Interim Guidelines

INTERIM GUIDELINES: Evaluation, Repair, Modification and Design of Steel Moment Frames Report No. SAC-95-02

SAC Joint Venture a partnership of: Structural Engineers Association of California Applied Technology Council California Universities for Research in Earthquake Engineering

Guidelines Development Committee Ronald O. Hamburger, EQE International, Inc., Chair Edward Beck, Law-Crandall, Inc. David Houghton, Myers, Nelson, Houghton, Inc. C. W. Pinkham, S. B. Barnes, Inc. Allan Porush, Dames & Moore Thomas Sabol, Englekirk and Sabol, Inc.

C. Mark Saunders, Rutherford & Chekene, Inc. Barry Schindler, John A. Martin & Associates Robert Schwein, Schwein-Christensen Laboratories Charles Thiel Jr., Telesis Consultants

SAC Management Committee Chairman - Arthur E. Ross Structural Engineers Association of California

Applied Technology Council

California Universities for Research in Earthquake Engineering

Maryann Phipps Arthur E. Ross

John Coil Christopher Rojahn

Robin Shepherd Charles Thiel Jr.

SAC Technical Committee Stephen A. Mahin Program Manager James O. Malley Project Director for Topical Research

Ronald O. Hamburger Project Director for Product Development

SAC Joint Venture 555 University Avenue, Suite 126 Sacramento, California 95825 916-427-3647

INTERIM GUIDELINES: Evaluation, Repair, Modification and Design of Steel Moment Frames SAC Program to Reduce Earthquake Hazards in Steel Moment Resisting Frame Structures SAC Project Oversight Committee Dr. William Hall, University of Illinois, Chair Susan Dowty, International Conference of Building Officials Roger Ferch, Herrick Corporation John Gross, National Institute of Standards and Technology Fred Herman, City of Palo Alto Richard Holguin, City of Los Angeles Nestor Iwankiw, American Institute of Steel Construction

Roy Johnston, Brandow & Johnston William Mosseker, WHM Consultants Joseph Nicoletti, URS/Blume Richard Ranous, California Office of Emergency Services M. P. Singh, National Science Foundation John Theiss, EQE International, Inc.

SAC Technical Advisory Board Robert Bachman, Fluor-Daniel Corp. Vitelmo Bertero, University of California at Berkeley John Fisher, Lehigh University Subash Goel, University of Michigan Thomas Heaton, United States Geologic Survey Thomas Henyey, Southern California Earthquake Consortium William Holmes, Rutherford & Chekene, Inc.

William Honeck, Forell/Elsesser Engineers, Inc. Stanley Lindsey, Stanley V. Lindsey Associates Harry Martin, American Iron and Steel Institute John Martin, Jr., John A Martin & Associates Duane Miller, Lincoln Electric Company Charles Thornton, Thornton-Tomasetti

Task Advisory Panel - Guidelines Development Robert Bachman, Fluor-Daniel Corp. Vitelmo Bertero, Univ. of Calif. at Berkeley David Bonneville, Degenkolb Engineers, Inc. Susan Dowty, International Conference of Building Officials Douglas Foutch, University of Illinois at Urbana Nancy Hamilton, Ove Arup & Partners Richard Holguin, City of Los Angeles William Holmes, Rutherford & Chekene, Inc.

John Hooper, RSP/EQE Henry Huang, County of Los Angeles Harry Martin, American Iron and Steel Institute John Nissen, John A. Martin & Associates Robert Pyle, American Institute of Steel Construction Jack Skiles, Omaha Public Power Corp. Charles Thornton, Thornton-Tomasetti Raymond Tide, Wiss, Janney, Elstner

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Interim Guidelines: Evaluation, Repair, Modification and Design of Steel Moment Frames

Foreword and Disclaimer The purpose of this document is to provide engineers and building officials with guidance on engineering procedures for evaluation, repair, modification and design of welded steel moment frame structures, to reduce the risks associated with earthquake-induced damage. The recommendations were developed by practicing engineers based on professional judgment and experience and a preliminary program of laboratory, field and analytical research. This preliminary research, known as the SAC Phase 1 program, commenced in November, 1994 and continued through the publication of these Interim Guidelines. Independent review and guidance was provided by an advisory panel comprised of experts from industry, practice and academia. Every reasonable effort has been made to assure the efficacy of the Interim Guidelines contained herein. However, users are cautioned that research into the behavior of these structures is continuing. The results of this research may invalidate or suggest the need for modification of recommendations contained herein. No warranty is offered with regard to the recommendations contained herein, either by the Federal Emergency Management Agency, the SAC Joint Venture, the individual joint venture partners, their directors, members or employees. These organizations and their employees do not assume any legal liability or responsibility for the accuracy, completeness, or usefulness of any of the information, products or processes included in this publication. The reader is cautioned to carefully review the material presented herein. Such information must be used together with sound engineering judgment when applied to specific engineering projects. These Interim Guidelines have been developed by the SAC Joint Venture with funding provided by the Federal Emergency Management Agency, under contract number EMW-95-K-4672.

Acknowledgment The SAC Joint Venture wishes to offer grateful acknowledgment to the Federal Emergency Management Agency (FEMA); FEMA’s project officer, Mr. Michael Mahoney; and technical advisor, Dr. Robert D. Hanson. Following the discovery of severe damage to steel moment-resisting frame buildings in the Northridge Earthquake, this agency recognized the significance of this issue to the engineering community as well as the public at large, and acted rapidly to provide the necessary funding to allow these Interim Guidelines to be developed, published and distributed. Without the support of this agency, the important information and material presented herein could not have been made available. SAC also wishes to recognize the American Institute of Steel Construction, the American Iron and Steel Institute, the American Welding Society, the California Office of Emergency Services, the Lincoln Electric Company, the Structural Shape Producers Council, and the many engineers, fabricators, inspectors and researchers who contributed services, materials, data and invaluable advice and assistance in the production of this document.

The SAC Joint Venture 555 University Avenue, Suite 126 Sacramento, CA 95825 phone 916-427-3647; facsimile 916-568-0677

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Interim Guidelines: Evaluation, Repair, Modification and Design of Steel Moment Frames

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Interim Guidelines: Evaluation, Repair, Modification andDesign of Steel Moment Frames

OVERVIEW The Northridge Earthquake of January 17, 1994, dramatically demonstrated that the prequalified, welded beam-to-column moment connection used for Special Moment Resisting Frames is much more susceptible to damage than was previously thought. The stability of moment frame structures in earthquakes is dependent on the capacity of the beam-column connection to remain intact and to resist tendencies to rotate, induced by the swaying of the building. These connections were believed to be ductile and capable of withstanding repeated cycles of large inelastic deformation. Although many affected connections were not damaged, a wide spectrum of unexpected brittle connection damage did occur, ranging from minor cracking observable only by detailed nondestructive testing (NDT) to completely severed columns. The most commonly observed damage occurred at the welds of girder bottom flanges to columns. Complete brittle fractures of the girder flange to column connections occurred in some cases. While no casualties or collapses occurred as a result of these connection failures, and some welded steel moment frame (WSMF) buildings were not damaged, the incidence of damage was sufficiently high in regions of strong motion to cause wide-spread concern by structural engineers and building officials. No comprehensive tabulation is yet available to determine how many steel buildings were damaged in the Northridge Earthquake. More than 100 damaged buildings have been identified so far, including hospitals and other health care facilities, government, civic and private offices, cultural facilities, residential structures, and commercial and industrial buildings. The effect of these observations has been a loss of confidence in the procedures used in the past to design and construct welded connections in steel moment frames, and a concern that structures incorporating these connections may not be adequately safe. It must be understood that the structural engineering community was surprised by the performance of these modern, code conforming structures. Prior to the discovery of this damage, many thought that WSMF structures were nearly invulnerable to earthquake damage. The unexpected brittle fracturing and attendant loss of connection strength resulted in serious degradation of the overall lateral-load-resisting capability of some affected buildings. Further, the ability of existing WSMF buildings to withstand earthquake-induced ground motion is now understood to be significantly less than that previously assumed. Research conducted to date has identified some, but probably not all, of the factors leading to this observed unsatisfactory behavior. At the same time, this research has indicated methods that can be used to improve the ability of these critical connections to more reliably withstand multiple, large, inelastic cycles. These include alterations in the basic design approach as well as improved practices for specification and control of materials and workmanship. While the work is not yet complete, and future research is likely to provide both more reliable and more economical methods of improving the performance of these structures, the current investigations have led to many design and retrofit measures that can be used today to provide more reliable and consistent performance of these buildings than occurred in the Northridge

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Interim Guidelines: Evaluation, Repair, Modification and Design of Steel Moment Frames

Earthquake. These are presented in these Interim Guidelines. They should not, however, be viewed as the only way of achieving these results, and the exercise of independent engineering judgment and alternative rational analytical approaches should be considered. It is anticipated that additional studies, planned by SAC and others, will lead to further improvements in our understanding of the problems, ability to predict probable earthquake performance and methods to design and construct more reliable structures. There are many complex issues involved in the evaluation, repair, modification and design of WSMF buildings for reliable earthquake performance. These include considerations of metallurgy, welding, fracture mechanics, systems behavior, and basic issues related to fabrication and erection practice. Much remains to be learned in each of these areas. Engineers not familiar with the issues involved are cautioned to obtain qualified advice and third party review when contemplating design decisions that represent significant departures from these Interim Guidelines. The current judgment given in these Interim Guidelines is that the historic practices used for the design and construction of WSMF connections do not provide adequate levels of building reliability and safety and should not continue to be used in the construction of new buildings intended to resist earthquake ground shaking through inelastic behavior. The risk to public safety associated with the continued use of existing WSMF buildings is probably no greater than that associated with many other types of existing buildings with known seismic vulnerabilities, which are not currently the subject of mandatory seismic rehabilitation programs. The earthquake risk of WSMF buildings, in general, may be evaluated in accordance with the following general principles: 1.

The historic practices and designs used for WSMF connections are no longer appropriate for design and construction of new steel buildings likely to experience large inelastic demands from earthquakes. Until research is completed, and better information becomes available, the procedures contained in these Interim Guidelines for the design of new buildings should be used in their place. The use of alternative systems, including bolted construction, braced construction, and moment-resisting frames incorporating partially restrained (PR) joints could also be considered, but are not directly addressed by these Interim Guidelines.

2.

As a class, existing undamaged WSMF buildings appear to have a lower risk of collapse than many other types of buildings with known seismic vulnerabilities, the performance of which is currently implicitly accepted. Consequently, mandated or emergency programs to upgrade the performance of these buildings does not appear necessary to achieve levels of life safety protection currently tolerated by society. However, the risk of collapse is definitely greater than previously thought. Individual owners should be made aware of the increased level of seismic risk and encouraged to perform modifications to provide more reliable seismic performance, particularly in building housing many persons, or in critical occupancies.

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Interim Guidelines: Evaluation, Repair, Modification andDesign of Steel Moment Frames

3.

Following strong earthquake-induced ground shaking, WSMF buildings incorporating the vulnerable welded moment-resisting connections should be subjected to rigorous evaluations to determine the extent and implications of any damage sustained. These Interim Guidelines may be used to determine which buildings should be evaluated, and for developing an appropriate program to perform such evaluations.

4.

Structural repair and modification programs for damaged WSMF buildings should consider the seismic risk inherent in the building including the local seismicity, site geologic conditions, the building’s individual construction characteristics, intended occupancy and the costs associated with alternative actions. The Interim Guidelines provided in this document for repair can restore a building’s pre-earthquake seismic resistance, but not significantly improve its original levels of safety or reduce the inherent seismic risk. The Interim Guidelines provided in this document for structural modification (upgrading) can be used both to improve building safety and reduce seismic risk. Except in those cases where regulation sets minimum acceptable standards for repair, the ultimate responsibility for deciding whether a building should be modified for improved performance lies with the building owner. It is the structural engineer’s responsibility to provide the owner with sufficient information upon which to base a decision. The following may be considered by engineers to provide such information: a)

When a WSMF has experienced damage to only a few of its moment-resisting connections this damage should be repaired in an expeditious manner. Repair to the original configuration, with proper materials and workmanship, will essentially restore the structure’s original earthquake-resisting capacity. However, it will not result in any significant improvement in the building’s future performance. The fact that the building experienced only light damage should not be considered a demonstration that the building has a high degree of earthquake resistance and in future earthquakes either more or less damage may be experienced, depending on the particular characteristics of the event. Connections which have been damaged can be economically modified at the same time that repairs are made. However, in buildings where damage is limited, modification of the few damaged connections will not result in any significant improvement in the future earthquake performance of the building. Modification of connections throughout the structure, or provision of an alternative lateral force resisting system should be considered as a method of substantially improving probable building performance; however, this will entail a significant cost premium over the basic repair project.

b)

When a WSMF has experienced damage to a significant percentage of its momentresisting connections (on the order of 25% in any direction of resistance), in addition to repair, consideration should be given to modifying the configuration of the individual damaged connections and possibly some or all of the undamaged connections to provide improved performance in the future. Modification of only vii

Interim Guidelines: Evaluation, Repair, Modification and Design of Steel Moment Frames

some connections, and not others, may cause an increase in vulnerability, due to unbalanced concentrations of stiffness and strength. Therefore, such partial modifications should be made with due consideration of the effect on overall system behavior. Repair and/or modification should be completed expeditiously by structural engineers who are experienced in the design of WSMF buildings and understand the features which caused the observed damage. c)

When a WSMF building has had many seriously damaged connections (on the order of 50% in direction of resistance), owners should be informed that this damage may have highlighted basic deficiencies in the existing structural system, or a geologic feature which unusually amplifies site motion. In such cases the existing system should be both repaired and modified to provide an acceptably reliable structural system. Modifications may consist either of local reinforcement of individual connections and/or alteration of the structure’s basic lateral-forceresisting system. Such modifications could include addition of braced frames, shear walls, energy dissipation devices, base isolation and similar measures.

These principles are for regular buildings that have good characteristics of design, materials, and construction workmanship. Buildings with clear and apparent seismic deficiencies pose substantial life safety hazards regardless of the type of structural system employed, or material type. Such deficiencies include incomplete load paths, incompatible structural systems, irregular configurations such as soft or weak stories or torsional irregularity, and improper construction practices. Any such deficiencies found in a WSMF should be corrected.

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Interim Guidelines: Evaluation, Repair, Modification andDesign of Steel Moment Frames

TABLE OF CONTENTS

1

2

3

4

FOREWORD AND DISCLAIMER ACKNOWLEDGMENT OVERVIEW

iii iii v

INTRODUCTION 1.1 Purpose 1.2 Scope 1.3 Background 1.4 The SAC Joint Venture 1.5 Sponsors 1.6 Summary of Phase 1 Research 1.7 Intent 1.8 Limitations 1.9 Use of the Guidelines

1-1 1-2 1-3 1-10 1-11 1-11 1-14 1-14 1-15

DEFINITIONS, ABBREVIATIONS & NOTATION 2.1 Definitions 2.1.1 Administrative 2.1.2 Technical 2.2 Abbreviations 2.3 Notations

2-1 2-1 2-3 2-9 2-11

CLASSIFICATION AND IMPLICATIONS OF DAMAGE 3.1 Summary of Earthquake Damage 3.2 Damage Types 3.2.1 Girder Damage 3.2.2 Column Flange Damage 3.2.3 Weld Damage, Defects and Discontinuities 3.2.4 Shear Tab Damage 3.2.5 Panel Zone Damage 3.2.6 Other Damage 3.3 Safety Implications 3.4 Economic Implications

3-1 3-2 3-3 3-5 3-7 3-9 3-10 3-11 3-12 3-14

POST-EARTHQUAKE EVALUATION 4.1 Scope 4.2 Preliminary Evaluation 4.2.1 Evaluation Process 4.2.1.1 Ground Motion 4.2.1.2 Additional Indicators 4.2.2 Evaluation Schedule

4-1 4-2 4-3 4-3 4-4 4-5

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Interim Guidelines: Evaluation, Repair, Modification and Design of Steel Moment Frames

4.3

4.4 4.5

5

4.2.3 Connection Inspections 4.2.3.1 Analytical Evaluation 4.2.3.2 Buildings with Enhanced Connections 4.2.4 Previous Evaluations and Inspections Detailed Evaluation Procedure 4.3.1 Eight Step Inspection and Evaluation Procedure 4.3.2 Step 1 - Categorize Connections By Group 4.3.3 Step 2 - Select Samples of Connections for Inspection 4.3.3.1 Method A - Random Selection 4.3.3.2 Method B - Deterministic Selection 4.3.3.3 Method C - Analytical Selection 4.3.4 Step 3- Inspect the Selected Samples of Connections 4.3.4.1 Characterization of Damage 4.3.5 Step 4 - Inspect Connections Adjacent to Damaged Connections 4.3.6 Step 5 - Determine Average Damage Index for the Group 4.3.7 Step 6 - Determine the Probability that the Connections in a Group at a Floor Level Sustained Excessive Damage 4.3.7.1 Some Connections In Group Not Inspected 4.3.7.2 All Connections in Group Inspected 4.3.8 Step 7 - Determine Recommended Recovery Strategies for the Building 4.3.9 Step 8 - Evaluation Report Alternative Group Selection for Torsional Response Qualified Independent Engineering Review 4.5.1 Timing of Independent Review 4.5.2 Qualifications and Terms of Employment 4.5.3 Scope of Review 4.5.4 Reports 4.5.5 Responses and Corrective Actions 4.5.6 Distribution of Reports 4.5.7 Engineer of Record 4.5.8 Resolution of Differences

POST-EARTHQUAKE INSPECTION 5.1 Connection Types Requiring Inspection 5.1.1 Welded Steel Moment Frame (WSMF) Connections 5.1.2 Gravity Connections 5.1.3 Other Connection Types 5.2 Preparation 5.2.1 Preliminary Document Review and Evaluation 5.2.1.1 Document Collection and Review 5.2.1.2 Preliminary Building Walk-Through 5.2.1.3 Structural Analysis 5.2.1.4 Vertical Plumbness Check x

4-6 4-7 4-7 4-8 4-10 4-11 4-12 4-13 4-14 4-16 4-17 4-18 4-18 4-21 4-23 4-23 4-23 4-25 4-26 4-28 4-30 4-32 4-33 4-33 4-33 4-34 4-34 4-34 4-34 4-35

5-1 5-1 5-3 5-3 5-4 5-4 5-4 5-4 5-4 5-5

Interim Guidelines: Evaluation, Repair, Modification andDesign of Steel Moment Frames

5.3

6

5.2.2 Connection Exposure Inspection Program 5.3.1 Visual Inspection (VI) 5.3.1.1 Top Flange 5.3.1.2 Bottom Flange 5.3.1.3 Column and Continuity Plates 5.3.1.4 Beam Web Shear Connection 5.3.2 Nondestructive Testing (NDT) 5.3.3 Inspector Qualification 5.3.4 Post-Earthquake Field Inspection Report 5.3.5 Written Report

POST-EARTHQUAKE REPAIR AND MODIFICATION 6.1 Scope 6.2 Shoring 6.2.1 Investigation 6.2.2 Special Requirements 6.3 Repair Details 6.3.1 Approach 6.3.2 Weld Fractures - Type W Damage 6.3.3 Column Fractures - Type C1 - C5 and P1 - P6 6.3.4 Column Splice Fractures - Type C7 6.3.5 Girder Flange Fractures - Type G3-G5 6.3.6 Buckled Girder Flanges - Type G1 6.3.7 Buckled Column Flanges - Type C6 6.3.8 Gravity Connections 6.3.9 Reuse of Bolts 6.3.10 Welding Specification 6.4 Preparation 6.4.1 Welding Procedure Specifications 6.4.2 Welder Training 6.4.3 Welder Qualifications 6.4.4 Joint Mock-ups 6.4.5 Repair Sequence 6.4.6 Concurrent Work 6.4.7 Quality Control/Quality Assurance 6.5 Execution 6.5.1 Introduction 6.5.2 Girder Repair 6.5.3 Weld Repair (Types W1, W2, or W3) 6.5.4 Column Flange Repairs - Type C2 6.6 Structural Modification 6.6.1 Definition of Modification 6.6.2 Damaged vs. Undamaged Connections xi

5-6 5-7 5-7 5-8 5-9 5-9 5-9 5-9 5-11 5-12 5-13

6-1 6-2 6-2 6-2 6-2 6-3 6-3 6-6 6-9 6-10 6-11 6-12 6-13 6-13 6-14 6-14 6-13 6-15 6-15 6-15 6-15 6-16 6-16 6-16 6-16 6-20 6-21 6-22 6-22 6-22 6-24

Interim Guidelines: Evaluation, Repair, Modification and Design of Steel Moment Frames

6.6.3 6.6.4 6.6.5 6.6.6

Criteria Strength Plastic Rotation Capacity Connection Qualification and Design 6.6.6.1 Qualification Test Protocol 6.6.6.2 Acceptance Criteria 6.6.6.3 Calculations 6.6.6.3.1 Material Strength Properties 6.6.6.3.2 Determine Plastic Hinge Location 6.6.6.3.3 Determine Probable Plastic Moment at Hinges 6.6.6.3.4 Determine Beam Shear 6.6.6.3.5 Determine Strength Demands on Connection 6.6.6.3.6 Check Strong Column - Weak Beam Conditions 6.6.6.3.7 Check Column Panel Zone 6.6.7 Modification Details 6.6.7.1 Haunch at Bottom Flange 6.6.7.2 Top and Bottom Haunch 6.6.7.3 Cover Plate Sections 6.6.7.4 Upstanding Ribs 6.6.7.5 Side-Plate Connections

7

NEW CONSTRUCTION 7.1 Scope 7.2 General - Welded Steel Frame Design Criteria 7.2.1 Criteria 7.2.2 Strength 7.2.3 Configuration 7.2.4 Plastic Rotation Capacity 7.2.5 Redundancy 7.2.6 System Performance 7.2.7 Special Systems 7.3 Connection Design and Qualification Procedures - General 7.3.1 Connection Performance Intent 7.3.2 Qualification by Testing 7.3.3 Design by Calculation 7.4 Guidelines for Connection Qualification by Testing 7.4.1 Testing Protocol 7.4.2 Acceptance Criteria 7.5 Guidelines for Connection Design by Calculation 7.5.1 Material Strength Properties 7.5.2 Design Procedure 7.5.2.1 Determine Plastic Hinge Locations 7.5.2.2 Determine Probable Plastic Moment at Hinge 7.5.2.3 Determine Shear at Plastic Hinge xii

6-25 6-27 6-28 6-30 6-30 6-32 6-32 6-33 6-35 6-35 6-36 6-37 6-38 6-38 6-39 6-39 6-41 6-42 6-44 6-45

7-1 7-3 7-3 7-4 7-4 7-7 7-9 7-10 7-10 7-11 7-11 7-11 7-11 7-13 7-13 7-14 7-15 7-15 7-17 7-17 7-18 7-20

Interim Guidelines: Evaluation, Repair, Modification andDesign of Steel Moment Frames

7.6 7.7 7.8

7.9

7.10

8

7.5.2.4 Determine Strength Demands at Critical Sections 7.5.2.5 Check for Strong Column - Weak Beam Condition 7.5.2.6 Check Column Panel Zone Metallurgy & Welding Quality Control / Quality Assurance Guidelines on Other Connection Design Issues 7.8.1 Design of Panel Zones 7.8.2 Design of Web Connections to Column Flanges 7.8.3 Design of Continuity Plates 7.8.4 Design of Weak Column and Weak Way Connections Moment Frame Connections for Consideration in New Construction 7.9.1 Cover Plate Connections 7.9.2 Flange Rib Connections 7.9.3 Bottom Haunch Connections 7.9.4 Top and Bottom Haunch Connections 7.9.5 Side-Plate Connections 7.9.6 Reduced Beam Section Connections 7.9.7 Slip-Friction Energy Dissipating Connections 7.9.8 Column Tree Connections 7.9.9 Slotted Web Connections Other Types of Welded Connection Structures 7.10.1 Eccentrically Braced Frames (EBF) 7.10.2 Dual Systems 7.10.3 Welded Base Plate Details 7.10.4 Vierendeel Truss Systems 7.10.5 Moment Frame Tubular Systems 7.10.6 Welded Connections of Collectors, Ties and Diaphragm Chords 7.10.7 Welded Column Splices 7.10.8 Built-up Moment Frame Members

METALLURGY & WELDING 8.1 Parent Materials 8.1.1 Steels 8.1.2 Chemistry 8.1.3 Tensile/Elongation Properties 8.1.4 Toughness Properties 8.1.5 Lamellar Discontinuities 8.2 Welding 8.2.1 Welding Process 8.2.2 Welding Procedures 8.2.3 Welding Filler Metals 8.2.4 Preheat and Interpass Temperatures 8.2.5 Postheat 8.2.6 Controlled Cooling xiii

7-20 7-21 7-22 7-22 7-23 7-23 7-23 7-24 7-24 7-25 7-26 7-27 7-29 7-30 7-31 7-32 7-35 7-36 7-37 7-38 7-39 7-40 7-40 7-41 7-41 7-42 7-42 7-43 7-43

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Interim Guidelines: Evaluation, Repair, Modification and Design of Steel Moment Frames

8.2.7 Metallurgical Stress Risers 8.2.8 Welding Preparation & Fit-up 9

10

11

12

8-17 8-17

QUALITY CONTROL/QUALITY ASSURANCE 9.1 Quality Control 9.1.1 General 9.1.2 Inspector Qualification 9.1.3 Duties 9.1.4 Records 9.1.5 Engineer Obligations 9.1.6 Contractor Obligations 9.1.7 Extent of Testing 9.2 Quality Assurance & Special Inspection 9.2.1 General 9.2.2 Inspector Qualifications 9.2.3 Duties 9.2.4 Records 9.2.5 Engineer Obligations 9.2.6 Contractor Obligations 9.2.7 Extent of QA Testing

9-1 9-1 9-1 9-1 9-1 9-2 9-2 9-3 9-4 9-4 9-4 9-4 9-4 9-5 9-5 9-5

VISUAL INSPECTION 10.1 Personnel Qualification 10.2 Written Practice 10.3 Duties

10-1 10-1 10-2

NONDESTRUCTIVE TESTING 11.1 Personnel 11.1.1 Qualification 11.1.2 Written Practice 11.1.3 Certification 11.1.4 Recertification 11.2 Execution 11.2.1 General 11.2.2 Magnetic Particle Testing (MT) 11.2.3 Liquid Penetrant Testing (PT) 11.2.4 Radiographic Testing (RT) 11.2.5 Ultrasonic Testing (UT)

11-1 11-1 11-2 11-2 11-2 11-2 11-2 11-3 11-4 11-4 11-4

REFERENCES

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Interim Guidelines: Evaluation, Repair, Modification and Design of Steel Moment Frames

Chapter 1 - Introduction

1. INTRODUCTION These Interim Guidelines apply to welded steel moment frame (WSMF) structures subject to large inelastic demands from earthquakes. They provide recommended methods for: determining which buildings should be subjected to detailed post-earthquake evaluations; developing a program for post-earthquake visual and non-destructive inspections of buildings suspected to have damage; evaluating the effect of discovered damage on residual building safety; identifying appropriate strategies for continued occupancy, structural repair and/or modification of damaged buildings; and designing and constructing new buildings. These recommendations are based on an initial, Phase 1, program of research that included collection and analysis of data on buildings damaged by the Northridge Earthquake; detailed structural analyses of damaged and undamaged buildings; review of past literature on relevant research; and laboratory testing of large-scale connection assemblies. They were developed by a group of researchers and practicing engineers, with assistance and consultation from experts in metallurgy, fracture mechanics, welding, design, structural steel production, fabrication erection and inspection. A significant body of valuable information is presented in these Interim Guidelines, which can be used today to provide improved reliability in welded steel moment frame structures. However, much additional research remains to be performed. The parameters controlling the performance of welded moment resisting connections are not yet fully understood, nor has consensus been obtained on all recommendations contained herein. Engineers engaged in the design of WSMF structures are advised to be watchful for new developments in the future. Although portions of this document are written in code-like language, it is not a building code, nor is it intended to be used as such. Rather, it is intended to provide engineers and building officials with information on what is known at the present time with regard to these structures, and to provide a series of recommendations that can be used on an interim basis to assist in practice. The use of these Interim Guidelines is not intended to serve as a substitute for the application of informed engineering judgment, nor should they be used to prevent the application of such judgment in particular engineering applications. 1.1 Purpose These Interim Guidelines have been prepared by the SAC Joint Venture to provide practicing engineers and building officials with: • understanding of the types of damage buildings incorporating fully restrained (FR) welded steel moment frame (WSMF) connections may experience in strong earthquakes, and the potential implications of such damage; • a methodology for post-earthquake inspection of existing WSMF buildings, to determine if significant structural damage has occurred;

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Interim Guidelines: Evaluation, Repair, Modification and Design of Steel Moment Frames

Chapter 1 - Introduction

• an approach for characterizing the relative severity of damage to a WSMF and to determine appropriate occupancy and repair strategies; • methods of repair for fractured, yielded and buckled elements in WSMF buildings and structures; • design approaches for modifications to existing WSMF buildings and structures with FR connections to improve performance in future earthquakes; and • design approaches for connections in new WSMF buildings and structures for improved performance in future strong earthquakes. Earlier publications by the SAC Joint Venture on this topic include a series of three Design Advisories and the proceedings of an International Workshop (SAC-1994-1). The International Workshop, held in October, 1994 was attended by more than 100 invited researchers, practicing engineers, representatives of industry, and government agencies, and provided an initial focus to the investigations of fractures sustained by welded steel moment-resisting buildings in the Northridge Earthquake. Design Advisory No. 1 (SAC-1994-2) and Design Advisory No. 2 (SAC-1994-3) contained collections of papers and topical reports prepared by practicing engineers, building officials, industry groups and researchers, suggesting factors which contributed to the observed damage, methods of repairing damage and designing new structures to avoid such damage in the future. Design Advisory No. 3 (SAC-1995) categorized the information presented in the previous advisories into a series of discrete engineering issues and presented the consensus opinions of a panel of practicing engineers, researchers and industry representatives with regard to appropriate response to these issues. Dissenting opinions and commentary were also provided as were specific recommendations for directed research required to provide resolution to a number of these issues. These Interim Guidelines provide specific engineering recommendations based on the results of an initial limited program of research. This research included evaluation of the characteristics of ground motion experienced throughout the Los Angeles area during the Northridge Earthquake, projection of potential ground motions resulting from future earthquakes in this region, analytical investigation of both damaged and undamaged structures affected by the Northridge Earthquake for their response to a range of ground motions, laboratory testing of representative beam-column connections in undamaged, damaged, repaired, and reinforced states, parametric studies on the effects of strain rate and toughness on connection performance, surveys of engineers and building owners to collect data on the extent of damage sustained in the Northridge Earthquake, and statistical evaluation of the data collected and engineering analysis of all of the above. 1.2 Scope These Interim Guidelines are applicable to steel moment-resisting frame structures incorporating fully restrained connections in which the girder flanges are welded to the columns and which are subject to significant inelastic demands from strong earthquake ground shaking. Recommendations are provided with regard to:

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Interim Guidelines: Evaluation, Repair, Modification and Design of Steel Moment Frames

Chapter 1 - Introduction

• Designation of buildings to be inspected following an earthquake producing strong ground motion; • Scope of inspection for buildings so designated; • Appropriate types of repairs for damaged buildings; • Methods to modify buildings to reduce the probability of connection fracture damage in future earthquake events; • Design of new Special Moment Resisting Frame (SMRF) buildings for seismic resistance; • Design of new Ordinary Moment Resisting Frame (OMRF) buildings located in Uniform Building Code (UBC) Seismic Zones 3 and 4 {National Earthquake Hazards Reduction Program (NEHRP) Map Areas 6 and 7}; and • Quality Assurance and Control in the repair, modification and construction of WSMF buildings. Commentary: The design recommendations contained in these Interim Guidelines are generally applicable to SMRF structures designed for earthquake resistance and to those OMRF structures located within UBC Seismic Zones 3 and 4 {NEHRP Map Areas 6 and 7}. The recommendations should be considered for the design of any welded steel moment frame structure that is desired to have a high degree of reliability for resisting earthquake induced forces. In particular, they should be considered for buildings occupied by a large number of people. Chapter 7 provides further guidelines on this applicability. 1.3 Background Following the January 17, 1994 Northridge, California Earthquake, more than 100 steel buildings with welded moment-resisting frames were found to have experienced beam-to-column connection fractures. The damaged structures cover a wide range of heights ranging from one story to 26 stories; and a wide range of ages spanning from buildings as old as 30 years of age to structures just being erected at the time of the earthquake. The damaged structures are spread over a large geographical area, including sites that experienced only moderate levels of ground shaking. Although relatively few such buildings were located on sites that experienced the strongest ground shaking, damage to these buildings was quite severe. Discovery of these extensive connection fractures, often with little associated architectural damage to the buildings, has been alarming. The discovery has also caused some concern that similar, but undiscovered damage may have occurred in other buildings affected by past earthquakes. Indeed, there are isolated reports of such damage. In particular, a publicly owned building at Big Bear Lake is known to have been damaged by the Landers-Big Bear, California sequence of earthquakes, and at least one building, under construction in Oakland, California at the time of the 1989 Loma Prieta Earthquake, was reported to have experienced such damage. 1-3

Interim Guidelines: Evaluation, Repair, Modification and Design of Steel Moment Frames

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WSMF construction is used commonly throughout the United States and the world, particularly for mid- and high-rise construction. Prior to the Northridge Earthquake, this type of construction was considered one of the most seismic-resistant structural systems, due to the fact that severe damage to such structures had rarely been reported in past earthquakes and there was no record of earthquakeinduced collapse of such buildings, constructed in accordance with contemporary US practice. However, the widespread severe structural damage which occurred to such structures in the Northridge Earthquake calls for re-examination of this premise. The basic intent of the earthquake resistive design provisions contained in the building codes is to protect the public safety, however, there is also an intent to control damage. The developers of the building code provisions have explicitly set forth three specific performance goals for buildings designed and constructed to the code provisions (SEAOC - 1990). These are to provide buildings with the capacity to • resist minor earthquake ground motion without damage; • resist moderate earthquake ground motion without structural damage but possibly some nonstructural damage; and • resist major levels of earthquake ground motion, having an intensity equal to the strongest either experienced or forecast for the building site, without collapse, but possibly with some structural as well as nonstructural damage. In general, WSMF buildings in the Northridge Earthquake met the basic intent of the building codes, to protect life safety. However, many of these buildings experienced significant damage that could be viewed as failing to meet the intended performance goals with respect to damage control. Further, some members of the engineering profession (SEAOC - 1995b) and government agencies (Seismic Safety Commission - 1995) have stated that even these performance goals, are inadequate for society’s current needs. WSMF buildings are designed to resist earthquake ground shaking, based on the assumption that they are capable of extensive yielding and plastic deformation, without loss of strength. The intended plastic deformation consists of plastic rotations developing within the beams, at their connections to the columns, and is theoretically capable of resulting in benign dissipation of the earthquake energy delivered to the building. Damage is expected to consist of moderate yielding and localized buckling of the steel elements, not brittle fractures. Based on this presumed behavior, building codes require a minimum lateral design strength for WSMF structures that is approximately 1/8 that which would be required for the structure to remain fully elastic. Supplemental provisions within the building code, intended to control the amount of interstory drift sustained by these flexible frame buildings, typically result in structures which are substantially stronger than this minimum requirement and in zones of moderate seismicity, substantial overstrength may be present to accommodate wind and gravity load design conditions. In zones of high seismicity, most such structures designed to minimum code criteria will not start to exhibit plastic behavior until ground motions are experienced that are 1/3 to 1/2 the severity anticipated as a design basis. This design approach has been developed based on historical precedent, the observation of steel building performance in past earthquakes, and limited research that 1-4

Interim Guidelines: Evaluation, Repair, Modification and Design of Steel Moment Frames

Chapter 1 - Introduction

has included laboratory testing of beam-column models, albeit with mixed results, and non-linear analytical studies. Observation of damage sustained by buildings in the Northridge Earthquake indicates that contrary to the intended behavior, in many cases brittle fractures initiated within the connections at very low levels of plastic demand, and in some cases, while the structures remained elastic. Typically, but not always, fractures initiated at, or near, the complete joint penetration (CJP) weld between the beam bottom flange and column flange (Figure 1-1). Once initiated, these fractures progressed along a number of different paths, depending on the individual joint conditions. Figure 1-1 indicates just one of these potential fracture growth patterns. Investigators initially identified a number of factors which may have contributed to the initiation of fractures at the weld root including: notch effects created by the backing bar which was commonly left in place following joint completion; sub-standard welding that included excessive porosity and slag inclusions as well as incomplete fusion; and potentially, preearthquake fractures resulting from initial shrinkage of the highly restrained weld during cool-down. Such problems could be minimized in future construction, with the application of appropriate welding procedures and more careful exercise of quality control during the construction process. However, it is now known that these were not the only causes of the fractures which occurred. Column flange Fused zone Beam flange

Backing bar Fracture

Figure 1-1 - Common Zone of Fracture Initiation in Beam -Column Connection Current production processes for structural steel shapes result in inconsistent strength and deformation capacities for the material in the through-thickness direction. Non-metallic inclusions in the material, together with anisotropic properties introduced by the rolling process can lead to lamellar weakness in the material. Further, the distribution of stress across the girder flange, at the connection to the column is not uniform. Even in connections stiffened by continuity plates across the panel zone, significantly higher stresses tend to occur at the center of the flange, where the column web produces a local stiffness concentration. Large secondary stresses are also induced into the girder flange to column flange joint by kinking of the column flanges resulting from shear deformation of the column panel zone. The dynamic loading experienced by the moment-resisting connections in earthquakes is characterized by high strain tension-compression cycling. Bridge engineers have long recognized that the dynamic loading associated with bridges necessitates different connection details in order to provide improved fatigue resistance, as compared to traditional building design that is subject to “static” 1-5

Interim Guidelines: Evaluation, Repair, Modification and Design of Steel Moment Frames

Chapter 1 - Introduction

loading due to gravity and wind loads. While the nature of the dynamic loads resulting from earthquakes is somewhat different than the high cycle dynamic loads for which fatigue-prone structures are designed, similar detailing may be desirable for buildings subject to seismic loading. In design and construction practice for welded steel bridges, mechanical and metallurgical notches should be avoided because they may be the initiators of fatigue cracking. As fatigue cracks grow under repetitive loading, a critical crack size may be reached whereupon the material toughness (which is a function of temperature) may be unable to resist the onset of brittle (unstable) crack growth. The beam-to-column connections in WSMF buildings are comparable to category C or D bridge details that have a reduced allowable stress range as opposed to category B details for which special metallurgical, inspection and testing requirements are applied. The rapid rate of loading imposed by seismic events, and the complete inelastic range of tension-compression-tension loading applied to these connections is much more severe than typical bridge loading applications. The mechanical and metallurgical notches or stress risers created by the beam-column weld joints are a logical point for fracture problems to initiate. This, coupled with the tri-axial restraint provided by the beam web and the column flange, is a recipe for brittle fracture. During the Northridge Earthquake, once fractures initiated in beam-column joints, they progressed in a number of different ways. In some cases, the fractures initiated but did not grow, and could not be detected by visual observation. In other cases, the fractures progressed completely through the thickness of the weld, and if fireproofing was removed, the fractures were evident as a crack through exposed faces of the weld, or the metal just behind the weld (Figure 1-2a). Other fracture patterns also developed. In some cases, the fracture developed into a through-thickness failure of the column flange material behind the CJP weld (Figure 1-2b). In these cases, a portion of the column flange remained bonded to the beam flange, but pulled free from the remainder of the column. This fracture pattern has sometimes been termed a “divot” or “nugget” failure. A number of fractures progressed completely through the column flange, along a near horizontal plane that aligns approximately with the beam lower flange (Figure 1-3a). In some cases, these fractures extended into the column web and progressed across the panel zone Figure (1-3b). Investigators have reported some instances where columns fractured entirely across the section.

a. Fracture at Fused Zone

b. Column Flange “Divot” Fracture

Figure 1-2 - Fractures of Beam to Column Joints 1-6

Interim Guidelines: Evaluation, Repair, Modification and Design of Steel Moment Frames

Chapter 1 - Introduction

a. Fractures through Column Flange

b. Fracture Progresses into Column Web

Figure 1-3 - Column Fractures Once these fractures have occurred, the beam - column connection has experienced a significant loss of flexural rigidity and capacity. Residual flexural strength and rigidity must be developed through a couple consisting of forces transmitted through the remaining top flange connection and the web bolts. Initial research suggests that residual stiffness is approximately 20% of that of the undamaged connection and that residual strength varies from 10% to 40% of the undamaged capacity, when loading results in tensile stress normal to the fracture plane. When loading produces compression across the fracture plane, much of the original strength and stiffness remain. However, in providing this residual strength and stiffness, the beam shear connections can themselves be subject to failures, consisting of fracturing of the welds of the shear plate to the column, fracturing of supplemental welds to the beam web or fracturing through the weak section of shear plate aligning with the bolt holes (Figure 1-4).

Figure 1-4 - Vertical Fracture through Beam Shear Plate Connection Despite the obvious local strength impairment resulting from these fractures, many damaged buildings did not display overt signs of structural damage, such as permanent drifts, or extreme damage to architectural elements. Until news of the discovery of connection fractures in some buildings began 7

Interim Guidelines: Evaluation, Repair, Modification and Design of Steel Moment Frames

Chapter 1 - Introduction

to spread through the engineering community, it was relatively common for engineers to perform cursory post-earthquake evaluations of WSMF buildings and declare that they were undamaged. In order to reliably determine if a building has sustained connection damage it is necessary to remove architectural finishes and fireproofing and perform nondestructive examination including visual inspection and ultrasonic testing. Even if no damage is found, this is a costly process. Repair of damaged connections is even more costly. A few WSMF buildings have sustained so much connection damage that it has been deemed more practical to demolish the structures rather than to repair them. Immediately following the Northridge Earthquake, a series of tests of beam-column subassemblies were performed at the University of Texas at Austin, under funding provided by the AISC as well as private sources. The test specimens used heavy W14 column sections and deep (W36) beam sections commonly employed in some California construction. Initial specimens were fabricated using the standard prequalified connection specified by the Uniform Building Code (UBC). Section 2211.7.1.2 of UBC-94 {NEHRP-91 Section 10.10.2.3} specified this prequalified connection as follows: “2211.7.1.2 Connection strength. The girder top column connection may be considered to be adequate to develop the flexural strength of the girder if it conforms to the following: 1. the flanges have full penetration butt welds to the columns. 2. the girder web to column connection shall be capable of resisting the girder shear determined for the combination of gravity loads and the seismic shear forces which result from compliance with Section 2211.7.2.1. This connection strength need not exceed that required to develop gravity loads plus 3(Rw/8) times the girder shear resulting from the prescribed seismic forces. Where the flexural strength of the girder flanges is greater than 70 percent of the flexural strength of the entire section, (i.e. btf/(d-tf)Fy>0.7ZxFy) the web connection may be made by means of welding or high-strength bolting. For girders not meeting the criteria in the paragraph above, the girder web-to-column connection shall be made by means of welding the web directly or through shear tabs to the column. That welding shall have a strength capable of developing at least 20 percent of the flexural strength of the girder web. The girder shear shall be resisted by means of additional welds or friction-type slip-critical high strength bolts or both. and: 2211.7.2.1 Strength. The panel zone of the joint shall be capable of resisting the shear induced by beam bending moments due to gravity loads plus 1.85 times the prescribed seismic forces, but the shear strength need not exceed that required to develop 0.8ΣMs of the girders framing into the column flanges at the joint...”

In order to investigate the effects that backing bars and weld tabs had on connection performance, these were removed from the specimens prior to testing. Despite these precautions, the test specimens failed at very low levels of plastic loading. Following these tests at the University of Texas at Austin, reviews of literature on historic tests of these connection types indicated a significant failure rate in past

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Interim Guidelines: Evaluation, Repair, Modification and Design of Steel Moment Frames

Chapter 1 - Introduction

tests as well, although these had often been ascribed to poor quality in the specimen fabrication. It was concluded that the prequalified connection, specified by the building code, was fundamentally flawed and should not be used for new construction in the future. In retrospect, this conclusion may have been premature. When the first test specimens for that series were fabricated, the welder failed to follow the intended welding procedures. Further, no special precautions were taken to assure that the materials incorporated in the work had specified toughness. Some engineers, with knowledge of fracture mechanics, have suggested that if materials with adequate toughness are used, and welding procedures are carefully specified and followed, adequate reliability can be obtained from the traditional connection details. Others believe that the conditions of high triaxial restraint present in the beam flange to column flange joint (Blodgett - 1995) would prevent ductile behavior of these joints regardless of the procedure used to make the welds. Further they point to the important influence of the relative yield and tensile strengths of beam and column materials, and other variables, that can affect connection behavior. To date, there has not been sufficient research conducted to resolve this issue. In reaction to the University of Texas tests as well as the widespread damage discovered following the Northridge Earthquake, and the urging of the California Seismic Safety Commission, in September, 1994 the International Conference of Building Officials (ICBO) adopted an emergency code change to the 1994 edition of the Uniform Building Code (UBC-94) {1994 NEHRP Recommended Provisions Section 5.2}. This code change, jointly developed by the Structural Engineers Association of California, AISI and ICBO staff, deleted the prequalified connection and substituted the following in its place: “2211.7.1.2 Connection Strength. Connection configurations utilizing welds or high-strength bolts shall demonstrate, by approved cyclic test results or calculation, the ability to sustain inelastic rotation and develop the strength criteria in Section 2211.7.1.1 considering the effect of steel overstrength and strain hardening.” “2211.7.1.1 Required strength. The girder-to-column connection shall be adequate to develop the lesser of the following: 1. The strength of the girder in flexure. 2. The moment corresponding to development of the panel zone shear strength as determined from formula 11-1.”

