des Arbeitskreises "Versuchstechnik Fels" der DGEG (German Geotechnical Society): Überbohr-Entlastungsversuche zur Bestimmung von Gebirgsspannungen. Bautechnik 67, No. 9, 308 - 314, 1990. Kleffner, H.-J.: Die Tunnel zur Querung des Thüringer Waldes. Proc. 14. Symp. für Felsmechanik und Tunnelbau/Eurock 2000. Aachen, Verlag Glückauf, Essen, 13 - 19, 2000. Kusterle, W.: Spritzbeton - Gegenwart und Zukunft; Technik Qualität, Arbeitsumfeld, Umweltbeeinflussung. Tunnelbau-Fachtagung 1998, 63 - 68, 1998. Maidl, B.: Handbuch des Tunnel- und Stollenbaus. Band I: Konstruktionen und Verfahren. Verlag Glückauf, Essen, 1984. Maidl, B.: Handbuch für Spritzbeton. Verlag Ernst & Sohn, Berlin, 1992. Manns, W.; Neubert, B.; Zimbelmann, R.: Spritzbeton im Test, Festigkeitsentwicklung und Verformungsverhalten. Beton 8/87, 1987. Modemann, H.-J.; Wittke, W.: Probleme bei einem flachliegenden Straßentunnel mit großen Querschnittsabmessungen in Spritzbetonbauweise. Vorträge der STUVA-Tagung '87 in Essen, Forschung + Praxis, Vol. 32, 79 - 88, 1988. WBI-PRINT 5

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New Austrian Torket System (NATS): Applikation von Spritzbeton, 1998. ÖBV (Austrian Concrete Association): Richtlinie Spritzbeton; Anwendung und Prüfung, 1998. Paul, A.; Gartung, E.: Empfehlung Nr. 15 des Arbeitskreises "Versuchstechnik Fels" der DGEG (German Geotechnical Society): Verschiebungsmessungen längs der Bohrlochachse - Extensometermessungen. Bautechnik 68, No. 2, 41 - 47, 1991. Reik, G.; Völter, U.: Empfehlung Nr. 18 des Arbeitskreises "Versuchstechnik Fels" der DGGT (German Geotechnical Society): Konvergenz- und Lagemessungen. Bautechnik 73, No. 10, 681 - 690, 1996. Salamon, M.D.G.: Elastic moduli of a stratified rock mass. Int. J. Rock Mech. Min. Sci., 512 - 527, 1968. Schmidtmann, W.; Erichsen, C.: Autobahntunnel Berg Bock - Spritzbetonvortrieb mit Höchstleistung. Geotechnik 24, No. 1, 37 - 41, 2001. Schuck, W.; Fecker, E.: Geotechnische Messungen in bestehenden Eisenbahntunneln. Taschenbuch für den Tunnelbau 1998, Verlag Glückauf, Essen, 44 - 84, 1997. Spang, K.; Egger, P.: Tragverhalten und Bemessung voll vermörtelter, schlaffer Anker im Diskontinuum. Felsbau 7, No. 4, 181 - 189, 1989. Tegelkamp, M.; Wittke-Gattermann, P.; Züchner, F.: S-Bahn Stuttgart - Tunnelvortrieb im wasserführenden Gebirge unter dem Stuttgarter Flughafen ohne Grundwasserabsenkung. Taschenbuch für den Tunnelbau 2001, Verlag Glückauf, Essen, 21 - 40, 2000. Weber, J. W.: Empirische Formeln zur Beschreibung der Festigkeitsentwicklung und der Entwicklung des E-Moduls von Beton. Betonwerk und Fertigteil-Technik, No. 12, 1979. Wittke, W.: Rock Mechanics - Theory and Applications with Case Histories. Springer Verlag, Berlin, Heidelberg, New York, Tokyo, 1990. WBI-PRINT 5

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Wittke, W.: Hohe Horizontalspannungen im Jura und ihre bautechnischen Konsequenzen. Proc. 9th Nat. Rock Mech. Symp., Aachen 1990, Special Edition Geotechnik, 174 - 184, 1991. Wittke, W.: Stability Analysis for Tunnels, Fundamentals. Geotechnical engineering in research and practice, WBI-PRINT 4, Verlag Glückauf, Essen, 2000. Wittke, W.; Feiser, J.; Krieger, J.; Rechtern, J.: Standsicherheitsnachweis für einen Kalottenvortrieb in einem horizontal geschichteten und vertikal geklüfteten Sedimentgestein. Vorträge der STUVA-Tagung '85 in Hannover, Forschung und Praxis, Vol. 30, 132144, 1986. Wittke, W.; Kiehl, J. R.: Ausbreitung sprengbedingter Erschütterungen und deren Auswirkungen auf Bauwerke. Taschenbuch für den Tunnelbau 2001, 41 - 64, 2000. Wittke, W.; Pierau, B.: Neubaustrecke Köln-Rhein/Main. Die Tunnel Niedernhausen und Limburg. ETR Bahnreport 2000, 20 - 24, 2000. Wittke, W.; Pierau, B.; Erichsen, C.: Anwendungsbereiche der Vortriebsklassen der Spritzbetonbauweise. Geotechnik 22, No. 2, 124 133, 1999. Wittke, W.; Pierau, B.; Erichsen, C.: Der Einsatz von Hochdruckinjektionen zur Baugrundverbesserung und für den Tunnelbau im Lokkergestein. 15. Veder Kolloquium. Graz, 155 - 182, 2000. Wittke, W.; Sternath, R.: 10 Autobahnunterfahrungen im Zuge der NBS Köln-Rhein/Main. Vorträge der Baugrundtagung 2000 in Hannover, DGGT, 357 - 364, 2000. Wittke-Gattermann, P.; Wittke, W.: Simulation der Absenkung des Grundwasserspiegels beim Tunnelvortrieb im Sedimentgestein mit Hilfe instationärer, räumlicher Berechnungen. Taschenbuch für den Tunnelbau 1998, Verlag Glückauf, Essen, 21 - 43, 1997.

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