Unfortunately, neither the required “inelastic rotation”, or calculation and test procedures are well defined by these code provisions. Design Advisory No. 3 (SAC-1995) included an Interim Recommendation (SEAOC-1995) that attempted to clarify the intent of this code change, and the preferred methods of design in the interim period until additional research could be performed and reliable acceptance criteria for designs re-established. The State of California similarly published a joint Interpretation of Regulations (DSA-OSHPD - 1994) indicating the interpretation of the current code requirements which would be enforced by the state for construction under its control. This applied only to the construction of schools and hospitals in the State of California. The intent of these Interim

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Interim Guidelines: Evaluation, Repair, Modification and Design of Steel Moment Frames

Chapter 1 - Introduction

Guidelines is to supplement these previously published documents and to provide updated recommendations based on the results of the limited directed research performed to date. 1.4 The SAC Joint Venture SAC is a joint venture of the Structural Engineers Association of California (SEAOC), the Applied Technology Council (ATC), and California Universities for Research in Earthquake Engineering (CUREe), formed specifically to address both immediate and long-term needs related to solving the problem of the WSMF connection. SEAOC is a professional organization comprised of more than 3,000 practicing structural engineers in California. The volunteer efforts of SEAOC’s members on various technical committees have been instrumental in the development of the earthquake design provisions contained in the Uniform Building Code as well as the National Earthquake Hazards Reduction Program (NEHRP) Recommended Provisions for Seismic Regulations for New Buildings. The Applied Technology Council is a non-profit organization founded specifically to perform problem-focused research related to structural engineering and to bridge the gap between civil engineering research and engineering practice. It has developed a number of publications of national significance including ATC 3-06, which served as the basis for the NEHRP Recommended Provisions. CUREe’s eight institutional members are: the University of California at Berkeley, the California Institute of Technology, the University of California at Davis, the University of California at Irvine, the University of California at Los Angeles, the University of California at San Diego, the University of Southern California, and Stanford University. this collection of university earthquake research laboratory, library, computer and faculty resources is the most extensive in the United States. The SAC Joint Venture allows these three organizations to combine their extensive and unique resources, augmented by subcontractor universities and organizations from around the nation, into an integrated team of practitioners and researchers, uniquely qualified to solve problems in earthquake engineering. The SAC Joint Venture developed a two phase program to solve the problem posed by the discovery of fractured steel moment connections following the Northridge Earthquake. Phase 1 of this program was intended to provide guidelines for the immediate post-Northridge problems of identifying damage in affected buildings and repairing this damage. In addition, Phase 1 included dissemination of the available design information to the professional community. It included convocation of a series of workshops and symposiums to define the problem; development and publication of a series of Design Advisories (SAC-1994-1, SAC-1994-2, SAC1995); limited statistical data collection, analytical evaluation of buildings and laboratory research; and the preparation of these Interim Guidelines. Phase 2 will consist of a longer term program of research and investigation to more carefully define the conditions which lead to the premature connection fractures and to develop sound guidelines for seismic design and detailing of improved or alternative WSMF connections for new buildings, as well as reliable retrofitting concepts for existing undamaged WSMF structures. The SAC Joint Venture’s unique capability to combine the efforts of researchers, industry representatives, code writers and practicing structural engineers is being applied to all major tasks. 1-10

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In addition, a Technical Oversight Committee and Technical Advisory Board with nationwide membership from the engineering, research and steel construction communities has been established to oversee the input of information, quality of technical investigations, and development of recommendations, and to assist in disseminating the information obtained. 1.5 Sponsors Funding for the Phase 1 SAC Steel Program was provided by the California Office of Emergency Services and the Federal Emergency Management Agency. Special efforts have been made to maintain a liaison with the engineering profession, researchers, the steel industry, fabricators, code writing organizations and model code groups, building officials, insurance and risk-management groups and federal and state agencies active in earthquake hazard mitigation efforts. SAC wishes to acknowledge the support and participation of each of the above groups as well as the American Iron and Steel Institute, the American Institute of Steel Construction, the Structural Shape Producers Council, the American Welding Society and the Lincoln Electric Company for the contribution of technical advice and assistance as well as material directly used in the research program. Acknowledgment is also made of the many engineers, fabricators, inspectors and researchers who contributed services and data for use in the development of these Guidelines. 1.6 Summary of Phase 1 Research These Interim Guidelines are based on the material presented in Design Advisory No. 3 (SAC1995), professional judgment and experience, a review of past relevant research, concurrent research being performed under grants provided by the National Science Foundation and supplemental information obtained in the SAC Phase 1 research program. This research included: • Collection of data on buildings damaged by the Northridge Earthquake. This consisted of the collection of detailed information on the configuration and detailing of WSMF buildings damaged by the Northridge Earthquake, together with data on the distribution, type and severity of damage within each structure. This work was conducted as an extension of an earlier survey, performed under funding from the National Institute of Standards and Technology (Youssef, et. al. - 1994). Data on a total of 89 buildings is available from these combined studies (Bonowitz & Youssef - 1995) • A telephone survey was conducted on a random sample of 200 steel framed buildings located within the zone which experienced estimated ground motion with a peak horizontal acceleration of 0.2g or greater during the Northridge Earthquake. The intent of this survey was to determine the geographic distribution of inspected, damaged and repaired structures in order to correlate damage with ground motion parameters and other factors. (Michael Durkin & Associates - 1995) • A series of interviews were conducted with engineers, inspectors, building officials and others engaged in the investigation and repair of a number of damaged WSMF buildings. The purpose of these interviews was to collect data on pertinent interpretations or trends noted by engineers and others engaged in this work. (Gates & Morden - 1995) 1-11

Interim Guidelines: Evaluation, Repair, Modification and Design of Steel Moment Frames

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• Maps of ground motion parameters(peak ground acceleration and pseudo spectral velocity at various periods) were developed for the San Fernando Valley and surrounding areas affected by strong ground motion in the Northridge Earthquake, based on fault rupture and ground motion propagation modeling techniques. Time histories of ground motion were developed for various discrete sites, using these same modeling techniques. These estimated ground motions were developed for use in comparing geographic distributions of damage with ground motion parameters, and as a basis for performing structural analyses of selected buildings. (Sommerville- 1995) • A fracture model element was developed for use with the DRAIN-2D, non-linear analysis software, to permit analytical simulation of the effect of beam-column connection fractures on overall structural behavior. (Campbell - 1995) • A series of linear and non-linear structural analyses were performed on eight WSMF buildings which were damaged by the Northridge Earthquake and on two WSMF buildings adjacent to two of these structures, which were not damaged. The purpose of these analyses was to explore the ability of analytical methods to predict the presence of damage within buildings as well as to predict specific locations within buildings where damage is likely to have occurred. In addition, these analyses were intended to indicate threshold demand levels at which damage is likely to have occurred, to provide information on the total demands developed in structures during response to various earthquake ground motions, and to explore the potential for earthquake induced collapse. (Krawinkler, et. al. 1995), (Engelhardt, et. al. - 1995a), (Hart, et. al. - 1995), (Kariotis & Eimani - 1995), (Anderson & Fillippou - 1995), (Naeim, et. al. - 1995), (Uang, et. al. - 1995), (Paret & Sasaki - 1995) • A series of parametric analytical investigations were performed to assess the influence of various ground motions and structural characteristics on seismic response of WSMF buildings. These included investigations involving hypothesized fractures of beam-column connections for various real and idealized frame structures subject to various intense ground motion records. The consequences of these ground motions were assessed as was the sensitivity of response to vertical ground motion and to various analytical modeling assumptions. (Iwan - 1995), (Hall - 1995), (Hart et. al. - 1995b), (Englehardt, et. al. 1995b), (Krawinkler, et. al. - 1995) • Four damaged beam-column connections were removed from a WSMF building which was affected by the Northridge Earthquake and subsequently demolished. These specimens were moved to a laboratory and subjected to testing to determine their residual strength and stiffness, for use in making assessments as to the consequences of fracture damage to overall building stability. Following this testing, the specimens were repaired and re-tested, to judge the effectiveness of the repair techniques. In addition, detailed building analyses were performed. (Anderson - 1995) • A total of 12 large scale beam-column assemblages were fabricated using typical preNorthridge detailing practice and following correct welding procedures. These were cycled inelastically, using a testing protocol similar to that indicated in ATC-24 (Applied

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Interim Guidelines: Evaluation, Repair, Modification and Design of Steel Moment Frames

Chapter 1 - Introduction

Technology Council - 1988) and experienced failure at low levels of plastic demand. Following initial testing and failure, the specimens were repaired using specifications followed by engineers in the Los Angeles area, or repaired and reinforced using details proposed by Los Angeles area engineers. The purpose of these tests was to explore whether initial structural capacity could be re-established in damaged structures by common repair techniques, and to determine the efficacy of proposed structural reinforcement techniques. (Popov et. al. - 1995), (Bertero and Whitaker- 1995), (Uang 1995b), (Engelhardt - 1995c) • Four large scale beam-column subassemblies were fabricated using selected details recommended in these guidelines for new construction and subjected to cyclic testing to failure. • A series of acoustic emission recordings were made on the large scale structural assemblages tested in the laboratory to assist in interpretation of the fracture sequence and to explore the ability of acoustic instrumentation techniques to identify damage in WSMF buildings affected by strong ground motion. (Thewalt - 1995), (Engelhardt, et. al. - 1995d) • A series of ambient vibration tests were performed on damaged buildings in order to determine the ability of low level vibration testing to be used as a method of detecting damage in WSMF buildings affected by strong ground motion, and to calibrate analytical models. (Beck - 1995) • Specimens from damaged connections in buildings affected by the Northridge Earthquake were removed from the buildings and subjected to metallurgic and fractographic analyses to determine the fracture mechanisms and effect of metallurgy on fracture behavior. (ATLSS - 1995a) • A series of moderate-scale “T” specimens were fabricated to simulate the connection of a beam bottom flange to a column flange in a major axis WSMF connection. These tests were performed to explore the ability to economically use moderate scale models to explore the behavior of large scale beam-column assemblages and also to perform parametric experimental studies on the effects of strain-rate on specimen behavior and the effects of weld metal notch-toughness and weld procedure on connection behavior. (ATLSS - 1995b) Additional information was collected from various other sources, including research performed under funding provided by the American Institute of Steel Construction, the National Science Foundation, and the National Institute of Standards and Technology, as well as testing performed as part of privately sponsored research (Allen, et. al. -1995, Jokerst - 1995) and lessons learned in the inspection, evaluation and repair of buildings which has taken place to date.

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1.7 Intent These Interim Guidelines are primarily intended for two different groups of potential users: a) Engineers engaged in evaluation, repair, and upgrade of existing WSMF buildings and in the design of new WSMF buildings incorporating either Special Moment-Resisting Frames or Ordinary Moment-Resisting Frames utilizing welded beam-column connections. The recommendations for new construction are applicable to all WSMF construction expected to resist earthquake demands through plastic behavior. b) Regulators and building departments responsible for control of the evaluation, repair, and occupancy of WSMF buildings that have been subjected to strong ground motion and for regulation of the design, construction, and inspection of new WSMF buildings. The fundamental goal of the information presented in these Interim Guidelines is to help identify and reduce the risks associated with earthquake-induced fractures in WSMF buildings through provision of timely information on how to inspect existing buildings for damage, repair damage if found, upgrade existing buildings and design new buildings. The information presented here primarily addresses the issue of beam-to-column connection integrity under the severe plastic demands that can be produced by building response to strong ground motion. Users are referred to the applicable provisions of the locally prevailing building code for information with regard to other aspects of building construction and earthquake damage control. 1.8 Limitations The information presented in these Interim Guidelines is based on limited research conducted since the Northridge Earthquake, review of past research and the considerable experience and judgment of the professionals engaged by SAC to prepare and review this document. Additional research on such topics as the effect of floor slabs on frame behavior, the effect of weld metal and base metal toughness, the efficacy of various beam-column connection details and the validity of current standard testing protocols for prediction of earthquake performance of structures are planned as part of the Phase 2 program and will likely provide important information not available at the time these Guidelines were formulated. Therefore, some recommendations cited herein may change as a result of forthcoming research results. Although the information presented is limited almost exclusively to technical engineering issues, it is well recognized that acceptable solutions to the steel WSMF problems must eventually address the non-technical concerns of building officials, owners, tenants, contractors, lenders, insurers, and legislators. It is hoped that by limiting the scope of this document to technical matters, this material can provide an objective basis for further discussion and debate. The information presented here is based on consideration of the typical building and WSMF frame configurations found in buildings today. Non-building structures (e.g. bridges, towers, or open frameworks) are not specifically addressed; however, to the extent that construction of these structures 1-14

Interim Guidelines: Evaluation, Repair, Modification and Design of Steel Moment Frames

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is similar to that for buildings, the information presented may be applicable. Beams and columns are assumed to be constructed of hot-rolled or built-up wide flange sections with beams framing into the column flange, although some recommendations should also apply to box columns and beams framing to column webs. The recommendations presented herein represent the group consensus of the committee of Guideline Writers employed by SAC following independent review by a technical advisory panel, Project Oversight Committee and Technical Advisory Board. They may not reflect the individual opinions of any single participant. They do not necessarily represent the opinions of the SAC Joint Venture, the Joint Venture partners, or the sponsoring agencies. Users are cautioned that available information on the nature of the WSMF problem is in a rapid stage of development and any information presented herein must be used with caution and sound engineering judgment. 1.9 Use of the Guidelines It is anticipated that the users of these Interim Guidelines will generally desire information in one or more of the following specific areas: 1. a general understanding of the performance of WSMF buildings in the Northridge Earthquake and the probable performance of such buildings in future earthquakes; 2. inspection, evaluation and repair of buildings which have been affected by the Northridge Earthquake or other earthquakes; 3. seismic upgrade of existing WSMF buildings to provide more reliable performance in future earthquakes; and 4. design of new WSMF buildings to provide more reliable performance in future earthquakes. In order to provide information useful to all such users, this document has been made quite broad. Table 1-1 provides a quick reference to the contents of this document.

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Table 1-1 - Quick Reference Guide User Need General Information

Section

Contents

Chapter 1

Introductory material

Chapter 2

Abbreviations, Notation & Terminology

Chapter 3

Damage Classification, Safety Issues, Economic Loss Data

Post-Earthquake Inspection,

Chapters 1-3

Background Information

Evaluation, and Repair

Chapters 4 and 5

Inspection

Chapter 6

Repair

Chapter 8

Metallurgy and Welding

Chapter 9, 10, 11

Inspection & Quality Control

Chapters 1-3

Background Information

Chapter 7

Design Criteria

Chapter 8

Metallurgy & Welding

Chapter 9, 10, 11

Inspection & Quality Control

New Building Design

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Interim Guidelines: Evaluation, Repair, Modification and Design of Steel Moment Frames

Chapter 2 - Definitions, Abbreviations & Notations

2. DEFINITIONS, ABBREVIATIONS & NOTATIONS This Chapter provides the definition of terms used throughout these Interim Guidelines. In addition, abbreviations and symbols, used in other sections of the Interim Guidelines are listed here, together with their typical usage. 2.1 Definitions As used in this document, the terms defined below shall be interpreted to have the meaning indicated, unless specifically indicated elsewhere in this document to have other meaning in a specific context. 2.1.1 Administrative

The definitions of this section apply to the titles of persons involved in the design, construction, regulation, or use of buildings and to the standards, codes and ordinances by which such use is regulated. Building Code

The locally enforced set of regulations governing the design, construction, alteration, occupancy and repair of building structures. Commentary: Although some municipalities and government agencies develop and maintain independent building codes, most building construction in the United States is regulated under locally adopted editions of one of three model building codes: the Uniform Building Code (UBC), the National Building Code (NBC) and the Standard Building Code (SBC). The UBC has been used as a model in this advisory because most buildings damaged by the Northridge Earthquake were designed under earlier editions of that code, and because the seismic design regulations contained in the other two codes, were until 1993, based on those contained in the UBC. In 1993, both the NBC and SBC adopted seismic design regulations based on the NEHRP Recommended Provisions for the Development of Seismic Regulations for New Buildings (Building Seismic Safety Council - 1991). Where references to the UBC provisions are contained in these Interim Guidelines, they are generally to the 1994 edition of that document, unless another edition is specifically identified. Where these Interim Guidelines make reference to specific provisions in the UBC, parallel provisions in the NEHRP Recommended Provisions are generally identified in {parentheses}, where parallel provisions exist. Note that the formulae and requirements contained in these parallel provisions are not always identical, and caution should be exercised when referencing the NEHRP Recommended Provisions from these Interim Guidelines.

Building Official

That officer or authorized representative who has been appointed with legal authority to regulate the construction, alteration, occupancy and use of building structures within a recognized state, county, or municipality.

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Building Owner

That person, corporation or agency holding legal title to the property being constructed, inspected, or repaired, or persons designated with authority to act on their behalf with regard to the building.

Contract Documents

The drawings, specifications and contractual terms under which the responsibilities of the various parties in a project to construct or modify a building are defined.

Contractor

That corporation, partnership, or person retained by the Building Owner to manage and/or perform construction work on a building.

Engineer of Record

The structural engineer in responsible charge of the preparation of drawings and specifications for the inspection, repair, modification or construction of a structure.

Erector

A contractor performing the erection, repair and/or modification of structural steel frames.

Evaluation

The process, including preliminary screening, on-site inspection, and structural analysis, of determining if a building has been structurally damaged, the effect of damage on the building’s integrity, and development of strategies for the occupancy, structural repair and/or modification of the building.

Fabricator

A contractor performing fabrication of structural steel elements to be incorporated in a structural steel frame.

Inspection

On-site investigation of the condition of a structure (or components of a structure) through direct visual observation, aided as necessary by special non-destructive testing techniques.

Owner’s Inspector

A welding inspector retained by the Building Owner to perform quality assurance inspections of weldments. The AWS D1.1 Code defines this individual as the “Verification Inspector.”

Peer Review

An independent technical review of project construction documents as well as supporting data, calculations and assumptions, conducted by structural engineers and intended to provide the Owner and Engineer of Record with an opinion as to the extent that the design complies with applicable standards of care and is likely to achieve its intended objectives.

Special Inspector

An Inspector employed by the Building Owner under the requirements of Section 1701 of the Building Code. When such person performs special inspections related to weldments, he/she shall possess the qualifications noted for a Welding Inspector.

Structural Engineer

A person holding professional engineering registration with the state having jurisdiction, for the practice of structural engineering. The person should have particular training, knowledge and expertise in the structural design of buildings and structures. In some states such a person may hold registration as a Civil Engineer or Professional Engineer.

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Welding Code

American Welding Society publication ANSI/AWS D1.1-94, Structural Welding Code - Steel, 1994 Edition.

Welding Engineer

A person with particular training, knowledge and expertise in metallurgy, the joining of metal elements to each other by the process of welding, and non-destructive testing techniques.

Welding Inspector

A person meeting the requirements of AWS D1.1, Section 6.1.3.1 (and certified by ICBO where applicable) to perform inspections of structural steel weldments. In AWS D1.1, this person is known as “Inspector.”

Welder

A person qualified to perform welding in accordance with the provisions of AWS D1.1.

2.1.2 Technical

The definitions of this section indicate the terms by which specific structural components and elements are indicated in this document. Assembly

The substructure of a steel frame that occurs at a floor level and consists of a single column and one or more floor girders and/or beams that attach directly to it.

Backing

A material or device placed against the back side of the joint, or at both sides of a weld in electroslag welding, to support and retain molten weld metal. The material may be partially fused or remain unfused during welding, and may be either metal or nonmetal.

Backup Bar

A non-preferred term, in common use, for a steel bar used as backing in a complete joint penetration weld. More appropriate terminology is “steel backing.”

Chord

A direct tension or compression element placed at diaphragm edges to resist flexural demands on the diaphragm.

Collector

A structural element used to accumulate shear forces from a diaphragm and distribute them to vertical elements such as frames or walls located along a common line. Also see Strut and Tie.

Connection

The attachment of one structural element, for example a beam, to another, for example a column. As typically used in this document, connection means the attachment of a beam to a column for moment resistance. Important components of this connection include the beam itself, the beam shear tab, the column and its associated panel zone, continuity and doubler plates, and any additional plates used to join these elements together. Other types of connections include bracing connections, gravity connections, base plate connections and column splice connections.

Damage

Degradation in the strength or stiffness of a structural element or alteration of the configuration of the structure or its elements resulting from structural loading, such as induced by an earthquake.

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Damage Index

A numerical index used to quantify the amount of degradation a moment resisting connection (or a group of moment resisting connections) has experienced. A value of 0 indicates no damage and a value of 10, total damage.

Design Basis Earthquake Earthquake ground motion with a probability of exceedance at a site of 10% in 50 years. Such ground motions has an average return period of 475 years. Diaphragm

A horizontal (or nearly horizontal) element of the lateral force resisting system used to distribute lateral loads to the vertical elements of the lateral force resisting system.

Drift

The total lateral deformation of a structure over its height.

Drift Index

Dimensionless quantity indicating the ratio of a structure’s lateral deformation to its height.

Dual System

A structural system in which lateral load resistance is provided by a moment resisting frame in parallel with one or more braced frames and/or shear walls, and meeting the criteria of UBC-94 Section 1627.6.5.

Ductility

The ability of a material, component, element or structure to deform inelastically beyond its yield strength without significant loss in load carrying ability.

Electrode

A component of the electrical circuit that terminates at the arc, molten conductive slag, or base metal.

End Dam

A small plate located at the edge of a beam flange to column flange joint, oriented perpendicular to the joint and intended to serve as a boundary for weld deposition. Commentary: End dams are a mis-application of the requirement for weld tabs that was adopted by some erectors in Southern California. End dams as such are not mentioned in the AWS D1.1 code and they do not constitute weld tabs as required and defined in the code.

Expected Yield Stress

The average stress at which material conforming to an ASTM specification will exhibit yield behavior, as determined by statistical evaluation of production samples.

Flux

A material used to hinder or prevent the formation of oxides and other undesirable substances in molten metal and on solid metal surfaces, and to dissolve or otherwise facilitate the removal of such substances.

Flux-Cored Arc Welding

An arc welding process that produces coalescence of metals by heating them with an arc between a continuous filler metal electrode and the work. Shielding is provided by a flux contained within the tubular electrode. Additional shielding may or may not be obtained from an externally supplied gas or gas mixture.

Fully Restrained Connection

A beam to column connection with sufficient rigidity to hold the original angles between the intersecting members virtually unchanged at loads approaching the strength of the weakest member.

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Gas Shielded FCAW

A flux-cored arc welding process variation in which additional shielding is obtained from an externally supplied gas or gas mixture.

Group

A set consisting of those moment resisting connections in a building primarily intended to resist lateral forces in a given direction of building response, and selected as having similar seismic response characteristics, and therefore, similar probability of being damaged in an earthquake

Gravity Connection

A connection designed to transmit gravity loads from one structural element to another, but not intended to participate in the lateral force resisting system for the structure.

Heat Affected Zone

The portion of the base metal whose mechanical properties or microstructure have been altered by the heat of welding, brazing, soldering, or thermal cutting.

Heat Treatment

A controlled heating and cooling of a metal, usually involving re-crystallization.

Incipient Root Crack

A small planar discontinuity or cracking at the root of a weld.

Interpass Temperature

In a multipass weld, the temperature of the weld area between weld passes.

Interstory Drift

The lateral deformation of a structure within a given story.

Interstory Drift Index

The drift index for a particular story of a structure.

Joint

The juncture of one piece of base metal (for example a beam flange) to another (for example a column flange).

Lamellar Discontinuities Defects in rolled structural shapes or plate, typically consisting of non-metallic sulfide and oxide inclusions which have been flattened by the rolling process and aligned parallel to the direction of rolling. Lamellar Tear

A subsurface terrace and step-like crack in the base metal with a basic orientation parallel to the wrought surface caused by tensile stresses in the through-thickness direction of the base metal weakened by the presence of small dispersed, planar shaped, nonmetallic inclusions parallel to the metal surface.

Lateral Force Resisting System

Those elements of a structure which are intended to provide lateral strength and stiffness for the resistance of lateral forces due to wind or earthquake.

Liquid Dye Penetrant Testing

A method of NDT in which a highly fluid, red dye penetrant is sprayed on the surface of a joint to detect open surface defects. (PT)

Magnetic Particle Testing

A method of NDT which uses a flux field and iron powder to detect surface and sub-surface discontinuities. (MT)

Magnitude

A scale indicating the energy released by an earthquake.

Maximum Capable

The most severe ground motion likely to be experienced at a site, given the

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Earthquake

known seismologic and geotectonic environment. This may be determined by deterministic methods in regions with well defined seismic sources, or by probabilistic methods. If probabilistic methods are used, it may be taken as that level of ground motion with a 10% probability of exceedance in 100 years. Such ground motion has an average return period of approximately 1,000 years.

Metallurgical Stress Riser

A significant deviation in the mechanical properties (usually hardness and micro-structure) between two adjacent regions in a weldment. These may result from arc strikes, improperly made tack welds, and improperly prepared thermally cut surfaces.

Minimum Specified Yield Strength

The lower bound of acceptable yield strength permitted by ASTM specifications, as measured by simple tensile test in accordance to ASTM requirements.

Modification

A structural alteration intended to improve the strength, stiffness, or energy dissipation capacity of a structure and/or its elements.

Moment Frame

A continuous plane of framing in which the beams are joined to the columns with moment resisting connections.

Moment Magnitude

A scale indicating the energy released by an earthquake. Moment magnitude can be calculated based on the surface area of fault rupture amount of slip across the surface, and the stress drop during the event. For moderate magnitude events ( 0.33

Dmax > 0.50

Recommended Strategy (Cumulative)

Note

Repair all connections discovered to have dj > 5 Repair all connections discovered to have dj > 2 Inspect all connections in the group. Repair all connections with dj > 2 A potentially unsafe condition may exist. Carefully evaluate the earthquake resistance of the building and the safety of its occupants and if not satisfied that adequate vertical stability, lateral strength and stiffness exists, notify the building owner of the potentially unsafe condition. Inspect all connections in the building. Repair all connections with dj > 1. Consider modification of all repaired connections and others as appropriate. An unsafe condition probably exists. Notify the building owner of this unless more detailed evaluations indicate otherwise. Inspect all connections in the building. Repair all damaged connections and modify all connections for better performance, or modify the building’s lateral-force-resisting system for improved performance.

1,2 1,2 2

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Notes to Table 4-5: 1. 2. 3.

4.

5.

6.

Includes damage discovered either as part of Step 2 or Step 3. Although repair is recommended only for the more seriously damaged connections, the repair of all connections that are damaged or otherwise deficient should be considered. The determination that an unsafe condition may exist should continue until either: a. full inspection reveals that the gravity system is not compromised, and that the damage index at any floor does not exceed 1/3, or b. detailed structural analyses indicate that a dangerous condition does not exist, or c. recommended repairs are completed for all connections having dj > 3. An unsafe condition probably exists. The building is almost certainly too severely damaged to provide adequate occupant safety in a strong earthquake. The structural engineer should either recommend that the building be vacated, or, alternatively, demonstrate by analysis that the risks to occupant safety, while repairs are conducted, are acceptable. If a decision is made to accept the shortterm risks of continued occupancy, an independent third party review of the basis of this decision is recommended. Repairs required to the building are extensive. In addition to repair, strong consideration should be given to performing systematic modifications of the building’s lateral-force-resisting system to provide more reliable performance in the future. The more restrictive observation governs the recommendation. If all connections in the group were inspected, than do not apply the criteria pertaining to P.

Commentary: The value of P (the probability that the connections on at least one floor have a cumulative damage index of 1/3 or more) and Dmax (the maximum damage index at a floor level within a group) were determined in Method A by using a random selection process, and thus represent a statistically valid basis for the characterization of the damage index for the group of connections, and thus for the building. Method B selects the connections by using a specified distribution throughout the building based on forcing selection of connections in every column line and floor. Method C selects the connections, based on engineering characterization of those most likely to have been damaged, modified to reflect a distribution throughout the structure. While the connections selected by Methods B and C are not truly random, they are widely distributed and have some characteristics of a random distribution. Such selections are judged to be sufficiently “random-like” to warrant processing as if the connections were selected randomly. Thus regardless of whether method A, B, or C was used, decisions on disposition of the building, and the need for repair measures can defensibly be based on the values of these two key parameters, as determined for each group of connections. For buildings that have experienced relatively limited levels of damage, Table 4-5 recommends repair of damaged connections, without further modification. This is not intended to indicate that buildings that experienced only slight damage have been demonstrated to be seismically rugged. In fact, if a building experienced light damage as a result of being subjected to relatively low levels of ground motion, it may have substantial vulnerability. This recommendation is made based on economic considerations and the fact that modification of buildings which are only slightly damaged entails a significant increase in the 4-27

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required investment. It should be made clear to the owner of such buildings that even an undamaged or fully repaired welded steel moment frame building still carries risk of damage, and to an uncertain extent, risk to life safety in subsequent large earthquakes. When damage is moderate (Dmax < 33%) consideration should be given to modification of those connections which are being repaired, to provide improved reliability in the future. However, the structural engineer is cautioned that modification of only those connections which have been damaged could unintentionally create an undesirable condition such as a weak story or torsional irregularity. Therefore, care should be taken that such conditions are not created by connection modifications. Modification of the entire structural lateral forceresisting system is strongly recommended when Dmax>0.50. This is not because the extent of damage indicates that the building is particularly vulnerable, although this may be the case, but because the work required to repair the building is extensive enough that a relatively small incremental investment will allow substantial improvement in the building’s future potential performance. If a decision to structurally modify a building is made, and it can be demonstrated that the structural modifications will reduce the earthquake demands on the existing WSMF connections from the original design levels, it may be acceptable not to repair some conditions. In such cases, analyses should be performed to demonstrate the adequacy of the modified structure assuming either that the affected connections have no moment-resisting capacity, or by including an estimate of their reliable post-elastic behavior in the damaged state. In no case should conditions that affect the gravity load-carrying capacity of the structure be left unrepaired. Recommendations to close a damaged building to occupancy should not be made lightly, as such decisions will have substantial economic impact, both on the building owner and tenants. A building should be closed to occupancy whenever, in the judgment of the structural engineer, damage is such that the building no longer has adequate lateral-force-resisting capacity to withstand additional strong ground shaking, or if gravity load carrying elements of the structure appear to be unstable. 4.3.9 Step 8 - Evaluation Report

When an evaluation of a WSMF building has been performed, the responsible structural engineer should prepare a written evaluation report and submit it to the owner, upon completion of the evaluation. When the building official has required evaluation of a WSMF building, this report should also be submitted to the building official. This report should directly or by attached references, document the inspection program that was performed, provide an interpretation of the

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results of the inspection program, and a recommendation as to appropriate repair and occupancy strategies. The report should include but not be limited to the following material: 1) Building Address 2) A narrative description of the building indicating plan dimensions, number of stories, total square feet, occupancy, the type and location of lateral-force-resisting elements. Include a description of the grade of steel specified for beams and columns, and if known, the type of welding (SMAW, FCAW, etc.) present. Indicate if moment connections are provided with continuity plates. The narrative description should be supplemented with sketches (plans and elevations) as necessary to provide a clear understanding of pertinent details of the building’s construction. The description should include an indication of any structural irregularities, as defined in the Building Code. 3) A description of nonstructural damage observed in the building, especially as relates to evidence of the drift or shaking severity experienced by the structure. 4) If a letter was submitted to the building official before the inspection process was initiated, indicating how the connections were divided into groups and the specific connections to be inspected; a copy of this letter should be included. 5) A description of the inspection and evaluation procedures used, including documentation of all instructions the inspectors, and of the signed inspection forms for each individual inspected connection. 6) A description, including engineering sketches, of the observed damage to the structure as a whole (e.g. - permanent drift) as well as at each connection, keyed to the damage types in Table 4-3; photographs should be included for all connections with damage index dj>5. (Refer to Section 5.3.5) 7) Calculations of davg, Di, and Dmax for each group, and if all connections in a group were not inspected, Pf and P. 8) Calculations demonstrating the safety of the building where Dmax > 33% and the structural engineer has determined that an unsafe condition does not exist. 9) A summary of the recommended actions (repair and modification measures and occupancy restrictions). Any recommendations which represent significant departures from the requirements of Section 4.3.8 should be carefully and completely explained. The report should include identification of any potentially hazardous conditions which were observed, including corrosion, deterioration, earthquake damage, pre-existing rejectable conditions, and evidence of poor workmanship or deviations from the approved drawings. In addition, the report should include an assessment of the potential impacts of observed conditions 4-29

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on future structural performance. The report should include the Field Inspection Reports of damaged connections, as an attachment and should bear the seal of the structural engineer in responsible charge of the evaluation. Commentary: Following completion of the detailed damage assessments, the structural engineer should prepare a written report. The report should include identification of any potentially hazardous conditions which were observed, including earthquake damage, pre-existing rejectable conditions, and evidence of poor workmanship or deviations from the approved drawings. In addition, the report should include an assessment of the potential impacts of observed conditions on future structural performance. The report should include the field inspection, visual inspection and NDT records, data sheets, and reports as attachments. The nature and scope of the evaluations performed should be clearly stated. If the scope of evaluation does not permit an informed judgment to be made as to the extent with which the building complies with the applicable building codes, or as to a statistical level of confidence that the damage has not exceeded an acceptable damage threshold, this should be stated. 4.4 Alternative Group Selection for Torsional Response This Section provides an optional procedure to that of Section 4.3.2, Step 1, that may be appropriate in selected situations where the structural engineer wants more reliable determination of the building’s susceptibility to excessive torsional response. If a building responds in a torsionally dominated manner, one side of the structure may experience substantially more damage than the other side. Such a situation would result in a building that is even more susceptible to torsional response in future strong ground shaking. In the group selection procedure of Section 4.3.2, the connections on opposite sides of a building are included in the same group. If the building responds torsionally, connections on one side will experience more damage and connections on the other side less damage, but the average damage statistics calculated for the group will mask this behavior. In this optional procedure a connection group is established on each side of the building’s center of lateral resistance so that if one side of the building has experienced greater damage, due to torsional response, this will be detected by the damage statistics calculated for the different groups. Typically, under this procedure, at least 4 groups of moment-resisting connections will be designated for the building, one on each of the north, south, east and west sides of the center of rigidity. Buildings with unusual plan shapes (triangular, hexagonal, etc.) may require more (or possibly fewer) groups of connections to adequately capture torsionally induced damage. For each group of connections, the following assumptions are made: 1. All of the connections in a group are expected to perform in the same statistical manner;

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2. The probability of damage to each connection is uniform over the group, that is, all connections have the same probability of failure; and, 3. Prior to inspection, whether an individual connection in the group is damaged or not is independent of the damage state of any other connection in the group. The number of groups should be increased as is required to meet these objectives. To reflect torsional response, resulting either from the structural response characteristics of the undamaged building or a chance concentration of damage that creates such an imbalance, each moment-resisting frame connection is assigned to a group according to the following procedure: a. Determine the approximate center of rigidity for torsional response of the first floor (assuming the building is in its pre-earthquake, original condition). Draw two orthogonal lines in the plan principal directions of the moment frames and extend these vertically as planes. These planes should be adjusted so that all of the connections along a given structural frame are assigned to the same group and that all frames on higher floors are unambiguously assigned to a group. Where the seismic system does not have an orthogonal system, the principal axes can be drawn skewed, or as appropriate to give approximately equal classes of connections assigned to one or the other directions. The following discussion assumes a building with principal orthogonal axes aligned with the north-south and east-west directions. b. All of the connections providing north-south lateral force resistance and located to the west of the center of resistance on all floors (and expected to perform in a similar manner) are assigned to the same group (No. 1). Both weak and strong axis connection connections are included. Similarly all of the connections providing northsouth lateral force resistance and located to the east of the center of resistance are assigned to a second group. A similar procedure is followed to assign connections providing east-west lateral force resistance to one of two additional groups. c. Sample selection from these groups may be made by any of Methods A, B, or C. In keeping with the suggestion in Section 4.3.3.1 paragraph 1, several of the connections in each group having the greatest distance from the assumed center of rotation should be included in each sample. Commentary: It is well known that torsion can play an important role in the distribution of loads on a building’s frame. The eccentricity of the damaged building, either by its design or the chance occurrence of damage to individual connections, has major implications for its response in future earthquakes. It is also clear that the building’s response in orthogonal directions is important. Therefore, for buildings with moment frames in both principal directions, it is recommended that the investigation procedure include at least four distinct groups of connections to reflect the torsional and orthogonal loading conditions.

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For buildings with moment frames in only one direction, it is recommended that the investigation procedure have at least two distinct groupings of connections. 4.5 Qualified Independent Engineering Review Independent third party review, by qualified professionals, is recommended throughout these Interim Guidelines when alternative approaches to evaluation or design are taken, or where approaches requiring high degrees of structural engineering knowledge and judgment are taken. Specifically, it is recommended that qualified engineering review be provided in any of the following cases: 1. Where an engineer elects to select connections for inspection by a method other than Methods A or B of Section 4.3 of these Interim Guidelines. 2. Where the calculated damage index Dmax exceeds 33% and the engineer has determined that an unsafe condition does not exist. 3. Where an engineer has decided not to repair damage otherwise recommended to be repaired by these Interim Guidelines. 4. When any story of the building has experienced a permanent lateral drift exceeding 1% of the story height and proposed repairs do not correct this condition. 5. When an engineer elects to design connections for plastic rotation capacities determined by analysis. 6. When an engineer elects to design connection configurations by calculations only, without the use of, or reference to, qualification tests for a connection prototype. Where independent review is recommended, the analysis and/or design should be subjected to an independent and objective technical review by a knowledgeable reviewer experienced in the design, analysis, and structural performance issues involved. The reviewer should examine the available information on the condition of the building, the basic engineering concepts, and the recommendations for proposed action. Commentary: The independent reviewer may be one or more persons whose collective experience spans the technical issues anticipated in the work. When more than one person is collectively performing the independent review, one of these should be designated the review chair, and should act on behalf of the team in presenting conclusions or recommendations. Independent third party review is not a substitute for plan checking. It is intended to provide the structural engineer of record with an independent opinion, by a qualified expert, on the adequacy of structural engineering decisions and approaches. The seismic behavior of WSMF structures is now 4-32

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understood to be an extremely complex issue. Proper understanding of the problem requires knowledge of structural mechanics, metallurgy, welding, fracture mechanics, earthquake engineering and statistics. Due to our limited current state of knowledge, even professionals who possess such knowledge face considerable uncertainty in making design judgments. Third party review should not be performed by unqualified individuals. 4.5.1 Timing of Independent Review

The independent reviewer(s) should be selected prior to the initiation of substantial portions of the design and/or analysis work that is to be reviewed, and review should start as soon as sufficient information to define the project is available. 4.5.2 Qualifications and Terms of Employment

The reviewer should be independent from the design and construction team. The reviewer should have no interest of any kind with the work being reviewed other than the performance of tasks required by this section. a. The reviewer should have no other involvement in the project before, during, or after the review. b. The reviewer should be selected and paid by the owner and should have an equal or higher level of technical expertise in the issues involved than the structural engineer of record. c. The reviewer (or in the case of peer review teams, the review chair) should be a structural engineer who is familiar with governing regulations for the work being reviewed. d. The reviewer should serve through completion of the project and should not be terminated except for failure to perform the duties specified herein. Such termination should be in writing with copies to the building official, owner, and the structural engineer-of-record. 4.5.3 Scope of Review

Review activities related to evaluation of the safety condition of a building should include a review of available construction documents for the building, all inspection and testing reports, any analyses prepared by the structural engineer of record, the method of connection sample selection and visual observation of the condition of the structure. Review should include consideration of the proposed design approach, methods, materials and details.

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4.5.4 Reports

The reviewer should prepare a written report to the owner and building official that covers all aspects of the structural engineering review performed including conclusions reached by the reviewer. Such reports should include statements on the following: a. Scope of engineering review performed with limitations defined. b. The status of the project documents at each review stage. c. Ability of selected materials and framing systems to meet performance criteria with given loads and configuration. d. Degree of structural system redundancy, ductility and compatibility, particularly in relation to lateral forces. e. Basic constructability of structural members and connections (or repairs and modifications of these elements). f. Other recommendations that would be appropriate to the specific project. g. Presentation of the conclusions of the reviewer identifying any areas which need further review, investigation and/or clarifications. h. Recommendations, if any. 4.5.5 Responses and Corrective Actions

The structural engineer-of-record should review the report from the reviewer and develop corrective actions and other responses as appropriate. Changes during the construction/field phases that affect the seismic resistance system should be reported to the reviewer in writing for action and recommendations. 4.5.6 Distribution of Reports

All reports, responses and corrective actions prepared pursuant to this section should be submitted to the building official and the owner along with other plans, specifications and calculations required. If the reviewer is terminated by the owner prior to completion of the project, then all reports prepared by the reviewer, prior to such termination, should be submitted to the building official, the owner, and the structural engineer-of-record within (10 ) ten working days of such termination. 4.5.7 Engineer of Record

The structural engineer-of-record should retain the full responsibility for the structural design as outlined in professional practice laws and regulations. 4-34

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4.5.8 Resolution of Differences

If the structural engineer-of-record does not agree with the recommendations of the reviewer, then such differences should be resolved by the building official in the manner specified in the applicable Building Code.

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Chapter 5 - Post-Earthquake Building Inspection

5. POST-EARTHQUAKE INSPECTION Post-earthquake inspection is that part of the post-earthquake evaluation process that is conducted at the building. It includes detailed visual observation of the condition of the entire structure as well as selected individual connections and elements. Visual observation is the primary tool for determining the damage sustained by the structure. It should be supplemented by non-destructive testing techniques as required to detect damage that is not directly observable. The moment-resisting connections to be inspected should be determined in accordance with Chapter 4. In addition, other potentially vulnerable connections should also be inspected, particularly when evidence of damage is found in the observation of overall building condition, or in the inspection of moment-resisting connections. Inspection should be conducted under the supervision of a structural engineer familiar with the issues involved. When lower tier personnel are used to perform the inspections, the structural engineer should ascertain that they have adequate knowledge of the types of damage likely to be encountered, and the indicators as to its existence. Careful recording and reporting of the results of inspections is critical to the process. Damage should be reported using the standard classification system of Section 3.1. Care must be taken to accurately report the location as well as the type and degree of damage, and since damage can increase as the building is subjected to additional loads, the date at which observations were made. When required by the building official, or recommended by the Interim Guidelines in Chapter 4, post-earthquake inspections of buildings may be conducted in accordance with the Interim Guidelines of this Chapter. An appropriate sample (or samples) of WSMF connections should be selected for inspection in accordance with the Chapter 4 Guidelines. These connections, and others deemed appropriate by the engineer, should be subjected to visual inspection (VI) and nondestructive testing (NDT) as required by this Chapter. 5.1 Connection Types Requiring Inspection 5.1.1 Welded Steel Moment Frame (WSMF) Connections

The inspection of a WSMF connection should include the complete joint penetration (CJP) groove welds connecting both top and bottom beam flanges to the column flange, including the backing bar and the weld access holes in the beam web; the shear tab connection, including the bolts, supplemental welds and beam web; the column's web panel zone, including doubler plates; and the continuity plates and continuity plate welds (See Figure 3-1). Commentary: The largest concentration of reported damage following the Northridge Earthquake occurred at the welded joint between the bottom girder 5-1

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flange and column, or in the immediate vicinity of this joint. To a much lesser extent, damage was also observed in some buildings at the joint between the top girder flange and column. If damage at either of these locations is substantial (dj per Chapter 4 greater than 5) then damage is also commonly found in the panel zone or shear tab areas. These Interim Guidelines recommend complete inspection, by visual and NDT assisted means, of all of these potential damage areas for a small representative sample of connections. This practice is consistent with that followed by most engineers in the Los Angeles area, following the Northridge Earthquake. It requires removal of fireproofing from a relatively large surface of the steel framing, which at most connections will be undamaged. Some engineers have suggested an alternative approach consisting of visual only inspections, limited to the girder bottom flange to column joint, but for a very large percentage of the total connections in the building. These bottom flange joint connections can be visually inspected with much less fireproofing removed from the framing surfaces. When significant damage is found at the exposed bottom connection, then additional fireproofing is removed to allow full exposure of the connection and inspection of the remaining surfaces. These engineers feel that by inspecting more connections, albeit to a lesser scope than recommended in these Interim Guidelines, their ability to locate the most severe occurrences of damage in a building is enhanced. These engineers use NDT assisted inspection on a very small sample of the total connections exposed to obtain an indication of the likelihood of hidden problems including damage types. If properly executed, such an approach can provide sufficient information to evaluate the post-earthquake condition of a building and to make appropriate occupancy, structural repair and/or modification decisions. It is important that the visual inspector be highly trained and that visual inspections be carefully performed, preferably by a structural engineer. Casual observation may miss clues that hidden damage exists. If, as a result of the partial visual inspection, there is any reason to believe that damage exists at a connection (such as small gaps between the CJP weld backing and column face), then complete inspection of the suspected connection, in accordance with the recommendations of these Interim Guidelines should be performed. If this approach is followed, it is recommended that a significantly larger sample of connections than otherwise recommended by these Interim Guidelines, perhaps nearly all of the connections, be inspected.

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5.1.2 Gravity Connections

In addition to the sample of moment-resisting connections recommended for inspection in Chapter 4, it may be appropriate to inspect selected gravity connections. These include gravity connections for: 1. beams framing orthogonally into a WSMF within the zone of influence of particular WSMF connections with significant damage, and 2. beams framing parallel to a WSMF where significant permanent drift has occurred. Inspection should include any shear tabs, clip angles, or similar elements and the welds and/or bolts attaching these elements to the beam and supporting framing member. Commentary: If little or no damage is found to the moment-resisting connections in a building, it is probable that the gravity connections have not sustained any significant damage. However, if substantial damage is found to moment-resisting connections, some inspection of the gravity connections in the zone of influence of the more heavily damaged moment-resisting connections is probably warranted. For beams framing orthogonally into a WSMF, the zone of influence includes those beams framing directly into columns with damaged connections, as shown in Figures 4-1 and 4-2. It also includes any other beams that could have experienced large torsional rotations as a result of flexural rotations experienced by the WSMF members they frame to. For beams aligned parallel with the WSMF, this zone of influence includes any portion of the structure likely to have experienced excessive drift, as indicated by the damaged moment connection. 5.1.3 Other Connection Types

The structural engineer should review the need to inspect a representative sample of other connection types that exhibit negative attributes similar to the CJP beam-to-column weld configuration. Commentary: These negative attributes include: the inherent residual stress concentrations caused by the welding sequence of highly restrained CJP groove welds used to connect WSMF beams and columns, and the particular care required during their execution to ensure that the welds have no material defects; the post-yield straining in the through-thickness direction of CJP welds used to join WSMF beams and columns; the post-yield straining in the through-thickness direction of WSMF column flanges in a tri-axial state of stress; the difficulty of executing the WSMF beam's bottom flange CJP weld through the restriction created by the web access hole; and the potential for creating a stress riser by leaving the steel backing (backing bar) in place after completing the CJP weld. Connections that are potential high priority candidates for inspection because of their similar connection and stiffness configuration, and because of their use of 5-3

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highly restrained CJP welds include certain eccentric braced frame (EBF) configurations, column-to-base plate connections, and certain drag and collector elements. In addition, selected column splices located such that stresses on the weld during the earthquake response likely approached the minimum specified yield strength should be inspected, including complete joint penetration welded splices in Group 4 and 5 shapes and partial penetration groove welded joints for all shape groups. Complete joint penetration flange welds in Group 4 and 5 Sections have demonstrated a vulnerability to brittle fracture under gravity load conditions. Partial joint penetration groove welds have an inherent “notch” or stress-riser condition which can serve as the initiation point for fracture under conditions of high tensile stress demands. 5.2 Preparation 5.2.1 Preliminary Document Review and Evaluation 5.2.1.1 Document Collection and Review

Prior to performing an inspection, the original construction drawings should be reviewed (if available) to identify the primary lateral and gravity load-resisting systems, typical detailing, presence of irregularities, etc. Pertinent available engineering and geotechnical reports, including previous damage survey reports and current ground motion estimates should also be reviewed. Specifications (including the original Welding Procedure Specifications), shop drawings, erection drawings, and construction records need not be reviewed. 5.2.1.2 Preliminary Building Walk-Through.

A walk-through should be conducted to note visible structural and nonstructural damage, deviations from the plans, and other conditions not evident from the document review. Commentary: If a preliminary post-earthquake evaluation has not previously been conducted, one should be performed at this time. A preliminary post-earthquake evaluation based on ATC-20 (Applied Technology Council 1989), or a similar standard, will not necessarily indicate that damage has been sustained. 5.2.1.3 Structural Analysis

A detailed structural analysis of the building need not be performed prior to performing building inspections. At the engineer’s discretion, such analyses may be performed, in order to develop an understanding as to which connections in the building are most critical and to the extent possible, an understanding of where damage may have concentrated. Analyses used for

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this purpose should be based on rational principles of engineering mechanics and to the extent possible, should use an actual representation of the ground motion experienced by the building. Commentary: Detailed analytical studies of buildings damaged by the Northridge Earthquake indicated some correlation between the actual occurrence of damage and predicted connections with high demands. However, this correlation was not large enough to warrant strong recommendations that analyses be performed prior to performing inspections. In fact, these analyses showed that it is important to inspect connections throughout the structure, regardless of the demands predicted by analysis. The Interim Guidelines for selecting a representative sample of connections for inspection, presented in Chapter 4, contain two methods, A and B, which do not require any prior analysis of the structure, other than to identify its structural system and the location of moment-resisting connections. Some engineers may feel that structural analyses are beneficial in developing a program of inspection, and will prefer to select a sample of connections for inspection based on such analyses. Sample selection Method C, in Chapter 4, is provided for engineers who prefer such an approach. Any rational method of analysis, including linear static, linear dynamic and nonlinear methods may be utilized. When performing dynamic analyses, it is important to use a representation of the ground motion that reasonably resembles that likely to have been experienced by the building, as opposed to a general smoothed response spectrum. The sharp peaks of response which occur over narrow bands of frequencies in actual ground motion recordings can accentuate higher mode response in some buildings, which may not be adequately detected using generic smoothed spectra. Analyses of the response of taller buildings affected by the Northridge Earthquake, as well as their damage patterns, suggest that higher mode effects had a significant impact on the locations of severe strength and deformation demands, as well as damage. The most reliable method of obtaining a representative ground motion is to use data directly recorded by instrumentation at the building site or a nearby site. Instruments located more than 1 km from the building site, or on sites with significantly different subsoil conditions should not be considered particularly representative. Seismologists have the capability to generate estimates of ground motion using fault rupture simulation and wave propagation modeling techniques that may be useful for these purposes as well. However, the engineer should be advised that great uncertainty is associated with such techniques and ground motion representations generated in this manner are only estimates. 5.2.1.4 Vertical Plumbness Check

A rigorous vertical plumbness check is not necessary unless signs of a permanent lateral drift (e.g., elevators are not functioning, door jambs are distorted, or the building is visibly tilted) are

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observed at one or more floors. In such cases, a vertical plumbness check should be conducted by a licensed Surveyor to determine the extent that the post-earthquake out-of-plumbness exceeds AISC Frame Tolerances as defined in the AISC Code of Standard Practice Section 7.11. If significant permanent lateral drift is determined to exist, the structural engineer should determine whether or not this drift, when superimposed with the postulated drift from a future earthquake, presents unacceptable P-∆ stress effects. Commentary: When the plumbness check is deemed advisable, preliminary checks can conveniently be made by the engineer from the interior of the building, through the elevator shaft with the use of a plumb bob as elevator appurtenances, such as sill plates on doors are typically constructed in close vertical alignment. When a more accurate evaluation of plumbness is required, surveying measurements should be made at each exterior principal corner of the building.. 5.2.2 Connection Exposure

Pre-inspection activities to expose and prepare a connection for inspection should include the local removal of suspended ceiling panels or (as applicable) local demolition of permanent ceiling finish to access the connection; and cleaning of the column panel zone, the column flange, continuity plates, beam web and flanges. The extent of the removal of fireproofing should be sufficient to allow adequate inspection of the surfaces to be inspected. Figure 5-1 suggests a pattern that will allow both visual and NDT inspection of the top and bottom beam flange to column joints, the beam web and shear connection, column panel zone and continuity plates, and column flanges in the areas of highest expected demands. The maximum extent of the removal of fireproofing need not be greater than a distance equal to the beam depth "d" into the beam span to expose evidence of any yielding.

6”

12” 6”

Figure 5-1 Recommended Zone for Removal of Fireproofing

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Commentary: Cleaning of weld areas and removal of mill scale and weld spatter should be done with care, preferably using a power wire brush, to ensure a clean surface that does not affect the accuracy of ultrasonic testing. The resulting surface finish should be clean, free of mill scale, rust and foreign matter. The use of a chisel should be avoided to preclude scratching the steel surfaces which could be mistaken for yield lines. Sprayed-on fireproofing on WSMFs erected prior to about 1980 is likely to contain asbestos and should be handled according to applicable standards for the removal of hazardous materials. To preclude physical exposure to hazardous materials and working conditions, the structural engineer should require by contractual agreement with the building owner, prior to the start of the inspection program, that the building owner deliver to the structural engineer for his/her review and files a laboratory certificate that confirms the absence of asbestos in structural steel fireproofing, local pipe insulation, ceiling tiles, and drywall joint compound. The pattern of fireproofing removal indicated in Figure 5-1 is adequate to allow visual and UT inspection of the top and bottom girder flange to column joints, the beam web and shear connection and the column panel zone. As discussed in the commentary to Section 5.1.1, some engineers prefer to initially inspect only the bottom beam flange to column joint. In such cases, the initial removal of fireproofing can be more limited than indicated in the figure. If after initial inspection, damage at a connection is suspected, then full removal, as indicated in the figure, should be performed to allow inspection of all areas of the connection. 5.3 Inspection Program 5.3.1 Visual Inspection (VI)

Visual Inspection is the primary means of determining the condition of the structure. It should be performed by, or under the direction of, a structural engineer, and in as many locations as is practical. As a minimum, it should be performed in those locations selected in accordance with one of the methods of Chapter 4. It may be performed and documented by other competent persons, but should be performed with a structural engineer's written instructions and guidance. When VI is performed by a testing agency, the agency and personnel performing the work should conform to the Interim Guidelines of Chapter 10. As a minimum, the structural engineer in charge should visit the site as needed during the performance of visual inspection to confirm that his/her instructions are understood and followed, and to provide a spot check of the adequacy of surface preparation of the connection for VI and NDT, that the recorded locations of damage are correct, and that damage is accurately reported. The presence or absence of damage should be recorded in a consistent and objective manner on a uniform data sheet that will allow later interpretation of the conditions and assessment of its severity and the types of repair which may be warranted. Severe damage should be documented 5-7

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with photographs. Data sheets should include a sketch of the connection and locations of any significant non-conforming or damage conditions noted. Damage should be classified in accordance with the system indicated in Section 3.1. Commentary: The presence or absence of the following conditions should be recorded for each inspected connection: a) Deviations from Construction Documents or Specifications. b) Continuity plates. c) Doubler plates (on one or both sides of the web). d) Supplemental web welds (from beam-web-to-shear-tab). e) Flange weld backing bar and runoff tabs. f) Flange weld end dams. g) Poor Fit-up of backing bar. h) Evidence of weld spatter (must be removed prior to performing Ultrasonic Testing). i) Smooth (or rough) beam web cope for weld access holes. j) Evidence of poor quality welding workmanship per AWS D1.1 Section 6.5.1 and 8.15.1. k) Undercut, underfill or excessive concavity/convexity of welds. l) Undersized fillet welds. The presence or absence of damage should be recorded. For purposes of visual inspection, backing bars need not be removed. If damage is discovered, it should be recorded by type, per the classification system of Chapter 3. When full inspection of a connection is conducted, both sides of the beam, column, and panel zone should be inspected. If one side of the connection is obstructed (e.g., by exterior walls), such obstructions need not be removed if the accessible side of the connection appears undamaged. Beam top flange connection welds may be inspected without local removal of the floor diaphragm finish if there is no apparent significant damage at the bottom beam flange, adjacent column flange, column web, or shear connection. If severe beam bottom flange damage is observed, removal of diaphragm materials to allow direct observation of the beam top flange is recommended. More information on VI may be found in AWS B1.11. 5.3.1.1 Top Flange

The exposed root of this "T" joint should be inspected to note any possible separation of the edge of the backup bar from the face of the column flange. The exposed surface of the beam flange and column flange should be observed to note any cracks which may have occurred. Beam 5-8

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flange base metal at the intersection of the weld access hole in the web to the beam flange should be inspected to note any visible cracks. 5.3.1.2 Bottom Flange

The possible separation of the backup bar to column flange should be inspected as above. The face of the weld should be inspected to note any possible cracks in the weld, the toe of the weld or in the adjacent base metal. The base metal of the column flange above, below and each side of the intersecting beam flange should be inspected to note any possible visible cracks as should the beam flange in the vicinity of the web access hole, as above. 5.3.1.3 Column and Continuity Plates

Column base metal and the continuity plates and their welds to the inside face of the column flanges should be visually inspected. The column web above and below the continuity plate should be inspected to note any visible cracks. 5.3.1.4 Beam Web Shear Connection

The shear connection plate, beam web, and corresponding bolts should be inspected to note any possible rotation. The base metal around the bolt head and nut including washers if used, may show signs of bright metal if rotation has occurred. Observation should include examination for bolts which may have loosened as well as any welds used in combination with the bolted connection. The exposed surface of the shear plate to column flange weld should be visually inspected with primary attention being paid to the termination of this weld near the beam's bottom flange. 5.3.2 Nondestructive Testing (NDT)

NDT should be used to supplement the visual inspection of connections selected in accordance with the Interim Guidelines of Chapter 4. The testing agency and NDT personnel performing this work should conform to the qualifications indicated in Chapter 11 of these Interim Guidelines. The following NDT techniques should be used at the top and bottom of each connection, where accessible, to supplement visual inspection: a) Magnetic particle testing (MT) of the beam flange - to column flange weld surfaces. All surfaces which were visually inspected should be tested using the magnetic particle technique. Commentary: The color of powder should be selected to achieve maximum contrast to the base and weld metal under examination. The test may be further enhanced by applying a white coating made specifically for MT or by applying penetrant developer prior to the MT examination. This background coating should be allowed to thoroughly dry before performing the MT. 5-9

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b) Ultrasonic testing (UT) of all faces at the beam flange welds and adjacent column flanges (extending at least 3 inches above and below the location of the CJP weld, along the face of the column, but not less than 1-1/2 times the column flange thickness). Commentary: The purpose of UT is to 1) locate and describe the extent of internal defects not visible on the surface and 2) to determine the extent of cracks observed visually and by MT. Requirements and acceptance criteria for NDT should be as given in AWS D1.1 Sections 6 and 8. Acceptance or rejection of planar weld discontinuity (cracks, slag inclusion, or lack of fusion), including root indications, should, as a minimum, be consistent with AWS Discontinuities Severity Class designations of cracks and defects per Table 8.2 of AWS D1.1 for Static Structures. Beam flange welds should be tested as "tension welds" per AWS D1.1 Table 8.15.3, Note 3. Backing bars need not be removed prior to performing UT. Commentary: The value of UT for locating small discontinuities at the root of beam flange to column flange welds when the backing is left in place is not universally accepted. The reliability of this technique is particularly questionable at the center of the joint, where the beam web obscures the signal. There have been a number of reported instances of UT detected indications which were not found upon removal of the backing, and similarly, there have been reported instances of defects which were missed by UT examination but were evident upon removal of the backing. The smaller the defect, the less likely it is that UT alone will reliably detect its presence. Despite the potential inaccuracies of this technique, it is the only method currently available, short of removal of the backing, to find subsurface damage in the welds. It is also the most reliable method for finding lamellar problems in the column flange (type C4 and C5 damage) opposite the girder flange. Removal of weld backing at these connections results in a significant cost increase that is probably not warranted unless UT indicates widespread, significant defects and/or damage in the building. The proper scanning techniques, beam angle(s) and transducer sizes should be used as specified in the written UT procedure contained in the Written Practice, prepared in accordance with Section 5.3.3 of these Interim Guidelines. The acceptance standard should be that specified in the original contract documents, but in no case should it be less than the acceptance criteria of AWS D1.1, Chapter 8, for Statically Loaded Structures. The base metal should be scanned with UT for cracks. Cracks which have propagated to the surface of the weld or beam and column base metal will probably have been detected by visual inspection and magnetic particle tests performed earlier. The purpose of ultrasonic testing of the base metal is to: 1. Locate and describe the extent of internal indications not apparent on the surface and, 5-10

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2. Determine the extent of cracks found visually and by magnetic particle test. Commentary: Liquid dye penetrant testing (PT) may be used where MT is precluded due to geometrical conditions or restricted access. Note that more stringent requirements for surface preparation are required for PT than MT, per AWS D1.1. If practical, NDT should be performed across the full width of the bottom beam flange joint. However, if there are no discontinuity signals from UT of accessible faces on one side of the bottom flange weld, obstructions on the other side of the connection need not be removed for testing of the bottom flange weld. Slabs, flooring and roofing need not be removed to permit NDT of the top flange joint unless there is significant visible damage at the bottom beam flange, adjacent column flange, column web, or shear connection. Unless such damage is present, NDT of the top flange should be performed as permitted, without local removal of the diaphragms or perimeter wall obstructions. It should be noted that UT is not 100% effective in locating discontinuities and defects in CJP beam flange to column flange welds. The ability of UT to reliably detect such defects is very dependent on the skill of the operator and the care taken in the inspection. Even under perfect conditions, it is difficult to obtain reliable readings of conditions at the center of the beam flange to column flange connection as return signals are obscured by the presence of the beam web. If backing is left in place on the welds, UT becomes even less reliable. There have been a number of reported instances in which UT indicated apparent defects, that were found not to exist upon removal of the backing. Similarly, UT has failed in some cases to locate defects that were later discovered upon removal of the backing. Additional information on UT may be found in AWS B1.10. 5.3.3 Inspector Qualification

Testing shall be performed and supervised by qualified and properly certified technicians. It is recommended that the structural engineer (or his/her agent) observe the inspection procedures directly until such a time that confidence is developed that the inspections are being made in accordance with the given instructions. It is the responsibility of the structural engineer in charge to confirm that only certified Level II technicians, certified in accordance with AWS D1.1-1994 Section 6.1.3.1 and 6.7.8 and SNT-TC-1A, are allowed to execute VI and NDT, under the supervision of a Level III technician with current certification by examination. All Level II certifications should be current, having been issued within 3 years prior to the start of the inspection program. The structural engineer should require that the inspector provide a Written Practice for his/her review and approval, and the building owner's file, in accordance with the requirements of SNT-TC-1A. The Written Practice should as a minimum provide 1) all certification records for all technicians executing VI and NDT on the project and 2) the detailed

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UT, MT and PT procedures used to perform NDT of WSMF connections, as well as other types of connections, in accordance with the instructions of the structural engineer. Commentary: Special care is recommended in the selection of inspectors because VI and NDT methods are highly dependent on the skill and integrity of the operator for proper interpretation of the results. 5.3.4 Post-Earthquake Field Inspection Report

A Field Inspection Report for both VI and NDT should be completed for each connection inspected and/or tested, regardless of whether damage or rejectable defects are detected. There are two (2) CJP welds for each WSMF beam-to-column connection. A standard form should be used to ensure complete documentation. Sample Field Inspection Report forms are provided in Figure 5-2 through 5-5. The technician should record the depth, length and location of observed indications, should characterize the discontinuities as planar (cracks or lack of fusion) or volumetric (porosity, slag, etc.), should classify the weld as acceptable or rejectable according to predetermined criteria, and should note any uncertainties. In addition, the form should record the date of testing, the person responsible, connection location and orientation, and descriptions of items not tested due to limited access. The Field Inspection Report form should, as a minimum, objectively identify for each CJP weld and shear tab tested the following information, as applicable: a) Damage/Defect type classification/description (per Section 3-1, and summarized for convenience at the rear of this Chapter as Table 5-1). b) AWS Discontinuity Severity Class of crack/defect per Table 8.2 of AWS D1.1-94. c) Depth of crack/defect. d) Length of crack/defect/damaged material. e) Location of crack/defect/damaged material. f) Identification of NDT procedure used. g) Possible inclusion of photographs. Commentary: To ensure a correct understanding and identification of reported connection damage, the format of the Field Inspection Report form should include an easy to understand graphic description of what face of the connection is being inspected (e.g., north face, south face, east or west) and at what framing elevation the connection is located (e.g., inspector is standing on the 4th floor and looking at a connection located at the 5th floor framing). In addition, any identified significant damage should be recorded on a generic sketch of the WSMF standard joint detail to facilitate consistent reporting and correct interpretation and assessment by the structural engineer.

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5.3.5 Written Report

Following completion of the detailed damage assessments, the structural engineer should prepare a written report, in accordance with Section 4.3.9.

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Table 5-1 - Connection Damage Classification Type G1 G2 G3 G4 G5 G6 G7 G8 C1 C2 C3 C4 C5 C6 C7 W1a

Location Girder Girder Girder Girder Girder Girder Girder Girder Column Column Column Column Column Column Column CJP weld

W1b

CJP weld

W2 W3 W4 W5 S1a S1b S2a S2b S3 S4 S5 S6 P1 P2 P3 P4 P5 P6 P7

CJP weld CJP weld CJP weld CJP weld Shear tab Shear tab Shear tab Shear tab Shear tab Shear tab Shear tab Shear tab Panel Zone Panel Zone Panel Zone Panel Zone Panel Zone Panel Zone Panel Zone

P8 P9

Panel Zone Panel Zone

Description Buckled Flange Yielded Flange Top or Bottom Flange fracture in HAZ Top or Bottom Flange fracture outside HAZ Top and Bottom Flange fracture Yielding or Buckling of Web Fracture of Web Lateral-torsional Buckling Incipient flange crack (detectable by UT) Flange tear-out or divot Full or partial flange crack outside HAZ Full or partial flange crack in HAZ Lamellar flange tearing Buckled Flange Fractured column splice Minor root indication, thickness < 3/16” or tf/4; width < bf/4 Root indication, thickness > 3/16” or tf/4; width > bf/4 Crack through weld metal thickness Fracture at girder interface Fracture at column interface UT detectable indication— non-rejectable Partial crack at weld to column (beam flanges sound) Partial crack at weld to column (beam flange cracked) Crack in Supplemental Weld (beam flanges sound) Crack in Supplemental Weld (beam flange cracked) Fracture through tab at bolt holes Yielding or buckling of tab Damaged, or missing bolts Full length fracture of weld to column Fracture, buckle, or yield of continuity plate Fracture of continuity plate welds Yielding or ductile deformation of web Fracture of doubler plate welds Partial depth fracture in doubler plate Partial depth fracture in web Full (or near full) depth fracture in web or doubler plate Web buckling Fully severed column

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Figure 5-2 Inspection Form - Major Axis Column Connection

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Figure 5-3 Inspection Form - Large Discontinuities - Major Axis

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Figure 5-4 Inspection Form - Major Axis Column Connection

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Figure 5-5 Inspection Form - Large Discontinuities - Minor Axis

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6. POST-EARTHQUAKE REPAIR AND MODIFICATION As used in these Interim Guidelines, repair means restoration of the strength, stiffness and deformation capacity of structural elements that have been damaged or have construction defects. Modification means actions taken to enhance the strength, stiffness or deformation capacity of either damaged or undamaged elements, or of the structure as a whole. Based on the observed behavior of actual buildings in the Northridge Earthquake, as well as recent test data, WSMF structures constructed with the typical pre-Northridge detailing and construction practice prevalent prior to the Northridge Earthquake do not have the same deformation capacity they were presumed to possess at the time of their design. The seismic risk associated with these structures is higher than typically judged as acceptable for buildings of new construction. When these buildings are damaged or have excessive construction defects, the risk is higher. Based on limited testing, it appears that the repair recommendations contained in this Chapter can be effective in restoring a building’s pre-earthquake condition. This does not imply, however, that the repaired building will be an acceptable seismic risk. As a minimum, it should be assumed that buildings that are repaired, but not modified, can sustain similar and possibly more severe damage in future earthquakes than they did in the present event. If this is unacceptable, either to the owner or the building official, then the building should be modified to provide improved future performance. Modification can consist of local reinforcement of individual moment connections as well as alteration of the basic lateral-force-resisting characteristics of the structure through addition of braced frames, shear walls, base isolation, energy dissipation devices, etc. 6.1 Scope This section provides interim guidelines for structural repair of earthquake damage and modification of structures to improve future earthquake performance. Repair constitutes any measure(s) taken to restore earthquake damaged joints, connections, elements of the building, or the building as whole, to their original strength, stiffness and deformation capacity. It does not include routine correction of non-conforming conditions during original fabrication. Interim Guidelines for acceptable methods of repair are provided in Sections 6.2 through 6.5 below. These Interim Guidelines are not intended to be used for the routine repairs of non-conformance commonly encountered in fabrication and erection work. Industry standard practices are acceptable for such repairs. Work that increases structural stiffness or strength of an element or the structure as a whole by more than 5% is classified as modification. Guidelines for some methods of modification are contained in Section 6.6.

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6.2 Shoring 6.2.1 Investigation

The structural engineer responsible for designing any damage repair should investigate the entire building for imminent collapse or life safety hazard conditions, regardless of joint considerations. Such conditions should be shored prior to commencement of any repairs. Commentary: In projects relating to construction of new buildings, it is common practice to delegate all responsibility for temporary shoring and bracing of the structure to the contractor. Such practice may not be appropriate for severely damaged buildings. The structural engineer should work closely with the contractor to define shoring and bracing requirements. Some structural engineers may wish to perform the design of temporary bracing systems. If the contractor performs such design, the structural engineer should review the designs for adequacy and potential effects on the structure prior to implementation. 6.2.2 Special Requirements.

Conditions which may become collapse or life safety hazards during the repair operations should be considered in the development of repair details and specifications, whether they involve the connection area directly or indirectly. These conditions should be brought to the attention of the contractor by the structural engineer, and adequate means of shoring these conditions should be provided. Consideration should be given to sequencing of repair procedures for proper design of any required shoring. For column repair details that require removal of 20% or more of the damaged cross section, consideration should be given to the need for shoring to prevent overstress of elements due to redistribution of loads. Commentary: In general, contractors will not have adequate resources to define when such shoring is necessary. Therefore, the Contract Documents should clearly indicate when and where shoring is required. Design of this shoring may be provided by the structural engineer, or the contractor may submit a shoring design to the structural engineer for review. 6.3 Repair Details The scope of repair work should be shown on drawings and specifications prepared by a structural engineer. The drawings should clearly indicate the areas requiring repair, as well as all repair procedures, details, and specifications necessary to properly implement the proposed repair. Sample repair details for various types of damage are included in these Interim Guidelines, for reference, only. Commentary: Examples of repair details are provided for some classes of damage, based on previous repairs performed in the field for specific projects. Limited testing indicates these repair methods can be effective. Details are not 6-2

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complete in all respects and should not be used verbatim, as construction documents. Many repairs will require the application of more than one operation, as represented by a given detail. The sample details indicated may not be directly applicable to specific repair conditions. The structural engineer is cautioned to thoroughly review the conditions at each damaged element, connection or joint, and to determine the applicability and suitability of these details based on sound structural engineering judgment, prior to employing them on projects. 6.3.1 Approach

Based on the nature and extent of damage several alternative approaches to repair should be considered. Repair approaches may include, but should not be limited to: a) replacement of portions of base metal (i.e. column and beam section), b) replacement of connection elements, c) replacement of connection weld, or d) repairs to portions of any of the aforementioned components. Any or all of these techniques may be appropriate. The approach(s) used should consider adjacent structural components which may be affected by the repair or the effects of the repair. Where base material is to be removed and replaced with plates, clear direction should be given to orient the plates with the direction of rolling of the plate parallel to the direction of application of major axial loads to be resisted by the plate. 6.3.2 Weld Fractures - Type W Damage

All fractures and rejectable defects found in weld material, either between girder and column or between connection element and structural member, should have sufficient material removed to completely eliminate any discontinuity or defect. NDT should be used to determine the extent of fracture or defect and sufficient material should be removed to encompass the damaged area. It is suggested that material removal extend 2 inches beyond the apparent end of the fracture or defect. Simple fillet welds may be repaired by backgouging to eliminate unsound weld material and replacing the damaged weld with sound material. Complete joint penetration (CJP) welds fractured through the full thickness should be replaced with sound material deposited in strict accordance with the Welding Procedure Specification (WPS) and project specifications. The use of weld dams on new welds is prohibited. Weld backing (backup bars), existing dams, and weld tabs should be removed from all welds that are being repaired. After backing is removed, the root should be backgouged to sound material, rewelded and a reinforcing fillet added. The structural engineer is cautioned to observe the provisions of AISC regarding intermixing of weld metals deposited by different weld processes (see AISC LRFD Manual of Steel 6-3

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Construction, second edition, page 6-77, and AISC ASD Steel Construction Manual, ninth edition, page 5-69). As an example, E7018 stick electrodes should not be used to weld over selfshielded flux cored arc welding deposits. Removed weld material from fractures not penetrating the full weld thickness should be replaced in the same manner as full thickness fractures. For other types of W damage, existing backing, end dams, and weld tabs should also be removed in a like manner to CJP weld replacement. Table 6-1 provides an index to suggested repair details for type W damage. Table 6-1 - Reference Details for Type W Damage Damage Class W1a, W1b W2 W3 W4 W5

Figure Figure 6-1, Figure 62 Figure 6-3 Figure 6-3 Figure 6-3 Figure 6-3

Commentary: FCAW-ss utilizes approximately 1-2% aluminum in the electrode to protect the weld from mixing with atmospheric nitrogen and oxygen. By itself, aluminum can reduce the toughness and ductility of weld metal. The design of FCAW-ss electrodes requires the balance of other alloys in the deposit to compensate for the effects of aluminum. Other welding processes rely on fluxes and/or gasses to protect the weld metal from the atmosphere, relieving them of any requirement to contain aluminum or other elements that offset the effects of aluminum. If the original weld that is being repaired consists of FCAW-ss and subsequent repair welds are made with SMAW (stick) using E7018, for example, the SMA arc will penetrate into the FCAW-ss deposit, resulting in the addition of some aluminum into the SMAW deposit. The notch toughness and/or ductility of the resultant weld metal may be substantially reduced as compared to pure E7018 weld metal, based on the depth of penetration into the FCAW-ss material. Various types of FCAW-ss electrodes may be mixed one with the other without potentially harmful effect. Further, FCAW-ss may be used to weld over other types of weld deposits without potentially harmful interaction. The structural engineer could specify all repairs on FCAW-ss deposits be made with FCAW-ss. Alternately, intermixing of FCAW-ss and other processes could be permitted provided the subsequent composition is demonstrated to meet material specification requirements.

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Existing column flange

Existing beam flange

Removed backing Reweld & reinforce w/ fillet

20o Min. Arc - Gouge

1/4” radius min.

Notes:

1 2 3 4 5

Remove existing backing. Taper the depth of grinding or air arc gouging at each end to the face of flange with a minimum 2:1 (horizontal/vertical) taper. Provide a minimum root radius of 1/4.” Grind all surfaces on which weld metal will be deposited. Surfaces should be smooth, uniform and free from fins, tears, fractures and other discontinuities which would adversely affect weld strength. A fillet weld should be applied to reinforce the joint. The size of the reinforcing fillet should be equal to 1/4 of the beam flange thickness, but not less than 1/4.” It need nor be more than 3/8.” On joints to be repaired, remove all remaining weld tabs and excess weld metal beyond the length of the joint and grind smooth. Imperfection less than 1/16" should be removed by grinding. Repair as necessary.

Figure 6-1 - Gouge & Re-weld of Root Defect or Damage - W1 Existing column flange Air-arc gouge Reweld Existing beam flange

Removed backing Backgouge , repair and reinforce per Figure 6-1.

Notes: 1. 2.

Remove the entire fracture plus 2” of sound metal beyond each end. For additional notes, refer to Figure 6-1

Figure 6-2 - Gouge & Re-weld of Fractured Weld - W1

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Existing column flange Air-arc gouge Reweld Existing beam flange

Remove backing after completing top welding, Backgouge , repair and reinforce, per Figure 6-1.

For notes see Figure 6-1 and 6-2.

Figure 6-3 - Backgouge and Reweld repair 6.3.3 Column fractures - Type C1 - C5 and P1 - P6

Any column fracture observable with the naked eye or found by NDT and classified as rejectable in accordance with the AWS D1.1 criteria for Static Structures should be repaired. Repairs should include removing the fracture such that no sign of rejectable discontinuity or defect within a six (6) inch radius around the fracture remains. Removal should include eliminating any zones of fracture propagation, with a minimum of heat used in the removal process. Following removal of material, MT and PT should be used to confirm that all fractured material has been removed. Repairs of removed material may consist of replacement of portions of column section, build-up with weld material where small portions of column were removed, or local replacement of removed base metal with weld material. Procedures of weld fracture repair should be applied to limit the heat affected area and to provide adequate ductility to the repaired joint. Tables 6-2 and 6-3 indicate representative details for these repairs. In many cases, it may be necessary to remove a portion of the girder framing to a column, in order to attain necessary access to perform repair work, per Figure 6-4. Refer to Section 6.3.5 for repair of girders. Remove Shear Tab Replace upon completion

New web plate thickness = tw +1/8”

tw Shore Beam Remove portion of existing beam. Provide minimum 2” radius.

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Figure 6-4 - Temporary Removal of Beam Section for Access When the size of divot (type C2) or transverse column fractures (types C1, C3, C4) dictate a total cut-out of a portion of a column flange or web (types P6, P7), the replacement material should be ultrasonically tested in accordance with ASTM A578-92, “Straight-Beam Ultrasonic Examination of Plane and Clad Steel Plates for Special Applications,” in conjunction with AWS K6.3 “Shearwave Calibration.” Acceptance criteria should be that of Level III. The replacement material should be aligned with the rolling direction matching that of the column. Table 6-2 - Reference Details for Type C and P Damage Damage Class Beam Access C1 C2 C3 C4 C5 P1 P2 P4 P5 P6 P7 P8

Figure Figure 6-4 Figure 6-4, 6-5 Figure 6-4, 6-6 Figure 6-4, 6-5 Figure 6-4, 6-5 Figure 6-4, 6-6 remove, prepare, replace arc-gouge and reweld arc-gouge and reweld Figure 6-7 Figure 6-7 Figure 6-7 Figure 6-8

Backgouge and reweld

Weld access hole in column web

Portion of E beam flange removed 45o

per AWS D1.1 section 3.2.5, and Figure 3.2

1

2 3

10o

Investigate extent of fracture by UT to confirm that fractures are contained with the 45 degree angle zone of a standard pre-qualified CJP groove weld as defined by AWS D1.1, Figure 2.4, Joint Designation BU4a-G Provide 10o bevel on lower flange plate, to channel slag out of joint. Grind all surfaces upon which weld metal will be deposited to smooth, uniform surface.

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Figure 6-5 - Backgouge and reweld of column flange

Weld access hole and backing

6” minimum

New flange splice plate

10o

Note:

Provide new flange plate material of the same strength, and width as the existing column flange. Align rolling direction of plate with that of column flange. New plate should be of the same thickness as the existing flange with a tolerance of -0”/+1/4.” The welding should be sequenced to connect the column flange to new flange plate welds prior to welding the column web to new flange plate. Bevel the lower edge of the column flange, and upper edge of the splice plate down 10o, to channel slag out of joint.

Figure 6-6 - Replacement of Column Flange Repair

Doubler Plate

Column web

Typical

Web with Doubler Plate

Notes: 1. 2.

Web without Doubler Plate

Prepare fractured section of doubler by air-arc gouging, grind and reweld, using web as backing Prepare fractured section of web by air-arc gouging, grind and reweld, using doubler as backing or backgouge and reweld from reverse side, if no doubler present.

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Figure 6-7 - Reweld Repair of Web plate and Doubler plate

Flange removal and replacment per Figure6-6, if required Weld access holes as required for weld terminations

Notes: 1. 2.

Sequence removal of portions of column and provide shoring as required to safely support existing column loads. Thickness of new web plate to match existing column web (tolerance +1/8”, -0”).

Figure 6-8 - Alternate Column Web Repair - Columns without Doubler Plates Commentary: Special attention should be given to conditions where more than 20% of the column cross section will be removed at one time, as special temporary shoring may be warranted. In addition, care should be taken when applying heat to a flange or web containing a fracture, as fractures have been observed to propagate with the application of heat. This can be prevented by drilling a small diameter hole at the end of the fracture, to prevent it from running. 6.3.4 Column splice fractures - Type C7

Any fractures detected in column splices should be repaired by removing the fractured material and replacing it with sound weld material. For partial joint penetration groove welds, remove up to one half of the material thickness from one side and replace with sound material. Where complete joint penetration groove welds are required, it may be preferable to provide a double bevel weld, repairing one half of the material thickness completely prior to preparing and repairing the other half. Alternatively, if calculations indicate that column loads may safely be resisted with the entire section of column flange removed, or if suitable shoring is provided, it may be preferable to use a single bevel weld. Commentary: Special attention should be given to these conditions, as the removal of material may require special temporary shoring. Also, since partial penetration groove welds can serve as fracture initiators in tension applications, 6-9

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consideration should be given to replacing such damaged splice areas with complete joint penetration welds. 6.3.5 Girder Flange Fractures - Type G3-G5

Repair of fractures in girder flanges may be performed by several methods. One method is to remove the fracture by air arc gouging such that no sign of discontinuity or defect within a six (6) inch radius around the fracture remains, preparing the surface by grinding and welding new material back. Alternatively, damaged portions of the girder flange may be removed and replaced with new plate as shown in Figure 6-9 or Figure 6-10.

Weld access hole

New web stiffeners, near side and far side New beam flange plate Typ.

Notes: 1. 2.

New plate thickness to match beam flange thickness + height of removed web fillet. Weld sequence - a) weld of new flange plate to column; b)weld of flange plate stiffeners to web and flange plate; c) weld of new flange plate to beam flange. d)weld of stiffener plate to beam flange and web

Figure 6-9 - Beam Flange Plate Replacement

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New beam flange plate

Figure 6-10 - Alternative Beam Flange Plate Replacement Commentary: Due to accessibility difficulties or excessive weld build-up requirements, it may become necessary to remove a portion of the girder flange to properly complete the joint repair. A minimum of six inches of girder flange may be removed to facilitate the joint repair, with the optimum length being equal to the flange width. After removal of the portion of flange, the face of column and cut edge of girder flange may then be prepared to receive a splice plate matching the flange in grade and width. Thickness should be adjusted as required to makeup the depth of the girder web and fillet removed as part of the preparation process. It is recommended that a double bevel joint be utilized in replacing the removed plate to eliminate the need for backup bars, consequently also eliminating the removal of these backup bars. A suggested joint detail is a BU3/TC-U5, per AWS D1.1, with 1/3 tflange-2/3 tflange bevels on the plate. The web of the girder should be prepared at the column and butt weld areas to allow welding access. Weld tabs may be used at the column and butt weld. The weld between the splice plate and the column flange should be completed first. If a double bevel weld is selected, the welder may choose to weld the first few passes from one face, then backgouge and weld from the second side. This may help to keep the interpass temperature below the maximum without down time often encountered in waiting for the weld to cool. 6.3.6 Buckled Girder Flanges - Type G1

Where the top or bottom flange of a girder has buckled, and the rotation between the flange and web is less than or equal to the mill rolling tolerance given in the AISC Manual of Steel Construction (AISC-1994 or AISC-1989) the flange need not be repaired. Where the angle is greater than mill rolling tolerance, repair should be performed and may consist of adding full 6-11

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height stiffener plates on the web over each portion of buckled flange, contacting the flange at the center of the buckle, (Figure 6-11) or using heat straightening procedures. Another available approach is to remove the buckled portion of flange and replace it with plate, similar to Figures 6-9 and 6-10.

New stiffener plates each side, tplate = tweb

Note:

Provide stiffeners at beginning of buckle and at center of buckle

Figure 6-11 - Addition of Stiffeners at Buckled Girder Flange Commentary: Should flange buckling occur on only one side of the web, and the buckle repair consists of adding stiffener plates, only the side that has buckled need be stiffened. In case of partial flange replacement, special shoring requirements should be considered by the design engineer. 6.3.7 Buckled column flanges - Type C6

Any column flange or portion of a flange that has buckled to the point where it exceeds the rolling tolerances given in the AISC Manual of Steel Construction should be repaired. Flange repair may consist either of flame straightening or of removing the entire buckled portion of flange and replacing it with material with yield properties similar to the actual yield properties of the damaged material similar to Figure 6-6. If workers with the appropriate skill to perform flame straightening are available, this is the preferred method. Commentary: For flange replacement, shoring is normally required. This shoring should be designed by the structural engineer, or may be designed by the contractor provided the design is reviewed by the structural engineer. Flame straightening can be an extremely effective method of repairing buckled members. It is performed by applying heat to the member in a triangular pattern, in order to induce thermal strains that straighten the member out. Very large bends can be straightened by this technique. However, the practice of this technique is not routine and there are no standard specifications available for controlling the work. Consequently, the success of the technique is dependent on the availability of workers who have the appropriate training and experience to perform the work. During the heat application process, the damaged member is locally heated to very high temperatures. Consequently, shoring may be required for members being straightened in this manner. 6-12

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A number of references are available that provide more information on this process and its applications, published by AISC and others (Avent - 1992), (Avent - 1995), (Shonafelt and Horn - 1984) 6.3.8 Gravity connections

Connections not part of the lateral load-resisting system may also be found to require repair due to excessive rotation or demand caused by distress of the lateral load-resisting system in the zone of influence. These connections should be repaired to a capacity at least equivalent to the pre-damaged connection capacity. Shear connections that are part of the lateral load resisting system should be repaired in a similar manner, with special consideration given to the nature and significance of the overall structural damage. In buildings which are repaired, but not modified, future earthquakes may cause moment connection failures with resulting large building deflections and high rotation demands at gravity connections. When repairing gravity connections, consideration should be given to providing connections with the ability to rotate with little or no reduction in vertical load carrying capacity, possibly by dissipating energy (through the use of slip critical bolts with horizontal short slotted holes). Commentary: In many cases, shear connections which were not a part of the lateral-force-resisting system provided an unanticipated redundancy after damage occurred to the primary WSMF lateral system. While repair details could provide for rotation to minimize damage, such details should not eliminate the beneficial effect of the extra strength and stiffness these shear connections provide. This is especially important in framing systems with low moment frame redundancy. The suggestion of providing gravity connections with slotted holes and slip critical bolts may be a reasonable compromise. Such a connection would be capable of providing some additional, unintended, strength and stiffness for the building but would also be able to withstand relatively large rotations without jeopardizing the gravity support the connection is actually intended to provide. 6.3.9 Reuse of Bolts

Bolts in a connection displaying bolt damage or plate slippage should not be re-used. As indicated in the AISC Specification for Structural Joints using ASTM A325 or A490 Bolts (American Institute of Steel Construction - 1985), A490 bolts and galvanized A325 bolts should not be retightened and re-used under any circumstances. Other A325 bolts may be reused if determined to be in good condition. Touching up or retightening previously tightened bolts which may have been loosened by the tightening of adjacent bolts need not be considered as reuse provided the snugging up continues from the initial position and does not require greater rotation, including the tolerance, than that required by Table 5 of the AISC Specification. Bolts in connections displaying bolt or plate slippage should not be reused. Commentary: Proper performance of high strength bolts used in slip critical applications requires proper tensioning of the bolt. Although a number of 6-13

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methods are available to ensure that bolts are correctly tensioned, the most common methods relate to torquing of the nut on the bolt. When a bolt has been damaged, the torquing characteristics will be altered. As a result, damaged bolts may either be over-tightened or under-tightened, if reinstalled. The threads of ASTM A-490 bolts and galvanized ASTM A-325 bolts become slightly damaged when tightened, and consequently, should not be reused. To determine if an ungalvanized ASTM A-325 bolt is suitable for re-use, a nut should be run up the threads of the bolt. If this can be done smoothly, without binding, then the bolt may be re-used. 6.3.10 Welding Specification

Welded repairs involving thick plates and conditions of high restraint should be specified with caution. These conditions can lead to large residual stresses and in some cases, initiation of cracking before the structure is loaded. The potential for problems can be reduced by specifying appropriate joint configurations, welding processes, control of preheat, heat input during welding and cooldown, as well as selecting electrodes appropriate to the application. Engineers who do not have adequate knowledge to confidently specify these parameters should seek consultation from a person with the required expertise. 6.4 Preparation 6.4.1 Welding Procedure Specifications

A separate Welding Procedure Specification (WPS) should be established for every different weld configuration, welding position, and material specification. Two categories of qualified welding procedures are given in AWS D1.1-94. The WPS should be reviewed by the structural engineer responsible for the repairs. The WPS is a set of focused instructions to the welders and inspectors stating how the welding is to be accomplished. Each type of weld should have its own WPS solely for the purpose of that weld. The WPS should include instructions for joint preparation based on material property and thickness, as well as welding parameters. Weld process, electrode type, diameter, stick-out, voltage, current, and interpass temperature should be clearly defined. In addition, joint preheat and postheat requirements should be specified as appropriate, including insulation guidelines if applicable. The WPS should also list appropriate interim specification requirements that are mandated by the project specification. Commentary: Preparation of the WPS is normally the responsibility of the fabricator/erector. Sample formats for WPS preparation and submission are included in AWS D1.1. Some contractors fill out the WPS by inserting references to the various AWS D1.1 tables rather than the actual data. This does not meet the intent of the WPS which is to provide specific instructions to the welder and inspector on how the weld is to be performed. The actual values of the parameters to be used should be included in the WPS submittal.

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6.4.2 Welder Training

Training of welders should take place at the outset of the repair operations. Welders and inspectors should be familiar with the WPS, and should be capable of demonstrating familiarity with each of its aspects. A copy of the WPS should be located on site, preferably at the connection under repair, accessible to all parties involved in the repair. 6.4.3 Welder Qualifications

Welders must be qualified and capable of successfully making the repair welds required. All welders should be qualified to the AWS D1.1 requirements for the particular welding process and position in which the welding is to be performed. Successful qualification to these requirements, however, does not automatically demonstrate a welder’s ability to make repair welds for all the configurations that may be encountered. Specific additional training and/or experience may be required for repair situations. Welders performing repairs should have a minimum of two years of verifiable field experience for the welding process that is employed, as well as experience in arc-gouging and thermal cutting of material. Inexperienced welders should demonstrate their ability to make proper repair welds. This may be done by welding on a mockup assembly (see Section 6.4.4) that duplicates the types of conditions that would be encountered on the actual project. Alternatively, the welder could demonstrate proficient performance on the actual project, providing this performance is continuously monitored, start to finish, during the construction of at least the first weld repair. This observation should be made by a qualified welding inspector or Welding Engineer. 6.4.4 Joint Mock-ups

A joint mock-up should be considered as a training and qualification tool for each type of repair the welder is to perform that is more challenging than work in which he/she has previously demonstrated competence, or at the discretion of the structural engineer. This will allow the welder to become familiar with atypical welds, and will give the inspector the opportunity to clearly observe the performance of each welder. An entire mock-up is recommended for each such case, rather than only a single pass or portion of the weld as all welding positions and types of weld would be experienced, thus showing the welder capable of successfully completing the weld in all required positions, and applying all heating requirements. 6.4.5 Repair Sequence

Repair sequence should be considered in the design of repairs, and any sequencing requirements should be clearly indicated on the drawings and WPS. Structural instabilities or high residual stresses could arise from improper sequencing. The order of repair of flanges, shear plates, fractured columns, etc. should be indicated on the drawings to reduce possible residual stresses.

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6.4.6 Concurrent Work

The maximum number of connections permitted to be repaired concurrently should be indicated on the drawings or in the project specifications. Commentary: Although a connection is damaged, it may still posses significant ability to participate in the structure’s lateral load resisting system. Consideration should be given to limiting the total number of connections being repaired at any one time, as the overall lateral load resistance of the structure may be temporarily reduced by some repair operations. If many connections are under repair simultaneously, the overall lateral resistance of the remaining frame connections may not be adequate to protect the structure’s stability. Although this appears to fall under the category of means and methods, the typical contractor would have no way of determining the maximum number of connections that can be repaired at any one time without requiring supplemental lateral bracing of the building during construction. Therefore, the structural engineer should take a pro-active role in determining this. 6.4.7 Quality Control/Quality Assurance:

Quality control and quality assurance should follow the guidelines set forth in Section 6.6 and Chapters 9, 10 and 11 of these Interim Guidelines. 6.5 Execution 6.5.1 Introduction

Recommended general requirements should include the following: 1. Strict enforcement of the welding requirements in AWS D1.1 as modified in 1994 UBC Chapter 22, Division VIII or IX. Commentary: Following the 1994 Northridge Earthquake, the AWS established a presidential task group to determine if deficiencies in the D1.1 code contributed to the unexpected damage, and to determine if modifications to the code should be made. That task group noted some areas of practice, related to steel moment frames in seismic zones, that could be improved relative to D1.1. These included the following recommendations: a) the root pass of the complete joint penetration welds of beam to column flanges should not exceed 1/4 inch in size, for prequalified procedures. b) where notch tough weld metal is desired, such as at the critical complete joint penetration welds of beam flanges to columns, the maximum interpass temperature should not exceed 550o. 6-16

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c) when a FCAW process is used, the welding procedure specification should conform to the electrode manufacturer’s recommendations. d) the criteria for joints loaded in tension should apply to both top and bottom flange connections in moment frames. Future editions of the AWS D1.1 code may adopt some or all of these recommendations. In the interim period, the structural engineer should consider including these recommendations in the project welding specifications, to supplement the standard AWS D1.1 requirements. 2. Implementation of the special inspection requirements in 1994 UBC section 1701 {NEHRP-91 Section 1.6.2.6} and AWS D1.1. Visual inspection means that the inspector inspects the welding periodically for adherence to the approved Welding Procedure Specification (WPS) and AWS D1.1 starting with preliminary tack welding and fit-up and proceeding through the welding process. Reliance on the use of nondestructive testing (NDT) at the end of the welding process alone should be avoided. Use visual inspection in conjunction with NDT to improve the chances of achieving a sound weld. 3. Require the fabricator to prepare and submit a WPS with at least the information required by AWS D1.1 as discussed in Section 4. 4. Welding electrodes should be capable of depositing weld metal with a minimum notch toughness as described in Chapter 8. 5. All welds for the frame girder-column joints should be started and ended on weld runoff tabs where practical. All weld tabs should be removed, the affected area ground smooth and tested for defects using the magnetic particle method. Acceptance criteria should be AWS D1.1, section 8.15.1. Imperfections less than 1/16” should be removed by grinding. Deeper gouges, areas of lack of fusion, slag inclusions, etc., should be removed by gouging or grinding and rewelding following the procedures outlined above. 6. Weld dams do not meet the intent of weld tabs, are not permitted by AWS D1.1, and should not be permitted in the work. Dams are not necessary when proper bead size limitations are observed. 7. Steel backing (backing bars), if used, should be removed from new and/or repaired welds at the girder bottom flange, the weld root back-gouged by air arcing and the area tested for defects using the magnetic particle method, as described above. The weld should be completed and reinforced with a fillet weld. Removal of the weld backing at repairs of the top girder flange weld may be considered, at the discretion of the structural engineer.

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Prior to removing weld backing, the contractor should prepare and submit a written WPS for review by the structural engineer. The WPS should conform to the requirements of AWS D1.1. In addition, a WPS should be prepared for each welding process to be used on the project and should include minimum preheat, maximum interpass temperatures, and the as-gouged cross section which must simulate a prequalified joint design of D1.1. If for any reason the WPS does not meet the prequalified limits of AWS it should be qualified by test, in accordance with Section 5.2 of AWS D1.1 In addition the contractor should propose the method(s) which will be used to remove the weld backing, back gouge to sound metal and when during this process he will apply preheat. Although project conditions may vary, the following general guidelines may be considered: The steel backing may be removed by either grinding or by the use of air arc, or oxyfuel gouging. The zone just beyond the theoretical 90 degree intersection of the beam to column flange should be removed by either air arc or oxy-fuel gouging followed by a thin grinding disk, or by a grinding disk alone. This shallow gouged depth of weld and base metal should then be tested by MT to determine if any linear indications remain. If the area is free of indications the area may then be re-welded. The preheat should be maintained and monitored throughout the process. If no further modification is to be made or if the modification will not be affected by a reinforcing fillet weld, the reinforcing fillet may be welded while the connection remains at or above the minimum preheat temperature and below the maximum interpass temperature. If weld tabs were used and are to be removed in conjunction with the removal of the weld backing, the tabs should be removed after the weld backing has been removed and fillet added. If cover plates are to be added, the removal of the weld tabs may occur before or after the plate is added depending on the width and configuration of the plate. This sequence should be submitted to the structural engineer for his/her approval prior to the beginning of the work. The weld tabs may be removed by air arc or oxy-fuel gouging followed by grinding or by grinding alone. The resulting contour should blend smoothly with the face of the column flange and the edge of the beam flange and should have a radius of 1/4-3/8 inch. The finished surface should be visually inspected for contour and any visually apparent indications. This should be followed by magnetic particle testing (MT). Linear indications found in this location of the weld may be detrimental. They may be the result of the final residue of defects commonly found in the weld tab area. Linear indications should be removed by lightly grinding or using a cutting tool until the indication is removed. If after removal of the defect the ground area can be tapered and is not beyond the theoretical 90 degree intersection of the beam flange edge and column flange, weld repair may not be necessary and should be avoided if possible. If the defect removal has extended into the theoretical weld section, then weld repair may be necessary. The weld repair should be performed in accordance with the contractor’s WPS, with strict adherence to the preheat requirements. 6-18

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The surface should receive a final visual inspection and MT after all repairs and surface conditioning has been completed. End dams, if present, should be removed if UT indicates rejectable flaws in the area of the end dam. Prior to removal of end dams, the contractor should submit a removal / repair plan which lists the method of dam removal, defect removal, welding procedure including, process, preheat, and joint configuration. The tab may be removed by grinding, air arc or oxy-fuel torch. Any weld defects should be removed by grinding or cutting tools, or by air arc gouging followed by grinding. The individual performing defect removal should be furnished the UT results which describe the location depth and extent of the defect(s). When the individual removing the defects has completed this operation, and has visually confirmed that no remnants remain, the surface should be tested by MT. Additional defect removal and MT may occur until the MT tests reveal that the defects have been removed. The contour of the surface at this point may be too irregular in profile to allow welding to begin. The surface should be conditioned by grinding or using a cutting tool to develop a joint profile which conforms to the WPS. Prior to welding MT should be performed to determine if any additional defects have been exposed. Based upon a satisfactory MT the joint may be prepared for welding. Weld tabs (and backing if necessary) should be added. The welding may begin and proceed in accordance with the WPS. The theoretical weld must be completed for its full height and length. Careful attention should be paid to ensure that weld bead size does not exceed that permitted by the WPS. If specified, the weld tabs and backing should be removed in accordance with the guideline section describing this technique. The final weld should be inspected by MT and UT. Commentary: Removal of the weld backing from the top flange may be difficult, particularly along perimeter frames where access to the outer side is restricted. Since the potential stress riser produced by the unwelded portions of the weld backing are not located on the extreme outer fiber of the frame girder, the benefits of removal may be limited in repair situations. Nevertheless, there may be benefits to providing a weld with a more favorable contour (i.e. that produced by the reinforcing fillet). Tests conducted to date have not been conclusive with regard to the benefit of top flange weld backing removal. At this time, there is no direct evidence that removal of weld backing from continuity plates in the column panel zone is required. The decision to remove end dams should be based upon the results of UT. Since numerous stop - starts have occurred in this section of the theoretical weld, rejectable edge indications may reduce the integrity of the weld, especially during dynamic or seismic loading. If, however the area is found acceptable by UT removal is not necessary. 6-19

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Excessive weaving of the weld bead, which can lead to unacceptable stresses at the toe of each weave, should not be allowed. However, some oscillation of the electrode may be required to obtain good fusion. 6.5.2 Girder Repair

If at bottom flange repairs back gouging removes sufficient material such that a weld backing is required for the repair, after welding the backing should be removed from the girder. Alternatively, a double-beveled joint may be used The weld root should be inspected and tested for imperfections, which if found, should be removed by back-gouging to sound material. A reinforcing fillet weld should be placed at “T” joints equal to one-quarter of the girder flange thickness. It need not exceed 3/8 inch (see Note J, Figure 2.4 of AWS D1.1.) If the bottom flange weld requires repair, the following procedure may be considered: 1.

The root pass should not exceed a 1/4 inch bead size.

2.

The first half-length root pass should be made with one of the following techniques, at the option of the contractor: a)

The root pass may be initiated near the center of the joint. If this approach is used, the welder should extend the electrode through the weld access hole, approximately 1” beyond the opposite side of the girder web. This is to allow adequate access for clearing and inspection of the initiation point of the weld before the second half-length of the root pass is applied. It is not desirable to initiate the arc in the exact center of the girder width since this will limit access to the start of the weld during post-weld operations. After the arc is initiated, travel should progress towards the end of the joint (outboard beam flange edge), and the weld should be terminated on a weld tab.

b)

The weld may be initiated on the weld tab, with travel progressing toward the center of the girder flange width. When this approach is used, the welder should stop the weld approximately 1” before the beam web. It is not advisable to leave the weld crater directly in the center of the beam flange width since this will hinder post-weld operations.

3.

The half length root pass should be thoroughly slagged and cleaned.

4.

The end of the half length root pass that is near the center of the beam flange should be visually inspected to ensure fusion, soundness, freedom from slag inclusions and excessive porosity. The resulting bead profile should be suitable for obtaining fusion by the subsequent pass to be initiated on the opposite side of the girder web. If the profile is not conducive to good fusion, the start of the first root pass should be ground, gouged, chipped or otherwise prepared to ensure adequate fusion.

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5.

The second half of the weld joint should have the root pass applied before any other weld passes are performed. The arc should be initiated at the end of the half length root pass that is near the center of the beam flange, and travel should progress to the outboard end of the joint, terminating on the weld tab.

6.

Each weld layer should be completed on both sides of the joint before a new layer is deposited.

7.

Weld tabs should be removed and ground flush to the beam flange. Imperfections less than 1/16” should be removed by grinding. Deeper gouges, areas of lack of fusion, slag inclusions, etc. should be removed by gouging or grinding and rewelding following the procedures outlined above.

6.5.3 Weld Repair (Types W1, W2, or W3)

When W1, W2, or W3 cracks are found, the column base metal should be evaluated using UT to determine if fractures have progressed into the flange. This testing should be performed both during the period of discovery and during repair. When a linear planar-type defect such as a crack or lack of fusion can be determined to extend beyond one-half the thickness of the beam flange, it is generally preferred to use a double-sided weld for repair (even though the fracture may not extend all the way to the opposite surface.) This is because the net volume of material that needs to be removed and restored is generally less when a double-sided joint is utilized. It also results in a better distribution of residual stresses since they are roughly balanced on either side of the center of the flange thickness. Repair of these cracks may warrant total removal of the original weld, particularly if multiple cracks are present. If the entire weld plus some base metal is removed care must be taken not to exceed the root opening and bevel limits of AWS D1.1 unless a qualified by test WPS is used. If this cannot be avoided one of two options is available: 1. The beveled face of the beam and/or the column face may be built up (buttered) until the desired root opening and angle is obtained. 2. A section of the flange may be removed and a splice plate inserted. Commentary: Building up base metal with welding is a less intrusive technique than removing large sections of the base metal and replacing with new plate. However, this technique should not be used if the length of build-up exceeds the thickness of the plate. 6.5.4 Column Flange Repairs - Type C2

Damage type C2 is a pullout type failure of the column flange material. The zone should be conditioned to a concave surface by grinding and inspected for soundness using MT. The 6-21

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concave area may then be built up by welding. The joint contour described in the WPS should specify a "boat shaped" section with a "U" shaped cross section and tapered ends. The weld passes should be horizontal stringers placed in accordance with the WPS. Since stop/starts will occur in the finished weld, care must be taken to condition each stop/start to remove discontinuities and provide an adequate contour for subsequent passes. The final surface should be ground smooth and flush with the column face. This surface and immediate surrounding area should be subjected to MT and UT. 6.6 STRUCTURAL MODIFICATION 6.6.1 Definition of Modification

Within the context of damage to WSMF connections, the term "structural modification" refers to alteration of the connection to improve its earthquake performance and that of the structure as a whole. This typically involves substantial changes to the connection's geometry, capacity, or relevant limit states (e.g. flexural or shear strength or stiffness). Work that includes removal of existing welds and replacement with welds of improved toughness and/or workmanship is not considered modification under these Interim Guidelines. Commentary: This term is contrasted with "repair," wherein the essential behavior of the connection is unchanged as a result of the repair effort. Geometrical or stiffness changes can involve spatial alterations to the elements of the connection, such as adding column stiffeners or the addition of new connection elements, such as cover plates, upstanding ribs, side plates or haunches. Changes to the connection's capacity, either in flexure or shear, may occur as a result of the addition of new connection components. Altering the connection's relevant limit states may occur, for example, when the location of the plastic hinge is shifted away from its original location or the shear capacity of the connection or one of its elements determines the behavior of the connection. Much of the damage that occurred in the Northridge Earthquake has been attributed to the presence of “crack like” conditions at the root of the complete joint penetration beam flange to column flange welds. These crack like conditions included lack of fusion at the weld root as well as the presence of partially fused weld backing. Some engineers believe that if these crack-like conditions are removed, substantial improvement in connection performance can be obtained. SAC conducted specific testing in the Phase 1 program in which such “dressing up” of these welds was performed. The performance of the connection in these tests was mixed, and often not substantially improved relative to that of connections in which the backing was left in place. Based on these tests, removal of weld backing, backgouging and repairing welds, and reinforcing with a fillet is not recommended as a means of connection modification, although it is an acceptable means of repair for joints with type W1 and W2 damage.

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Several engineers and researchers knowledgeable in fracture mechanics have suggested that the standard, unreinforced moment connection could perform acceptably if weld metal and base metal with adequate toughness were incorporated, and beam flange to column joints are executed in such a manner that large crack-like discontinuities are not present (removal of backing and weld tabs, backgouging, and reinforcing with a fillet). Other engineers knowledgeable in mechanics of materials (Blodgett - 1994) believe that regardless of the toughness of the weld metal employed, the connection configuration is such that reliable performance is unlikely. If joints with adequate weld metal toughness can provide substantially more reliable performance, then, removal of existing low-toughness welds and replacement with new tough material may be an acceptable means of modification. To date, only limited testing of such assemblies have been conducted. In one test (Popov - 1995) an assembly consisting of a W36 x 150 beam connected to a W14 x 257 column and originally fabricated using E70T4 electrodes (not having rated notch toughness) was repaired following initial testing by completely removing the complete joint penetration welds of the beam flanges to column flanges and replacing them with new welds made with electrodes having specified notch toughness. Weld backing and weld tabs were removed and the welds were reinforced with a fillet. The specimen was successfully tested to a plastic rotation of 0.04 radians. However, until additional research can be performed to quantify the reliability obtained through the use of notch tough weld metal, this is not recommended by itself as a method of modification in these Interim Guidelines. Modification of the structure as a whole, as opposed to individual connection modifications, can be an effective means of obtaining more reliable performance. The addition of braced frames, shear walls, energy dissipation systems, base isolation, etc., can be used to reduce the total deformation demand induced in the structure by earthquakes, and consequently the need for the moment-resisting connections to resist large plastic rotation demands. Interim Guidelines for these types of modifications are not directly included in this document. However, sections on connection qualification presented below provide information that can be used to determine the plastic rotation capacity of existing connections in the building. Once this is determined, the effectiveness of proposed global modification measures can be assessed, as part of the design process. 6.6.2 Damaged vs. Undamaged Connections

Engineers should inform building owners that substantial improvements in the reliability of future earthquake performance of a WSMF building can be obtained by structural modification. Modification can be made at connections that have sustained damage as well as those that are undamaged. On the basis of cost, some owners may elect to modify those connections which have been damaged, and which will be repaired, but not other, undamaged connections. If a 6-23

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building has had only a few scattered connections damaged, such an approach will not result in any significant improvement in future building performance, and is not recommended. If a substantial number of connections in a building have been damaged and will be repaired, modification of these damaged connections may improve future building performance, depending on the distribution of damaged connections, throughout the building. Therefore, consideration of such an approach has been recommended in Chapter 4 of these Interim Guidelines. If possible, it is recommended that the modification of connections follow a rational spatial distribution, so a to distribute the enhanced energy dissipation capacity (and ductility) throughout the building. As a minimum, structural modification should consider the effect of those modifications on the performance of the lateral system as well as on the performance of individual components of the frames. An appropriate analysis should be performed of the building, considering the modifications, to ensure that undesirable stiffness irregularities are not introduced or made more severe, and that excessive demand is not concentrated in connections unable to resist the applied loads or deformations. The effects of connection modifications on inelastic demands in adjacent columns and panel zones should be considered. Commentary: Structural modification of connections will normally be performed as a means of enhancing the expected performance of the building in future earthquakes, by minimizing the potential for fractures. The intent of modification is to make the connection sufficiently strong that inelastic behavior of the frame will be controlled by the formation of plastic hinges within the girder spans. Evaluation of statistical data on the types and distribution of damage experienced by 89 buildings affected by the Northridge Earthquake (Bonowitz & Youssef - 1995) indicates that the spatial distribution of damage other than small root indications (Type W1) has modest correlation with the distribution of high seismic demands predicted by traditional analytical approaches. The distribution of type W1 indications appear to be random. A modification scheme that selects connections on the basis of existing damage could therefore result in a random distribution of connections with improved performance characteristics. In such an approach, connections that may undergo high plastic rotation demands or may be part of a lateral system with limited redundancy might not be modified in favor of connections damaged as a result of poor workmanship. The result of this could be a modified system with only marginally improved behavior. Connections that have not been modified can be expected to have a significant failure rate in subsequent earthquakes, at near-elastic demand levels. Therefore, the amount of improvement obtained by modifying only the damaged connections is not directly quantifiable. Generally, as more connections in the building are modified, the potential performance of the building should improve. An alternative approach, and one that appears to represent a more reliable method of ensuring that the earthquake performance of the lateral system is equivalent to that assumed at the time the WSMF was designed, is to modify all of

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the connections. Tests on girder-column connections similar to those found in many buildings suggest that the traditional welded flange/bolted web connection cannot develop the rotational demands implicit in building code designs. Modified connections appear to represent one approach to achieve the required level of deformation capacity. Modification of only selected connections may be a cost-effective approach if the analysis can accurately predict the demand on the connections as well as the consequences of future connection failures in the modified and unmodified connections. The structural engineer should inform the building owner of the assumed benefits as well as the potential disadvantages of a scheme that modifies only a selected number of the connections. The reliability of analyses used to justify such a partial modification scheme is sensitive to the modeling assumptions and the ground motion input. 6.6.3 Criteria

Connection modification intended to permit inelastic frame behavior should be proportioned so that the required plastic deformation of the frame may be accommodated through the development of plastic hinges at pre-determined locations within the girder spans, as indicated in Figure 6-12. Beam-column connections should be designed with sufficient strength (through the use of cover plates, haunches, side plates, etc.) to force development of the plastic hinge away from the column face. This condition may also be attained through local weakening of the beam section, at the desired location for plastic hinge formation. All elements of the connection should have adequate strength to develop the forces resulting from the formation of the plastic hinge at the predetermined location, together with forces resulting from gravity loads.

h

Undeformed frame

Deformed frame shape

Plastic Hinges

drift angle - θ

L’ L

Figure 6-12 - Desired Plastic Frame Behavior Commentary: Nonlinear deformation of frame structures is accommodated through the development of inelastic flexural or shear strains within discrete regions of the structure. At large inelastic strains these regions can develop into 6-25

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plastic hinges, which can accommodate significant concentrated rotations at constant (or nearly constant) load through yielding at tensile fibers and buckling at compressive fibers. If a sufficient number of plastic hinges develop in a frame, a mechanism is formed and the frame can deform laterally in a plastic manner. This behavior is accompanied by significant energy dissipation, particularly if a number of members are involved in the plastic behavior, as well as substantial local damage to the highly strained elements. The formation of hinges in columns, as opposed to beams, is undesirable, as this results in the formation of mechanisms with relatively few elements participating, and consequently little energy dissipation occurring. In addition, such mechanisms also result in local damage to critical gravity load bearing elements. The prescriptive connection contained in the UBC and NEHRP Recommended Provisions prior to the Northridge Earthquake was based on the development of plastic hinges within the beams at the face of the column, or within the column panel zone itself. If the plastic hinge develops in the column panel zone, the resulting column deformation results in very large secondary stresses on the beam flange to column flange joint, a condition which can contribute to brittle failure. If the plastic hinge forms in the beam, at the face of the column, this can result in very large through-thickness strain demands on the column flange material and large inelastic strain demands on the weld metal and surrounding heat affected zones. These conditions can also lead to brittle joint failure. In order to achieve more reliable performance, it is recommended that the connection of the beam to the column be modified to be sufficiently strong to force the inelastic action (plastic hinge) away from the column face. Plastic hinges in steel beams have finite length, typically on the order of half the beam depth. Therefore, the location for the plastic hinge should be shifted at least that distance away from the face of the column. When this is done, the flexural demands on the columns are increased. Care must be taken to assure that weak column conditions are not inadvertently created by local strengthening of the connections. It should be noted that connection modifications of the type described above, while believed to be effective in preventing brittle connection fractures, will not prevent structural damage from occurring. Brittle connection fractures are undesirable because they result in a substantial reduction in the lateral-forceresisting strength of the structure which, in extreme cases, can result in instability and collapse. Connections modified as described in these Interim Guidelines should experience many fewer such brittle fractures than unmodified connections. However, the formation of a plastic hinge within the span of a beam is not a completely benign event. Beams which have formed such hinges may exhibit large buckling and yielding deformation, damage which typically must be repaired. The cost of such repairs could be comparable to the costs incurred in repairing fracture damage experienced in the Northridge Earthquake. The primary difference is that life safety protection will be significantly enhanced and 6-26

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most structures that have experienced such plastic deformation damage should continue to be safe for occupancy, while repairs are made. If the types of damage described above are unacceptable for a given building, then alternative methods of structural modification should be considered, that will reduce the plastic deformation demands on the structure during a strong earthquake. Appropriate methods of achieving such goals include the installation of supplemental braced frames, energy dissipation systems, and similar systematic modifications of the building’s basic lateral force resisting system. 6.6.4 Strength

When these Interim Guidelines require determination of the strength of a framing element or component, this shall be calculated in accordance with the criteria contained in UBC-94, Section 2211.4.2 {NEHRP-91 Section 10.2, except that the factor φshould be taken as 1.0}, restated as follows: 2211.4.1 Member strength. Where this section requires that the strength of the member be developed, the following shall be used: Flexure Shear Axial compression Axial tension Connectors Full Penetration welds Partial Penetration welds Bolts and fillet welds

M s = Z Fy Vs = 0.55 Fy d t Psc = 1.7 Fa A Pst = Fy A Fy A 1.7 allowable (see commentary) 1.7 allowable

Commentary: Partial penetration welds are not recommended for tension applications in critical connections resisting seismic induced stresses. The geometry of partial penetration welds creates a notch-like condition that can initiate brittle fracture under conditions of high tensile strain. 6.6.5 Plastic Rotation Capacity

The plastic rotation capacity of modified connections should reflect realistic estimates of the required level of plastic rotation demand. In the absence of detailed calculations of rotation demand, connections should be shown to be capable of developing a minimum plastic rotation capacity on the order of 0.025 to 0.030 radian. The demand may be lower when braced frames, supplemental damping, base isolation, or other elements are introduced into the moment frame system, to control its lateral deformation; when the design ground motion is relatively low in the range of predominant periods for the structure; and when the frame is sufficiently strong. If calculations are performed to determine the required connection plastic rotation capacity, the capacity should be taken somewhat greater than the calculated deformation demand, due to the high variability and uncertainty inherent in predictions of inelastic seismic response. Until

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better guidelines become available, a required plastic rotation capacity on the order of 0.005 radians greater than the demand calculated for the design basis earthquake (or if greater conservatism is desired - the maximum capable earthquake) is recommended. Rotation demand calculations should consider the effect of plastic hinge location within the beam span, as indicated in Figure 6-12, on plastic rotation demand. Calculations should be performed to the same level of detail specified for nonlinear dynamic analysis for base isolated structures in UBC94 Section 1655 {NEHRP-94 Section 2.6.4.4}. Ground motion time histories utilized for these nonlinear analyses should satisfy the scaling requirements of UBC-94 Section 1655.4.2 {NEHRP-94 Section 2.6.4.4} except that instead of the base isolated period, TI, the structure period, T, calculated in accordance with UBC-94 Section 1628 {NEHRP-94 Section 2.3.3.1} should be used. Commentary. Traditionally, structural engineers have calculated demand in moment frames by sizing the members for strength and drift using code forces (either equivalent static or reduced dynamic forces) and then "developing the strength of the members." Since 1988, "developing the strength" has been accomplished by prescriptive means. It was assumed that the prescribed connections would be strong enough so that the girder would yield (in bending), or the panel zone would yield (in shear) in a nearly perfectly plastic manner producing the plastic rotations necessary to dissipate the energy of the earthquake. It is now known that the prescriptive connection is often incapable of behaving in this manner. In the 1994 Northridge earthquake, many moment-frame connections fractured with little evidence of plastic hinging of the girders or yielding of the column panel zones. Testing of moment frame connections both prior to and subsequent to the earthquake suggests that the standard welded flange-bolted web connection is unable to reliably provide plastic rotations beyond about 0.005 radian for all ranges of girder depths and often fails below that level. Thus, for frames designed for code forces and for the code drift limits, new connection configurations must be developed to reliably accommodate such rotation without brittle fracture. In order to develop reasonable estimates of the plastic rotation demands on a frame’s connections, it is necessary to perform inelastic time history analyses. For regular structures, approximations of the plastic rotation demands can be obtained from linear elastic analyses. Analytical research (Newmark and Hall 1982) suggests that for structures having the dynamic characteristics of most WSMF buildings, and for the ground motions typical of western US earthquakes, the total frame deflections obtained from an unreduced (no R or Rw factor) dynamic analysis provide an approximate estimate of those which would be experienced by the inelastic structure. For the typical spectra contained in the building code, this would indicate expected drift ratios on the order of 1%. The drift demands in a real structure, responding inelastically tend to concentrate in a few stories, rather than being uniformly distributed throughout the structure’s 6-28

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height. Therefore, it is reasonable to expect typical drift demands in individual stories on the order of 1.5% to 2% of the story height. As a rough approximation, the drift demand may be equated to the joint rotation demand, yielding expected rotation demands on the order of perhaps 2%. Since there is considerable variation in ground motion intensity and spectra, as well as the inelastic response of buildings to these ground motions, conservatism in selection of an appropriate connection rotation demand is warranted. In recent testing of large scale subassemblies incorporating modified connection details, conducted by SAC and others, when the connection design was able to achieve a plastic rotation demand of 0.025 radians or more for several cycles, the ultimate failure of the subassembly generally did not occur in the connection, but rather in the members themselves. Therefore, the stated connection capacity criteria would appear to result in connections capable of providing reliable performance. It should be noted that the connection assembly capacity criteria for the modification of existing buildings, recommended by these Interim Guidelines, is somewhat reduced compared to that recommended for new buildings (Chapter 7). This is typical of approaches normally taken for existing structures. For new buildings, these Interim Guidelines discourage building-specific calculation of required plastic rotation capacity for connections and instead, encourage the development of highly ductile connection designs. For existing buildings, such an approach may lead to modification designs that are excessively costly, as well as the modification of structures which do not require such modification. Consequently, an approach which permits the development of semi-ductile connection designs, with sufficient plastic rotation capacity to withstand the expected demands from a design earthquake is adopted. It should be understood that buildings modified to this reduced criteria will not have the same reliability as new buildings, designed in accordance with the recommendations of Chapter 7. The criteria of Chapter 7 could be applied to existing buildings, if superior reliability is desired. When performing inelastic frame analysis, in order to determine the required connection plastic rotation capacity, it is important to accurately account for the locations at which the plastic hinges will occur. Simplified models, which represent the hinge as occurring at the face of the column, will underestimate the plastic rotation demand. This problem becomes more severe as the column spacing, L, becomes shorter and the distance between plastic hinges, L’, a greater portion of the total beam span. In extreme cases, the girder will not form plastic hinges at all, but instead, will develop a shear yield, similar to an eccentric braced frame.

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6.6.6 Connection Qualification and Design

Modified girder-column connections may be qualified by testing or designed using calculations. Qualification by testing is the preferred approach. Preliminary designs of connections to be qualified by test may be obtained using the calculation procedures of Section 6.6.6.3. The procedures of that section may also be used to calibrate previous tests of similar connection configurations to slightly different applications, by extrapolation. Extrapolation of test results should be limited to connections of elements having similar geometries and material specifications as the tested connections. Designs based on calculation alone should be subject to qualified independent third party review. 6.6.6.1 Qualification Test Protocol

Unless future testing programs reveal significant effects of dynamic loading rate or time history loading, a testing protocol similar to ATC-24, Guidelines for Cyclic Seismic Testing of Components of Steel Structures (Applied Technology Council - 1992), is recommended as the basis for qualification tests. The testing program should replicate as closely as practical the anticipated conditions in the field, including such factors as: a) Member sizes. b) Material specifications. c) Welding process, details and construction conditions. d) Cover-plates, continuity plates, web tabs, bolts, and doubler plates. e) Connection configuration (e.g., beams on both sides). f) Induced stresses because of restraint conditions on the welds and connection members. g) Axial load, where pertinent. h) Gravity load, where significant. The testing program should be organized to provide as much information as possible about the capability of the connections selected. The following minimum program is recommended: a) Test two full size specimens of the largest representative beam/column assembly in the project. b) Test one additional full size specimen for each beam/column assembly with significantly different interaction properties, such as beam flange width-thickness (b/t) ratio, panel zone stress/distortion, etc. If any of the specimens fails to meet the qualification criteria, the connection should be redesigned and retested.

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Where two-sided connections are used in the structure, and the type of connection being used can be expected to perform differently in a two-sided use than in one-sided use, it should be tested in the two-sided configuration as well as the one-sided. Two-sided connection assemblies can be expected to behave differently than one-sided assemblies, for example, when panel zone distortions will be significantly different, or when systems involve transfer of stress to the column by plates, welds, or other elements which are connected to the beams on both sides of the column. Testing to include axial load should be considered when analysis indicates that significant tension can be expected to occur in a significant number of the columns represented by the specimen and where the connection type relies on the through-thickness strength of the column flanges. If the presence of a floor slab is anticipated to have significant influence on either the location or mechanism of the plastic hinge formed, than this should also be included in the test specimen. Commentary: The use of connection configurations that have been qualified by test is the preferred approach. While the testing of all connection geometries and member combinations in any given building is not practical, the number of tests must be large enough to be meaningful yet small enough to not be unreasonably costly. Testing, within the limitations of test specimen simplification, has the advantage of being able to replicate fabrication and welding procedures, joint geometry and member size, and potential modes of failure. If the testing is done in a manner consistent with other testing programs, reasonable comparisons can be made. On the other hand, testing is expensive and it is difficult to realistically test the girder-column connection using actual restraint conditions and earthquake loading rates. Calculations offer an economical alternative to testing that can accommodate different girder and column sizes, altered connection geometries, and member properties. Nevertheless, recent testing on girdercolumn connections from WSMFs casts doubt on some fundamental assumptions upon which the calculations are based and therefore, they should be used with caution. Since the level of confidence in connections developed strictly on the basis of calculations may not be as high as those based on tests, the use of testing is encouraged. Tests are, however, relatively expensive and a reasonable degree of flexibility in interpreting the results of limited testing programs must be acknowledged. How much extrapolation should be accepted is a difficult decision. As additional testing is done, more information may be available on what constitutes "conservative" testing conditions, thereby allowing easier decisions relative to extrapolating tests to actual conditions which are likely to be less demanding than the tests. For example, it is hypothesized that connections of shallower, thinner flanged members are likely to be more reliable than similar connections consisting of deeper, thicker flanged members. Thus, it may be possible to test the

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largest assemblages of similar details and extrapolate to the smaller member sizes -- at least in comparable member group families. 6.6.6.2 Acceptance Criteria

The minimum acceptance criteria for connection qualification for specimens tested in accordance with these Interim Guidelines should be as follows: a) The connection should develop beam plastic rotations as indicated in Section 6.6.5, for at least one complete cycle. b) The connection should develop a minimum strength equal to 80% of the plastic strength of the girder, calculated using minimum specified yield strength Fy, throughout the loading history required to achieve the required plastic rotation capacity, as indicated in a), above. Commentary: Many connection configurations will be able to withstand plastic rotations on the order of 0.025 radians or more, but will have sustained significant damage and degradation of stiffness and strength in achieving this deformation. The intent of the acceptance criteria presented in this Section is to assure that when connections experience the required plastic rotation demand, they will still have significant remaining ability to participate in the structure’s lateral load resisting system. 6.6.6.3 Calculations

All connections designs should be based on test data and the use of connections based upon calculations only is not recommended. An approved program of variations on the tested proto-typical connections may use calculations to assist in extrapolation of results. Calculations should be correlated to tested material properties for base metals and welds. The properties should be those corresponding to the axes of loading of the base metal and weld in the joints and to the welding processes and materials intended for use. The tested properties may be specific to the materials and processes to be used in the project, or based on a statistically-based testing program. Use of properties inferred from other testing programs must be done with appropriate care and, where such inferred properties are used, designs should reflect the uncertainty inherent in such an indirect approach. Calculations should initiate with the selection of a connection configuration, such as one of those indicated in Section 6.6.7, that will permit the formation of a plastic hinge within the beam span, away from the face of the column, when the frame is subjected to gravity and lateral loads. 6.6.6.3.1 Material Strength Properties In the absence of project specific material property information (for example, mill test reports), the values listed in Table 6-3 should be used to determine the strength of steel shape and 6-32

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plate for purposes of calculation. The permissible strength for weld metal should be taken in accordance with the building code. Table 6-3 - Properties for Use in Connection Modification Design Material Fy (ksi) Fy m (ksi) 1 A36 Beam 36 Dual Certified Beam 50 Axial, Flexural 552 Shape Group 1 Shape Group 2 582 Shape Group 3 572 Shape Group 4 542 Through-Thickness A572 Column/Beam 50 Axial, Flexural 582 Shape Group 1 582 Shape Group 2 572 Shape Group 3 572 Shape Group 4 552 Shape Group 5 Through-Thickness Notes: 1. See Commentary 2. Based on coupons from web. For thick flanges, the Fy flange is approximately 0.95 Fy web. 3. See Commentary

Fu (ksi) 1

65 min.

Note 3 65 min.

Note 3,

Commentary: Table 6-3, Note 1 - The material properties for steel nominally designated on the construction documents as ASTM A36 can be highly variable and in recent years, steel meeting the specified requirements for both ASTM A36 and A572 has routinely been incorporated in projects calling for A36 steel. Consequently, unless project specific data is available to indicate the actual strength of material incorporated into the project, the properties for ASTM A572 steel should be assumed when ASTM A36 is indicated on the drawings, and the assumption of a higher yield stress results in a more severe design condition. Table 6-3, Note 3 - The causes for through-thickness failures of column flanges (types C2, C4, and C5), observed both in buildings damaged by the Northridge Earthquake and in some test specimens, are not well understood. They are thought to be a function of the metallurgy and “purity” of the steel; conditions of loading including the presence of axial load and rate of loading application; conditions of tri-axial restraint; conditions of local hardening and embrittlement within the weld’s heat affected zone; and by the relationship of the connection components as they may affect flange bending stresses and flange curvature induced by panel zone yielding. Given the many complex factors which can affect the through-thickness strength of the column flange, determination of a reliable basis upon which to set permissible design stresses will require significant research. 6-33

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Interim Recommendation No. 2 (SEAOC-1995) included a value of 40 ksi, applied to the projected area of beam flange attachment, for the throughthickness strength to be used in calculations. This value was selected because it was consistent with the successful tests of cover plated assemblies conducted at the University of Texas at Austin (Engelhardt and Sabol - 1994). However, because of the probable influence of all the factors noted above, this value can only be considered to reflect the specific conditions of those tests and specimens. Although reduced stresses at the column face produced acceptable results in the University of Texas tests, the key to that success was more likely the result of forcing the plastic hinge away from the column than reduction of the throughthickness stress by the cover plates. Reduction of through-thickness column flange stress to ever lower levels by the use of thicker cover plates is not recommended, since such cover plates will result in ever higher forces on the face of the column flange. Notwithstanding all of the above, successful tests using cover plates and other measures of moving hinges (and coincidentally reducing through-thickness stress) continue to be performed. In the interim, structural engineers choosing to utilize connections relying on through-thickness strength should recognize that despite the successful testing, connections relying on through-thickness strength can not be considered to be fully reliable until the influence of the other parameters discussed above can be fully understood. A high amount of structural redundancy is recommended for frames employing connections which rely on throughthickness strength of the column flange. 6.6.6.3.2 Determine Plastic Hinge Location The desired location for the formation of plastic hinges should be determined as a basic parameter for the calculations. For beams with gravity loads representing a small portion of the total flexural demand, the plastic hinge may be assumed to occur at a distance equal to 1/3 of the beam depth from the edge of the reinforced connection (or start of the weakened beam section), unless specific test data for the connection indicates that a different value is appropriate. Refer to Figure 6-13.

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Edge of reinforced connection

Plastic hinge

L’

L

Connection reinforcement

d/3

Edge of reinforced connection

Beam depth - d

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Figure 6-13 - Location of Plastic Hinge Commentary: The suggested location for the plastic hinge, at a distance d/3 away from the end of the reinforced section is based on the observed behavior of test specimens, with no significant gravity load present. If the significant gravity load is present, this can shift the locations of the plastic hinges, and in the extreme case, even change the form of the collapse mechanism. If flexural demand on the girder due to gravity load is less than about 30% of the girder plastic capacity, this effect can safely be neglected, and the plastic hinge locations taken as indicated. If gravity demands significantly exceed this level then plastic analysis of the girder should be performed to determine the appropriate hinge locations. Note that in zones of high seismicity (UBC Zones 3 and 4, and NEHRP Map Areas 6 and 7) gravity loading on the girders of earthquake resisting frames typically has a very small effect. 6.6.6.3.3 Determine Probable Plastic Moment at Hinges The probable value of the plastic moment, Mpr, at the location of the plastic hinges should be determined from the equation: M pr = 0.95αZ b Fya where: α

(6-1)

is a coefficient that accounts for the effects of strain hardening and modeling uncertainty, taken as: 1.1

when qualification testing is performed or calculations are correlated with previous qualification testing

1.3

when design is based on calculations, alone.

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Fya

is the actual yield stress of the material, as identified from mill test reports. Where mill test data for the project is not traceable to specific framing elements, the average of mill test data for the project for the given shape may be used. When mill test data for the project is not available, the value of Fym, from table 6-3 may be used.

Zb

is the plastic modulus of the section Commentary: The 0.95 factor, in equation 6-1, is used to adjust the yield stress in the beam web, where coupons for mill certification tests are normally extracted, to the value in the beam flange. Beam flanges, being comprised of thicker material, typically have somewhat lower yield strengths than do beam web material. The factor of 1.1 recommended to account for strain hardening, or other sources of strength above yield, agrees fairly well with available test results. It should be noted that the 1.1 factor could underestimate the over-strength where significant flange buckling does not act as the gradual limit on the connection. Nevertheless, the 1.1 factor seems a reasonable expectation of over-strength considering the complexities involved. Connection designs that result in excessive strength in the girder connection relative to the column or excessive demands on the column panel zone are not expected to produce superior performance. There is a careful balance that must be maintained between developing connections that provide for an appropriate allowance for girder overstrength and those that arbitrarily increase connection demand in the quest for a “conservative” connection design. The factors suggested above were chosen in an attempt to achieve this balance, and arbitrary increases in these values are not recommended.

6.6.6.3.4 Determine Beam Shear The shear in the beam, at the location of the plastic hinge should be determined. A free body diagram of that portion of the beam located between plastic hinges is a useful tool for obtaining the shear at each plastic hinge. Figure 6-14 provides an example of such a calculation.

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L/2 Plastic hinge

P

Note: if 2Mpr /L’is less then the gravity shear in the free body (in this case P/2 + wL’/2), then the plastic hinge location will shift and L’ must be adjusted, accordingly

L’ d/3 L

P

VA

w Mpr

“A”

Vp

Mpr

L’

taking the sum of moments about “A” = 0 Vp ={Mpr + Mpr + P L’/2 + wL’2/2}/L’

Figure 6-14 - Sample Calculation of Shear at Plastic Hinge 6.6.6.3.5 Determine Strength Demands on Connection In order to complete the design of the connection, including sizing the various plates and joining welds which make up the connection, it is necessary to determine the shear and flexural strength demands at each critical section. These demands may be calculated by taking a free body of that portion of the connection assembly located between the critical section and the plastic hinge. Figure 6-15 demonstrates this procedure for two critical sections, for the beam shown in Figure 6-14. Plastic hinge

Plastic hinge

Mpr

Mf

Vp

dc

x

Mf=Mpr +Vpx Critical Section at Column Face

Mpr

Mc

Vp x+dc/2

Mf=Mpr +Vp(x+dc/2) Critical Section at Column Centerline

Figure 6-15 - Calculation of Demands at Critical Sections Commentary: Each unique connection configuration may have different critical sections. The vertical plane that passes through the joint between the beam flanges and column (if such joining occurs) will typically define at least one such 6-37

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critical section, used for designing the joint of the beam flanges to the column, as well as evaluating shear demands on the column panel zone. A second critical section occurs at the center line of the column. Moments calculated at this point are used to check weak beam - strong column conditions. Other critical sections should be selected as appropriate to the connection configuration. 6.6.6.3.6 Check for Strong Column - Weak Beam Condition Buildings which form sidesway mechanisms through the formation of plastic hinges in the beams can dissipate more energy than buildings that develop mechanisms consisting primarily of plastic hinges in the columns. Therefore, if an existing building’s original design was such that hinging would occur in the beams rather than the columns, care should be taken not to alter this behavior with the addition of connection reinforcement. To determine if the desired strong column - weak beam condition exists, the connection assembly should be checked to determine if the following equation is satisfied:

∑Z where:

c

(Fyc − f a )

∑M

c

> 1.0

(6-2)

Zc is the plastic modulus of the column section above and below the connection Fyc is the minimum specified yield stress for the column above and below fa is the axial load in the column above and below Mc is the moment calculated at the center of the column in accordance with Section 6.6.6.3.5 Commentary: Equation 6-2 is based on the building code provisions for strong column - weak beam design. The building code provisions for evaluating strong column - weak beam conditions presume that the flexural stiffness of the columns above and below the beam are approximately equal. If non-symmetrical connection configurations are used, such as a haunch on the bottom side of the beam, this can result in an uneven distribution of stiffness between the two column segments. In such cases, a plastic analysis should be considered to determine if an undesirable story mechanism is likely to form in the building.

6.6.6.3.7 Check Column Panel Zone The adequacy of the shear strength of the column panel zone should be checked. For this purpose, the term 0.8ΣMf should be substituted for the term 0.8ΣMs in UBC-94 Section 2211.7.2.1 {0.9ΣφbMp in NEHRP-91 Section 10.10.3.1} repeated below for convenience of reference. Mf is the calculated moment at the face of the column, when the beam mechanism forms, calculated as indicated in Section 6.6.6.3.5, above. 2211.7.2.1 Strength. The panel zone of the joint shall be capable of resisting the shear induced by beam bending moments due to gravity loads plus 1.85 times the prescribed seismic forces, but the shear strength need not exceed that required to develop 0.8ΣMs of the girders framing into the column flanges at the joint. The joint panel zone shear strength may be obtained from the following formula:

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 3b c t c f 2  V = 0.55Fy d c t 1 +  dbdct  

(11-1)

where: bc = width of column flange db = the depth of the beam (including haunches or cover plates) dc = the depth of the column t = the total thickness of the panel zone including doubler plates tcf = the thickness of the column flange

6.6.7 Modification Details

There are many potential details that can be used to modify the performance of girdercolumn joints in existing WSMF structures. Several of these have been tested as part of the SAC Phase 1 effort. While these repair and modification configurations do not represent all potential geometries and the number of replicates is very limited, these tests do provide important insight into the behavior of the modified connection configurations. Figures shown below present conceptual connection configurations that have been subjected to limited testing and have shown an acceptable level of performance. Reference to laboratory testing is provided for those connection configurations for which research has been reported. However, it should be noted that none of these connections has been tested sufficiently at this time to permit unqualified use of the connection. The figures provided in the following sections are schematic, indicating the general type of connection configuration being described. When designing connections patterned after the reported test data, the test specimen details included in the references should be reviewed to determine specific details not shown. The SAC Joint Venture does not endorse or specifically recommend any of the connection details shown in this Section. These are presented only to acquaint the reader with available information on representative testing of different connection configurations that have been performed by various parties. 6.6.7.1 Haunch at Bottom Flange

Figure 6-16 illustrates two alternative configurations of this detail that have been tested (Uang - 1995). The basic concept is to reinforce the connection with the provision of a triangular haunch at the bottom flange. The intended behavior of both configurations is to shift the plastic hinge from the face of the column and to reduce the demand on the CJP weld by increasing the effective depth of the section. In one test, shown on the left of Figure 6-16, the joint between the girder bottom flange and column was cut free, to simulate a condition which might occur if the bottom joint had been damaged, but not repaired. In a second tested configuration, the bottom flange joint was repaired and the top flange was replaced with a locally thickened plate, similar to the detail shown in Figure 6-9.

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d

Thickened flange

d/4

WT, trimmed

2

1

or Bottom Flange Not Attached Top Flange Not Reinforced

or Bottom Flange Attached Top Flange Reinforced

Figure 6-16 - Bottom Haunch Connection Modification Design Issues: This approach developed acceptable levels of plastic rotation. Acceptable levels of connection strength were also maintained during large inelastic deformations of the plastic hinge. This approach does not require that the top flange be modified, or slab disturbed, unless other conditions require repair of the top flange, as in the detail on the left of Figure 6-16. The bottom flange is generally far more accessible than the top flange because a slab does not have to be removed. In addition, the haunch can be installed at perimeter frames without removal of the exterior building cladding. There did not appear to be any appreciable degradation in performance when the bottom beam flange was not re-welded to the face of the column. Eliminating this additional welding should help reduce the cost of the repair. Performance is dependent on properly executed complete joint penetration welds at the column face and at the attachment of the haunch to the girder bottom flange. The joint can be subject to through-thickness flaws in the column flange; however, this connection may not be as sensitive to this potential problem because of the significant increase in the effective depth of the beam section which can be achieved. Welding of the bottom haunch requires overhead welding. The skewed groove welds of the haunch flanges to the girder and column flanges may be difficult to execute. Experimental Results: This approach developed excellent levels of plastic rotation. In Specimen 1, the bottom flange CJP weld was damaged in a prior test but was not repaired: only the bottom haunch was added. During the test of specimen 1, a slowly growing crack developed at the underside of the top flange-web intersection, perhaps exacerbated by significant local buckling of the top flange. Some of the buckling may be attributed to lateral torsional buckling that occurred because the bottom flange was not restrained by a CJP weld. A significant portion of the flexural strength was lost during the cycles of large plastic rotation. In the second specimen, the bottom girder flange weld was intact during the haunch testing, and its performance was significantly improved compared with the first specimen. The test was stopped when significant local buckling led to a slowly growing crack at the beam flange and web intersection. At this time, it appears that repairing damaged bottom flange welds in this configuration can produce

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better performance. Acceptable levels of flexural strength were maintained during large inelastic deformations of the plastic hinge for both specimens. Quantitative Results: No. of specimens tested: 2 Girder Size: W30 x 99 Column Size: W14 x 176 Plastic Rotation achievedSpecimen 1:0.04 radian (w/o bottom flange weld) Specimen 2:0.05 radian (with bottom flange weld) 6.6.7.2 Top and Bottom Haunch

Figure 6-17 illustrates the basic geometry of the top and bottom haunch detail. The intended behavior of this modification is to shift the plastic hinge from the face of the column and to reduce the demand on the CJP weld to the column flange by increasing the effective depth of the section. As opposed to the bottom-only haunch, of Section 6.6.7.1, this detail further reduces the demand on all CJP welds and allows for the structural engineer to introduce filler metal with better toughness properties into all critical joints, without necessarily having to remove the top flange CJP weld.

d/3

d

WT

1 2

or

Figure 6-17 - Top and Bottom Haunch Modification Detail Two specimens for this detail have been tested to date, with excellent results. Possible variations that have not yet been tested include using a shallower haunch at the top flange, substitution of a flat cover plate for the top haunch, and not rewelding either of the original girder flanges to the column, if these have been damaged. Design Issues: The haunches can be installed at perimeter frames without removal of the exterior building cladding. Performance is dependent on the proper execution of the CJP welds from the haunch to the girder and column flanges, which can be difficult. The joint at the column flange is subject to through-thickness flaws in the column flange, however, due to the additional depth of the section at this joint, and the resulting reduced stresses, this design may not be particularly sensitive to this. 6-41

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Experimental Results: This approach developed excellent levels of plastic rotation in two specimens. The tests were terminated when fractures across the width of the column flanges developed at the locations of severe buckling in these flanges. Acceptable levels of connection strength were maintained throughout the test. Quantitative Results: No. of specimens tested: 2 Girder Size: W30 x 99 Column Size: W14 x 176 Plastic Rotation achievedSpecimen 1:0.07 radian Specimen 2:0.07 radian 6.6.7.3 Cover Plate Sections

Figure 6-18 illustrates the basic configurations of cover plate connections. The assumption behind the cover plate is that it reduces the demand on the weld at the column flange and shifts the plastic hinge away from the column face. Only the connection with cover plates on the top of the top flange has been tested. There are no quantitative results for cover plates on the bottom side of the top flange, such as might be used in repair. It is likely that thicker plates would be required where the plates are installed on the underside of the top flange. The implications of this deviation from the tested configuration should be considered.

d

Near and Far Sides

d/2, typical Top &Bottom

Top &Bottom

Figure 6-18 - Cover Plate Connection Modification Design Issues: Approximately eight connections similar that shown in Figure 6-18 have been tested (Engelhardt & Sabol - 1994), and have demonstrated the ability to achieve acceptable levels of plastic rotation provided that the beam flange to column flange welding is correctly executed and through-thickness problems in the column flange are avoided. The option with the top flange cover plate located on top of the flange can be used on perimeter frames where access to the outer side of the beam is restricted by existing building cladding. The option with the cover plate for the top flange located beneath the flange can be installed without requiring modification of the slab. In the figures shown, the bottom cover plate is rectangular, and sized 6-42

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slightly wider than the beam flange to allow downhand fillet welding of the joint between the two plates. Some configurations using triangular plates at the bottom flange, similar to the top flange have also been tested. Designers using this detail are cautioned to be mindful of not making cover plates so thick that excessively large welds of the beam flange combination to column flange result. As the cover plates increase in size, the weld size must also increase. Larger welds invariably result in greater shrinkage stresses and increased potential for cracking prior to actual loading. In addition, larger welds will lead to larger heat affected zones in the column flange, a potentially brittle area. Performance is dependent on properly executed girder flange welds. The joint can be subject to through-thickness failures in the column flange. Access to the top of the top flange requires demolition of the existing slab. Access to the bottom of the top flange requires overhead welding and may be problematic for perimeter frames. Costs are greater than those associated with approaches that concentrate modifications on the bottom flange Experimental Results: Six of eight connections tested by the University of Texas at Austin were able to achieve plastic rotations of at least 0.025 radians, or better. These tests were performed using heavy column sections which forced nearly all of the plastic deformation into the beam plastic hinge; very little column panel zone deformation occurred. Strength loss at the extreme levels of plastic rotation did not reduce the flexural capacity to less than the plastic moment capacity of the section based on minimum specified yield strength. One specimen achieved plastic rotations of 0.015 radians when a brittle fracture of the CJP weld (type W2 failure) occurred. This may partially be the result of a weld that was not executed in conformance with the specified welding procedure specification. The second unsuccessful test specimen achieved plastic rotations of 0.005 radian when a section of the column flange pulled out (type C2 failure). The successful tests were terminated either when twisting of the specimen threatened to damage the test setup or the maximum stroke of the loading ram was achieved. Quantitative Results: No. of specimens tested: 8 Girder Size: W36 x 150 Column Size: W14 x 455, and 426 Plastic Rotation achieved6 Specimens : >.025 radian to 0.05 radian 1 Specimen: 0.015 radian (W2 failure) 1 Specimen: 0.005 radian (C2 failure) 6.6.7.4 Upstanding Ribs

Figures 6-19 illustrates the basic configuration of connections with upstanding ribs. The assumption behind the rib plate is that it reduces the demand on the weld at the column flange and shifts the plastic hinge from the column face. The figure indicates alternative configurations using either one centered rib, or two spaced ribs on each flange. Test data is available only for the case with two ribs.

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2 1

tested configuration

d

d/2

alternate configuration

Typical

Typical

Figure 6-19 - Upstanding Rib Connection Modification Design Issues: Two connections similar to Figure 6-19, with two spaced ribs at each flange have been tested (Engelhardt & Sabol - 1994), and demonstrated the ability to achieve acceptable levels of plastic rotation provided that the girder flange welding is correctly executed. This modification can be used on perimeter frames where access to the outer side of the girder is restricted by existing building cladding. Performance is dependent on properly executed girder flange welds. The joint can be subject to through-thickness failures in the column flange. Access to the top of the top flange requires demolition of the existing slab. Access to the bottom of the top flange requires overhead welding and may be problematic for perimeter frames. The size of the specimens tested required the use of two upstanding ribs per flange. This increased the costs significantly above those designs that use only one rib per flange, located above the girder center line. However, limited testing of the design with one rib at the girder centerline, performed as part of a program related to eccentric braced frames, indicated the potential for premature failure of the weld of the rib to the girder at the outstanding edge. Experimental Results: Two connections have been tested (Engelhardt & Sabol - 1994) using two plates on the top and bottom flanges. The columns used in the test were very heavy and the flanges were able to resist the applied loads from the ribs without distorting. Similar performance might not occur with lighter column sections. In addition, the size of the columns forced all of the plastic deformation into the beam plastic hinge; very little column panel zone deformation occurred. Strength loss at the extreme levels of plastic rotation did not reduce the flexural capacity to less than the plastic moment capacity of the section based on minimum specified yield strength, however, strength loss occurred more quickly than with the cover plated specimens. The tests were terminated when a slow tear of the beam bottom flange occurred at the tips of the ribs. Quantitative Results: No. of specimens tested: 2 Girder Size: W36 x 150 Column Size: W14 x 426 6-44

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Plastic Rotation achieved2 Specimens : >.025 radian 6.6.7.5 Side-Plate Connections

This approach eliminates loading the column in the through-thickness direction by removing the CJP welds at the girder flange and by shifting the plastic hinge from the column face. The tension and compression forces are transferred from the girder flanges into the column through fillet welds. A mechanism to provide a direct connection between the column panel zone and the beam flanges is required; the difficulty appears to be equalizing the width of the beam and column flanges. Experimental Results: At least two configurations of side-plated connections have been tested. One set, shown in Figure 6-20, utilized flat bars at the top and bottom girder flanges, to transfer flange forces to the column (Engelhardt & Sabol - 1994). The girder was widened to the width of the column with the use of filler plates. The specimens achieved plastic rotations of 0.015 radians, however, fractures developed within the welds connecting the beam flange to the transfer plates. Failure of the shear tab, and finally the side plates themselves followed the initiation of these fractures. It is believed that the unsuccessful behavior of this particular specimen was related to the method used to increase the width of the beam flange to equal that of the column flange, using a combination of a filler bar and welding. Other approaches that rely on a flat filler plate to transfer the forces may perform better.

Possible Alternative

Tested Configuration

Figure 6-20 - Side Plate Connection Modification Quantitative Results: Separate Top & Bottom Side Plates No. of specimens tested: 2 Girder Size: W36 x 150 Column Size: W14 x 426 Plastic Rotation achieved2 Specimens : >.015 radian

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A second, proprietary configuration, is shown in Figure 6-21. Three specimens representative of the “new structure” configuration have undergone full-scale testing to date and achieved large plastic rotations. Loss of strength at large plastic rotation demands was comparable to that of other successful connections. No tests have yet been conducted of the repair configuration. The developer of this connection has applied for US and foreign patents. Further information on technical data for this configuration, and license fees, may be obtained from the developer.

New Building Configuration

Repair Configuration

WARNING: The information presented in this figure is PROPRIETARY. US and Foreign Patents have been applied for. Use of this information is strictly prohibited except as authorized in writing by the developer. Violators shall be prosecuted in accordance with US and Foreign Patent Intellectual Property Laws.

Figure 6-21 - Proprietary Side Plate Connection Modification Design Issues: Testing of three prototype specimens (Uang & Latham - 1995) indicates that this connection has the ability to achieve very satisfactory levels of plastic rotation without relying on sensitive CJP welds between the column and girder flanges, or requiring specification of notch-tough weld material. The elimination of the through-thickness loading of the flange may result in higher levels of connection reliability. Due to the exclusive use of fillet welds, special inspection requirements for welding and bolting can be reduced significantly with this connection. This connection is proprietary (patent pending) and not in the public domain. It has not been tested in a repair condition. Access to the top of the top flange of the girder might require demolition of the existing slab. The cost of the connection may be greater than some of the other modification methods discussed above; however, this cost differential may not be as great on double-sided connections because much of the cost is associated with the side plates which are similar for both single-sided and double-sided connections. Publicly bid projects may have to develop performance specifications to permit other connections to be considered for use unless a strong case for sole-sourcing the connection can be made. Quantitative Results: No. of specimens tested: 3 Girder Size: W36 x 150 Column Size: W14 x 426

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Plastic Rotation achieved3 Specimens : >.042 to 0.06 radian

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7. NEW CONSTRUCTION The building code provisions for earthquake resistive design of Special Moment-Resisting Frames (SMRFs) assume that these structures are extremely ductile and therefore are capable of large plastic rotations at, or near to, their beam-column connections. Based on limited research, and observations of damage experienced in the Northridge Earthquake, it appears that conventionally designed connection assemblies configured such that plastic deformation concentrates at the beam-column connection are not capable of reliably withstanding large plastic rotation demands. The reliability appears to decrease as the size of the connected members increases. Other factors affecting this reliability appear to include the quality of workmanship, joint detailing, toughness of the base and weld metals, relative strengths of the connection elements, and the combined stresses present on these elements. Unfortunately, the quantitative relationship between these factors and connection reliability is not well defined at this time. In order to attain frames that can reliably perform in a ductile manner, these Interim Guidelines recommend that SMRF connections be configured with sufficient strength so that plastic hinges occur within the beam span and away from the face of the column. All elements of the frame, and the connection itself, should be designed with adequate strength to develop these plastic hinges. The resulting connection assemblies are somewhat complex and the factors limiting their behavior not always evident. Therefore, qualification of connection designs through prototype testing, or by reference to tests of similar connection configurations is recommended. These procedures should also be applied to the design of Ordinary Moment-Resisting Frames (OMRFs) located in zones of higher seismicity, or for which highly reliable earthquake performance is desired, unless it can be demonstrated that the connections can resist the actual demands from a design earthquake and remain elastic. Interim Guidelines for determining if a design meets this condition are provided. Light, single-story, frame structures, the design of which is predominated by wind loads, have performed well in past earthquakes and may continue to be designed using conventional approaches, regardless of the seismic zone they are located in. Materials and workmanship are critical to frame behavior and careful specification and control of these factors is essential. Interim Guidelines for the specification of materials and control of workmanship are provided in this Chapter, as well as in Chapters 8, 9, 10 and 11. 7.1 Scope This Chapter presents interim design guidelines for new welded steel moment frames (WSMFs) intended to resist seismic demands through inelastic behavior. The criteria apply to all SMRF structures designed for earthquake resistance and those OMRF structures located in Uniform Building Code (UBC) Seismic Zones 3 and 4 {National Earthquake Hazards Reduction Program (NEHRP) Map Areas 6 and 7}. Light, single-story buildings, the design of which is governed by wind, need not consider these Interim Guidelines. Frames with bolted connections, either fully restrained (type FR) or partially restrained (type PR), are beyond the scope of this 7-1

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document. However, the acceptance criteria for connections may be applied to type FR bolted connections as well. Commentary: Observation of damage experienced by WSMF buildings in the Northridge Earthquake and subsequent laboratory testing of large scale beamcolumn assemblies has demonstrated that the standard details for WSMF connections commonly used in the past are not capable of providing reliable service in the post-elastic range. Therefore, structures which are expected to experience significant post-elastic demands from design earthquakes, or for which highly reliable seismic performance is desired, should be designed using the Interim Guidelines presented herein. In order to determine if a structure will experience significant inelastic behavior in a design earthquake, it is necessary to perform strength checks of the frame components for the combination of dead and live loads expected to be present, together with the full earthquake load. Except for structures with special performance goals, or structures located within the near field (within 10 kilometers) of known active earthquake faults, the full earthquake load may be taken as the minimum design earthquake load specified in the building code, but calculated using a lateral force reduction coefficient (Rw or R) of unity. If all components of the structure and its connections have adequate strength to resist these loads, or nearly so, then the structure may be considered to be able to resist the design earthquake, elastically. Design of frames to remain elastic under unreduced (Rw {R} taken as unity) earthquake forces may not be an overly oppressive requirement, particularly in more moderate seismic zones. Most frame designs are currently controlled by drift considerations and have substantially more strength than the minimum specified for design by the building code. As part of the SAC Phase 1 research, a number of modern frame buildings designed with large lateral force reduction coefficients (Rw = 12, {R = 8}) were evaluated for unreduced forces calculated using the standard building code spectra. It was determined that despite the nominally large lateral force reduction coefficients used in the original design, the maximum computed demands from the dynamic analyses were only on the order of 2 to 3 times those which would cause yielding of the real structures (Krawinkler, et. al. - 1995; Uang, et. al. - 1995; Engelhardt, et. al. - 1995, Hart, et. al. - 1995; Kariotis and Eimani - 1995). Therefore, it is not unreasonable to expect that OMRF structures (nominally designed with a lateral force reduction coefficient Rw = 6 {R = 4.5}) could resist the design earthquakes with near elastic behavior. Regardless of these considerations, better seismic performance can be expected by designing structures with greater ductility rather than less and engineers are not encouraged to design structures for elastic behavior using brittle or unreliable details..

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For structures designed to meet special performance goals, and buildings located within the near field of major active faults, full earthquake loads calculated in accordance with the above procedure may not be adequate. For such structures, the full earthquake load should be determined using a site specific ground motion characterization and a suitable analysis procedure. Recent research (Heaton, et. al. - 1995) suggests that the elastic response spectrum technique, typically used for determining seismic forces for structural design, may not provide an adequate indication of the true earthquake demands produced by the large impulsive ground motions common in the near field of large earthquake events. Further, this research indicates that frame structures, subjected to such impulsive ground motions can experience very large drifts, and potential collapse. Direct nonlinear time history analysis, using an appropriate ground motion representation would be one method of more accurately determining the demands on structures located in the near field. Additional research on these effects is required. As an alternative to use of the criteria contained in these Interim Guidelines, OMRF structures in zones of high seismicity (UBC seismic zones 3 and 4 and NEHRP map areas 6 and 7) may be designed for the connections to remain elastic (Rw or R taken as 1.0) while the beams and columns are designed using the standard lateral force reduction coefficients specified by the building code. Although this is an acceptable approach, it may result in much larger connections than would be obtained by following these Interim Guidelines. The use of partially restrained connections may be an attractive and economical alternative to the design of frames with fully restrained connections. However, the design of frames with partially restrained connections is beyond the scope of this document. The AISC is currently working on development of practical design guidelines for frames with partially restrained connections. 7.2 General - Welded Steel Frame Design Criteria 7.2.1 Criteria

Welded Steel Moment Frame (WSMF) systems should, as a minimum, be designed for the provisions of the prevailing building code and these Interim Guidelines. Special MomentResisting Frames (SMRF)s and Ordinary Moment-Resisting Frames (OMRF)s with FR connections, should additionally be designed in accordance with the emergency code change to the 1994 UBC {NEHRP-1994}, restated as follows: 2211.7.1.1. Required Strength {NEHRP-1994 Section 5.2, revision to Ref. 8.2c of Ref. 5.3} The girder-to-column connections shall be adequate to develop the lesser of the following: 1.

The strength of the girder in flexure.

2.

The moment corresponding to development of the panel zone shear strength as determined by Formula (11-1).

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Connection configurations utilizing welds and high strength bolts shall demonstrate, by approved cyclic test results or calculation, the ability to sustain inelastic rotations and to develop the strength criteria in Section 2211.7.1.1 considering the effects of steel overstrength and strain hardening.

Commentary: At this time, no recommendations are made to change the minimum lateral forces, drift limitations or strength calculations which determine member sizing and overall performance of moment frame systems, except as recommended in Sections 7.2.2, 7.2.3 and 7.2.4. The design of joints and connections is discussed in Section 7.3. The UBC permits OMRF structures with FR connections, designed for 3/8Rw times the earthquake forces otherwise required, to be designed without conforming to Section 2211.7.1. However, this is not recommended. 7.2.2 Strength

When these Interim Guidelines require determination of the strength of a framing element or component, this shall be calculated in accordance with the criteria contained in UBC-94, Section 2211.4.2 {NEHRP-91 Section 10.2, except that the factor φshould be taken as 1.0}, restated as follows: 2211.4.1 Member strength. Where this section requires that the strength of the member be developed, the following shall be used: Flexure Shear Axial compression Axial tension Connectors Full Penetration welds Partial Penetration welds Bolts and fillet welds

M s = Z Fy Vs = 0.55 Fy d t Psc = 1.7 Fa A Pst = Fy A Fy A 1.7 allowable (see commentary) 1.7 allowable

Commentary: Partial penetration welds are not recommended for tension applications in critical connections resisting seismic induced stresses. The geometry of partial penetration welds creates a notch-like condition that can initiate brittle fracture under conditions of high tensile strain. 7.2.3 Configuration

Frames should be proportioned so that the required plastic deformation of the frame may be accommodated through the development of plastic hinges at pre-determined locations within the girder spans, as indicated in Figure 7-1. Beam-column connections should be designed with sufficient strength (through the use of cover plates, haunches, side plates, etc.) to force development of the plastic hinge away from the column face. This condition may also be attained through local weakening of the beam section at the desired location for plastic hinge formation.

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h

Undeformed frame

Chapter 7 - New Construction

Deformed frame shape

Plastic Hinges

drift angle - θ

L’ L

Figure 7-1 - Desired Plastic Frame Behavior Commentary: Nonlinear deformation of frame structures is accommodated through the development of inelastic flexural or shear strains within discrete regions of the structure. At large inelastic strains these regions can develop into plastic hinges, which can accommodate significant concentrated rotations at constant (or nearly constant) load through yielding at tensile fibers and buckling at compressive fibers. If a sufficient number of plastic hinges develop in a frame, a mechanism is formed and the frame can deform laterally in a plastic manner. This behavior is accompanied by significant energy dissipation, particularly if a number of members are involved in the plastic behavior, as well as substantial local damage to the highly strained elements. The formation of hinges in columns, as opposed to beams, is undesirable, as this results in the formation of mechanisms with relatively few elements participating, so called “story mechanisms” and consequently little energy dissipation occurring. In addition, such mechanisms also result in local damage to critical gravity load bearing elements. The prescriptive connection contained in the UBC and NEHRP Recommended Provisions prior to the Northridge Earthquake was based on the development of plastic hinges within the beams at the face of the column, or within the column panel zone itself. If the plastic hinge develops in the column panel zone, the resulting column deformation results in very large secondary stresses on the beam flange to column flange joint, a condition which can contribute to brittle failure. If the plastic hinge forms in the beam, at the face of the column, this can result in very large through-thickness strain demands on the column flange material and large inelastic strain demands on the weld metal and surrounding heat affected zones. These conditions can also lead to brittle joint failure. In order to achieve more reliable performance, it is recommended that the connection of the beam to the column be configured to force the inelastic action (plastic hinge) away from the column face. This can be done either by local reinforcement of the 7-5

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connection, or locally reducing the cross section of the beam, at a distance away from the connection. Plastic hinges in steel beams have finite length, typically on the order of half the beam depth. Therefore, the location for the plastic hinge should be shifted at least that distance away from the face of the column. When this is done through reinforcement of the connection, the flexural demands on the columns, for a given beam size, are increased. Care must be taken to assure that weak column conditions are not inadvertently created by local strengthening of the connections. It should be noted that some professionals and researchers believe that configurations which permit plastic hinging to occur adjacent to the column face may still provide reliable service under some conditions. These conditions may include limitations on the size of the connected sections, the use of base and weld metals with adequate notch toughness, joint detailing that minimizes notch effects, and appropriate control of the relative strength of the beam and column materials. Sufficient research has not been performed to date either to confirm these suggestions or define the conditions in which they are valid. Research however does indicate that reliable performance can be attained if the plastic hinge is shifted away from the column face, as suggested above. Consequently, these Interim Guidelines make a general recommendation that this approach be taken. Additional research should be performed to determine the acceptability of other approaches. It should also be noted that reinforced connection (or reduced beam section) configurations of the type described above, while believed to be effective in preventing brittle connection fractures, will not prevent structural damage from occurring. Brittle connection fractures are undesirable because they result in a substantial reduction in the lateral-force-resisting strength of the structure which, in extreme cases, can result in instability and collapse. Connections configured as described in these Interim Guidelines should experience many fewer such brittle fractures than unmodified connections. However, the formation of a plastic hinge within the span of a beam is not a completely benign event. Beams which have formed such hinges may exhibit large buckling and yielding deformation, damage which typically must be repaired. The cost of such repairs could be comparable to the costs incurred in repairing fracture damage experienced in the Northridge Earthquake. The primary difference is that life safety protection will be significantly enhanced and most structures that have experienced such plastic deformation damage should continue to be safe for occupancy, while repairs are made. If the types of damage described above are unacceptable for a given building, then alternative structural systems should be considered, that will reduce the plastic deformation demands on the structure during a strong earthquake. Appropriate methods of achieving such goals include the installation of supplemental braced frames, energy dissipation systems, base isolation systems 7-6

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and similar structural systems. Framing systems incorporating partially restrained connections may also be quite effective in resisting large earthquake induced deformation with limited damage. 7.2.4 Plastic Rotation Capacity

The plastic rotation capacity of connection assemblies should reflect realistic estimates of the total (elastic and plastic) drift likely to be induced in the frame by earthquake ground shaking, and the geometric configuration of the frame. For frames of typical configuration, and for ground shaking of the levels anticipated by the building code, a minimum plastic rotation capacity of 0.03 radian is recommended. When the configuration of a frame is such that the ratio L/L’is greater than 1.25, the plastic rotation demand should be taken as follows: θ = 0.025(1 + ( L − L' ) L' )

(7-1)

where: L is the center to center spacing of columns, and L’is the center to center spacing of plastic hinges in the bay under consideration The indicated rotation demands may be reduced when positive means, such as the use of base isolation or energy dissipation devices, are introduced into the design, to control the building’s response. When such measurers are taken, nonlinear dynamic analyses should be performed and the connection demands taken as 0.005 radians greater than the rotations calculated in the analyses. The nonlinear analyses should conform to the criteria specified in UBC-94 Section 1655 {NEHRP-94 Section 2.6.4.2} for nonlinear dynamic analysis of base isolated structures. Ground motion time histories utilized for these nonlinear analyses should satisfy the scaling requirements of UBC-94 Section 1655.4.2 {NEHRP-94 Section 2.6.4.4}, except that if the building is not base isolated, the structure period T, calculated in accordance with UBC-94 Section 1628 {NEHRP-94 Section 2.3.3.1} should be substituted for TI. Commentary: Traditionally, engineers have calculated demand in moment frames by sizing the members for strength and drift using code forces (either equivalent static or reduced dynamic forces) and then "developing the strength of the members." Since 1988, "developing the strength" has been accomplished by prescriptive means based on a review of testing of moment frame connections to that date. It was assumed that the prescribed connections would be strong enough that the beam or girder would yield (in bending), or the panel zone would yield (in shear) in a nearly perfectly plastic manner producing the plastic rotations necessary to dissipate the energy of the earthquake. A realistic estimate of the interstory drift demand for most structures and most earthquakes is on the order of 0.015 to 0.025 times the story height for WSMF structures designed to code allowable drift limits. In such frames, a portion of the drift will be due to elastic deformations of the frame, while the balance must 7-7

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be provided by inelastic rotations of the beam plastic hinges, by yielding of the column panel zone, or by a combination of the two. In the 1994 Northridge Earthquake, many moment-frame connections fractured with little evidence of plastic hinging of the beams or yielding of the column panel zones. Testing of moment frame connections both prior to and subsequent to the earthquake suggests that the standard, pre-Northridge, welded flange-bolted web connection is unable to reliably provide plastic rotations beyond about 0.005 radian for all ranges of beam depths and often fails below that level. Since the elastic contribution to drift may approach 0.01 radian, the necessary inelastic contributions will exceed the capability of the standard connection in many cases. For frames designed for code forces and for the code drift, the necessary plastic rotational demand may be expected to be on the order of 0.02 radian or more and new connection configurations should be developed to accommodate such rotation without brittle fracture. The recommended connection demand of 0.03 radians was selected both to provide a comfortable margin against the demands actually expected in most cases and because in recent testing of connection assemblies, specimens capable of achieving this demand behaved in a ductile manner through the formation of plastic hinges. For a given building design, and known earthquake hazard, it is possible to more accurately estimate plastic rotation demands on frame connections. This requires the use of nonlinear analysis techniques. Analysis software, capable of performing such analyses is becoming more available and many design offices will have the ability to perform such analyses and develop more accurate estimates of inelastic demands for specific building designs. However, when performing such analyses, care should be taken to evaluate building response for multiple earthquake time histories, representative of realistic ground motions for sites having similar geologic characteristics and proximity to faults, as the actual building site. Relatively minor differences in the ground motion time history used as input in such an analysis can significantly alter the results. Since there is significant uncertainty involved in any ground motion estimate, it is recommended that analysis not be used to justify the design of structures with non-ductile connections, unless positive measures such as the use of base isolation or energy dissipation devices are taken, to provide reliable behavior of the structure. It has been pointed out that it is not only the total plastic rotation demand that is important to connection and frame performance, but also the connection mechanism (for example - panel zone yielding, girder flange yielding/buckling, etc.) and hysteretic loading history. These are matters for further study in the continuing research on connection and joint performance.

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7.2.5 Redundancy

The frame system should be designed and arranged to incorporate as many moment-resisting connections as is reasonable into the moment frame. Commentary: Early moment frame designs were highly redundant and nearly every column was designed to participate in the lateral-force-resisting system. In an attempt to produce economical designs, recent practice often produced designs which utilized only a few large columns and beams in a small proportion of the building’s frames for lateral resistance, with the balance of the building columns designed not to participate in lateral resistance. This practice led to the need for large welds at the connections and to reliance on only a few connections for the lateral stability of the building. The resulting large framing elements and connections are believed to have exacerbated the poor performance of the preNorthridge connection. Further, if only a few framing elements are available to resist lateral demands, then failure of only a few connections has the potential to result in a significant loss of earthquake resisting strength. Together, these effects are not beneficial to building performance. The importance of redundancy to building performance can not be overemphasized. Even connections designed and constructed according to the improved procedures recommended by these Interim Guidelines will have some potential, albeit greatly reduced, for brittle failures. As the number of individual beams and columns incorporated into the lateral-force-resisting system is increased, the consequences of isolated connection failures significantly reduces. Further, as more framing elements are activated in the building’s response to earthquake ground motion, the building develops greater potential for energy absorption and dissipation, and ability to control earthquake induced deformations to acceptable levels. Incorporation of more of the building framing into the lateral-force-resisting system will lead to smaller members and therefore an anticipated increase in the reliability of individual connections. It will almost certainly lead to improved overall system reliability. Further, recent studies conducted by designers indicate that under some conditions, redundant framing systems can be constructed as economically as non-redundant systems. In these studies, the additional costs incurred in making a greater number of field-welded moment-resisting connections in the more redundant frame were balanced by a reduced total tonnage of steel in the lateral-force-resisting systems and sometimes, reduced foundation costs as well. The 1994 UBC requirements limit the relative number of weak column/strong beam connections in the moment frame system. There is a divergence of opinion among structural engineers on the desirability of frames in which all beamcolumn connections are made moment-resisting, including those of beams 7-9

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framing to the minor axis of columns. Use of such systems as a means of satisfying these Interim Guidelines requires careful consideration by the structural engineer. Limited testing in the past has indicated that moment connections made to the minor axis of wide flange columns are subject to the same types of fracture damage experienced by major axis connections. As of this time, there has not been sufficient research to suggest methods of making reliable connections to the column minor axis. 7.2.6 System Performance

WSMF design should consider all effects of connection modifications on the response and performance of the frame. Commentary: Methods developed thus far for improving performance of beam/column connections involve shifting of the hinge point away from the column face either by reinforcement of the connection (e.g. haunches, cover plates, etc.), or reducing the relative strength of the beam locally. These modifications affect the overall stiffness of the frame and, therefore, its seismic response. In fact, it can be shown that the use of smaller beam sizes and haunched connections will result in the same overall frame stiffness as the use of larger beams and unstiffened connections. Additionally, haunching or reinforcement results in magnified moments and shears at the column face which should be included in the strong column/weak beam calculations, panel zone and web connection calculations, and column axial demand calculations. Unsymmetrical haunches, placed on only the bottom (or top) of the beam can also change the relative stiffness of columns above and below the beam resulting in unexpected formation of plastic hinges in one of the columns. In addition, if plastic hinges are forced out into the beam span, away from the column face, the local lateral stability of beams at plastic hinges away from the column should be considered. 7.2.7 Special Systems

When WSMFs are used as components of "Tube" type buildings with beams yielding in shear rather than bending, or in Dual System structures, appropriate consideration should be given to the differences in plastic rotation demands expected (as compared to pure moment frame designs) when applying these provisions. (See discussion in Section 7.10.) Commentary: Moment frames which are employed in dual systems in low-rise buildings or in the lower levels of taller buildings may have significantly lower rotation demands than those in pure frame buildings. Engineers may consider it appropriate to use less conservative connection designs or qualification requirements for such frames, or for portions of such frames. Appropriate analytical substantiation should be provided for any alternative criteria utilized.

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For tube frames with shear-yielding beams, qualification by testing is recommended, but designs and requirements may differ from those presented in these Interim Guidelines. Again appropriate analytical substantiation should be provided for the selected criteria. 7.3 Connection Design & Qualification Procedures - General 7.3.1 Connection Performance Intent

The intent of connection design should be to force the plastic hinge away from the face of the column to a pre-determined location within the beam span. This may be accomplished by local reinforcement of the connection itself (cover plates, haunches, side plates, etc.) or by local reductions of the beam section (drilled holes, trimmed flanges, etc.). All elements of the connection should have adequate strength to develop the forces resulting from the formation of the plastic hinge at the predetermined location, together with forces resulting from gravity loads. 7.3.2 Qualification by Testing

Connection strength and plastic rotation capacity should be demonstrated by approved cyclic testing as described in Section 7.4, except as indicated in paragraph 7.3.3. It is recommended that preliminary design of specimens to be tested be developed using the Interim Guidelines of Section 7.5. Extrapolation and interpolation of test results using the calculation procedures of Section 7.5 is acceptable for connections of elements having similar geometries and material specifications as tested connections. Commentary: Cyclic testing of connections matching the essential features of those to be used in the actual design is the most reliable method of assuring that the expected connection performance can be attained. Section 7.4 describes testing guidelines in detail. Guidelines for extrapolation by calculation are given in Section 7.5. 7.3.3 Design by Calculation

Connection design by calculations alone may be acceptable under the following conditions: a)

Calculations are based on comparison with previously tested assemblies, or with prototype connections tested for the project;

b)

Conditions of the calculated detail, including member property relationships, material properties, welding materials, processes and procedures, and construction sequence, mirror those of the tested detail as closely as possible; or

c)

Qualified third party review, in accordance with Section 4.5 is performed.

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Commentary: Use of calculations based on engineering principles alone, or to extrapolate data from tests performed on assemblies which do not precisely mirror the conditions of the calculated assembly, requires caution and judgment. Subjective factors affecting the acceptability of such an approach should include: a)

The importance of the structure: Greater caution in applying a calculation-only approach should be exercised for more important facilities, particularly when the facility is expected to remain functional after a major earthquake.

b)

Confidence in the lateral forces used in design: Projects which carefully apply seismic forces based on extensively researched, site-specific seismic hazard studies, and having resulting designs with low calculated rotation demands, may warrant more confidence in the application of connection designs using calculations only, as opposed to those that do not use this type of information. Most structures are designed to satisfy the minimum code seismic forces. Structures that are designed assuming higher levels of seismic demand (both strength and stiffness) than found in typical projects, could also possibly be demonstrated to warrant greater latitude in applying a calculation-only approach.

c)

The degree of redundancy, regularity and potential over-strength in the structure: Greater care in applying a calculation-only approach should be considered in structures with a limited number of lateral-force-resisting elements in each direction or those with unusual building geometries. Structures with a high degree of redundancy may be demonstrated to be better able to tolerate limited instances of marginal connection performance. Frames designed to limit the rotational demand by relying on elastic or near-elastic behavior may also be more amenable to a calculation-only approach than those that depend on high levels of plastic rotation to dissipate anticipated seismic demands. However, it has not been shown that superior seismic performance results when strength is substituted for ductility, and overly strong, frames with non ductile connections are not the intent of these guidelines.

d)

Proximity to active faults: Ground motion records from recent earthquakes clearly demonstrate that sites located close to a fault rupture experience substantially more severe ground motion than is explicitly provided for in current code design provisions. When a building is located within 5 km of an active fault, the plastic rotation demands on connections may exceed those provided for in these Guidelines, and additional caution in design procedures is warranted.

For structures that are essential, contain hazardous materials, are designed with a low degree of conservatism or redundancy, connections qualification by test (either through reference to tests from other projects or project-specific 7-12

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testing of connections) is strongly recommended. This recommendation should be considered until such time as SAC or other research develops sufficient data to allow formulation of analytical design guidelines for general application. For non-essential structures designed with a reasonable degree of redundancy or overstrength and incorporating enhanced welding requirements and quality control, calculations as described above, using proportioning and stress levels compatible with previously completed test programs, may provide sufficient assurance of reliability. 7.4 Guidelines for Connection Qualification by Testing 7.4.1 Testing Protocol

Unless future testing programs reveal significant effects of dynamic loading rate or time history loading, and unless the effects of other factors (e.g., restraint conditions and composite slab effects) are found to be compelling, a testing protocol similar to ATC-24, Guidelines for Cyclic Seismic Testing of Components of Steel Structures (Applied Technology Council - 1992), is recommended as the basis for qualification tests. The testing program should replicate as closely as practical the anticipated conditions in the field, including such factors as: a) Member sizes. b) Material specifications. c) Welding process, details and construction conditions. d) Cover plates, continuity plates, web tabs, bolts, and doubler plates. e) Connection configuration (e.g., beams on both sides). f) Induced stresses because of restraint conditions on the welds and connection members. g) Axial load, where pertinent. h) Gravity load, where significant. The testing program should be organized to provide as much information as possible about the capability of the connections selected. The following program is recommended: a) Test at least two full size specimens representative of the larger beam/column assemblies in the project.

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b) Test one additional full size specimen representative of other beam/column assemblies with significantly different interaction properties, such as beam b/t, panel zone stress/distortion, etc. If any of the specimens fails to meet the qualification criteria, the connection should be redesigned and retested. Where two-sided connections are used in the structure, and the type of connection being used can be expected to perform differently in a two-sided use than in one-sided use, it should be tested in the two-sided configuration as well as the one-sided. Two-sided connection assemblies can be expected to behave differently than one-sided assemblies, for example, when panel zone distortions will be significantly different, or when systems involve transfer of stress to the column by plates, welds, or other elements which are connected to the beams on both sides of the column. The inclusion of axial load should be considered when analysis indicates that significant tension can be expected to occur in a significant number of the columns represented by the specimen and where the connection type relies on the through-thickness strength of the column flanges. If the presence of a floor slab is anticipated to have significant influence on either the location or mechanism of the plastic hinge formed, than this should also be included in the test specimen. 7.4.2 Acceptance Criteria

The minimum acceptance criteria for connection qualification for specimens tested in accordance with these Interim Guidelines should be as follows: a) The connection should develop beam plastic rotations as indicated in Section 7.2.4, for at least one complete cycle. b) The connection should develop a minimum strength equal to the plastic strength of the girder, calculated using minimum specified yield strength Fy, throughout the loading history required to achieve the required plastic rotation capacity, as indicated in a), above. If the load limiting mechanism in the test is buckling of the girder flanges, the engineer, upon consideration of the effect of strength degradation on the structure, may consider a minimum of 80% of the nominal strength as acceptable. Commentary: While the testing of all connection geometries and member combinations in any given building might be desirable, it would not be very practical nor necessary. Test specimens should replicate, within the limitations associated with test specimen simplification, the fabrication and welding procedures, connection geometry and member size, and potential modes of failure. If the testing is done in a manner consistent with other testing programs, reasonable comparisons can be made. On the other hand, testing is expensive and it is difficult to realistically test the beam-column connection using actual boundary conditions and earthquake loading histories and rates. 7-14

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It was suggested in Interim Recommendation No. 2 by the SEAOC Seismology Committee that three tested specimens be the minimum for qualification of a connection. Further consideration has led to the recognition that while three tests may be desirable, the actual testing program selected should consider the conditions of the project. Since the purpose of the testing program is to "qualify the connection", and since it is not practical for a given project to do enough tests to be statistically meaningful considering random factors such as material, welder skills, and other variables, arguments can be made for fewer tests of identical specimens, and concentration on testing specimens which represent the range of different properties which may occur in the project. Once a connection is qualified, that is, once it has been confirmed that the connection can work, monitoring of actual materials and quality control to assure emulation of the tested design becomes most important. Because of the cost of testing, use of calculations for interpolation or extrapolation of test results is desirable. How much extrapolation should be accepted is a difficult decision. As additional testing is done, more information may be available on what constitutes "conservative" testing conditions, thereby allowing easier decisions relative to extrapolating tests to actual conditions which are likely to be less demanding than the tests. For example, it is hypothesized that connections of shallower, thinner flanged members are likely to be more reliable than similar connections consisting of deeper, thicker flanged members. Thus, it may be possible to test the largest assemblages of similar details and extrapolate to the smaller member sizes - at least within comparable member group families. Extrapolation or interpolation of results with differences in welding procedures, details or material properties is more difficult. 7.5 Guidelines for Connection Design by Calculation In conditions where it has been determined that design of connections by calculation is sufficient, or when calculations are used for interpolation or extrapolation, the following guidelines should be used. 7.5.1 Material Strength Properties

In the absence of project specific material property information, the values listed in Table 7-1 should be used to determine the strength of steel shape and plate for purposes of calculation. The permissible strength for weld metal should be taken in accordance with the building code. Additional information on material properties may be found in the Interim Guidelines of Chapter 8.

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Table 7-1 - Properties for Use in Connection Design Material A36

Fy (ksi) 36

Fy m (ksi) use values for Dual Certified

Dual Certified Beam Axial, Flexural3 50 Shape Group 1 551 Shape Group 2 581 Shape Group 3 571 Shape Group 4 541 Through-Thickness A572 Column/Beam Axial, Flexural3 50 Shape Group 1 581 Shape Group 2 581 Shape Group 3 571 Shape Group 4 571 Shape Group 5 551 Through-Thickness A913-50 Axial, Flexural 50 581 Through-thickness A913--65 Axial, Flexural 65 751 Notes: 1. Based on coupons from web. For thick flanges, the Fy flange is approximately 0.95 Fy web. 2. See Commentary 3. Values based on (SSPC-1994)

Fu (ksi) 58

65 min.

Note 2 65 min.

Note 2, 65 min. Note 2, 80 min.

Commentary: The causes for through-thickness failures of column flanges (types C2, C4, and C5), observed both in buildings damaged by the Northridge Earthquake and in some test specimens, are not well understood. They are thought to be a function of the metallurgy and “purity” of the steel; conditions of loading including the presence of axial load and rate of loading application; conditions of tri-axial restraint; conditions of local hardening and embrittlement within the weld’s heat affected zone; and by the relationship of the connection components as they may affect flange bending stresses and flange curvature induced by panel zone yielding. Given the many complex factors which can affect the through-thickness strength of the column flange, determination of a reliable basis upon which to set permissible design stresses will require significant research. Interim Recommendation No. 2 (SEAOC-1995) included a value of 40 ksi, applied to the projected area of beam flange attachment, for the throughthickness strength to be used in calculations. This value was selected because it was consistent with the successful tests of assemblies with cover plates conducted 7-16

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at the University of Texas at Austin (Engelhardt and Sabol - 1994). However, because of the probable influence of all the factors noted above, this value can only be considered to reflect the specific conditions of those tests and specimens. Although reduced stresses at the column face produced acceptable results in the University of Texas tests, the key to that success was more likely the result of forcing the plastic hinge away from the column than reduction of the throughthickness stress by the cover plates. Reduction of through-thickness column flange stress to ever lower levels by the use of thicker cover plates is not recommended, since such cover plates will result in ever higher forces on the face of the column flange. Notwithstanding all of the above, successful tests using cover plates and other measures of moving hinges (and coincidentally reducing through-thickness stress) continue to be performed. In the interim, engineers choosing to utilize connections relying on through-thickness strength should recognize that despite the successful testing, connections relying on through-thickness strength can not be considered to be fully reliable until the influence of the other parameters discussed above can be fully understood. A high amount of structural redundancy is recommended for frames employing connections which rely on through-thickness strength of the column flange. 7.5.2 Design Procedure

Select a connection configuration, such as one of those indicated in Section 7.9, that will permit the formation of a plastic hinge within the beam span, away from the face of the column, when the frame is subjected to gravity and lateral loads. The following procedure should be followed to size the various elements of the connection assembly: 7.5.2.1 Determine Plastic Hinge Locations

For beams with gravity loads representing a small portion of the total flexural demand, the plastic hinge may be assumed to occur at a distance equal to 1/3 of the beam depth from the edge of the reinforced connection (or start of the reduced beam section), unless specific test data for the connection indicates that a different value is appropriate. Refer to Figure 7-2.

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Interim Guidelines: Evaluation, Repair, Modification and Design of Steel Moment Frames

Edge of reinforced connection

Plastic hinge

d/3

L’

Edge of reinforced connection

Beam depth - d

Chapter 7 - New Construction

L

Figure 7-2 - Location of Plastic Hinge Commentary: The suggested location for the plastic hinge, at a distance d/3 away from the end of the reinforced section (or beginning of reduced section) is based on the observed behavior of test specimens, with no significant gravity load present. If significant gravity load is present, this can shift the locations of the plastic hinges, and in the extreme case, even change the form of the collapse mechanism. If flexural demand on the girder due to gravity load is less than about 30% of the girder plastic capacity, this effect can safely be neglected, and the plastic hinge locations taken as indicated. If gravity demands significantly exceed this level then plastic analysis of the girder should be performed to determine the appropriate hinge locations. In zones of high seismicity (UBC Zones 3 and 4, and NEHRP Map Areas 6 and 7) gravity loading on the girders of earthquake resisting frames typically has a very small effect. 7.5.2.2 Determine Probable Plastic Moment at Hinges

Determine the probable value of the plastic moment, Mpr, at the location of the plastic hinges as: M pr = βM p = βZ b Fy where: ß

(7-2)

is a coefficient that adjusts the nominal plastic moment to the estimated hinge moment based on the mean yield stress of the beam material and the estimated strain hardening. When designs are based upon calculations alone, an additional factor is recommended to account for uncertainty. In the absence of adequate testing of the type described above, ß should be taken as 1.4 for ASTM A572 and for A913, Grades 50 and 65 steels. Where adequate testing has been performed ß should be permitted to be taken as 1.2 for these materials. 7-18

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Zb

Chapter 7 - New Construction

is the plastic modulus of the section

Commentary: In order to compute β, the expected yield strength, strain hardening and an appropriate uncertainty factor need to be determined. The following assumed strengths are recommended: Expected Yield:

The expected yield strength, for purposes of computing (Mpr) may be taken as: Fye = 0.95 Fym

(7-3)

The 0.95 factor is used to adjust the yield stress in the beam web, where coupons for mill certification tests are normally extracted, to the value in the beam flange. Beam flanges, being comprised of thicker material, typically have somewhat lower yield strengths than do beam web material. Fy m for various steels are as shown in Table 7-1, based on a survey of web coupon tensile tests (Steel Shape Producers Council - 1994). The engineer is cautioned that there is no upper limit on the yield point for ASTM A36 steel and consequently, dual-certification steel having properties consistent with ASTM A572, Grade 50 is routinely supplied when ASTM A36 is specified. Consequently, it is the recommendation here that the design of connections be based on an assumption of Grade 50 properties, even when A36 steel is specified for beams. It should be noted that at least one producer offers A36 steel with a maximum yield point of 50 ksi in shape sizes ranging up to W 24x62. Strain Hardening: A factor of 1.1 is recommended for use with the mean yield stress in the foregoing table when calculating the probable plastic moment capacity Mpr.. The 1.1 factor for strain hardening, or other sources of strength above yield, agrees fairly well with available test results. The 1.1 factor could underestimate the over-strength where significant flange buckling does not act as a gradual limit on the beam strength. Nevertheless, the 1.1 factor seems a reasonable expectation of over-strength considering the complexities involved. Modeling Uncertainty: Where a design is based on approved cyclic testing, the modeling uncertainty may be taken as 1.0, otherwise the recommended value is 1.2. In summary, for Grade 50 steel, we have: β = [0.95 (54 ksi to 58 ksi)/50 ksi] (1.1) 1.2) = 1.35 t0 1.45, say 1.4 7.5.2.3 Determine Shear at the Plastic Hinge

The shear at the plastic hinge should be determined by statics, considering gravity loads acting on the beam. A free body diagram of that portion of the beam between plastic hinges, is a useful 7-19

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tool for obtaining the shear at each plastic hinge. Figure 7-3 provides an example of such a calculation. For the purposes of such calculations, gravity load should be based on the load combinations required by the building code in use. L/2 Plastic hinge

P

Note: if 2Mpr /L’is less then the gravity shear in the free body (in this case P/2 + wL’/2), then the plastic hinge location will shift and L’ must be adjusted, accordingly

L’ d/3 L

P w

VA

Mpr “A”

Vp

Mpr

L’

taking the sum of moments about “A” = 0 Vp ={Mpr + Mpr + P L’/2 + wL’2/2}/L’

Figure 7-3 - Sample Calculation of Shear at Plastic Hinge Commentary: The UBC gives no specific guidance on the load combinations to use with strength level calculations while the NEHRP Recommended Provisions do specify specific load factors for the various dead, live and earthquake components of load. For designs performed in accordance with the UBC it is customary to use unfactored gravity loads when checking the strength of elements. 7.5.2.4 Determine Strength Demands at Each Critical Section

In order to complete the design of the connection, including sizing the various plates and joining welds which make up the connection, it is necessary to determine the shear and flexural strength demands at each critical section. These demands may be calculated by taking a free body of that portion of the connection assembly located between the critical section and the plastic hinge. Figure 7-4 demonstrates this procedure for two critical sections, for the beam shown in Figure 7-3.

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Plastic hinge

Plastic hinge

Mpr

Mf

Vp

Mpr

Mc dc

Vp

x

x+dc/2

Mf=Mpr +Vpx Critical Section at Column Face

M c=Mpr +Vp(x+dc/2) Critical Section at Column Centerline

Figure 7-4 - Calculation of Demands at Critical Sections Commentary: Each unique connection configuration may have different critical sections. The vertical plane that passes through the joint between the beam flanges and column (if such joining occurs) will typically define at least one such critical section, used for designing the joint of the beam flanges to the column, as well as evaluating shear demands on the column panel zone. A second critical section occurs at the center line of the column. Moments calculated at this point are used to check strong column - weak beam conditions. Other critical sections should be selected as appropriate. 7.5.2.5 Check for Strong Column - Weak Beam Condition

When required by the building code, the connection assembly should be checked to determine if strong column - weak beam conditions are satisfied. In lieu of UBC-94 equation 11-3.1 {NEHRP-91 equation 10-3}, the following equation should be used:

∑Z where:

c

(Fyc − f a )

∑M

c

> 1.0

(7-4)

Zc is the plastic modulus of the column section above and below the connection Fyc is the minimum specified yield stress for the column above and below fa is the axial load in the column above and below Mc is the moment calculated at the center of the column in accordance with Section 7.5.2.4 Commentary: The building code provisions for evaluating strong column - weak beam conditions presume that the flexural stiffness of the columns above and below the beam are approximately equal. If non-symmetrical connection configurations are used, such as a haunch on the bottom side of the beam, this can result in an uneven distribution of stiffness between the two column segments.

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7.5.2.6 Check Column Panel Zone

The adequacy of the shear strength of the column panel zone should be checked. For this purpose, the term 0.8ΣMf should be substituted for the term 0.8ΣMs in UBC-94 Section 2211.7.2.1 {0.9ΣφbMp in NEHRP-91 Section 10.10.3.1}, repeated below for convenience of reference. Mf is the calculated moment at the face of the column, when the beam mechanism forms, calculated as indicated in Section 7.5.2.4 above. 2211.7.2.1 Strength. The panel zone of the joint shall be capable of resisting the shear induced by beam bending moments due to gravity loads plus 1.85 times the prescribed seismic forces, but the shear strength need not exceed that required to develop 0.8ΣMs of the girders framing into the column flanges at the joint. The joint panel zone shear strength may be obtained from the following formula:

 3b c t c f 2  V = 0.55Fy d c t 1 +  dbdct  

(11-1)

where: bc = width of column flange db = the depth of the beam (including any haunches or cover plates) dc = the depth of the column t = the total thickness of the panel zone including doubler plates tcf = the thickness of the column flange

7.6 Metallurgy and Welding For Guidelines on Metallurgy and Welding for New Structures, see Chapter 8 of these Interim Guidelines. The recommendation for welding electrodes capable of depositing weld metal with specified notch toughness, as described therein, should apply to the critical beam flange to column flange field welded joints. It need not apply to shop welds of continuity plates, etc. Commentary: This is an area of continuing controversy in the community, requiring additional research for resolution. Some professionals and researchers knowledgeable in fracture mechanics believe it is essential that all weld metal in the beam column connection, including both field and shop welds, welds of continuity plates, doubler plates, etc., as well as the welds of beam to column flanges, should have minimum specified notch toughness. Some of these same professionals believe that the notch toughness requirement should apply to the combined metal, consisting of deposited electrode metal and fused base metal. The current recommendations, which are less restrictive than this position, are based on recommendations of members of the AWS D1.1 committee. These recommendations are consistent with the observation of damage in the Northridge Earthquake, in which most fractures initiated at the root of the beam flange to column flange weld. It is of course possible that if notch tough material is used at this joint and not at others, fractures will initiate in future events at the next critical section, which may be a welded joint using material with low toughness at continuity plates, or other locations. 7-22

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While there is a lack of agreement as to the extent to which notch toughness specifications should apply to welded joints in the moment connection, there is general agreement that previously acceptable electrodes that had no reported notch toughness values should no longer be used for the critical beam flange to column flange field welded joints. Most of the electrodes that are currently commercially available and have specified notch toughness requirements will meet the notch toughness recommendations contained in Chapter 8 of these Interim Guidelines. Additional research may indicate that alternate criteria are appropriate. A similar level of disagreement exists with regard to the need for specifying notch toughness in base metals. Most of the fractures which have been investigated have initiated in the weld metal rather than in the base metal. Once these fractures extend into the base metal, they have already reached significant size and material toughness alone may not be able to arrest them. Additional research into the benefits of tough material, both in welds and base metals is clearly called for. 7.7 Quality Control/Quality Assurance Refer to Chapters 9, 10 and 11 of these Interim Guidelines. 7.8 Guidelines on Other Connection Design Issues The emphasis thus far in testing of connection assemblages has been on the beam flange/column flange joint. The other components of the connection such as panel zones, web connections and continuity plates have not been studied significantly as independent parameters in the available testing programs to date. It is assumed that the variation of these components will have effects on the performance of the connection and thus on the flange joints, and that an as yet undetermined balance of the sizes and details of these significant components will result in the optimum performance of a particular connection and its various joints. Interim Guidelines for these other critical portions of the connection assembly are presented below. 7.8.1 Design of Panel Zones

No current recommendations are made to supplement or modify the UBC-1994 {NEHRP-91} provisions for the design of panel zones, other than as indicated in Section 7.5.2.6, above. Panel zone demands should be calculated in accordance with Section 7.5.2.6. As with other elements of the connection, available panel zone strength should be computed using minimum specified yield stress for the material, except when the panel zone strength is used as a limit on the required connection strength, in which case Fym should be used. Where connection design for two-sided connection assemblies is relying on test data for onesided connection assemblies, consideration should be given to maintaining the level of panel zone

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deformation in the design to a level consistent with that of the test, or at least assume that the panel zone must remain elastic, under the maximum expected shear demands. Commentary: At present, no changes are recommended to the code requirements governing the design of panel zones, other than in the calculation of the demand. There is evidence that panel zone yielding may contribute to the plastic rotation capability of a connection. However, there is also concern and some evidence that if the deformation is excessive, a kink will develop in the column flange at the joint with the beam flange and, if the local curvature induced in the beam and column flanges is significant, can contribute to failure of the joint. This would suggest that greater conservatism in column panel zone design may be warranted. In addition to the influence of the deformation of the panel zone on the connection performance, it should be recognized that the use of doubler plates and especially the welding associated with them is likely to be detrimental to the connection performance. It is recommended that the Engineer consider use of column sizes which will not require addition of doubler plates, where practical. 7.8.2 Design of Web Connections to Column Flanges

Specific modifications to the code requirements for design of shear connections are not made at this time. It should be noted that the emergency code change to the UBC-94 {NEHRP-94} deleted the former requirements for supplemental web welds on shear connections. This is felt to be appropriate since these welds can apparently contribute to the potential for shear tab failure at large induced rotations. When designing shear connections for moment-resisting assemblies, the designer should calculate shear demands on the web connection in accordance with Section 7.5.2.4, above. Commentary: Some engineers consider that it is desirable to develop as much bending strength in the web as possible. Additionally, it has been observed in some laboratory testing that pre-mature slip of the bolted web connection can result in large secondary flexural stresses in the beam flanges and the welded joints to the column flange. However, there is some evidence to suggest that if flange connections should fail, welding of shear tabs to the beam web may promote tearing of the tab weld to the column flange or the tab itself through the bolt holes, and some have suggested that welding be avoided and that web connections should incorporate horizontally slotted holes to limit the moment which can be developed in the shear tab, thereby protecting its ability to resist gravity loads on the beam in the event of flexural connection failure. 7.8.3 Design of Continuity Plates

Contrary to current code requirements, it is recommended that continuity plates be provided in all cases and that the thickness be at least equal to the thickness of the beam flange (not 7-24

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including cover plates) or one half the total effective flange thickness (flange plus cover plate). Welds of the continuity plate to the column should develop the strength of the continuity plate. Where two-sided connection assemblies are designed based on one-sided connection assembly test data, consideration should be given to the effect of the greater distortion of the continuity plates expected in the two sided case. For reinforced connections using vertical ribs or other configurations of reinforcement, continuity plate sizing should be based on engineering principles and consideration of stress patterns which may occur due to column flange distortion. For connections incorporating haunches, continuity plates should be provided opposite the joint of the haunch flange with the column flange. Commentary: The determination of continuity plate thickness requires, in addition to code conformance requirements, engineering judgment based on recognition of two competing factors: a)

Overly thick continuity plates and their welding will contribute to restraint and consequent residual stresses in the column, as well as to the other usual detrimental effects of large welds. Conditions of high restraint tend to be conducive to the initiation of fracture.

b)

Omission of continuity plates or the use of overly thin continuity plates will permit column flange distortions which will, in turn, lead to higher stress concentrations in the beam flange joint opposite the column web.

Testing to date has not firmly established an appropriate design criteria for continuity plates, or even that these are definitely needed to obtain good connection performance in all cases. However, tests of specimens reinforced with cover plates to date, have been most successful when continuity plates were present (Engelhardt & Sabol - 1994). Tests using otherwise similar designs but with different continuity plate thicknesses have not been performed. This is an area where further research would be beneficial. 7.8.4 Design of Weak Column and Weak Way Connections

The code permits the use of strong beam/weak column designs under certain circumstances. There is some question as to what should be required for the connections at such conditions. While testing has demonstrated little capability of the pre-Northridge prescriptive connection to develop significant beam yielding without failure, it should be recognized that if the beam is stronger than the column, considering conservative estimates of the column strength including strain hardening, then the beam and its connection can be expected to remain below even this low failure threshold, and it would appear to be unnecessary to provide strengthened connections. 7-25

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When beam connections are made to the web of columns (weak way) which are stronger than the beams, then connection design should be treated similarly to that of strong direction connections with additional consideration for the unique features of weak direction connections (see Tsai and Popov - 1988). Note that the question of column flange through-thickness strength is not a consideration for this type of connection, but that development of the strength of a cover plated flange through welds in shear to the inside face of the column flange may be difficult. Unless the members so connected represent a very small part of lateral resistance of the structure, testing of such connections should be considered as mandatory. Extrapolation of results from strong way connection testing should not be done. The effect of weak way connection action on the strength and behavior of companion strong way connections, for columns participating in orthogonal lateral-force-resisting frames, has not been tested. Commentary: Since 1985, the strong column/weak beam principle has been required, but exceptions have been provided which permit weak columns in some instances. These exceptions have not been revoked, and, in fact, the interest in redundancy generated by the Northridge failures has actually increased interest in their use, to the extent permitted, in moment frame systems, where all beamcolumn connections in the structure are connected for moment resistance and made part of the lateral-force-resisting system. Considering the fact that columns resisting flexural demands about their minor axes will not generally be capable of developing the beam flexural yield strength should permit consideration of the pre-Northridge connections for this use. On the other hand, where specific code exceptions permit use of weak column systems for all or a large part of the lateral resistance a more conservative approach is merited. Use of weak column systems as the primary lateral resistance is strongly discouraged and should not be considered as a desirable or acceptable method of avoiding beam flange connection concerns and reinforcing requirements. Further, although logic would indicate that the strength demand on connections in weak column structures would be limited by column hinging, and that therefore the beam-column connection should be protected, evidence suggests that this may not be the case. It has been reported that a hospital structure affected by the Northridge Earthquake experienced failure of almost all of its beam-column connections, despite having all or many weak column conditions. 7.9 Moment Frame Connections for Consideration in New Construction The moment frame connection formerly prescribed by the code was configured to require development of a plastic hinge in the beam adjacent to the beam-to-column connection. The Northridge experience and subsequent testing have shown that as the possible result of a number of factors, it is not reasonable to expect reliable development of plastic hinges at this location, at least within the range of design parameters explored to date. Therefore, connections should be configured to encourage plastic hinging action to other locations.

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The types of connections described in the following subsections are felt to offer some promise of providing more reliable inelastic action in WSMFs, consistent with that assumed in the design of such frames. It is of course assumed that the required joints, both welded and bolted, have been installed with appropriate quality control as described previously. Reference to laboratory testing is provided for those connection configurations for which research has been reported. However, it should be noted that none of these connections has been tested sufficiently at this time to permit unqualified use of the connection. The figures provided in the following sections are schematic, indicating the general type of connection configuration being described. When designing connections patterned after the reported test data, the test specimen details included in the references should be reviewed to determine specific details not shown. The SAC Joint Venture does not endorse or specifically recommend any of the connection details shown in this Section. These are presented only to acquaint the reader with available information on representative testing of different connection configurations that have been performed by various parties. Commentary: With the large interest and availability of funding for research on steel moment frame connections, any lists of connection concepts, such as the above will necessarily become at least partially obsolete by the time they are published. With this in mind, it is very important that there be a publicly accessible center to accumulate testing results as they become available. It is the recommendation of this guideline that as efforts in this area progress, SAC become the repository and distribution group for such information. It is hoped that all engineers, researchers, and contractors responsible for tested connections will willingly share all information on the tests and designs with SAC, with the structural engineering profession, and with the building construction industry. The various connections suggested in this section were all nominally fully restrained (FR) connections. It has been suggested that partially restrained (PR) connections may be a cost-effective and reliable alternative to these connections. AISC and NSF are currently conducting research into the use of this system and it may become an attractive alternative in the future. 7.9.1 Cover Plate Connections

Figure 7-5 illustrates the basic configuration of cover plated connections. Short cover plates are added to the top and bottom flanges of the beam with fillet welds adequate to transfer the cover plate forces to the beam flanges. The bottom flange cover plate is shop welded to the column flange and the beam bottom flange is field welded to the column flange and to the cover plate. The top flange and the top flange cover plate are both field welded to the column flange with a common weld. The web connection may be either welded or high strength (slip critical) bolted. Limited testing of these connections (Engelhardt & Sabol - 1994), (Tsai & Popov -1988) has been performed. 7-27

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A variation of this concept which has been tested successfully very recently (Forrel/Elsesser Engineers -1995), uses cover plates sized to take the full flange force, without direct welding of the beam flanges themselves to the column. In this version of the detail, the cover plate provides a cross sectional area at the column face about 1.7 times that of the beam flange area. In the detail which has been tested, a welded shear tab is used, and is designed to resist a significant portion of the plastic bending strength of the beam web.

T&B

Figure 7-5 - Cover Plate Connection Design Issues: Approximately eight connections similar to that shown in Figure 7-5 have been recently tested (Engelhardt & Sabol - 1994), and they have demonstrated the ability to achieve acceptable levels of plastic rotation provided that the beam flange to column flange welding is correctly executed and through-thickness problems in the column flange are avoided. This configuration is relatively economical, compared to some other reinforced configurations, and has limited architectural impact. Six of eight connections tested by the University of Texas at Austin were able to achieve plastic rotations of at least 0.025 radians, or better. Strength loss at the extreme levels of plastic rotation did not reduce the flexural capacity to less than the plastic moment capacity of the section based on minimum specified yield strength. One specimen achieved plastic rotations of 0.015 radians when a brittle fracture of the CJP weld (type W2 failure) occurred. This may partially be the result of a weld that was not executed in conformance with the specified welding procedure specification. The second unsuccessful test specimen achieved plastic rotations of 0.005 radian when a section of the column flange (type C2 failure) occurred. A similar failure occurred in recent testing by Popov of a specimen with cover plates having a somewhat modified plan shape.

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Quantitative Results: No. of specimens tested: 8 Girder Size: W36 x 150 Column Size: W14 x 455 Plastic Rotation achieved6 Specimens : >0.025 radian 1 Specimen: 0.015 radian 1 Specimen: 0.005 radian Although apparently more reliable than the former prescriptive connection, this configuration is subject to some of the same flaws including dependence on properly executed beam flange to column flange welds, and through-thickness behavior of the column flange. Further these effects are somewhat exacerbated as the added effective thickness of the beam flange results in a much larger groove weld at the joint, and therefore potentially more severe problems with brittle heat affected zones and lamellar defects in the column. Indeed, a significant percentage of connections of this configuration have failed to produce the desired amount of plastic rotation. 7.9.2 Flange Rib Connections

Figure 7-6 demonstrates the basic configuration for connections with flange ribs. The intent of the rib plates is to reduce the demand on the weld at the column flange and to shift the plastic hinge from the column face.

2 1

Typ.

Typ.

Figure 7-6 - Flange Rib Connection Design Issues: There is a limited body of testing of connections similar to these (Engelhardt & Sabol - 1994), (Tsai & Popov - 1988), and they have demonstrated the ability to achieve acceptable levels of plastic rotation provided that the girder flange welding is correctly executed.

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Quantitative Results: No. of specimens tested: 2 Girder Size: W36 x 150 Column Size: W14 x 455 Plastic Rotation achieved2 Specimens : >0.025 radian Performance is dependent on properly executed girder flange welds. The joint can be subject to through-thickness failures in the column flange, although it should be somewhat more resistant to such failures than connections reinforced with cover plates, as the weld size is reduced. The size of the specimens tested required the use of two upstanding ribs per flange. This increased the costs significantly above those designs that use only one rib per flange, located above the girder center line. However, limited testing of the design with one rib at the girder centerline (Tsai & Popov) indicated the potential for premature failure of the weld of the rib to the girder at the outstanding edge. It should also be noted that the specimens tested by Engelhardt & Sabol, and reported above, incorporated columns with particularly heavy flanges. The ribs have the potential to cause high local stresses in the column flanges and this configuration may not behave acceptably when used with lighter section. Preliminary reports from fabricators and erectors indicate that the cost of this connection is quite high, relative to other configurations. 7.9.3 Bottom Haunch Connections

d

Figure 7-7 indicates several potential configurations for single, haunched beam-column connections. As with the cover plated and ribbed connections, the intent is to shift the plastic hinge away from the column face and to reduce the demand on the CJP weld by increasing the depth of the section. To date, the configuration incorporating the triangular haunch has been subjected to limited testing. Testing of configurations incorporating the straight haunch are currently planned, but have not yet been performed.

WT

d/3

WT

1 2 or or

Figure 7-7 - Bottom Haunch Connection Modification Two tests have been performed to date, both successfully. Both tests were conducted in a repair/modification configuration. In one test, a portion of the girder top flange, adjacent to the 7-30

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column, was replaced with a thicker plate. In addition, the bottom flange and haunch were both welded to the column. This specimen developed a plastic hinge within the beam span, outside the haunched area and behaved acceptably. A second specimen did not have a thickened top flange and the bottom girder flange was not welded to the column. Plastic behavior in this specimen occurred outside the haunch at the bottom flange and adjacent to the column face at the top flange. Failure initiated in the girder at the juncture between the top flange and web, possibly contributed to by buckling of the flange as well as lateral torsional buckling of the section. Fracture progressed slowly along the top fillet of the girder and eventually, traveled into the flange itself. Design Issues: The haunch can be attached to the girder in the shop, reducing field erection costs. Weld sizes are smaller than in cover plated connections. The top flange is free of obstructions. Quantitative Results: No. of specimens tested: 2 Girder Size: W30 x 99 Column Size: W14 x 176 Plastic Rotation achievedSpecimen 1:0.04 radian (w/o bottom flange weld and reinforced top flange) Specimen 2:0.05 radian (with bottom flange weld and reinforced top flange) Performance is dependent on properly executed complete joint penetration welds at the column face. The joint can be subject to through-thickness flaws in the column flange; however, this connection may not be as sensitive to this potential problem because of the significant increase in the effective depth of the beam section which can be achieved. Welding of the bottom haunch requires overhead welding when relatively shallow haunches are used. The skewed groove welds of the haunch flanges to the girder and column flanges may be difficult to execute. The increased depth of the beam, resulting from the haunch may have undesirable impact on architectural design. Unless the top flange is prevented from buckling at the face of the column, performance may not be adequate. For configurations incorporating straight haunches, the haunch must be long, in order to adequately develop stress into the haunch, through the web. This tends to increase demands at the column face. Additional testing of all these configurations is recommended. 7.9.4 Top and Bottom Haunch Connections

Figure 7-8 illustrates this connection configuration. Haunches are placed on both the top and bottom flanges. Two tests have been performed on connections utilizing this configuration; both were highly successful.

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d/3

d

WT

1 2

or

Figure 7-8 - Top and Bottom Haunch Connection Design Issues: In two tests of this connection configuration performed to date, it has exhibited extremely ductile behavior. Plastic rotations as large as 0.07 radians were obtained. In addition to having very good plastic capacity, the connection is highly redundant. If failure should occur at one of the complete joint penetration welds of the haunch plate, significant residual strength would be available from the remaining girder flange welds. This is one of the more costly connection configurations. Some of this cost could be reduced by eliminating the welds between the girder flanges and columns, however, the performance of the connection in that configuration has not been tested. The presence of the haunch at the top of the girder could be an architectural problem. Quantitative Results: No. of Specimens Tested: 2 Girder Size: W30 x 99 Column Size W14 x 176 Plastic Rotation achieved - 0.07 radians 7.9.5 Side-Plate Connections

This approach eliminates loading the column in the through-thickness direction by removing the CJP welds at the girder flange and by shifting the plastic hinge from the column face. The tension and compression forces are transferred from the girder flanges into the column through fillet welds. A mechanism to provide a direct connection between the column panel zone and the beam flanges is required; the difficulty appears to be equalizing the width of the beam and column flanges. At least two configurations of side-plated connections have been tested. One set, shown in Figure 7-9, utilized flat bars at the top and bottom girder flanges, to transfer flange forces to the column (Engelhardt & Sabol - 1994). The girder was widened to the width of the column with 7-32

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the use of filler plates. The specimens achieved plastic rotations of 0.015 radians, however, fractures developed within the welds connecting the beam flange to the transfer plates. Failure of the shear tab, and finally the side plates themselves followed the initiation of these fractures. It is believed that the unsuccessful behavior of this particular specimen was related to the method used to increase the width of the beam flange to equal that of the column flange, using a combination of a filler bar and welding. Other approaches, such as providing a full width cover plate for the girder flanges, may provide better performance.

Possible Alternative

Tested Configuration

Figure 7-9 Side Plate Connection Design Issues: This connection avoids both the large complete joint penetration welds of the beam flange to the column and the potential for through-thickness failure of the column flange. Much of the additional fabrication can be performed in the shop. This connection did not demonstrate adequate plastic rotation capacity in the configurations tested to date. Additional testing is required to determine if modified configurations will perform in a more acceptable manner. Quantitative Results: Separate Top & Bottom Side Plates No. of specimens tested: 2 Girder Size: W36 x 150 Column Size: W14 x 455 Plastic Rotation achieved2 Specimens :0.015 radian A second, proprietary configuration, is shown in Figure 7-10. Three specimens have undergone full-scale testing to date and achieved large plastic rotations. Loss of strength at large plastic rotation demands was comparable to that of other successful connections. The developer 7-33

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of this connection has applied for US and foreign patents. Further information on technical data for this configuration, may be obtained from the developer.

NOTICE OF CONFIDENTIAL INFORMATION: WARNING: The information presented in this figure is PROPRIETARY. US and Foreign Patents have been applied for. Use of this information is strictly prohibited except as authorized in writing by the developer. Violators shall be prosecuted in accordance with US and Foreign Patent Intellectual Property Laws.

Figure 7-10 - Proprietary Side Plate Connection Design Issues: Testing of three prototype specimens (Uang & Latham - 1995) indicates that this proprietary connection has the ability to achieve very satisfactory levels of plastic rotation without relying on sensitive CJP welds between the column and girder flanges or specifying weld material with notch toughness. The elimination of the through-thickness loading of the flange may result in higher levels of connection reliability. Due to the exclusive use of fillet welds, special inspection requirements for welding and bolting can be reduced significantly with this connection. This connection is proprietary and license fees are associated with its use. The cost of the connection may be greater than some of the other modification methods discussed above; however, this cost differential may not be as great on double-sided connections because much of the cost is associated with the side plates which are similar for both single-sided and doublesided connections. However, double sided connections will require doubling the sizes of the welds which deliver the forces to the columns, and potentially increasing plate thickness as well. The connection of beams framing into the minor axis of the column are made more difficult by this connection, particularly if they must be connected for moment resistance. Publicly bid projects will have to develop performance specifications to permit other connections to be considered for use unless a strong case for sole-sourcing the connection can be made. Quantitative Results: No. of specimens tested: 3 Girder Size: W36 x 150 7-34

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Column Size: W14 x 426 Plastic Rotation achieved3 Specimens : >0.042 radian to 0.06 radian 7.9.6 Reduced Beam Section Connections

In this connection, the cross section of the beam is intentionally reduced within a segment, to produce an intended plastic hinge zone or fuse, located within the beam span, away from the column face. Several ways of performing this cross section reduction have been proposed. One method includes removal of a portion of the flanges, symmetrical about the beam centerline, in a so-called “dog bone” profile. Care should be taken with this approach to provide for smoothly contoured transitions to avoid the creation of stress risers which could initiate fracture. It has also been proposed to create the reduced section of beam by drilling a series of holes in the beam flanges. Figure 7-11 illustrates both concepts. The most successful configurations taper the reduced section, through the use of unsymmetrical cut-outs, or variable size holes, to balance the cross section and the flexural demand. Testing of this concept was first performed by a private party, and US patents were applied for and granted. These patents have now been released. Limited testing of both “dog-bone” and drilled hole configurations have been performed in Taiwan (Chen and Yeh - 1995). The American Institute of Steel Construction is currently performing additional tests of this configuration (Smith-Emery - 1995), however the full results of this testing are not yet available. There is a concern that the presence of a concrete slab at the beam top flange would tend to limit the effectiveness of the reduced section of that flange, particularly when loading places the top flange into compression. It may be possible to mitigate this effect with proper detailing of the slab.

Symmetrical

Unsymmetrical

Weakened Segment

Figure 7-11 - Reduced Beam Section Connection Design Issues: This connection type is potentially the most economical of the several types which have been suggested. The reliability of this connection type is dependent on the quality of the complete joint penetration weld of the beam to column flange, and the through-thickness 7-35

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behavior of the column flange. If the slab is not appropriately detailed, it may inhibit the intended “fuse” behavior of the reduced section beam segment. It is not clear at this time whether it would be necessary to use larger beams with this detail to attain the same overall system strength and stiffness obtained with other configurations. In limited testing conducted to date of the unsymmetrical “dog-bone” configuration (Smith-Emery - 1995), the plastic hinging which occurred at the reduced section was less prone to buckling of the flanges than in some of the other configurations which have been tested, due to the very compact nature of the flange in the region of the plastic hinge. Quantitative Results: No. of specimens tested: 2 Girder Size: W30 x 99 Column Size: W14 x 176 Plastic Rotation achieved- 0.03 radian 7.9.7 Slip - Friction Energy Dissipating Connection

This connection uses high strength bolts and slotted holes to develop the flange forces into the columns. A brass shim in the shear transfer plane provides for controlled friction force. In concept, slip along this bolted connection limits the amount of force which can be transferred to the column and allows plastic deformation to occur in a benign manner. Two alternative configurations have been suggested for attachment of the flanges to the column. One incorporates bolted “T” sections and the other welded plates. Figure 7-12 shows the bolted “T” configuration. To date, two tests have been performed on the bolted “T” configuration (Popov & Yang1995). Results were excellent with large inelastic displacements obtained without strength or stiffness degradation. Type “X” bolts

Steel Shims

Tee Section w/ slotted holes

Brass Shims

Figure 7-12 - Slip Friction Energy Dissipation Connection 7-36

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Design Issues: In the limited testing performed, this connection was able to accommodate large inelastic displacements without damage to the connection or beams. This connection can be assembled in the field without welding and can accommodate large plastic rotations without permanent damage to the structure. The connection is sensitive to fit-up, cleanliness of the faying surfaces and tension in the highstrength bolts, and therefore will require careful field quality control. As with the “reduced section beam” connections, this connection may be sensitive to the presence of a slab and careful detailing of the slab to permit the expected connection rotations to occur may be required. The strength that can be developed by this connection is limited by the number of bolts that can be practically placed. It may not be suitable for use with larger members with high strength demands. The brass shims, used at the slip plane interface are quite costly. Metal parts kept in contact under pressure over a period of years may tend to become partially welded together, potentially reducing the effectiveness of the connection with time. Additional research is required on this effect. 7.9.8 Column-Tree Connection

This concept has been widely used in Japan, with mixed success. Short stubs of girders are fabricated and shop welded to the column. Field connection to the balance of the girder is made with bolted connections. The girder stubs can be intentionally fabricated stronger than the balance of the girder, to force yielding and formation of a plastic hinge away from the column. Figure 7-13 demonstrates the basic concept. Extensive testing of this connection has not been performed in the US Some variations of this connection, in common use in Japan, performed very poorly in the recent Kobe Earthquake (Watabe-1995). In at least one version of this connection, the beam stub ran continuously through the connection and the columns were shop welded to the top and bottom of this stub. A number of these connections experienced fracture of the shop weld of the column to the beam stub. However, it is reasonable to expect that configurations for this concept can be developed that would permit more favorable behavior.

Figure 7-13 - Column Tree Connection 7-37

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Design Issues: The basic advantage to this connection is that critical welding can be performed in the fabrication shop where it should be possible to attain better quality control. In addition, field erection costs are reduced through the use of bolted field connections. Testing of this connection in configurations similar to US construction practice has not been performed. Some configurations utilized in Japan performed poorly in the Kobe Earthquake. The connection is dependent on the quality of beam flange to column flange welding and the through-thickness behavior of the column flange. Transportation and handling of tree columns is probably somewhat more difficult and expensive than for standard columns. 7.9.9 Slotted Web Connections

In the former prescriptive connection, in which the beam flanges were welded directly to the column flanges, beam flexural stress was transferred into the column web through the combined action of direct tension across the column flange, opposite the column web, and through flexure of the column flange. This stress transfer mechanism results in a large stress concentration at the center of the beam flange, opposite the column web. Recent research (Allen, et. al. - 1995) indicates that the provision of continuity plates within the column panel zone reduces this stress concentration somewhat, but not completely. The intent of slotted web connections is to further reduce this stress concentration and to achieve a uniform distribution of flexural stress across the beam flange at the connection. Figure 7-14 indicates one configuration for this connection type that has been successfully tested. In this configuration, vertical plates are placed between the column flanges, opposite the edges of the top and bottom beam flanges to stiffen the outstanding column flanges and draw flexural stress away from the center of the beam flange. Horizontal plates are placed between these vertical plates and the column web to transfer shear stresses to the panel zone. The web itself is softened with the cutting of a vertical slot in the column web, opposite the beam flange. High fidelity finite element models were utilized to confirm that a nearly uniform distribution of stress occurs across the beam flange.

PP

1/4” Slot Typical 3/4” Hole

Figure 7-14 - Slotted Web Connection

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Design Issues: This detail is potentially quite economical, entailing somewhat more shop fabrication than the former prescriptive connection, but similar levels of field erection work. Contrary to the recommendations contained in these Interim Guidelines, this connection does not shift the location of plastic hinging away from the column face. However, two connections similar to that shown in Figure 7-14 have recently been tested succesfully (Allen. - 1995). The connection detail is sensitive to the quality of welding employed in the critical welds, including those between the beam and column flanges, and between the vertical and horizontal plates and the column elements. It has been reported that one specimen, with a known defect in the beam flange to column flange weld was informally tested and failed at low levels of loading. The detail is also sensitive to the balance in stiffness of the various plates and flanges. For configurations other than those tested, detailed finite element analyses may be necessary to confirm that the desired uniform stress distribution is achieved. The developer of this detail indicates that for certain column profiles, it may be possible to omit the vertical slots in the column web and still achieve the desired uniform beam flange stress distribution. This detail may also be sensitive to the toughness of the column base metal at the region of the fillet between the web and flanges. In heavy shapes produced by some rolling processes the metal in this region may have substantially reduced toughness properties relative to the balance of the section. This condition, coupled with local stress concentrations induced by the slot in the web may have the potential to initiate premature fracture. The developer believes that it is essential to perform detailed analyses of the connection configuration, in order to avoid such problems. Popov tested one specimen incorporating a locally softened web, but without the vertical and horizontal stiffener plates contained in the detail shown in Figure 7-14. That specimen failed by brittle fracture through the column flange which progressed into the holes cut into the web. The stress patterns induced in that specimen, however, were significantly different than those which occur in the detail shown in the figure. Quantitative Results: Number of specimens tested: 2 Girder Size: W 27x94 Column Size: W 14x176 Plastic Rotation Achieved: Specimen 1: 0.025 radian Specimen 2: 0.030 radian 7.10 Other Types of Welded Connection Structures These Interim Guidelines have focused on the design of the moment and shear resisting FR connections in moment frame systems in which the lateral forces are resisted by bending in beams and columns. In addition to moment frame systems, there are a number of other system types which conceivably could exhibit connection or joint distress similar to that seen in moment frames in Northridge when deformed under high intensity earthquake motions. Except for one detail of a welded base plate at the base of a braced frame, there has been no reported damage to the following systems from the Northridge earthquake. The response to earthquake motions, however, presents similar potential conditions to those found in moment frame connections. 7-39

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Therefore when designing new construction, close attention should be given to these structural systems and details so that damage similar to that observed on WSMF systems can be avoided in future earthquakes. 7.10.1 Eccentrically Braced Frames (EBF)

EBF provisions in the code require the use of "link beam" elements. Link beams are usually designed to yield in shear in the web, but can be designed to yield in flexure. In some configurations, the link beam is connected to the column flange in a manner nearly identical to the connection of the WSMF. The connection of the brace element to the beam also connects to the beam flange, but the connection has additional design requirements which modify the type of connection. In addition to connection concerns for EBFs, the currently recognized variability of steel strengths should be considered in designing the components of the EBF. It is recommended that connections in EBFs intended to resist plastic rotation demands be designed the same as WSMF connections, as previously defined in these Interim Guidelines, with due consideration given to the additional shear forces which may be induced in the link beams. Commentary: Although the code provisions are intended to cause the link beams connected to columns to yield in shear (length of link limited to 1.6MS/VS), the link may be at or near its bending strength when shear yielding would occur. Thus the connection to the column flanges may be vulnerable to the same or similar problems as those exhibited by WSMF connections during the Northridge earthquake. Recognition of the probable strength of steel in the link beam could be critical to the performance of these structures. The inelastic behavior of EBF structures is intended to be controlled through yielding of these links. If link beams are fabricated from excessively strong material, they may not yield before other parts of the frame become damaged. 7.10.2 Dual Systems

The provisions for Dual Systems in the code require that the system include a Special Moment-Resisting Frame (SMRF) designed according to the same provisions as if it were the primary system but capable of resisting at least 25% of the required lateral forces. In addition, it is required to have a primary system consisting of either concrete shear walls, Special Concentrically Braced Frames (SCBF), Concentrically Braced Frames (CBF), or Eccentrically Braced Frames (EBFs). Connection design for moment frames used in Dual Systems should conform to the recommendations of these Interim Guidelines for SMRF systems. Commentary: Prior to the 1967 UBC, Dual Systems design required a primary system (shear walls or CBFs) capable of resisting 100% of the required lateral forces in conjunction with a "back-up" SMRF capable of resisting at least 25% of the total forces. The assumption for this type of system was that the SMRF would take over and prevent collapse of the structure in the event of failure of the stiffer, 7-40

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but ostensibly less ductile, primary system. In this concept of design, the SMRF was there solely for redundancy. In the 1967 UBC, an additional provision was added in which the primary system and the SMRF were required to be designed to share the total required lateral force according to their elastic stiffness. In the 1988 UBC, the requirement that the primary system (shear walls or braced frames) be designed to resist 100% of the required total force was eliminated, but the other two requirements remain. This potentially makes it even more important that the SMRF portion of a dual system have adequate ductility to survive a major event. In general, dual systems have been a somewhat controversial system. Some engineers believe that the added redundancy provided by the backup system is quite beneficial while others do not believe that the relatively weak and often flexible back-up system improves building performance significantly. Little analytical research of these systems has been performed. Such research would be beneficial, however, in providing guidance as to the amount of ductility required of the backup frame system. 7.10.3 Welded Base Plate Details

The detail of concern is in any system of steel framing where a column, which is subject to high axial tension or flexure, or both, is directly welded to its baseplate in a manner similar to that used for beam-to-column moment connections. Additional concerns occur when anchor bolts are fastened to the base plate in close proximity to the bottom of the column. Commentary: When a column is welded directly to the base plate and has the potential for being loaded with significant tension or tension in combination with flexure, CJP welds and the through-thickness strength of the base plate are required to resist the tensile forces. The combination of uncertainty of the through-thickness strength and the uncertainty of the axial loading suggests that another type of connection detail should be chosen. Frequently, the anchor bolts are placed close to the face of the column flanges. If the anchor bolts are strong enough so that the mechanism of failure is flexure in the base plate, the short flexural span makes it impossible for flexural yielding to occur and may result in a brittle fracture of the plate or of the CJP welds. 7.10.4 Vierendeel Truss Systems

A Vierendeel Truss (VT) is a type of truss without diagonals in which shear forces are resisted by the vertical members and chords, acting together as moment-resisting frames. VT's may have diagonals in some bays in some designs, but may also be designed to rely totally on the verticals. Where both chords and verticals of VT's are wide flange shapes, the connections of the verticals to the chords and the chords to the columns are often detailed in the same manner as the beam-tocolumn flange connection of WSMFs. A variation on the conventional horizontal Vierendeel Truss which deserves similar attention is a system where vertical loads in a discontinuous column 7-41

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are supported by moment connected beams at several floors, rather than by a single transfer girder above the location of the column discontinuity. Commentary: Considering the brittle nature of the damage to steel structures due to the Northridge earthquake, Engineers should have some concerns about VT systems as described above even when they are designed to carry vertical loads only, particularly if the loadings are variable and could significantly exceed design loads in extreme cases. Where VT's are a part of the lateral system, either serving simply as a moment frame girder or as a transfer girder, seismic deformations potentially could lead to yielding at the connections of the truss verticals to the chords. Such connections should be designed in the same manner as the beam-to-column connections of WSMFs. If such yielding is possible, the effect of such yielding on the vertical load capacity and deformation should be investigated. 7.10.5 Moment Frame Tubular Systems

This type of lateral-force-resisting system is common in very tall buildings. The moment frame is arranged with relatively short spacing of columns around the perimeter of the structure. The system is actually a special type of WSMF which has very stiff beams so that the chord forces at the end of a moment frame can be distributed to adjacent columns perpendicular to the plane or around the corner of the moment frame. The system is defined as a three-dimensional space frame structure composed of three or more frames connected at the corners (or intersections) to form a vertical tube-like structure (or a structure composed of several adjacent tubes). Of particular concern is the short beam span which renders some of the solutions for local strengthening of beams difficult to achieve. On the other hand, plastic rotational demands due to high seismic forces may be shown to be very low in some designs. Commentary: Moment frame tube structures are normally very redundant systems with many moment-resisting members and connections. A thorough analysis of the structural system should be made to determine what potential plastic rotation demand would be required on the connections. With very tall buildings, seismic response becomes more heavily influenced by the higher modes of vibration, and design of members and connections might be controlled by wind forces. 7.10.6 Welded Connections of Collectors, Ties and Diaphragm Chords

These members are part of a building’s lateral-force-resisting system. They are usually horizontal members which, in addition to supporting adjacent gravity load, are also required to transmit large axial tensile and compressive forces. If development of the tensile and flexural forces at the connections to the columns requires welding of the member flanges to the column, all of the recommendations for WSMF connections should be followed.

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Commentary: Where chord, collector and time members are nearly pure axial members, such as would occur in a building with shear walls or braced frames that is laterally very stiff, the former prescriptive connection may be found to be sufficient, depending on the size of the member.. 7.10.7 Welded Column Splices

Even though no column splice damage has been reported from the Northridge earthquake, column splices incorporating partial penetration flange welds should be used cautiously, particularly if the potential for large tensile and/or flexural forces are present. Partial penetration welds result in a crack-like feature, which can initiate fracture under conditions of high stress. A number of structures experienced failure of column splices in the 1995 Kobe Earthquake, with some such failures leading to structural collapse. Commentary: Of particular concern would be the use of partial penetration butt welds on the column flanges. The configuration of partial penetration welds provides a notch on the inner edge of the weld. Thus other methods of effecting a column splice should be used if significant or unpredictable tensile or flexural forces are possible. When considering bending in column splices of moment frames, it has been shown by inelastic time-history analyses that reliance should not be placed on the inflection point occurring at the mid height of the column. The studies show that the location of the hinge can change significantly as the structure deforms, both due to higher mode effects and due to the inelastic response of the members. 7.10.8 Built-up Moment Frame Members

Built-up beams and columns used in moment frame systems have the same concerns in the design of connections as the rolled shape systems previously discussed. The welds connecting the various component parts of the built-up members should be designed to be capable of resisting the effects of potential plastic behavior and connections of built-up members should be designed to preclude reliance on yielding of steel in areas of confined or restrained joints. Commentary: In beams, the effect of the shape of the components, including the relative thickness of flanges and web can be significant in determining the forces required to be developed in the various joints in the connection of the beam to the column. Also the joint between the web and flanges of the built-up beams, particularly in the areas of potential plastic hinges, should be designed to be capable of permitting flange buckling without weld failure. In H shaped built-up columns, the welded joints between the web and flanges should be designed to develop the panel zone shears based on the probable location of the plastic hinges. In tubular or box shaped columns, the placing of the plates and the selection of the type of weld connecting webs to flanges is

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important in providing adequate joints to resist the forces in the beam-to-column connection zone. Some testing has been performed Taiwan (Tsai - 1995).

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Chapter 8 - Metallurgy & Welding

8. METALLURGY & WELDING Standard industry specifications for construction materials and processes permit wide variation in strength, toughness and other properties that can be critical to structural performance. This Chapter provides basic information on the variations in properties that occur, practical steps an engineer can take to control critical properties to acceptable levels of tolerance, and the specific instances when such measures may be appropriate. 8.1 Parent Materials 8.1.1 Steels

Designers should specify materials which are readily available for building construction and which will provide suitable ductility and weldability for seismic applications. Structural steels which may be used in the lateral-force-resisting systems for structures designed for seismic resistance without special qualification include those contained in Table 8-1. Refer to the applicable ASTM reference standard for detailed information. Table 8-1 - Structural Steel Prequalified for Use in Seismic Lateral-Force-Resisting Systems ASTM Specification ASTM A36 ASTM A283 Grade D ASTM A500 (Grades B & C) ASTM A501 ASTM A572 (Grades 42 & 50) ASTM A588

Description Carbon Structural Steel Low and Intermediate Tensile Strength Carbon Steel Plates Cold-Formed Welded & Seamless Carbon Steel Structural Tubing in Rounds & Shapes Hot-Formed Welded & Seamless Carbon Steel Structural Tubing High-Strength Low-Alloy Columbium-Vanadium Steels of Structural Quality High-Strength Low-Alloy Structural Steel (weathering steel)

Structural steels which may be used in the lateral-force-resisting systems of structures designed for seismic resistance with special permission of the building official are those listed in Table 8-2. Steel meeting these specifications has not been demonstrated to have adequate weldability or ductility for general purpose application in seismic-force-resisting systems, although it may well possess such characteristics. In order to demonstrate the acceptability of these materials for such use in WSMF construction it is recommended that connections be qualified by test, in accordance with the guidelines of Chapter 7. The test specimens should be fabricated out of the steel using those welding procedures proposed for use in the actual work.

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Table 8-2 - Non-prequalified Structural Steel ASTM Specification Description ASTM A242 High-Strength Low-Alloy Structural Steel ASTM A709 Structural Steel for Bridges ASTM A913 High-Strength Low-Alloy Steel Shapes of Structural Quality, Produced by Quenching & Self-Tempering Process Commentary: Many WSMF structures designed in the last 10 years incorporated ASTM A36 steel for the beams and ASTM A572 grade 50 steel for the columns. This provided an economical way to design structures for the strong column weak beam provisions contained in the building code. Recent studies conducted by the Structural Shape Producers Council (SSPC), however, indicate that material produced to the A36 specification has wide variation in strength properties with actual yield strengths that often exceed 50 ksi. This wide variation makes prediction of connection and frame behavior difficult. Some have postulated that one of the contributing causes to damage experienced in the Northridge Earthquake was inadvertent pairing of overly strong beams with average strength columns. The AISC and SSPC have been working for several years to develop a new specification for structural steel that would have both minimum and maximum yield values defined and provide for a margin between maximum yield and minimum ultimate tensile stress. AISC recently submitted such a specification, for a material with 50 ksi specified yield strength, to ASTM for development into a standard specification. It is anticipated that domestic mills will begin producing structural shapes to this specification within a few years and that eventually, this new material will replace A36 as the standard structural material for incorporation into lateral-force-resisting systems. Under certain circumstances it may be desirable to specify steels that are not recognized under the UBC for use in lateral-force-resisting systems. For instance, ASTM A709 might be specified if the designer wanted to place limits on toughness for fracture-critical applications. In addition, designers may wish to begin incorporating ASTM A913, Grade 65 steel, as well as other higher strength materials, into projects, in order to again be able to economically design for strong column - weak beam conditions. Designers should be aware, however, that these alternative steel materials may not be readily available. It is also important when using such non-prequalified steel materials, that precautions be taken to ensure adequate weldability of the material and that it has sufficient ductility to perform under the severe loadings produced by earthquakes. The cyclic test program recommended by these Interim Guidelines for qualification of connection designs, by test, is believed to be an adequate approach to qualify alternative steel material for such use as well.

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Note that ASTM A709 steel, although not listed in the building code as prequalified for use in lateral-force-resisting systems, actually meets all of the requirements for ASTM A36 and ASTM A572. Consequently, special qualification of the use of this steel should not be required. 8.1.2 Chemistry

ASTM specifications define chemical requirements for each steel. A chemical analysis is performed by the producer on each heat of steel. End product analyses can also be specified on certain products. A certified mill test report is furnished to the customer with the material. The designer should specify that copies of the mill test reports be submitted for his/her conformance review. In general, ASTM specifications for structural steels include maximum limits on carbon, manganese, silicon, phosphorous and sulfur. Ranges and minimums are also limited on other elements in certain steels. Chromium, columbium, copper, molybdenum, nickel and vanadium may be added to enhance strength, toughness, weldability and corrosion resistance. These chemical requirements may vary with the specific product and shape within any given specification. Commentary: Some concern has been expressed with respect to the movement in the steel producing industry of utilizing more recycled steel in its processes. This results in added trace elements not limited by current specifications. Although these have not been shown quantitatively to be detrimental to the performance of welding on the above steels, a new specification for structural steel proposed by AISC does place more control on these trace elements. Mill test reports now include elements not limited in some or all of the specifications. They include copper, columbium, chromium, nickel, molybdenum, silicon and vanadium. The analysis and reporting of an expanded set of elements should be possible, and could be beneficial in the preparation of welding procedure specifications (WPSs) by the welding engineer if critical welding parameters are required. Modern spectrographs used by the mills are capable of automated analyses. When required by the engineer, a request for special supplemental requests should be noted in the contract documents. 8.1.3 Tensile/Elongation Properties

Mechanical property test specimens are taken from rolled shapes or plates at the rolling mill in the manner and location prescribed by ASTM A6 and ASTM A370. Table 8-3 gives the basic mechanical requirements for commonly used structural steels. Properties specified, and controlled by the mills, in current practice include minimum yield strength, ultimate tensile strength and minimum elongation. However, there can be considerable variability in the actual properties of steel meeting these specifications. SSPC, in cooperation with SEAOC, has collected statistical data on the strength characteristics of two grades (ASTM A36 and ASTM A572 Grade 50) of structural steels, based on mill test reports from selected domestic producers for the 1992 production year. Data were also collected for "Dual

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Grade" material that was certified by the producers as complying with both ASTM A36 and ASTM A572 Grade 50. Table 8-4 summarizes these results as well as data provided by a single producer for ASTM A913 material. Table 8-3 - Typical Tensile Requirements for Structural Shapes Minimum Yield Strength, Ksi

ASTM

Ultimate Tensile Strength, Ksi

Minimum Elongation % in 2 inches 212 213 212 213 212 21 21 17

A36 36 58-801 4 A242 42 63 MIN. A572, GR50 50 65 MIN. A588 50 70 MIN. A709, GR36 36 58-80 A709, GR50 50 65 MIN. A913, GR50 50 65 MIN. A913, GR65 65 80 MIN. Notes: 1No maximum for shapes greater than 426 lb./ft. 2Minimum is 19% for shapes greater than 426 lb. /ft. 3Minimum is 18% for shapes greater than 426 lb./ft. 4. Minimum is 50 ksi for Shape Groups 1 and 2, 46 ksi for Shape Group 3

Minimum Elongation % in 8 inches 20 18 18 18 20 18 18 15

Unless special precautions are taken to limit the actual strength of material incorporated into the work to defined levels, new material specified as ASTM A36 should be assumed to be the dual grade for connection demand calculations, whenever the assumption of a higher strength will result in a more conservative design condition. Commentary: The data given in Table 8-4 for A36 and A572 Grade 50 is somewhat weighted by the lighter, Group 1 shapes that will not ordinarily be used in WSMF applications. Excluding Group 1 shapes and combining the Dual Grade and A572 Grade 50 data results in a mean yield strength of 48 ksi for A36 and 57 ksi for A572 Grade 50 steel. It should also be noted that 50% of the material actually incorporated in a project will have yield strengths that exceed these mean values. For the design of facilities with stringent requirements for limiting post-earthquake damage, consideration of more conservative estimates of the actual yield strength may be warranted. In wide flange sections the tensile test coupons are currently taken from the web. The amount of reduction rolling, finish rolling temperatures and cooling conditions affect the tensile and impact properties in different areas of the member. Typically, the web exhibits about five percent higher strength than the flanges due to faster cooling. Table 8-4 - Statistics for Structural Shapes1 Statistic

A 36

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Dual Grade GRADE

A572 GR50

A913 GR65

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Chapter 8 - Metallurgy & Welding

A 36

Dual Grade GRADE

A572 GR50

A913 GR65

Mean Minimum Maximum Standard Deviation [ s ] Mean + 1 s

Yield Point (ksi) 49.2 36.0 72.4 4.9 54.1

55.2 50.0 71.1 3.7 58.9

57.6 50.0 79.5 5.1 62.7

75.3 68.2 84.1 4.0 79.3

Mean Minimum Maximum Standard Deviation [ s ] Mean + 1 s

Tensile Strength (ksi) 68.5 58.0 88.5 4.6 73.1

73.2 65.0 80.0 3.3 76.5

75.6 65.0 104.0 6.2 81.8

89.7 83.4 99.6 3.5 93.2

Mean Minimum Maximum Standard Deviation [ s ] Mean + 1 s Mean - 1 s

Yield/Tensile Ratio 0.72 0.51 0.93 0.06 0.78 0.66

0.75 0.65 0.92 0.04 0.79 0.71

0.76 0.62 0.95 0.05 0.81 0.71

0.84 0.75 0.90 0.03 0.87 0.81

1: The data presented for ASTM A36, “Dual Grade” and ASTM A572 Grade 50 were included as part of the SSPC study (SSPC-1994). The data for ASTM A913 were derived from a single producer and may not be available from all producers.

Design professionals should be aware of the variation in actual properties permitted by the ASTM specifications. This is especially important for yield strength. Yield strengths for ASTM A36 material have consistently increased over the last 15 years so that several grades of steel may have the same properties or reversed properties, with respect to beams and columns, from those the designer intended. Investigations of structures damaged by the Northridge earthquake found some WSMF connections in which beam yield strength exceeded column yield strength despite the opposite intent of the designer. As an example of the variations which can be found, Table 8-5 presents the variation in material properties found within a single building affected by the Northridge earthquake. Properties shown include measured yield strength (Fya,), measured tensile strength (Fua ) and Charpy V-Notch energy rating (CVN).

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Table 8-5 - Sample Steel Properties from a Building Affected by the Northridge Earthquake Shape

Fya1 ksi

Fua, ksi

CVN, ft-lb.

W36 X 182

38.0

69.3

18

W36 X 230

49.3

71.7

195

Note 1 - ASTM A36 material was specified for both structures.

The practice of dual certification of A36 and A572, Grade 50 can result in mean yield strengths that are fifty percent higher than the specified yield of A36. Since there is no practical way to discern whether dual grade steel will be supplied, unless direct purchase of steel from specific suppliers is made, in the absence of such procurement practices, the prudent action for determining connection requirements, where higher strengths could be detrimental to the design, would be to assume the dual grade material whenever A36 or A572 Grade 50 is specified. 8.1.4 Toughness Properties

For critical connections, non-redundant components and unusual or difficult geometries involving Group 3 (with flanges 11/2 inches or thicker) 4 and 5 shapes and plates and built-up sections over two inches thick with welded connections, the designer should consider specifying toughness requirements on the parent materials. A Charpy V-Notch (CVN) value of 20 ft.-lb. at 70 degrees F. should be specified when toughness is deemed necessary for an application. Refer to Figure 8-1 for typical CVN test specimen locations. The impact test should be conducted in accordance with ASTM A673, frequency H, with the following exceptions: a) The center longitudinal axis of the specimens should be located as near as practicable to midway between the inner flange surface and the center of the flange thickness at the intersection of the web mid-thickness. Refer to AISC LRFD specification, Section A3-1c, Heavy Shapes (American Institute of Steel Construction - 1993) b) Tests should be conducted by the producer on material selected from a location representing the top of each ingot or part of an ingot used to produce the product represented by these tests. For the continuous casting process, the sample may be taken at random.

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bf/2 bf/3

CL

tf/2

Typical CVN Specimen ASTM A673

CVN Specimen AISC- LRFD A3-1c

Figure 8-1 - Standard Locations for Charpy V-Notch Specimen Extraction, Longitudinal Only Commentary: Many variables are recognized in analyzing the metallurgy of WSMF members. Until more research is available on the through-thickness properties of members thicker than two inches, a conservative approach is indicated. Specifying toughness properties in critical, unusual or non-redundant connections should be considered. As temperature decreases or strain rate increases, toughness properties decrease. Charpy V-notch impact (CVN) tests, pre-cracked CVN tests and other fracture toughness tests can identify the nil ductility temperature (NDT) - the temperature below which a material loses all ductility and fractures in a brittle manner. On a microscopic level, this equates to a change in the fracture mechanism from shear to cleavage. Fracture that occurs by cleavage at a nominal tensile stress below yield is referred to as a brittle fracture. A brittle fracture can occur in structural steel when a particular combination of low temperature, tensile stress, high strain rate and a metallurgical or mechanical notch is present. Plastic deformation can only occur through shear stress. Shear stress is generated when uniaxial or bi-axial straining occurs. In tri-axial stress states, the maximum shear stress approaches zero as the principal stresses increase. When these stresses approach equality, a cleavage failure can occur. Welding and other sources of residual stresses can set up a state of tri-axial stress leading to brittle fractures. The necessity for minimum toughness requirements is not agreed to by all. There is also disagreement as to how much toughness should be required. The

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AWS Presidential Task Group recommended toughness values of 15 ft-lb. at different temperatures, depending on the anticipated service conditions. A temperature of 70 degrees F was recommended for enclosed structures and 40 degrees F for exposed structures. The 1993 AISC LRFD Specification, Section A3-1c, Heavy Shapes, requires toughness testing [Charpy V-Notch] under the following conditions for Group 4 and 5 shapes and plates exceeding 2 inches in thickness: a) When spliced using complete joint penetration welds; b) when complete joint penetration welds through the thickness are used in connections subjected to primary tensile stress due to tension or flexure of such members.” Where toughness is required, the minimum value should be 20 ft-lb. at 70°F. Plates thicker than two inches and sections with flanges thicker than two inches can be expected to have significantly variable grain sizes across the section. The slower cooling rate of the web-flange intersection in thick sections produces a larger grain size which exhibits less ductility and notch toughness. ANSI/ASTM A673 and A370 establish the procedure for longitudinal Charpy V-notch testing. The impact properties of steel can vary within the same heat and piece, be it as-rolled, controlled rolled, or heat treated. Normalizing or quenching and tempering will reduce the degree of variation. Three specimens are taken from a single test coupon or location. The average must exceed the specified minimum, but one value may be less than the specified minimum but must be greater than the larger of two thirds of the specified minimum or 5 ft-lb. The longitudinal axis of the specimen is parallel to the longitudinal axis of the shape or final rolling direction for plate. For shapes, the specimen is taken from the flange 1/3 the distance from the edge of the flange to the web. The frequency of testing [heat or piece], the test temperature, and the absorbed energy are specified by the user. [NOTE: heat testing (frequency H) for shapes, means one CVN test set of samples from at least each 50 tons of the same shape size, excluding length, from each heat in the as-rolled condition. Piece testing (frequency P) for shapes, means one CVN test set of specimens from at least each 15 tons or each single length of 15 tons of the same shape size, excluding length, from each heat in the as-rolled condition.] Heat testing is probably adequate in most circumstances. The specimen location required by ASTM A673 is not at the least tough part of a W shape. For a W shape, the volume at the flange web intersection has the lowest ratio of surface area to volume and hence cools the slowest. This slow cooling causes grain growth and reduced toughness. The finer the grain, the tougher the material. Also, ASTM A673 does not specify where in the product run of an ingot to sample. Impurities tend to rise to the upper portion of the ingot during cooling from molten metal. Impurities reduce the toughness of the finished metal. Hence, shapes produced from the upper portions of an ingot can be expected to have lower toughness, and samples should be taken from shapes

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produced from this portion of the ingot. In the continuous casting process, impurities tend to be more evenly distributed; hence, samples taken anywhere should suffice. The AISC LRFD specification requires testing from the upper portion of the ingot and near the web flange intersection. Even though the AISC LRFD specification does not require toughness testing for the typical WSMF connection, i.e., a Group 2 beam to a Group 4 column, it appears that there may be inadequate through thickness toughness in the Group 4 and 5 column flanges. In response to concerns raised following the Northridge Earthquake, the AISC conducted a statistical survey of the toughness of material produced in structural shapes, based on data provided by six producers for a production period of approximately one year (American Institute of Steel Construction - 1995). This survey showed a mean value of Charpy V notch toughness for all shape groups that was well in excess of 20 ft-lb. at 70 degrees F. However, not all of the samples upon which these data are based were taken from the core area, recommended by these Interim Guidelines. Consequently, this survey does not provide definitive information on the extent to which standard material produced by the mills participating in this survey will meet the recommended values. 8.1.5 Lamellar Discontinuities

For critical joints (beam to column CJP welds or other tension applications where Z-axis or tri-axial stress states exist), ultrasonic testing (UT) should be specified for the member loaded in the Z axis direction, in the area of the connection. A distance 3 inches above and below the location to be welded to the girder flange is recommended. The test procedure and acceptance criteria given in ASTM A898-91, Standard Specification for Straight Beam Ultrasonic Examination of Rolled Steel Structural Shapes, Level I, should be applied. This testing should be done in the mill or fabrication shop for new construction. For repair welding, the same procedure should be applied in the field, as access permits. Commentary: Very little test data exist on the through thickness properties of structural shapes nor are there any standard test methods for determining these properties. Nevertheless, the typical beam-column joints in WSMFs rely heavily on the through-thickness properties of column flanges. Some of the proposed strengthening and reinforcing solutions will transmit even more forces into the Z axis of the column flanges. Laminations (pre-existing planes of weakness) and lamellar tearing (cracks parallel to the surface) will impair the Z axis strength and toughness properties. These defects are mainly caused by non-metallic sulfides and oxides which begin as almost spherical in shape, and become elongated in the rolling process. When Z axis loading occurs from weld shrinkage strains or external loading, microscopic cracks may form between the discrete, elongated nonmetallic inclusions. As they link up, lamellar tearing occurs.

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Longitudinal wave ultrasonic testing is very effective in mapping serious lamellar discontinuities. Improved quality steel does not eliminate weld shrinkage and, by itself, will not necessarily avoid lamellar tearing in highly restrained joints. Ultrasonic testing should not be specified without due regard for design and fabrication considerations. In cases where lamellar defects or tearing are discovered in erection or on existing buildings, the designer must consider the consequences of making repairs to these areas. Gouging and repair welding will add additional cycles of weld shrinkage to the connection and may promote crack extensions or new lamellar tearing. When secondary cracking is discovered, a welding engineer should be consulted to generate a special WPS for the repair. 8.2 Welding 8.2.1 Welding Process

The welding process to be used to execute the joint weld [e.g. shielded metal (SMAW), flux cored (FCAW), submerged (SAW), gas metal arc weld (GMAW), or electroslag (requires qualification of the welding procedure specification)] should be specified in the Contract Documents for weld repairs. Contract documents for new construction should state any restrictions on weld parameters or processes. Most pre-Northridge production welding was executed using FCAW using a self-shielding process (FCAW-SS). Shielded metal arc welding (stick welding) is often used for damage repairs, in tight conditions and in some shop applications. Commentary: At this time there is no clear evidence that one method can produce uniformly superior welds although poor welds can be produced with any of the methods. 8.2.2 Welding Procedures

Welding should be performed within the parameters established by the electrode manufacturer and the Welding Procedure Specification (WPS), required under AWS D1.1. Commentary: For example, the position (if applicable), electrode diameter, amperage or wire feed speed range, voltage range, travel speed range and electrode stickout (e.g. all passes, 0.072 in. diameter, 248 to 302 amps, 19 to 23 volts, 6 to 10 inches/minute travel speed, 170 to 245 inches/minute wire feed speed, 1/2" to 1" electrode stickout) should be established. This information is generally submitted by the fabricator as part of the Welding Procedure Specification. Its importance in producing a high quality weld is essential. The following information is presented to help the engineer understand some of the issues surrounding these parameters.

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The amperage, voltage, travel speed, electrical stickout and wire feed speed are functions of each electrode. If prequalified WPSs are utilized, these parameters must be in compliance with the AWS D1.1 requirements. For FCAW and SMAW, the parameters required for an individual electrode vary from manufacturer to manufacturer. Therefore, for these processes, it is essential that the fabricator/erector utilize parameters that are within the range of recommended operation published by the filler metal manufacturer. Alternately, the fabricator/erector could qualify the welding procedure by test in accordance with the provisions of AWS D1.1 and base the WPS parameters on the test results. For submerged arc welding, the AWS D1.1 code provides specific amperage limitations since the solid steel electrodes used by this process operate essentially the same regardless of manufacture. The filler metal manufacturer’s guideline should supply data on amperage or wire feed speed, voltage, polarity, and electrical stickout. The guidelines will not, however, include information on travel speed which is a function of the joint detail. The contractor should select a balanced combination of parameters, including travel speed, that will ensure that the code mandated weld-bead sizes (width and height) are not exceeded. 8.2.3 Welding Filler Metals

The current AWS D1.1 requirements should be incorporated as written in the Code. The welding parameters should be clearly specified using a combination of the Project Specifications, the Project Drawings, the Shop Drawings and the welding procedure specifications, as required by AWS D1.1. For welding on ASTM A572 steel, the AWS D1.1 code requires the use of low-hydrogen electrodes. With SMAW welding, a variety of non-low hydrogen electrodes are commercially available. These electrodes are not appropriate for welding on the higher strength steels used in building construction today, although they were popular in the past when lower strength steels were employed. All of the electrodes that are employed for flux cored arc welding (both gas shielded and self-shielded), as well as submerged arc welding, are considered low hydrogen. For critical joints (beam to column CJP welds or other tension applications where Z-axis loading or tri-axial stress states exist), toughness requirements for the filler metals should be specified. A minimum CVN value of 20 ft.-lb. at a temperature of 0 degrees F. should be required, unless more stringent requirements are indicated by the service conditions and/or the Contract Documents. The filler metal should be tested in accordance with the AWS A5 filler metal specification to ensure it is capable of achieving this level of notch toughness. The filler metal manufacturers Typical Certificate of Conformance, or a suitably documented test performed by the contractor, should be used to document the suitability of the electrode used. These tests should be performed for each filler metal by AWS classification, filler metal manufacturer and filler metal manufacturer’s trade name. The sizes as specified by the AWS A5 document should be tested, although the exact diameter used in production need not be specifically tested. This requirement should not be construed to imply lot or heat testing of filler metals.

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Electrode specification sheets should be provided by the Fabricator/Erector prior to commencing fabrication/erection. Commentary: Currently, there are no notch toughness requirements for weld metal used in welding ASTM A 36 or A 572, Grade 50, steel in AWS D1.1. This topic has been extensively discussed by the Welding Group at the Joint SAC/AISC/AISI/NIST Invitational Workshop on September 8 and 9, 1994, and by all participants of the SAC Invitational Workshop on October 28 and 29, 1994. The topic was also considered by the AWS Presidential Task Group, which decided that additional research was required to determine the need for toughness in weld metal. There is general agreement that adding a toughness requirement for filler metal would be desirable and easily achievable. Most filler metals are fairly tough, but some will not achieve even a modest requirement such as 5 ft-lb. at + 70°F. What is not in unanimous agreement is what level of toughness should be required. The recommendation from the Joint Workshop was 20 ft-lb. at -20°F per Charpy V-Notch [CVN] testing. The recommendation from the SAC Workshop was 20 ft-lb. at 30°F lower than the Lowest Ambient Service Temperature (LAST) and not above 0°F. The AWS Presidential Task Group provided an interim recommendation for different toughness values depending on the climatic zone, referenced to ASTM A709. Specifically, the recommendation was for 20 ft-lb. at temperatures of 70 degrees F for Zone 1, 40 degrees F for Zone 2, and 10 degrees F for Zone 3. The AWS also suggested toughness values for base metals used in these applications. Some fractured surfaces in the Northridge and Kobe Earthquakes revealed evidence of improper use of electrodes and welding procedures. Prominent among the misuses were high production deposition rates. Pass widths of up to 11/2 inches and pass heights of 1/2 inch were common. The kind of heat input associated with such large passes promotes grain growth in the HAZ and attendant low notch toughness. Root gap, access capability, electrode diameter, stick-out, pass thickness, pass width, travel speed, wire feed rate, current and voltage were found to be the significant problems in evaluation of welds in buildings affected by the Northridge earthquake. Welding electrodes for common welding processes include: AWS A5.20: AWS A5.29: AWS A5.1: AWS A5.5: AWS A5.17: AWS A5.23: AWS A5.25:

Carbon Steel Electrodes for FCAW Low Alloy Steel Electrodes for FCAW Carbon Steel Electrodes for SMAW Low Alloy Steel Covered Arc Welding Electrodes (for SMAW) Carbon Steel Electrodes and Fluxes for SAW Low Alloy Steel Electrodes and Fluxes for SAW Carbon and Low Alloy Steel Electrodes and Fluxes for Electroslag Welding

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In flux cored arc welding, one would expect the use of electrodes that meet either AWS A5.20 or AWS A5.29 provided they meet the toughness requirements specified below. Except to the extent that one requires Charpy V-Notch toughness and minimum yield strength, the filler metal classification is typically selected by the Fabricator. Compatibility between different filler metals must be confirmed by the Fabricator, particularly when SMAW and FCAW-SS processes are mixed. Generally speaking, SMAW-type filler metals may not be applied to FCAW-SS type filler metals (e.g. when a weld has been partially removed) while FCAW-type filler metals may be applied to SMAW-type filler metals. This recommendation considers the use of aluminum as a killing agent in FCAW-SS electrodes that can be incorporated into the SMAW filler metal with a reduction in impact toughness properties. As an aid to the engineer, the following interpretation of filler metal classifications is provided below: E1X2X3T4X5 For electrodes specified under AWS A5.20 1 2 3 4 5 6 E X X T X X For electrodes specified under AWS A5.29 E1XX7X8X9X10 For electrodes specified under AWS A5.1 or AWS A5.5. NOTES: 1.

Indicates an electrode.

2.

Indicates minimum tensile strength of deposited weld metal (in tens of ksi, e.g., 7 = 70 ksi).

3.

Indicates primary welding position for which the electrode is designed (0 = flat and horizontal and 1 = all positions).

4.

Indicates a flux cored electrode. Absence of a letter indicates a "stick" electrode for SMAW.

5.

Describes usability and performance capabilities. For our purposes, it conveys whether or not Charpy V-Notch toughness is required (1, 5, 6 and 8 have impact strength requirements while 2, 4, 7, 10 and 11 do not). A "G" signifies that the properties are not defined by AWS and are to be agreed upon between the manufacturer and the specifier. Impact strength is specified in terms of the number of foot-pounds at a given temperature (e.g., 20 ft-lb. at 0 degrees F). Note that for electrodes specified under AWS A5.20, the format for usage is "T-X".

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6.

Designates the chemical composition of deposited metal for electrodes specified under AWS A5.29. Note that there is no equivalent format for chemical composition for electrodes specified under AWS A5.20.

7.

The first two digits (or three digits in a five digit number) designate the minimum tensile strength in ksi.

8.

The third digit (or fourth digit in a five digit number) indicates the primary welding position for which the electrode is designed (1 = all positions, 2 = flat position and fillet welds in the horizontal position, 4 = vertical welding with downward progression and for other positions.)

9.

The last two digits, taken together, indicate the type of current with which the electrode can be used and the type of covering on the electrode.

10.

Indicates a suffix (e.g., A1, A2, B1, etc.) designating the chemical composition of the deposited metal.

Electrode Diameter: (See AWS D1.1 Section 4.14.1.2) The issue of maximum electrode diameter has not been studied sufficiently to determine whether or not electrode diameter is a critical variable. Recent tests have produced modified frame joints with acceptable test results using the previous standard-of-practice 0.120 in. diameter wire. The use of smaller diameter electrodes will slow the rate of deposition (as measured by volume) but will not, in and of itself, produce an acceptable weld. The following lists the maximum allowable electrode diameters for prequalified FCAW WPS’s according to D1.1: • • • • •

Horizontal, complete or partial penetration welds: 1/8 inch (0.125")* Vertical, complete or partial penetration welds: 5/64 inch (0.078") Horizontal, fillet welds: 1/8 inch (0.125") Vertical, fillet welds: 5/64 inch (0.078") Overhead, reinforcing fillet welds: 5/64 inch (0.078") * This value is not part of D1.1-94, but will be part of D1.1-96.

For a given electrode diameter, there is an optimum range of weld bead sizes that may be deposited. Weld bead sizes that are outside the acceptable size range (either too large or too small) may result in unacceptable weld quality. The D1.1 code controls both maximum electrode diameters and maximum bead sizes (width and thickness). Prequalified WPS’s are required to meet these code requirements. Further restrictions on suitable electrode diameters are not recommended.

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8.2.4 Preheat and Interpass Temperatures

The preheat temperatures and conditions given in AWS D1.1, Chapter 4 should be strictly observed with special attention given to Section 4.2, for the thickness of metal to be welded. For repair welding of earthquake damage, the AASHTO/AWS D1.5 Bridge Welding Code preheat requirements for fracture-critical, non-redundant applications should be considered. Cracking of welds and heat affected zones should be avoided. One type of weld cracking is hydrogen induced cracking (HIC). For a given steel, variables that reduce HIC tendencies are prioritized as follows: 1. Lower levels of hydrogen. 2. Higher preheat and interpass temperatures. 3. Postheat. 4. Retarded cooling (insulating blankets). Only low hydrogen electrodes should be used for fabrication and/or repair of seismically loaded structures. Proper preheat and interpass temperatures should be maintained. AWS D1.1 requirements are generally adequate for new construction. Highly restrained repair welds may require higher preheat levels. Control of hydrogen and proper preheat and interpass temperature is much more powerful for overcoming HIC than postheat or retarded cooling methods. Retarded cooling has limited benefit if the entire piece is not preheated - obviously impractical for structural applications. The engineer is encouraged to emphasize proper preheat and the use of low hydrogen electrodes and practice. If these measures are insufficient to prevent cracking, a welding engineer should be consulted to determine appropriate measures that should be incorporated to eliminate cracking. These measures may or may not call for additional preheat, postheat, or retarded cooling. While low hydrogen electrodes and proper preheat is essential, postheat and retarded cooing is not generally required and should not be used for routine construction or repair. Commentary: There are two primary purposes for preheating and interpass temperature requirements: (1) To drive off any surface moisture or condensation which may be present on the steel so as to lessen the possibility of hydrogen being introduced into the weld metal and HAZ, and (2) To prevent the steel mass surrounding the weld from quenching the HAZ as cooling occurs after welding.

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Chapter 8 - Metallurgy & Welding

Virtually all weld repairs are made under conditions of high restraint. Consequently, higher preheat/interpass temperatures may be required for repair applications. As steel is cooled from the austenitic range (above about 1330 degrees F), it goes through a critical transition temperature. If it goes through that temperature range too fast, a hard, brittle phase called martensite forms (quenching). If it passes through that temperature range at a slower rate, ductile, tougher phases called bainite or ferrite/pearlite form. Preheating of the surrounding mass provides a slower cooling rate for the weld metal and HAZ. The American Association of State Highway and Transportation Officials (AASHTO) recognizes repair welding as more critical in its guidelines for the repair of fracture-critical bridge members. The purpose, in part, is to allow more plastic flow and yielding, at welding temperatures, in the area near the weld. The requirements are given in Table 8-6: Table 8-6 - AASHTO Preheat Requirements for Fracture Critical Repairs1 Steel Thickness, in. Minimum Preheat/Interpass Temp., °F A36/A572 to 1-1/2 325 A36/A572 >1-1/2 375 1- Reference AASHTO/AWS D1.5-95 Bridge Welding Code Preheat temperatures should be measured at a distance from the weld equal to the thickness of the part being welded, but not less than three inches, in any direction including the through thickness of the piece. Where plates are of different thicknesses, the pre-heat requirement for the thicker plate should govern. Maintenance of these temperatures through the execution of the weld (i.e. the interpass temperature) is essential. Maximum interpass temperatures should be limited to 550 degrees F for prequalified WPSs, for fracture-critical applications. Higher interpass temperatures could be employed if those higher temperature limits are qualified by test. 8.2.5 Postheat

Postheat is the application of heat in the 400 degrees F to 600 degrees F range after completion of welding. It may be helpful in mitigating some cracking tendencies. Commentary: A postheat specification might require that complete joint penetration groove welds in existing buildings be postheated at 450 degrees F for two hours. The purpose of this postheat is to accelerate the removal of hydrogen from the weld metal and HAZ and reduce the probability of cracking due to hydrogen embrittlement. Hydrogen will migrate within the weld metal at approximately 1 inch per hour at 450 degrees F, and at about 1 inch per month at 70 degrees F. To the extent that hydrogen embrittlement is of concern, postheat

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is one method of mitigating cracking. The use of low hydrogen electrodes, proper welding procedures, and uniformly applied and maintained preheat may represent a cost-effective method of addressing the problem of hydrogen embrittlement in lieu of postheat. When postheat is required, AASHTO/AWS D1.5-95 specifications require this to be done immediately upon completion of welding. The postheat is between 400 to 500 degrees F for one hour minimum, for each inch of the thickest member or for two hours, whichever is less. 8.2.6 Controlled Cooling

Most of the weldment cooling is effected by conductance within the steel rather than radiation. Retarded cooling should only be specified in cases where large weldments subject to significant residual stresses due to restraint (e.g. multiple members framing into one connection with Z axis loading) or ambient temperatures that would result in rapid cooling of large weldments. The length of time to cool down the weld and the level of insulation required are a function of weldment temperature, thickness of base metal and ambient temperature. Commentary: Active systems of ramp-down cooling are generally not required; however, in highly restrained conditions they may offer an added advantage. 8.2.7 Metallurgical Stress Risers

Metallurgical discontinuities such as tack welds, air-arc gouging and flame cutting without preheating or incorporation into the final weld should not be permitted. Inadvertent damage of this type should be repaired by methods approved by the engineer, following the AWS D1.1 criteria and a specific WPS covering repairs of this type. Commentary: Metallurgical stress risers may result from tack welds, air-arc gouging and flame cutting performed without adequate preheat. However, preheating is not necessarily required for air arc gouging or flame cutting used in the preparation of a surface to receive later welding. The subsequent heat input during the welding process should adequately anneal the affected area. The AWS D1.1 code requires the same preheating for tack welding operations as normal welding, with the exception of tack welds that are incorporated into subsequent submerged arc weld deposits. Arc strikes can also be a source of metallurgical stress risers and should not be indiscriminately made. AWS D1.1 Section 3.10 indicates that “arc strikes outside the area of permanent welds should be avoided on any base metal. Cracks or blemishes caused by arc strikes should be ground to a smooth contour and checked to ensure soundness.”

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Chapter 8 - Metallurgy & Welding

8.2.8 Welding Preparation & Fit-up

Any cracked columns, welds, or beam flanges should be prepared to receive the welding contemplated by the engineer. AWS D1.1 provides guidance on the precise nature of the fit-up requirements and tolerances.

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Chapter 9 - Quality Control/Quality Assurance

9. QUALITY CONTROL/QUALITY ASSURANCE Quality control is principally the responsibility of the contractor, while Quality Assurance is performed at the prerogative of the owner and as mandated by the Building Code. Key parts of the Quality Control program include assuring that all parties understand what is to be constructed, and the standards that apply. All workers and inspectors should be adequately qualified to perform the required work, and should have written procedures, approved by the engineer, for the work that is to be performed. 9.1 Quality Control Fabrication/erection inspection and testing should be the responsibility of the contractor, unless otherwise provided for in the Contract Documents. 9.1.1 General

A pre-job meeting or series of meetings should be held with the owner's representatives, the engineer, the Fabricator/Erector's production and QC personnel to plan and discuss the project and fabrication procedures. Welders and welding operators should also be involved at some level, either by a meeting or direct dissemination of the information. Fabrication/erection inspection and testing should be performed prior to assembly, during assembly, fit-up, tacking, welding and after welding to ensure that the materials and workmanship meet the requirements of the Contract Documents. The fitters and welders should have the applicable WPS document and drawings for each connection and joint at their assembly station. 9.1.2 Inspector Qualification

Inspectors responsible for acceptance or rejection of materials and workmanship should be qualified in accordance with Sections 10 and 11 of these Guidelines. The engineer should have the authority and duty to verify the qualifications of the inspectors. 9.1.3 Duties

The inspector should ascertain that all materials comply with the Contract Documents, either by mill certifications or testing. The inspector should verify that all fabrication and erection welding is performed in accordance with the Contract Documents. Detailed duties are further described in Section 10 of these Guidelines. 9.1.4 Records

The QC inspector should insure that each welder has a unique identification mark or die stamp to identify his or her welds. The inspector should also mark the welds/parts/ joints that have been inspected and accepted with a distinguishing mark or die stamp, or alternatively, maintain records

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indicating the specific welds inspected by each person. The NDT technician should use the weld identification system given in AWS D1.1, Sections 6.19.1 and 6.19.2. The inspector should keep a record of all welders, welding operators and tack welders; all procedure and operator qualifications; all accepted parts; the status of all rejected joints; NDT test reports; and other such information as may be required. 9.1.5 Engineer Obligations

The structural engineer or designated welding engineer should perform a review of the Fabricator/Erector’s Quality Control program, equipment condition, and availability of equipment and qualified personnel. The review should include the following: a)

Interview with Fabricator/Erector’s designated Quality Control personnel.

b)

Review of Fabricator/Erector’s written quality procedure manual.

c)

Review of Fabricator/Erector’s Procedure Qualification Records (PQR’s) and WPS applicable to the specific project.

d)

Review of Welder Performance Records.

e)

Review of the Fabricator/Erector’s NDT procedures, equipment calibration records, and personnel training records. Alternatively, the Fabricator/Erector may contract with an outside Quality Control company for NDT services; however, this should not take the place of the owner’s QA responsibility for NDT.

f)

Designate any specific NDT requirements which apply to the project and which are beyond those required by the Code.

g)

A meeting with the owner’s representative, fabricator/erector’s Quality Control personnel and the welder, to review the WPS.

9.1.6 Contractor Obligations

The contractor should make available to the inspector and NDT Technician all drawings, project specifications, mill certifications, welder qualifications, WPSs and PQRs applicable to the project. The contractor should cooperate fully with requests from inspection and testing personnel for access to the connections and joints to be inspected or tested. This includes beam and column turning in the shop, weld backing removal and access platforms or scaffolding as required to perform the work safely. The contractor should be responsible for all necessary corrections of deficiencies in materials and workmanship. The contractor should comply with all requests of the inspector to correct deficiencies. The NDT Technician should be apprised of any repairs made by the contractor. Inspections should be performed in a timely manner. Disputes should be resolved by the structural engineer of record, or by a welding engineer.

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9.1.7 Extent of Testing

Information furnished to the bidding contractors should clearly identify the extent of inspection and testing to be performed by the contractor. Weld joints requiring testing by Contract Documents shall be tested for their full length, unless partial or spot testing is specified. When partial or spot testing is specified, the location and lengths of welds or categories of weld to be tested should be clearly designated in the Contract Documents. Each spot test should cover at least 4 inches of the weld length. When spot testing reveals indications of rejectable discontinuities that require repair, the extent of those discontinuities should be explored. Two additional spots in the same segment of weld joint should be taken at locations away from the original spot. When either of the two additional spots show defects that require repair, the entire segment of weld represented by the original weld should be completely tested. Where spot testing or percentage sampling is specified on certain welds, the contract drawings and shop drawings should so state using NDT symbols in conjunction with the welding symbols. On projects where a sliding sampling scale is specified, based on the UT reject level of individual welders, the inspector should keep records on each welder or welding operator. These records will be used as a basis for sampling rate reduction. Commentary: AWS D1.1 uses the term "Fabrication/Erection Inspection" synonymously with the classical "Quality Control" function of other industries. A basic premise of Quality Control is to have the production, engineering and Quality Control departments independent of one another. The contractor should be responsible for establishing the Quality Control program and for in-progress Quality Control of work. Part of this effort is to require that welders meet established minimum requirements. Execution of critical welds requires skilled welders who will follow the project welding requirements. An important part of any Quality Control program is assuring that the workers have the appropriate qualifications to perform the work. Welds executed by welders who do not satisfy the welder performance qualifications should be considered rejectable. Important aspects of a QC program should include as a minimum: 1.

Welders shall be qualified for the work they will be doing per AWS D1.1, Section 5, Part C.

2.

The qualifications of each welder should be certified by an appropriate authority and verified by the contractor and Special Inspector. The engineer should establish whether there are certifications from selected jurisdictions that will or will not be accepted as acceptable substitutions.

The Quality Control function of the contractor should be isolated from the production department and the QC Manager should report directly to a high level company officer to avoid conflicts of interest with production.

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9.2 Quality Assurance & Special Inspection Verification inspection and testing should be the responsibility of the owner and/or the engineer unless otherwise provided for in the Contract Documents. The Quality Assurance designate should act for and in behalf of the owner or engineer on all inspection, NDT and quality matters that are within the scope of the Contract Documents. 9.2.1 General

Verification inspection and testing are the prerogatives of the owner who may perform this function or, when provided for in the Contract Documents, waive independent verification, or stipulate that both inspection and verification shall be performed by the contractor. In municipalities that have adopted the UBC, verification inspection and testing is mandated for structural welding, and is designated as “Special Inspection.” The QA inspector should be included in the pre-job meetings for fabrication and erection discussions referenced in 9.1.1. Fabrication/erection verification inspection and testing should be performed concurrently with the Quality Control inspection and testing to ensure that the contractor's QC program is meeting the requirements of the Contract Documents. The QA inspector should ensure that the fitters and welders have the applicable WPS document and required information for each connection and joint at their assembly station. 9.2.2 Inspector Qualification

Inspectors responsible for acceptance or rejection of materials and workmanship should be qualified in accordance with Chapters 10 and 11 of these Interim Guidelines. The engineer should have the authority and duty to verify the qualifications of the inspectors. The inspector may use assistants who are formally designated, aware of their assigned responsibility and the acceptance criteria, and work under the direct supervision and monitoring of a qualified inspector. 9.2.3 Duties

The QA inspector should verify the qualifications of the QC inspectors and the NDT technicians. The inspector should verify that the mill certifications for all materials are being checked by the QC inspector and that they comply with the Contract Documents. The inspector should verify that all fabrication and erection welding is performed in accordance with the Contract Documents. Detailed duties are further described in Chapter 10 of these Interim Guidelines. 9.2.4 Records

The inspector should ensure that each welder, NDT technician and QC inspector has a unique identification mark or die stamp to identify his or her welds/weld tests/weld inspections. The QA Inspector should ensure that the QA and NDT personnel are keeping the proper records of all welders, welding operators and tack welders; all procedure and operator qualifications; all 9-4

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accepted parts; the status of all rejected joints; NDT test reports; and other such information as may be required. 9.2.5 Engineer Obligations

The structural engineer or designated welding engineer should perform a complete review of the QA Agency. This review should encompass personnel qualification, written procedures manual, and availability of equipment and qualified personnel. The Agency should employ an American Society for Nondestructive Testing (ASNT) Level III qualified person who oversees equipment calibration and personnel certification and training for the project on a full time basis. Reviews should be performed in a timely manner. Disputes should be resolved by a qualified welding engineer. 9.2.6 Contractor Obligations

The contractor should make available to the QA Inspector and QA NDT Technician (if applicable) all drawings, project specifications, mill certifications, welder qualifications, WPSs and PQRs applicable to the project. The contractor should cooperate fully with requests from inspection and testing personnel for access to the connections and joints to be inspected or tested. This includes beam and column turning in the shop, weld backing removal and access platforms or scaffolding as required to perform the work safely. The contractor should be responsible for all necessary corrections of deficiencies in materials and workmanship. The contractor should comply with all requests of the QA Inspector to correct deficiencies. The QA NDT Technician should be apprised of any repairs made by the contractor. 9.2.7 Extent of QA Testing

The QA representative may perform independent inspecting and testing to the extent established in the contract documents. When conditions exist that make further testing advisable, the QA representative, with the concurrence of the structural engineer of record, may perform additional independent inspection and testing, to the degree his/her judgment suggests as appropriate. Acceptance criteria should be mutually agreeable to the inspector and contractor. Discrepancies between the QC and QA decisions should be resolved by the engineer. Commentary: AWS D1.1 uses the term "Verification Inspection" synonymously with the "Quality Assurance" function of other industries. The purpose of QA programs is to provide an oversight to the contractor's QC program. This may range from simple records/report reviews to a full testing and inspection program, depending on the effectiveness of the Fabricator/Erector's QC program, and the requirements of the building code. Often this cannot be established until the contractor is selected. The owner must ensure that an adequate Quality Control program is in place, and is responsible for the Quality Assurance program. The use of “licensed” or

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“qualified” fabrication shops in lieu of requiring independent Quality Assurance provided by the owner is not recommended. However, a fabrication shop that is licensed or qualified by a recognized program, such as the AISC Quality Certification Program, does provide a minimum assurance of capability of good performance. The owner is responsible for establishing the Quality Assurance program. Elements in an acceptable Quality Assurance program should conform to those required by the UBC. Since most owners have little expertise or knowledge related to construction, this often means that the engineer must advise the owner, and, in many cases, establish the program. Example Quality Assurance requirements might include the following: 1.

The lead welding inspector should be a Certified Welding Inspector (CWI) per AWSQC-1 Standards, and, where applicable, should be certified by the responsible jurisdiction as a qualified inspector for structural steel welding. Other welding inspectors performing visual inspection under the supervision of the lead welding inspector should hold an active and appropriate certification. Not more than four non-CWIs should be under the supervision of a CWI.

2.

All welding should be inspected visually as required by AWS D1.1 (See AWS D1.1 Section 8.15.1).

3.

All complete and partial joint penetration welds should be inspected ultrasonically as required by AWS D1.1 (See AWS D1.1 Section 8.15.4) after the weld is completed and has cooled down. The inspector and NDT technician should perform the following tasks for each weld. a.

Verify material identification per approved shop drawings and specifications.

b.

Perform a UT lamination check of the column and beam as required by AWS D1.1 or at least within a 6 in. radius around the weld. As a minimum this check should be performed after welding, however, if performed before welding as well, this may save some rework effort.

c.

Verify that an approved welding procedure specification (WPS) has been provided and that the WPS has been reviewed with each welder performing the weld. A copy of the appropriate WPSs should be at each joint. Welds not executed in conformance with the WPS should be considered rejectable (See AWS D1.1 Section 6.3.1).

d.

Identify welding consumables per approved shop drawings and approved WPS (See AWS D1.1 Sections 6.2 and 6.5.3).

e.

Verify welder identification and certification. Verify that required supplemental qualification tests have been passed (See AWS D1.1 Section 6.4) and mock-ups, if required by the Contract Documents, have been executed. 9-6

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f.

Verify proper amperage and voltage of the welding process by using a hand held calibrated amp and volt meter. (Similar equipment should also be used by the fabricator.) Amperage and voltage should be measured at the arc with this equipment.

g.

Visually inspect all required welds in accordance with AWS D1.1. Verify and document the fabrication sequence including the following per approved shop drawings and approved WPS (See AWS D1.1 Section 6.5.4):

h.

1.

Fit-up;

2.

Preheat and interpass temperatures;

3.

Welding machine settings. Voltage should be determined at the arc and amperage on the cables. Welds executed outside of the parameters contained in the approved WPS should be considered rejectable;

4.

Weld sequence;

5.

Weld pass sequence and size of weld bead;

6.

Peening, if required;

7.

Removal of backup and weld (extension) tabs, preparatory grinding and cleaning, and execution of reinforcing fillet weld, as required by the WPS;

8.

Application and maintenance of postheat or insulation to completed weld as required by the WPS.

Ultrasonically inspect in accordance with AWS D1.1. Attempt to pass sound through the entire weld volume from two crossing directions where possible. In particular, inspect the beam bottom flange from both "A" and "B" faces. This will require adequate staging to be provided by the contractor to permit safe access by the inspector. This is normally not a problem in existing buildings; however, it may be more difficult on buildings under construction.

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Chapter 10 - Visual Inspection

10. VISUAL INSPECTION Visual inspection is the primary method which should be used to confirm that the procedures, materials and workmanship incorporated in the Work are those that have been specified and approved for the project. Visual inspection should be conducted by appropriately qualified personnel, in accordance with a written practice. 10.1 Personnel Qualification Visual inspection personnel should be qualified under AWS D1.1, Chapter 6. The basis of qualification should be specified by the Engineer. Acceptable qualification bases are : a)

Current or previous certification as an AWS Certified Welding Inspector (CWI) in accordance with the provisions of AWS QC1, Standard and Guide for Qualification and Certification of Welding Inspectors, or

b)

Current or previous qualification by the Canadian Welding Bureau (CWB) to the requirements of the Canadian Standard Association (CSA) Standard W178.2, Certification of Welding Inspectors, or

c)

An engineer or technician who, by training or experience, or both, in metals fabrication, inspection and testing, is competent to perform inspection work.

The qualification of an inspector will remain in effect indefinitely, provided the inspector remains active in the inspection of welded steel fabrication, unless there is a specific reason to question the inspector's ability. The Engineer should have the authority to verify the qualification of inspectors. 10.2 Written Practice a)

The employer (Testing Agency or Fabricator/Erector) should maintain a written practice for the control and administration of inspection personnel training and qualification.

b)

The written practice should describe the employer's procedures for visual welding inspection and material controls for determining the acceptability of materials and weldments in accordance with the applicable codes, standards, specifications and procedures.

c)

The employer's written practice should describe the training and experience and requirements for qualification.

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10.3 Duties a)

The inspector should review and understand the applicable portions of the Specifications, the Contract Drawings and the Shop Drawings for the project.

b)

The inspector should verify that all applicable welding Procedure Qualification Records (PQR)s, welder and welding operator qualifications and welding procedure specifications (WPS) are available, current and accurate.

c)

The inspector should require requalification of any welder, welding operator or tack welder who has, for a period of six months, not used the process for which the person was qualified.

d)

The inspector should check all mill certificates for material compliance with the project requirements.

e)

The inspector should verify the electrode/wire specification sheets for compliance with the Contract Documents.

f)

The inspector should make certain that all electrodes are used only in the positions and with the type of welding parameters specified in the WPS.

g)

The inspector should, at suitable intervals, observe joint preparation, assembly practice, preheat temperatures, interpass temperatures, welding techniques, welder performance and post-weld dressing to make certain that the applicable requirements of the WPS and Code are met.

h)

The inspector should inspect the work to ensure compliance with AWS D1.1, Sections 3 and 8.15. Size and contour of welds should be measured with suitable gauges. Visual inspection may be aided by a strong light, magnifiers, or other devices which may be helpful.

i)

The QC inspector should be responsible for scheduling the NDT technicians in a timely manner, after the visual inspection is complete and the assembly has cooled. For repair welding, the NDT should not be performed sooner than 48 hours after the welding is complete and cooled to ambient temperature.

j)

Inspectors should identify the inspected and accepted welds, assemblies and connections with a personal mark or stamp, or maintain adequate records to indicate the status of inspection work. The accepted and rejected items should be documented in a written report. The report should be transmitted to the designated recipients in a timely manner.

Commentary: Depending on how the QA and QC functions are structured for any particular project, the role of the visual inspector may vary considerably. Ideally, the QC inspector is an employee of the contractor and answers to a QA 10-2

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department head who is not connected with production. If this is not the case, an inherent conflict of interest may be present. The level of involvement of the QA agency is highly dependent on the structure of the contractor's QC program. If the contractor's QC program is well organized, has competent inspection and testing personnel and is truly independent of production, the QA function can operate in the classical manner as an overseer wherein random spot inspection and testing suffice. In the opposite case where the QC department is being run by production, the QA agency must take a very active role and perform many of the QC duties. The definitions of these roles can directly affect the project structure and associated budgets. The Owner cannot accurately budget for QA testing and inspection until the contractor is selected and the QC program established. Alleviating this dilemma requires the designer to tightly specify the QC and QA programs. Although AWS D1.1 allows inspector qualification without the CWI certification under the QC1 criteria, it is strongly recommended that the inspection personnel be CWI certified (or previously certified), by experience and written examination.

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Chapter 11 - Nondestructive Testing

11. NONDESTRUCTIVE TESTING Nondestructive testing includes magnetic particle testing (MT), Liquid Dye Penetrant testing (PT), Radiographic Testing (RT) and Ultrasonic Testing (UT). The purpose of nondestructive testing is to serve as a backup to Visual Inspection and to detect flaws and defects that are not visible. Nondestructive examination is not a replacement for an adequate program of Visual Inspection, and should not be used as such. 11.1 Personnel 11.1.1 Qualification

Nondestructive testing personnel shall be qualified under The American Society for Nondestructive Testing, Inc., Recommended Practice No. SNT-TC-1A, in one of the three following levels: a)

NDT Level I - An NDT Level I individual should be qualified to properly perform specific calibrations, specific NDT, and specific evaluations for acceptance or rejection determinations according to written instructions and to record results. The NDT Level I should receive the necessary instruction or supervision from a certified NDT Level III individual or designee.

b)

NDT Level II - An NDT Level II individual should be qualified to set up and calibrate equipment and to interpret and evaluate results with respect to applicable codes, standards and specifications. The NDT Level II should be thoroughly familiar with the scope and limitations of the methods for which he/she is qualified and should exercise assigned responsibility for on-the-job training and guidance of trainees and NDT Level I personnel. The NDT Level II should be able to organize and report the results of NDT.

c)

NDT Level III - An NDT Level III individual should be capable of establishing techniques and procedures; interpreting codes, standards, specifications and procedures; and designating the particular NDT methods, techniques, and procedures to be used. The NDT Level III should be responsible for the NDT operations for which he/she is qualified and assigned and should be capable of interpreting and evaluating results in terms of existing codes, standards, and specifications. The NDT Level III should have sufficient practical background in applicable materials, fabrication, and product technology to establish techniques and to assist in establishing acceptance criteria when none are otherwise available. The NDT Level III should have general familiarity with other appropriate NDT methods, as demonstrated by the ASNT Level III Basic examination or other means. The NDT Level III, in the methods in which certified, should be capable of training and examining NDT Level I and II personnel for certification in those methods.

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11.1.2 Written Practice

a)

The employer (Testing Agency or Fabricator/Erector) should maintain a written practice for the control and administration of NDT personnel training, examination and certification.

b)

The employer's written practice should describe the responsibility of each level of certification for determining the acceptability of materials and weldments in accordance with the applicable codes, standards, specifications and procedures.

c)

The employer's written practice should describe the training, experience and examination requirements for each level of certification.

11.1.3 Certification

a)

Certification of all levels of NDT personnel is the responsibility of the employer.

b)

Certification of NDT personnel should be based on demonstration of satisfactory qualification in accordance with Sections 6, 7 and 8 of SNT-TC-1A, as modified by the employer's written practice.

c)

Personnel certifications should be maintained on file by the employer and a copy should be carried by the technician.

11.1.4 Recertification

a)

b)

c)

All levels of NDT Personnel should be recertified periodically in accordance with one of the following criteria: i)

Evidence of continuing satisfactory performance

ii)

Reexamination in those portions of the examinations in Section 8 deemed necessary by the employer's NDT Level III

Recommended maximum recertification intervals are: i)

Levels I and II - 3 years

ii)

Level III - 5 years

The employer's written practice should include rules covering the duration of interrupted service that requires reexamination and recertification.

11.2 Execution 11.2.1 General

Nondestructive testing should not be used in lieu of visual inspection. Commentary: Visual inspection and NDT should be used as a complement to one another. There are four basic testing methods beyond visual inspection which are commonly used: magnetic particle (MT), liquid penetrant (PT), radiographic 11-2

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testing (RT) and ultrasonic testing (UT). The uses of the methods are described in detail in AWS B1.0, Guide for Nondestructive Inspection of Welds. When nondestructive testing other than visual is to be required, it should be so stated in the bid documents. This information should designate the categories of welds to be examined, the extent of examination in each category and the methods of testing. The designer should require that the testing laboratory employing the NDT technicians be certified by the National Institute of Standards and Technology, NAVLAP program and that the technicians are qualified under ASTM E543. Additionally, the laboratory should employ a Level III NDT supervisor under the requirements of SNT-TC-1A. The designer or his/her designated welding engineer should be familiar with the strengths and limitations of each NDT method. Incorrect selection of the methods has caused false reliance on the results. Each method has its own strengths and weaknesses. Magnetic particle and liquid penetrant testing require the least amount of training; radiographic and ultrasonic testing require a higher level of training and background. NDT technicians are not generally required to be certified welding inspectors under the QC1 requirements; however, it is highly recommended that at least one NDT technician active on the project site be so qualified. 11.2.2 Magnetic Particle Testing (MT)

MT may be used for surface and near-surface linear defect flaw detection. It is essential that for linear indications to respond to MT, they must be oriented at an angle between 45o and 90o , with the maximum influence occurring at 90o to the flux field. Therefore each area tested should have the electromagnetic yoke positioned at 0o then at 90o . Commentary: MT’s depth limitation is less than 1/8 inch for typical flaws. The instrument consists of an electro-magnetic yoke which sets up a magnetic flux field around a weld. A very fine magnetic powder dust is applied to the area being tested. As the flux lines cross a linear defect the field is interrupted and the powder aligns with the defect. Spurious indications are sometimes encountered along areas of poor weld bead contour, undercut or overlap. The use of a white background paint to improve contrast can improve the reliability of this method. A key use of this method is during air-arc gouging to determine if a crack has been totally removed. Root pass testing is also commonly done with MT. These tests, of course, require that the NDT technician be continually present during welding.

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Design of Steel Moment Frames

11.2.3 Liquid Penetrant Testing (PT)

PT may be used to locate defects which are open to the surface. Commentary: In PT, a highly fluid, red dye penetrant is sprayed on the surface of the joint and allowed to soak into any open surface defect by gravity and capillary action. The surface is then wiped clean and a white developer with a powder consistency is applied. The red dye bleeds back out of the defect highlighting the flaw. The method is typically used on completed welds. Due to the problems associated with additional surface preparations and the time involved with PT, it is recommended that MT be applied when ever possible. There may be situations where, because of geometrical conditions or restricted access, MT cannot be performed. PT is an allowable option keeping in mind that additional surface preparation may be necessary. 11.2.4 Radiographic Testing (RT)

RT may be specified for internal flaw detection. Commentary: The RT procedure consists of using an X-ray or gamma ray source to expose a film similar to that used in medical applications. The most common shop and field technique uses an iridium 192 source of gamma rays on one side of the member being inspected and a film cassette on the opposite side. An exposure is made and the film developed much the way photographic negatives are produced. Areas of different film density relate to flaws in the weldment. RT is sensitive to cracks, lack of fusion, lack of penetration, slag inclusions and porosity defects. RT is rather insensitive to lamellar type defects perpendicular to the path of radiation. It does produce a permanent film record. Due to its two dimensional capability, it gives limited information about the depth of the defect or the angular orientation of a crack. RT has limited application in WSMFs because groove welds in T-joints and the associated geometry of beam-column connections make it impractical. Additionally, the surrounding area must be cleared of personnel for radiation safety requirements. RT is a very useful tool for inspection of groove welds in butt splices in plate applications. 11.2.5 Ultrasonic Testing (UT)

UT should be specified as the main form of NDT used in support of VI for the testing of WSMFs. The bottom beam flange to column flange weld should be inspected in accordance with the requirements of AWS D1.1. The proper scanning techniques beam angle(s) and transducer should be used as specified in a written ultrasonic test procedure. The acceptance standard should be that specified in the original contract documents. If these documents are unavailable 11-4

Interim Guidelines: Evaluation, Repair, Modification and Design of Steel Moment Frames

Chapter 11 - Nondestructive Testing

the acceptance criteria of D1.1 Chapter 8 Statically Loaded Structures should be used. The shear wave scan should be preceded by a scan for laminations in the base metal as specified in D1.1. Rejectable discontinuities should be reported on a standard format as recommended by D1.1, i.e.; length, amplitude and classification. Reflections generated from the root and backing bar area of the weld may be cause for further exploration when: 1. the operator is unable to determine if the signal is from a crack or the weld backing. 2. a reflection can be detected in the web zone but the received signal is not great enough to cause rejection. Although different angles, transducer sizes and scanning methods may be used to further evaluate the root area, the removal of the backing bar may be just as cost effective and will always yield more positive results. After the backing has been thoroughly removed, the root should be tested with MT to detect any linear indication. Typically, on existing buildings being inspected for damage, only the inside face of the top flange of the beam to column weld is accessible. This will require the lower portion including the root to be tested in the second leg of the ultrasonic sound path. This increases the difficulty of evaluating the root and weld backing which is difficult enough to evaluate in the first leg of sound travel. As in the bottom flange, all rejectable discontinuities should be recorded. If root defects are found or discontinuities which are difficult to interpret, it should be the engineers decision whether or not to do further exploration by UT and/or remove the steel backing. Access may become a problem at perimeter columns where one half of the top beam flange is inaccessible. Commentary: The UT test involves sending ultrasonic frequency sound waves into a weldment. Any reflector within the weld or parent metal sends back a reflected signal to the instrument. The sent and received signals are presented on an oscilloscope for interpretation. Unlike RT, MT and PT, the interpretation of the received signal is highly dependent on the skill and training of the technician. The location and depth of the flaw can be accurately determined. The shape and type can also be interpreted to some degree by competent operators. The scanning surfaces must be clean and free from fireproofing, upset metal and weld spatter for proper transducer contact. AWS D1.1, Section 6.19, requires that the entire area to be scanned by shear wave for weld flaw detection be first scanned by longitudinal wave to detect any lamellar defects. These defects can mask indications from the weld areas, if present, and are not favorably oriented for shear wave testing. UT is highly sensitive to planar defects if they are favorably oriented to the sound beam. The primary testing is done by utilizing a shear wave transducer from the flange faces of the beams. The key to detection is to select the proper testing angle which will intercept the flaw perpendicular to its orientation. The amplitude of the received signal is directly related to the flaw orientation and, 11-5

Interim Guidelines: Evaluation, Repair, Modification and Chapter 11 - Nondestructive Testing

Design of Steel Moment Frames

hence, the rejection criteria. In the typical T-joint configuration of WSMF connections, defects in the HAZ of the prepared bevel and root area are favorably oriented to the sound path. This is not the case for the column face HAZ which is not optimally oriented. Sometimes this area can be inspected by using a longitudinal wave transducer from the back side of the column face if no continuity plates are present; however, AWS has no rejection criteria for this method. UT technicians are prone to skipping the lamination check when pressed for production. Recalibration of the instrument is required each time the transducer is changed. The intent of D1.1, 6.19.6.2 is to achieve shear wave testing from both the top of the beam flange (A surface) and from the bottom of the beam flange (B surface). High production pressures sometimes force premature movement of scaffolding, allowing the UT technician access to only the top of the bottom flange (A surface). This precludes proper testing of the weld area below the beam web. Another area of concern is back-up bar removal. Removal of backing is left as an option in D1.1 which defers to the Contract Documents. It is strongly recommended that back-up bar removal be required in the Contract Documents to enhance visual and UT inspection. A common problem with rejects identified by UT technicians occurs during the air-arc gouging of the defect area. If too large of a carbon arc electrode is used or if too large a pass is taken, the defect can easily be gouged out without ever being observed by the welder or the UT inspector. For typical WSMF welds, a 1/4 or 3/16 inch maximum size electrode should be used and light skim passes taken. The UT technician should observe the process through a welding shield. A technician can be falsely lured into reducing his/her rejection criteria if no defect is found during gouging.

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Interim Guidelines: Evaluation, Repair, Modification and Design of Steel Moment Frames

References

12. REFERENCES ATLSS, Fractorgrahpic Analysis of Speciments from Failed Moment Connections, (publication pending, title not exact), SAC, 1995. ATLSS, Testing of Welded “T” Specimens, (publication pending, title not exact), SAC, 1995. Allen J., Personal Correspondence, Test Reports for New Detail, July 30, 1995. Allen J., Partridge, J.E., and Richard, R.M., Stress Distribution in Welded/Bolted Beam to Column Moment Connections. The Allen Company, March, 1995. American Association of State Highway and Transportaion Officials, Bridge Welding Code AASHTO/AWS D1.5, 1995. American Institute of Steel Construction, Statistical Analysis of Charpy V-notch Toughness For Steel Wide Flange Structural Shapes, July, 1995. American Institute of Steel Construction, Manual of Steel Construction, ASD, Ninth Edition, 1989. American Institute of Steel Construction, Manual of Steel Construction, LRFD, Second Edition, 1994. American Institute of Steel Construction, Load and Resistance Factor Design Specification for Structural Steel Buildings, December 1, 1993. American Institute of Steel Construction, Specification for Structural Joints using ASTM A325 or A490 Bolts. 1985. American Institute of Steel Construction, AISC Northridge Steel Update I, October, 1994. American Welding Society, Guide for Nondestructive Inspection of Welds, AWS B1.10-86, 1986. American Welding Society, Guide for Visual Inspection of Welds, AWS B1.11-88, 1988. American Welding Society, Structural Welding Code - Steel AWS, D1.1-94, 1994. Anderson, J.C., Johnson, R.G., Partridge, J.E., “Post Earthquake Studies of A Damaged Low Rise Office Building” Technical Report: Case Studies of Steel Moment Frame Building Performance in the Northridge Earthquake of January 17, 1994 SAC 95-07. SAC, December, 1995. Anderson, J.C., Filippou, F.C., Dynamic Response Analysis of the 18 Story Canoga Building, SAC, March, 1995.

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Interim Guidelines: Evaluation, Repair, Modification and Design of Steel Moment Frames

References

Anderson, J.C., Test Results for Repaired Specimen NSF#1, Report to AISC Steel Advisory Committee, June, 1995. Applied Technology Council, Earthquake Damage Evaluation Data for California ATC-13, Redwood City, CA 1985. Applied Technology Counicl, Procedures for Post Earthquake Safety Evaluations of Buildings ATC-20, Redwood City, CA, 1989. Applied Technology Council, Guidelines for Cyclic Seismic Testing of Components of Steel Structures, ATC-24, Redwood City, CA, 1992. Astaneh-Asl, A. Post-Earthquake Stability of Steel Moment rames with Damage Connections. Proceedings of the Third International Workshop on Connections in Steel Structures, University of Trento, Trento, Italy, 1995. Avent, R., “Designing Heat-Straightening Repairs,” National Steel Construction Conference Proceedings, Las Vegas, NV, AISC, 1992. Avent, R., “Engineered Heat Straightening,” National Steel Construction Conference Proceedings, San Antonio, TX, AISC, 1995. Beck, J.L., May, B.S., Polidori, D.C., Vanik, M.W., “Ambient Vibration Surveys of Three SteelFrame Buildings Strongly Shaken by the 1994 Northridge Earthquake”, Analytical and Field Investigations of Buildings Affected by the Northridge Earthquake of January 17, 1994, SAC 9504 Part 2, SAC, December, 1995. Bertero, V.V., and Whittaker, A., Testing of Repaired Welded Beam Column Assemblies, SAC, publication pending (title not exact), 1995. Blodgett, O., “Evaluation of Beam to Column Connections”, SAC Steel Moment Frame Connection Advisory No. 3, Feb. 1995. Bonowitz, D, and Youssef, N. “SAC Survey of Steel-Moment Frames Affected by the 1994 Northridge Earthquake”, Surveys and Assessment of Damage to Buildings Affected by the Northridge Earthquake of January 17, 1994 SAC 95-06, SAC, 1995. Building Seismic Safety Council. NEHRP Recommended Provisions for Seismic Regulations for New Buildings -1991 Edition FEMA 222, FEMA 223, Washington D.C., January, 1992. Building Seismic Safety Council. NEHRP Recommended Provisions for Seismic Regulations for New Buildings -1994 Edition FEMA 222A, FEMA223A, Washington D.C., July, 1995. Campbell, K.W. and Bazorgnia, Y., “Near Source Attentuation of Peak Horizontal Acceleration from World Wide Accelerogram Records from 1957 - 1993,” Proceedings of the Fifth National Conference on Earthquake Engineering, Chicago, Ill, 1994. 12-2

Interim Guidelines: Evaluation, Repair, Modification and Design of Steel Moment Frames

References

Campbell, S., “Modeling of Weld Fractures Using the Drain Programs”, Technical Report: Parametric Analytical Investigations of Ground Motion and Structural Response, Northridge Earthquake of January 17, 1994 SAC95-05. SAC, 1995. Chen, S.J. and Yeh, C.H., Enhancement of Ductility of Steel Beam-to-Column Connections for Seismic Resistance, Department of Construction Engineering, National Taiwan University, May, 1995. Diererlein, G. “Summary of Building Analysis Studies” Analytical and Field Investigations of Buildings Affected by the Northridge Earthquake of January 17, 1994, SAC 95-04 Part 1, SAC, December, 1995 Durkin, M. E., “Inspection, Damage, and Repair of Steel Frame Buildings Following the Northridge Earthquake”, Technical Report: Surveys and Assessment of Damage to Buildings Affected by the Northridge Earthquake of January 17, 1994 SAC 95-06, SAC, December, 1995. Engelhardt, M.D., and Sabol, T.A. Testing of Welded Steel Moment Connections In Response to the Northridge Earthquake, Progress Report to the AISC Advisory Subcommittee on Special Moment Resisting Frame Research, October, 1994. Engelhardt, M. D., Keedong, K.M. Sabol T. A., Ho, L., Kim, H. Uzarski, J. and Abunnasar, H. “Analysis of a Six Story Steel Moment Frame Building in Santa Monica”, Analytical and Field Investigations of Buildings Affected by the Northridge Earthquake of January 17, 1994, SAC 9504 Part 1 SAC, December, 1995. Engelhardt, M. D., Keedong, K.M., Uzarski, J., Abunassar, H., Sabol, T.A., Ho, L., and Kim, H. “Parametric Studies on Inelastic Modeling of Steel Moment Frames”, Technical Report: Parametric Analytical Investigations of Ground Motion and Structural Response, Northridge Earthquake of January 17, 1994 SAC95-05. SAC, 1995. Engelhardt, M.D., et. al. Testing of Repaired Welded Beam Column Assemblies, SAC, publication pending (title not exact), 1995. Englehardt, M.D., et. al. Accoustic Emission Recordings for Welded Beam Column Assembly Tests, SAC, publication pending (title not exact), 1995. Frank, K.H. “The Physical and Metallurgical Properties Of Structural Steels” State of Art Papers: Metallurgy, Fracture Mechanics, Welding, Moment Connections and Frame System Behavior SAC 95-09. SAC, September, 1996 Fillippou, F.C. “Nonlinear Static and Dynamic Analysis of Canoga Park Towers with FEAPSTRUC”, Analytical and Field Investigations of Buildings Affected by the Northridge Earthquake of January 17, 1994, SAC 95-04 Part 2, SAC., December, 1995.

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Interim Guidelines: Evaluation, Repair, Modification and Design of Steel Moment Frames

References

Fisher, J.W., Dexter, R.J., and Kauffman, E.J., “Fracture Mechanics of Welded Structural Steel Connections.” State of Art Papers: Metallurgy, Fracture Mechanics, Welding, Moment Connections and Frame System Behavior SAC 95-09. SAC, September, 1996 Forrel/Elsesser Engineers, Inc., Lawrence Berkeley National Labs Steel Joint Test - Technical Brief, San Francisco, CA, July 17, 1995. Gates, W.E., and Morden, M., “Lessons from Inspection, Evaluation, Repair and Construction of Welded Steel Moment Frames Following the Northridge Earthquake”, Surveys and Assessment of Damage to Buildings Affected by the Northridge Earthquake of January 17, 1994 SAC 95-06 SAC, December, 1995. Gates, W.E. “Interpretation of SAC Survey Data on Damaged Welded Steel Moment Frames Following the Northridge Earthquake”, Surveys and Assessment of Damage to Buildings Affected by the Northridge Earthquake of January 17, 1994 SAC 95-06, SAC, December, 1995. Green, M. “Santa Clarita City Hall; Northridge Earthquake Damage” Technical Report: Case Studies of Steel Moment Frame Building Performance in the Northridge Earthquake of January 17, 1994 SAC 95-07. SAC, December, 1995. Hall, J.F., “Parameter Study of the Response of Moment-Resisting Steel Frame Buildings to Near-Source Ground Motions”, Technical Report: Parametric Analytical Investigations of Ground Motion and Structural Response, Northridge Earthquake of January 17, 1994 SAC9505. SAC, 1995. Hajjar, J.F., O’Sullivan D.P., Leon, R. T., Gourley, B.C. “Evaluation of the Damage to the Borax Corporate Headquarters Building As A Result of the Northridge Earthquake”, Technical Report: Case Studies of Steel Moment Frame Building Performance in the Northridge Earthquake of January 17, 1994 SAC 95-07. SAC, December, 1995. Harrison, P.L. and Webster, S.E., Examination of Two Moment Resisting Frame Connectors Utilizing a Cover-Plate Design, Brittish Steel Technical, Swinden Laboratories, Moorgate, Rotherham, 1995. Hart, G.C., Huang, S.C., Lobo, R.F., Van Winkle, M., Jain, A., “Earthquake Response of Strengthened Steel Special moment Resisting Frames” Analytical and Field Investigations of Buildings Affected by the Northridge Earthquake of January 17, 1994, SAC 95-04 Part 1, SAC., December, 1995 Hart, G.C., Huang, S., Lobo, R., and Stewart, J., “Elastic and Inelastic Analysis for Weld Failure Prediction of Two Adjacent Steel Buildings”, ” Analytical and Field Investigations of Buildings Affected by the Northridge Earthquake of January 17, 1994, SAC 95-04 Part 1, SAC, December, 1995.

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Interim Guidelines: Evaluation, Repair, Modification and Design of Steel Moment Frames

References

Hart, G.C., Huang, S., Lobo, R., and Stewart, J., “Influence of Vertical Ground Motion on Special Moment-Resisting Frames”, Technical Report: Parametric Analytical Investigations of Ground Motion and Structural Response, Northridge Earthquake of January 17, 1994 SAC9505. SAC, 1995. Heaton, T.H., Hall, J.F., Wald, D.J., and Halling, M.W. “Response of High-Rise and BaseIsolated Buildings to a Hypothetical Mw 7.0 Blind Thrust Earthquake” Science Vol. 26, pp 206211, January, 1995. International Conference of Building Officials, Uniform Building Code UBC-94. Whittier, CA, 1994. Iwan, W.D., “Drift Demand Spectra for Selected Northridge Sites”, Technical Report: Parametric Analytical Investigations of Ground Motion and Structural Response, Northridge Earthquake of January 17, 1994 SAC95-05. SAC, 1995. Joyner, W.B., and Boore, D.M., “Ground Motion Parameters for Seismic Design,”Bulletin of the Sesimological Society of America, 1994. Kariotis, J. and Eimani, T.J., “Analysis of a Sixteen Story Steel Frame Building at Site 5, for the Northridge Earthquake”, Analytical and Field Investigations of Buildings Affected by the Northridge Earthquake of January 17, 1994, SAC 95-04 Part 2, SAC, December, 1995. Krawinkler, H.K., “Systems Behavior of Structural Steel Frames Subjected to Earthquake Ground Motions” State of Art Papers: Metallurgy, Fracture Mechanics, Welding, Moment Connections and Frame System Behavior SAC 95-09. SAC, September, 1996 Krawinkler, H.K., Ali, A.A., Thiel, C.C., Dunlea, J.M., “Analysis of a Damaged 4-Story Building and an Undamaged 2- Story Building”, Analytical and Field Investigations of Buildings Affected by the Northridge Earthquake of January 17, 1994, SAC 95-04 Part 1, SAC, December, 1995. Leon, R. T., “Seismic Performance of Bolted and Riveted Connections” State of Art Papers: Metallurgy, Fracture Mechanics, Welding, Moment Connections and Frame System Behavior SAC 95-09. SAC, September, 1996 Miller, D.K. “Welding of Seismically Resistant Steel Structures” State of Art Papers: Metallurgy, Fracture Mechanics, Welding, Moment Connections and Frame System Behavior SAC 95-09. SAC, September, 1996 Naeim F., DiJulio, R., Benuska, K., Reinhorn, A. M., and Chen, L. “Evaluation of Seismic Performance of an 11 Story Steel Moment Frame Building During the 1994 Northridge Earthquake”, ” Analytical and Field Investigations of Buildings Affected by the Northridge Earthquake of January 17, 1994, SAC 95-04 Part 2 SAC, December, 1995.

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References

Newmark, N.M. and Hall W.J., Earthquake Spectra and Design. Earthquake Engineering Research Institute, 1982. Paret, T.F., Sasaki, K.K., “Analysis of a 17 Story Steel Moment Frame Building Damaged by the Northridge Earthquake”, Analytical and Field Investigations of Buildings Affected by the Northridge Earthquake of January 17, 1994, SAC 95-04 Part 2, SAC, December, 1995. Popov, E.P. and Yang, T.S. Steel Seismic Moment Resisting Connections. University of California at Berkeley, May, 1995. Popov, E.P., et. al. Testing of Repaired Welded Beam Column Assemblies, SAC, publication pending (title not exact), 1995. SAC, Proceedings of the International Workshop on Steel Moment Frames, October 23-24, 1994 SAC-94-01. Sacramento, CA, December, 1994. SAC . Steel Moment Frame Advisory No. 1. September, Sacramento, CA, 1994. SAC . Steel Moment Frame Advisory No. 2. October, Sacramento, CA, 1994. SAC . Steel Moment Frame Advisory No. 3 SAC-95-01, February, Sacramento, CA, 1995. Shonafelt, G.O., and Horn, W.B.. Guidelines for Evaluation and Repair of Damaged Steel Bridge Members, NCHRP Report 271, Transportation Research Board, 1984. Skiles, J.L. and Campbell, H.H., “Why Steel Fractured in the Northridge Earthquake” SAC Advisory No. 1, October, 1994. Seismic Safety Commission, Northridge Earthquake Turning Loss to Gain, Report to the Governor, Sacramento, CA, 1995. Smith-Emry Company. Report of Test, July, 1995. Sommerville, P, Graves, R., Chandan, S. Technical Report: Characterization of Ground Motion During the Northridge Earthquake of January 17, 1994, SAC 95-03, SAC, December, 1995. State of California. Division of the State Architect (DSA) and Office of Statewide Health Planning and Development (OSHPD). Interpretation of Regulations Steel Moment Resisting Frames, Sacramento, CA, 1994. Structural Engineers Association of California (SEAOC), Seismology Committee, Recommended Lateral Force Requirements and Commentary, Sacramento, CA. 1990. Structural Engineers Association of California (SEAOC), Seismology Committee, Interim Recommendations for Design of Steel Moment Resisting Connection,. Sacramento, CA, January, 1995.

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Interim Guidelines: Evaluation, Repair, Modification and Design of Steel Moment Frames

References

Structural Engineers Association of California (SEAOC), Vision 2000: A Framework for Performance Based Engineering of Buildings, Sacramento, CA, April, 1995. Structural Shape Producers Council, Statistical Analysis of Tensile Data for Wide Flange Structural Shapes, 1994. Thiel, C.C., and Zsutty, T.C., “Earthquake Characteristics and Damage Statistics,” Earthquake Spectra, Volume 3, No. 4., Earthquake Engineering Research Institute, Oakland, Ca. 1987. Tsai, K.C. and Popov, E. P. “Seismic Steel Beam-Column Moment Connections” State of Art Papers: Metallurgy, Fracture Mechanics, Welding, Moment Connections and Frame System Behavior SAC 95-09. SAC, September, 1996 Uang, C.M. and Latham, C.T. Cyclic Testing of Full-Scale MNH-SMRF Moment Connections, Structural Systems Research, University of California, San Diego, March, 1995. Tsai, K.C. and Popov, E.P., Steel Beam - Column Joints In Seismic Moment Resisting Frames, Report No. UCB/EERC-88/19, Earthquake Engineering Research Center, University of California, Berkeley, Nov., 1988. Uang, C.M., Yu, Q.S., Sadre, A., Bonowitz, D., Youssef, N. “Performance of a 13 Story Steel Moment-Resisting Frame Damaged in the 1994 Northridge Earthquake”, ” Analytical and Field Investigations of Buildings Affected by the Northridge Earthquake of January 17, 1994, SAC 9504 Part 2 SAC, December, 1995. Uang, C.M. and Bondad, D. Progress Report on Cyclic Testing of Three Repaired UCSD Specimens, SAC, 1995. Uang, C.M. and Lee, C.H. “Seismic Response of Haunch Repaired Steel SMRFs: Analytical Modelling and Case Studies” ” Analytical and Field Investigations of Buildings Affected by the Northridge Earthquake of January 17, 1994, SAC 95-04 Part 2, SAC., December, 1995 Wald, D.J., Heaton, T.H., and Hudnut, K.W., The Slip History of the 1994 Northridge, California, Earthquake Determined from Strong-Motion, Teleseismic, GPS, and Leveling Data, United Sates Geologic Survey, 1995. Watabe, M. Peformance of Wooden Houses and Steel Buildings during the Great Hanshin Earthquake, Architectural Institute of Japan, May, 1995. Youssef, N.F.G, Bonowitz, D., and Gross, J.L., A Survey of Steel Moment-Resisting Frame Buildings Affected by the 1994 Northridge Earthquake, NISTR 5625, Gaithersburg Md, April, 1995.

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FEDERAL EMERGENCY MANAGEMENT AGENCY

FEMA 267b / June, 1999

Interim Guidelines Advisory No. 2

Steel Moment Frame Structures

Program to Reduce the Earthquake Hazards of

Supplement to FEMA-267

INTERIM GUIDELINES ADVISORY NO. 2 Supplement to FEMA-267 Interim Guidelines: Evaluation, Repair, Modification and Design of Welded Steel Moment Frame Structures Report No. SAC-99-01

SAC Joint Venture a partnership of: Structural Engineers Association of California (SEAOC) Applied Technology Council (ATC) California Universities for Research in Earthquake Engineering (CUREe) Prepared for SAC Joint Venture Partnership by Guidelines Development Committee Ronald O. Hamburger, Chair Thomas Sabol John D. Hooper C. Mark Saunders Robert E. Shaw Raymond H.R. Tide Lawrence D. Reaveley

Project Oversight Committee William J. Hall, Chair John N. Barsom Shirin Ader John Barsom Roger Ferch Theodore V. Galambos John Gross James R. Harris

Richard Holguin Nestor Iwankiw Roy G. Johnston Len Joseph Duane K. Miller John Theiss John H. Wiggins

SAC Project Management Committee SEAOC: William T. Holmes ATC: Christoper Rojahn CUREe: Robin Shepherd

Program Manager: Stephen A. Mahin Investigations Director: James O. Malley Product Director: Ronald O. Hamburger Federal Emergency Management Agency Project Officer: Michael Mahoney Technical Advisor: Robert D. Hanson SAC Joint Venture 555 University Avenue, Suite 126 Sacramento, California 95825 916-427-3647 June, 1999 i

THE SAC JOINT VENTURE SAC is a joint venture of the Structural Engineers Association of California (SEAOC), the Applied Technology Council (ATC), and California Universities for Research in Earthquake Engineering (CUREe,) formed specifically to address both immediate and long-term needs related to solving problems of the Welded Steel Moment Frame (WSMF) connection that became apparent as a result of the 1994 Northridge earthquake. SEAOC is a professional organization composed of more than 3,000 practicing structural engineers in California. The volunteer efforts of SEAOC’s members on various technical committees have been instrumental in the development of the earthquake design provisions contained in the Uniform Building Code as well as the National Earthquake Hazards Reduction Program (NEHRP) Provisions for Seismic Regulations for New Buildings. The Applied Technology Council is a non-profit organization founded specifically to perform problem-focused research related to structural engineering and to bridge the gap between civil engineering research and engineering practice. It has developed a number of publications of national significance including ATC 306, which serves as the basis for the NEHRP Recommended Provisions. CUREe is a nonprofit organization formed to promote and conduct research and educational activities related to earthquake hazard mitigation. CUREe’s eight institutional members are: the California Institute of Technology, Stanford University, the University of California at Berkeley, the University of California at Davis, the University of California at Irvine, the University of California at Los Angeles, the University of California at San Diego, and the University of Southern California. This collection of university earthquake research laboratory, library, computer and faculty resources is among the most extensive in the United States. The SAC Joint Venture allows these three organizations to combine their extensive and unique resources, augmented by subcontractor universities and organizations from around the nation, into an integrated team of practitioners and researchers, uniquely qualified to solve problems related to the seismic performance of WSMF structures.

DISCLAIMER The purpose of this document is to serve as a supplement to the FEMA-267 publication Interim Guidelines: Evaluation, Repair, Modification and Design of Welded Steel Moment Frame Structures. This Advisory, which is intended to be used in conjunction with FEMA-267, supercedes and entirely replaces Interim Guidelines Advisory No. 1 (FEMA 267a). FEMA-267 was published to provide engineers and building officials with guidance on engineering procedures for evaluation, repair, modification and design of welded steel moment frame structures, to reduce the risks associated with earthquake-induced damage. The recommendations were developed by practicing engineers based on professional judgment and experience and a preliminary program of laboratory, field and analytical research. This preliminary research, known as the SAC Phase 1 program, commenced in November, 1994 and continued through the publication of the Interim Guidelines document. This Interim Guidelines Advisory No. 2, which updates and replaces Interim Guidelines Advisory No. 1, is based on supplementary data developed under a program of continuing research, known as the SAC Phase 2 program, as well as findings developed by other, independent researchers. Final design recommendations, superceding both FEMA-267 and this document are scheduled for publication in early 2000. Independent review and guidance in the production of both the FEMA-267, Interim Guidelines and the advisories was provided by a project oversight panel comprised of experts from industry, practice and academia. Users are cautioned that research into the behavior of these structures is continuing. Interpretation of the results of this research may invalidate or suggest the need for modification of recommendations contained herein. No warranty is offered with regard to the recommendations contained herein, either by the Federal Emergency Management Agency, the SAC Joint Venture, the individual joint venture partners, their directors, members or employees. These organizations and their employees do not assume any legal liability or responsibility for the accuracy, completeness, or usefulness of any of the information, products or processes included in this publication. The reader is cautioned to carefully review the material presented herein. Such information must be used together with sound engineering judgment when applied to specific engineering projects. This Interim Guidelines Advisory has been prepared by the SAC Joint Venture with funding provided by the Federal Emergency Management Agency, under contract number EMW-95-C-4770. The SAC Joint Venture gratefully acknowledges the support of FEMA and the leadership of Michael Mahoney and Robert Hanson, Project Officer and Technical Advisor, respectively. The SAC Joint Venture also wishes to express its gratitude to the large numbers of engineers, building officials, organizations and firms that provided substantial efforts, materials, and advice and who have contributed significantly to the progress of the Phase 2 effort.

ii

Interim Guidelines Advisory No. 2

SAC 99-01

PREFACE Purpose The purpose of the Interim Guidelines Advisory series is to provide engineers and building officials with timely information and guidance resulting from ongoing problem-focused studies of the seismic behavior of moment-resisting steel frame structures. These advisories are intended to be supplements to FEMA-267 Interim Guidelines: Evaluation, Repair, Modification and Design of Welded Steel Moment Frame Structures first published in August 1995. The first Interim Guidelines Advisory, FEMA-267a, was published in January 1997. The specific revisions and updates to the Interim Guidelines contained in FEMA-267a were developed based on input obtained from a group of engineers and building officials actively engaged in the use of the FEMA-267 document, in the period since its initial publication in August 1995. That input was obtained during a workshop held in August 1996, in Los Angeles, California. This second Interim Guidelines Advisory has been prepared as a series of updates and revisions both to the FEMA-267, Interim Guidelines which it supplements and to the FEMA267a, Interim Guidelines Advisory publication, which it supercedes. The material contained in this Interim Guidelines Advisory No. 2 is based on the extensive analytical and laboratory research that has been conducted by the SAC Joint Venture and other researchers during the intervening period, along with recent developments in the steel construction industry. The material contained in this Advisory has been formatted to match that contained in the original Interim Guidelines, to permit the user to insert this material directly into appropriate sections of that document. This Advisory is not intended to serve as a self-contained text and should not be used as such. It does, however, completely replace the material contained in FEMA-267a. A new set of recommendations for the design, analysis, evaluation repair, retrofit and construction of moment-resisting steel frames is currently being prepared as part of the Phase 2 Program to Reduce Earthquake Hazards in Steel Moment Frame Structures. These new Seismic Design Criteria, which are anticipated to be completed early in the year 2000, will replace in their entirety the FEMA-267 Interim Guidelines and this Interim Guidelines Advisory No. 2. Background The Northridge earthquake of January 17, 1994, dramatically demonstrated that the prequalified, welded beam-to-column moment connection commonly used in the construction of welded steel moment resisting frames (WSMFs) in the period 1965-1994 was much more susceptible to damage than previously thought. The stability of moment frame structures in earthquakes is dependent on the capacity of the beam-column connection to remain intact and to resist tendencies of the beams and columns to rotate with respect to each other under the influence of lateral deflection of the structure. The prequalified connections were believed to be ductile and capable of withstanding the repeated cycles of large inelastic deformation explicitly relied upon in the building code provisions for the design of these structures. Although many affected connections were not damaged, a wide spectrum of unexpected brittle connection iii

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fractures did occur, ranging from isolated fractures through or adjacent to the welds of beam flanges to columns, to large fractures extending across the full depth of the columns. At the time this damage was discovered, the structural steel industry and engineering profession had little understanding of the specific causes of this damage, the implications of this damage for building safety, or even if reliable methods existed to repair the damage which had been discovered. Although the connection failures did not result in any casualties or collapses, and many WSMF buildings were not damaged, the incidence of damage was sufficiently pervasive in regions of strong ground motion to cause wide-spread concern by structural engineers and building officials with regard to the safety of these structures in future earthquakes. In response to these concerns, the Federal Emergency Management Agency (FEMA) entered into a cooperative agreement with the SAC Joint Venture to perform problem-focused study of the seismic performance of welded steel moment connections and to develop interim recommendations for professional practice. Specifically, these recommendations were intended to address the inspection of earthquake affected buildings to determine if they had sustained significant damage; the repair of damaged buildings; the upgrade of existing buildings to improve their probable future performance; and the design of new structures to provide more reliable seismic performance. Within weeks of receipt of notification of FEMA’s intent to enter into this agreement, the SAC Joint Venture published a series of two design advisories (SAC, 1994a; SAC, 1994b). These design advisories presented a series of papers, prepared by engineers and researchers engaged in the investigation of the damaged structures and presenting individual opinions as to the causes of the damage, potential methods of repair, and possible designs for more reliable connections in the future. In February 1995, Design Advisory No. 3 (SAC, 1995a) was published. This third advisory presented a synthesis of the data presented in the earlier publications, together with the preliminary recommendations developed in an industry workshop, attended by more than 50 practicing engineers, industry representatives and researchers, on methods of inspecting, repairing and designing WSMF structures. At the time this third advisory was published, significant disagreement remained within the industry and the profession as to the specific causes of the damage observed and appropriate methods of repair given that the damage had occurred. Consequently, the preliminary recommendations were presented as a series of issue statements, followed by the consensus opinions of the workshop attendees, where consensus existed, and by majority and dissenting opinions where such consensus could not be formed. During the first half of 1995, an intensive program of research was conducted to more definitively explore the pertinent issues. This research included literature surveys, data collection on affected structures, statistical evaluation of the collected data, analytical studies of damaged and undamaged buildings and laboratory testing of a series of full-scale beam-column assemblies representing typical pre-Northridge design and construction practice as well as various repair, upgrade and alternative design details. The findings of this research (SAC 1995c, SAC 1995d, SAC 1995e, SAC 1995f, SAC 1995g, SAC 1996) formed the basis for the development of FEMA 267 - Interim Guidelines: Evaluation, Repair, Modification, and Design of Welded Steel Moment Frame Structures (SAC, 1995b), which was published in August, 1995. FEMA 267 provided the first definite, albeit interim, recommendations for practice, following the discovery of connection damage in the Northridge earthquake.

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As a result of these and supplemental studies conducted by the SAC Joint Venture, as well as independent research conducted by others, it is now known that a large number of factors contributed to the damage sustained by steel frame buildings in the Northridge earthquake. These included: • design practice that favored the use of relatively few frame bays to resist lateral seismic demands, resulting in much larger member and connection geometries than had previously been tested; • standard detailing practice which resulted in the development of large inelastic demands at the beam to column connections; • detailing practice that often resulted in large stress concentrations in the beam-column connection, as well as inherent stress risers and notches in zones of high stress; • the common use of welding procedures that resulted in deposition of low toughness weld metal in the critical beam flange to column flange joints; • relatively poor levels of quality control and assurance in the construction process, resulting in welded joints that did not conform to the applicable quality standards; • excessively weak and flexible column panel zones that resulted in large secondary stresses in the beam flange to column flange joints; • large variations in the strengths of rolled shape members relative to specified values; • an inherent inability of material to yield under conditions of high tri-axial restraint such as exist at the center of the beam flange to column flange joints. With the identification of these factors it was possible for FEMA 267 to present a recommended methodology for the design and construction of moment-resisting steel frames to provide connections capable of more reliable seismic performance. This methodology included the following recommendations: • proportion the beam-column connection such that inelastic behavior occurs at a distance remote from the column face, minimizing demands on the highly restrained column material and the welded joints; • specify weld filler metals with rated toughness values for critical welded joints; • detail connections to incorporate beam flange continuity plates, to minimize stress concentrations; • remove backing bars and weld tabs from critical joints to minimize the potential for stress risers and notch effects and also to improve the reliability with which flaws at the weld root can be observed and repaired;

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• qualify connection configurations through a program of full-scale inelastic testing of representative beam-column assemblies, fabricated in the same manner as is proposed for use in the structure; • increased participation of the design professional in the specification and surveillance of welding procedures and the quality assurance process for welded joints. In the time since the publication of FEMA-267, SAC has continued, under funding provided by FEMA, to perform problem-focused study of the performance of moment resisting connections of various configurations. This work, which is generally referred to as the SAC Phase II program, includes detailed analytical evaluations of buildings and connections, parametric studies into the effects on connection performance of connection configuration, base and weld metal strength, toughness and ductility, as well as additional large scale testing of connection assemblies. The intent of this study is to support development of final guidelines that will present more reliable and economical performance-based methods for: • identification of damaged structures following an earthquake and determination of the extent, severity and consequences of such damage; • design of effective repairs for damaged structures; • identification of existing structures that are vulnerable to unacceptable levels of damage in future earthquakes; • design of structural upgrades for existing vulnerable structures; • design of new structures that are suitably resistant to earthquake induced damage; • procedures for construction quality assurance that are consistent with the levels of reliability intended by the design criteria. This Phase II program of research, which is being conducted by the SAC Joint Venture in parallel and coordination with work by other researchers, is anticipated to be complete in late 1999. It is the intent of FEMA and the SAC Joint Venture to ensure that pertinent information and findings from this program are made available to the user community in a timely manner through the publication of this series of design advisory documents. This Interim Guidelines Advisory No. 2 is the second such publication. Format This Advisory has been prepared as a series of updates and revisions to the FEMA-267, Interim Guidelines publication. It has been formatted in a manner intended to facilitate the identification of changes to the original FEMA-267 text. Only those sections of FEMA-267 that are being revised at this time are included. Other sections of FEMA-267 remain in effect as the current best recommendations of the SAC Joint Venture. This Advisory replaces the earlier Interim Guidelines Advisory, FEMA-267a, in its entirety.

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To facilitate coordination of this Advisory with FEMA-267, the existing system of chapter and section numbering has been retained. The Table of Contents lists all sections of the chapters being revised, including those sections for which no revisions are included. Within the body of this document, a section heading is provided for each section of the chapter; however, if no revision to the section is currently being made, this is indicated immediately beneath the section heading. To facilitate reading of this document, where a revision is made to a section in FEMA 267, the entire text of that section is included herein. Where existing text from FEMA-267 is reproduced in this document, without edit, it is shown in normal face type for guidelines, and in italicized type for commentary. Where existing text is being deleted, this is shown in strike through format. A single strikethrough indicates text deleted in the first advisory, FEMA-267a. A double strikethrough indicates text deleted in this current advisory. New text is shown in underline format. A single underline identifies text added in the first advisory, FEMA-267a. A double underline identifies text added in this current advisory. When a modification has been made to a portion of text, relative to FEMA-267, this will also be noted by the presence of a vertical line at the outside margin of the page. The following two paragraphs illustrate these conventions for guideline and commentary text, respectively. This sentence is representative of typical guideline text, that has been reprinted from FEMA-267 without change.This sentence, is representative of the way in which text being deleted from FEMA-267 in this Interim Guidelines Advisory is identified. This sentence illustrates the way in which text deleted from FEMA-267 in the previous Interim Guidelines Advisory is identified. This sentence illustrates the way in which text being added to FEMA-267 in this Interim Guidelines Advisory is identified.This sentence illustrates the way in which text added to FEMA-267 in the previous Interim Guidelines Advisory is identified. Commentary: This sentence is representative of typical commentary text, that has been reprinted from FEMA-267 without change. This sentence is representative of the way in which commentary text being deleted from FEMA-267 in this Interim Guidelines Advisory is identified. However, this sentence, is representative of the way in which text being deleted from FEMA-267 commentary in the previous Advisory is identified. This sentence indicates the way in which text added to the FEMA-267 commentary in this Advisory is shown.This final sentence illustrates the way in which text added in previous advisory, FEMA-267a, is identified. Intent This Interim Guidelines Advisory, together with the Interim Guidelines they modify, are primarily intended for two different groups of potential users: a) Engineers engaged in evaluation, repair, and upgrade of existing WSMF buildings and in the design of new WSMF buildings incorporating either Special Moment-Resisting Frames or Ordinary Moment-Resisting Frames utilizing welded beam-column connections. The

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recommendations for new construction are applicable to all WSMF construction expected to resist earthquake demands through plastic behavior. b) Regulators and building departments responsible for control of the evaluation, repair, and occupancy of WSMF buildings that have been subjected to strong ground motion and for regulation of the design, construction, and inspection of new WSMF buildings. The fundamental goal of the information presented in the Interim Guidelines as modified by this Advisory is to help identify and reduce the risks associated with earthquake-induced fractures in WSMF buildings through provision of timely information on how to inspect existing buildings for damage, repair damage if found, upgrade existing buildings and design new buildings. The information presented here primarily addresses the issue of beam-to-column connection integrity under the severe inelastic demands that can be produced by building response to strong ground motion. Users are referred to the applicable provisions of the locally prevailing building code for information with regard to other aspects of building construction and earthquake damage control. Limitations The information presented in this Interim Guidelines Advisory, together with that contained in the Interim Guidelines it modifies, is based on limited research conducted since the Northridge Earthquake, review of past research and the considerable experience and judgment of the professionals engaged by SAC to prepare and review this document. Additional research on such topics as the effect of floor slabs on frame behavior, the effect of weld metal and base metal toughness, the efficacy of various beam-column connection details and the validity of current standard testing protocols for prediction of earthquake performance of structures is continuing as part of the Phase 2 program and is expected to provide important information not available at the time this Advisory was formulated. Therefore, many of the recommendations cited herein may change as a result of forthcoming research results. The recommendations presented herein represent the group consensus of the committee of Guideline Writers retained by SAC following independent review by the Project Oversight Committee. They may not reflect the individual opinions of any single participant. They do not necessarily represent the opinions of the SAC Joint Venture, the Joint Venture partners, or the sponsoring agencies. Users are cautioned that available information on the nature of the WSMF problem is in a rapid stage of development and any information presented herein must be used with caution and sound engineering judgment.

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TABLE OF CONTENTS

1

3

4

THE SAC JOINT VENTURE DISCLAIMER PREFACE Purpose Background Format Intent Limitations

ii ii iii iii iii vi vii viii

INTRODUCTION 1.1 Purpose 1.2 Scope 1.3 Background 1.4 The SAC Joint Venture 1.5 Sponsors 1.6 Summary of Phase I Research 1.7 Intent 1.8 Limitations 1.9 Use of the Guidelines

1-1 1-1 1-1 1-8 1-8 1-8 1-8 1-9 1-9

CLASSIFICATIONS AND IMPLICATIONS OF DAMAGE 3.1 Summary of Earthquake Damage 3.2 Damage Types 3.2.1 Girder Damage 3.2.2 Column Flange Damage 3.2.3 Weld Damage, Defects and Discontinuities 3.2.4 Shear Tab Damage 3.2.5 Panel Zone Damage 3.2.6 Other Damage 3.3 Safety Implications 3.4 Economic Implications

3-1 3-1 3-1 3-1 3-1 3-4 3-4 3-4 3-5 3-7

POST-EARTHQUAKE EVALUATION 4.1 Scope 4.2 Preliminary Evaluation 4.2.1 Evaluation Process 4.2.1.1 Ground Motion 4.2.1.2 Additional Indicators 4.2.2 Evaluation Schedule 4.2.3 Connection Inspections 4.2.3.1 Analytical Evaluation 4.2.3.2 Buildings with Enhanced Connections 4.2.4 Previous Evaluations and Inspections

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4.4 4.5

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Detailed Evaluation Procedure 4.3.1 Eight Step Inspection and Evaluation Procedure 4.3.2 Step 1 - Categorize Connections By Group 4.3.3 Step 2 - Select Samples of Connections for Inspection 4.3.3.1 Method A - Random Selection 4.3.3.2 Method B - Deterministic Selection 4.3.3.3 Method C - Analytical Selection 4.3.4 Step 3- Inspect the Selected Samples of Connections 4.3.4.1 Damage Characterization 4.3.5 Step 4 - Inspect Connections Adjacent to Damaged Connections 4.3.6 Step 5 - Determine Average Damage Index for the Group 4.3.7 Step 6 - Determine the Probability that the Connections in a Group at a Floor Level Sustained Excessive Damage 4.3.7.1 Some Connections In Group Not Inspected 4.3.7.2 All Connections in Group Inspected 4.3.8 Step 7 - Determine Recommended Recovery Strategies for the Building 4.3.9 Step 8 - Evaluation Report Alternative Group Selection for Torsional Response Qualified Independent Engineering Review 4.5.1 Timing of Independent Review 4.5.2 Qualifications and Terms of Employment 4.5.3 Scope of Review 4.5.4 Reports 4.5.5 Responses and Corrective Actions 4.5.6 Distribution of Reports 4.5.7 Engineer of Record 4.5.8 Resolution of Differences

POST-EARTHQUAKE INSPECTION 5.1 Connection Types Requiring Inspection 5.1.1 Welded Steel Moment Frame (WSMF) Connections 5.1.2 Gravity Connections 5.1.3 Other Connection Types 5.2 Preparation 5.2.1 Preliminary Document Review and Evaluation 5.2.1.1 Document Collection and Review 5.2.1.2 Preliminary Building Walk-Through 5.2.1.3 Structural Analysis 5.2.1.4 Vertical Plumbness Check 5.2.2 Connection Exposure 5.3 Inspection Program 5.3.1 Visual Inspection (VI) 5.3.1.1 Top Flange 5.3.1.2 Bottom Flange

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5.3.2 5.3.3 5.3.4 5.3.5 6

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5.3.1.3 Column and Continuity Plates 5.3.1.4 Beam Web Shear Connection Nondestructive Testing (NDT) Inspector Qualification Post-Earthquake Field Inspection Report Written Report

5-6 5-7 5-7 5-9 5-9 5-9

POST-EARTHQUAKE REPAIR AND MODIFICATION 6.1 Scope 6.2 Shoring 6.3 Repair Details 6.4 Preparation 6.5 Execution 6.6 Structural Modification 6.6.1 Definition of Modification 6.6.2 Damaged vs. Undamaged Connections 6.6.3 Criteria 6.6.4 Strength and Stiffness 6.6.4.1 Strength 6.6.4.2 Stiffness 6.6.5 Plastic Rotation Capacity 6.6.6 Connection Qualification and Design 6.6.6.1 Qualification Test Protocol 6.6.6.2 Acceptance Criteria 6.6.6.3 Calculations 6.6.6.3.1 Material Strength Properties 6.6.6.3.2 Determine Plastic Hinge Location 6.6.6.3.3 Determine Probable Plastic Moment at Hinges 6.6.6.3.4 Determine Beam Shear 6.6.6.3.5 Determine Strength Demands on Connection 6.6.6.3.6 Check Strong Column - Weak Beam Conditions 6.6.6.3.7 Check Column Panel Zone 6.6.7 Modification Details 6.6.7.1 Haunch at Bottom Flange 6.6.7.2 Top and Bottom Haunch 6.6.7.3 Cover Plate Sections 6.6.7.4 Upstanding Ribs 6.6.7.5 Side-Plate Connections 6.6.7.6 Bolted Brackets

6-1 6-1 6-1 6-1 6-1 6-1 6-1 6-1 6-1 6-4 6-4 6-6 6-7 6-10 6-11 6-11 6-12 6-13 6-16 6-18 6-19 6-20 6-21 6-23 6-24 6-24 6-26 6-26 6-28 6-29 6-29

NEW CONSTRUCTION 7.1 Scope 7.2 General - Welded Steel Frame Design Criteria 7.2.1 Criteria 7.2.2 Strength and Stiffness

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7.4

7.5

7.6 7.7 7.8

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7.2.2.1 Strength 7.2.2.2 Stiffness 7.2.3 Configuration 7.2.4 Plastic Rotation Capacity 7.2.5 Redundancy 7.2.6 System Performance 7.2.7 Special Systems Connection Design and Qualification Procedures - General 7.3.1 Connection Performance Intent 7.3.2 Qualification by Testing 7.3.3 Design by Calculation Guidelines for Connection Qualification by Testing 7.4.1 Testing Protocol 7.4.2 Acceptance Criteria Guidelines for Connection Design by Calculation 7.5.1 Material Strength Properties 7.5.2 Design Procedure - Strengthened Connections 7.5.2.1 Determine Plastic Hinge Locations 7.5.2.2 Determine Probable Plastic Moment at Hinge 7.5.2.3 Determine Shear at Plastic Hinge 7.5.2.4 Determine Strength Demands at Critical Sections 7.5.2.5 Check for Strong Column - Weak Beam Condition 7.5.2.6 Check Column Panel Zone 7.5.3 Design Procedure - Reduced Beam Section Connections 7.5.3.1 Determine Reduced Section and Plastic Hinge Locations 7.5.3.2 Determine Strength and Probable Plastic Moment in RBS 7.5.3.3 Strong Column - Weak Beam Condition 7.5.3.4 Column Panel Zone 7.5.3.5 Lateral Bracing 7.5.3.6 Welded Attachments Metallurgy & Welding Quality Control / Quality Assurance Guidelines on Other Connection Design Issues 7.8.1 Design of Panel Zones 7.8.2 Design of Web Connections to Column Flanges 7.8.3 Design of Continuity Plates 7.8.4 Design of Weak Column and Weak Way Connections Moment Frame Connections for Consideration in New Construction 7.9.1 Cover Plate Connections 7.9.2 Flange Rib Connections 7.9.3 Bottom Haunch Connections 7.9.4 Top and Bottom Haunch Connections 7.9.5 Side-Plate Connections 7.9.6 Reduced Beam Section Connections 7.9.7 Slip-Friction Energy Dissipating Connections

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7.9.8 Column Tree Connections 7.9.9 Slotted Web Connections 7.9.10 Bolted Bracket Connections Other Types of Welded Connection Structures 7.10.1 Eccentrically Braced Frames (EBF) 7.10.2 Dual Systems 7.10.3 Welded Base Plate Details 7.10.4 Vierendeel Truss Systems 7.10.5 Moment Frame Tubular Systems 7.10.6 Welded Connections of Collectors, Ties and Diaphragm Chords 7.10.7 Welded Column Splices 7.10.8 Built-up Moment Frame Members

METALLURGY & WELDING 8.1 Parent Materials 8.1.1 Steels 8.1.2 Chemistry 8.1.3 Tensile/Elongation Properties 8.1.4 Toughness Properties 8.1.5 Lamellar Discontinuities 8.1.6 K-Area Fractures 8.2 Welding 8.2.1 Welding Process 8.2.2 Welding Procedures 8.2.3 Welding Filler Metals 8.2.4 Preheat and Interpass Temperatures 8.2.5 Postheat 8.2.6 Controlled Cooling 8.2.7 Metallurgical Stress Risers 8.2.8 Welding Preparation & Fit-up

REFERENCES

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1. INTRODUCTION 1.1 Purpose There are no modifications to the Guidelines or Commentary of Section 1.1 at this time. 1.2 Scope There are no modifications to the Guidelines or Commentary of Section 1.2 at this time. 1.3 Background Following the January 17, 1994 Northridge, California Earthquake, more than 100 steel buildings with welded moment-resisting frames were found to have experienced beam-to-column connection fractures. The damaged structures cover a wide range of heights ranging from one story to 26 stories; and a wide range of ages spanning from buildings as old as 30 years of age to structures just being erected at the time of the earthquake. The damaged structures are were spread over a large geographical area, including sites that experienced only moderate levels of ground shaking. Although relatively few such buildings were located on sites that experienced the strongest ground shaking, damage to these buildings was quite severe. Discovery of these extensive connection fractures, often with little associated architectural damage to the buildings, was has been alarming. The discovery has also caused some concern that similar, but undiscovered damage may have occurred in other buildings affected by past earthquakes. Indeed, there are now confirmed isolated reports of such damage. In particular, a publicly owned building at Big Bear Lake is known to have been was damaged by the Landers-Big Bear, California sequence of earthquakes, and at least one building, under construction in Oakland, California at the time fo the several buildings were damaged during the 1989 Loma Prieta Earthquake, was reported to have experienced such damage in the San Francisco Bay Area. WSMF construction is used commonly throughout the United States and the world, particularly for mid- and high-rise construction. Prior to the Northridge Earthquake, this type of construction was considered one of the most seismic-resistant structural systems, due to the fact that severe damage to such structures had rarely been reported in past earthquakes and there was no record of earthquakeinduced collapse of such buildings, constructed in accordance with contemporary US practice. However, the widespread severe structural damage which occurred to such structures in the Northridge Earthquake calleds for re-examination of this premise. The basic intent of the earthquake resistive design provisions contained in the building codes is to protect the public safety, however, there is also an intent to control damage. The developers of the building code provisions have explicitly set forth three specific performance goals for buildings designed and constructed to the code provisions (SEAOC - 1990). These are to provide buildings with the capacity to • resist minor earthquake ground motion without damage; Introduction 1-1

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• resist moderate earthquake ground motion without structural damage but possibly some nonstructural damage; and • resist major levels of earthquake ground motion, having an intensity equal to the strongest either experienced or forecast for the building site, without collapse, but possibly with some structural as well as nonstructural damage. In general, WSMF buildings in the Northridge Earthquake met the basic intent of the building codes, to protect life safety. However, the ground shaking intensity experienced by most of these buildings was significantly less than that anticipated by the building codes. Many buildings that experienced moderate intensity ground shaking experienced significant damage that could be viewed as failing to meet the intended performance goals with respect to damage control. Further, some members of the engineering profession (SEAOC - 1995b) and government agencies (Seismic Safety Commission - 1995) have stated that even these performance goals are inadequate for society’s current needs. WSMF buildings are designed to resist earthquake ground shaking based on the assumption that they are capable of extensive yielding and plastic deformation, without loss of strength. The intended plastic deformation is intended to be developed through a combination of consists of plastic rotations developing within the beams, at their connections to the columns, and plastic shear yielding of the column panel zones,. and is tTheoretically these mechanisms should be capable of resulting in benign dissipation of the earthquake energy delivered to the building. Damage is expected to consist of moderate yielding and localized buckling of the steel elements, not brittle fractures. Based on this presumed behavior, building codes require a minimum lateral design strength for WSMF structures that is approximately 1/8 that which would be required for the structure to remain fully elastic. Supplemental provisions within the building code, intended to control the amount of interstory drift sustained by these flexible frame buildings, typically result in structures which are substantially stronger than this minimum requirement and in zones of moderate seismicity, substantial overstrength may be present to accommodate wind and gravity load design conditions. In zones of high seismicity, most such structures designed to minimum code criteria will not start to exhibit plastic behavior until ground motions are experienced that are 1/3 to 1/2 the severity anticipated as a design basis. This design approach has been developed based on historical precedent, the observation of steel building performance in past earthquakes, and limited research that has included laboratory testing of beamcolumn models, albeit with mixed results, and non-linear analytical studies. Observation of damage sustained by buildings in the Northridge Earthquake indicates that contrary to the intended behavior, in some many cases brittle fractures initiated within the connections at very low levels of plastic demand, and in some cases, while the structures remained essentially elastic. Typically, but not always, fractures initiated at, or near, the complete joint penetration (CJP) weld between the beam bottom flange and column flange (Figure 1-1). Once initiated, these fractures progressed along a number of different paths, depending on the individual joint and stress conditions. Figure 1-1 indicates just one of these potential fracture growth patterns. Investigators initially identified a number of factors which may have contributed to the initiation of fractures at the weld root including: notch effects created by the backing bar which was commonly left in place following joint completion; sub-standard welding that included excessive porosity and slag inclusions as well as incomplete fusion; Introduction 1-2

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and potentially, pre-earthquake fractures resulting from initial shrinkage of the highly restrained weld during cool-down. Such problems could be minimized in future construction, with the application of appropriate welding procedures and more careful exercise of quality control during the construction process. However, it is now known that these were not the only causes of the fractures which occurred. Column flange Fused zone Beam flange

Backing bar Fracture

Figure 1-1 - Common Zone of Fracture Initiation in Beam -Column Connection Current production processes for structural steel shapes result in inconsistent strength and deformation capacities for the material in the through-thickness direction. Non-metallic inclusions in the material, together with anisotropic properties introduced by the rolling process can lead to lamellar weakness in the material. Further, the distribution of stress across the girder flange, at the connection to the column is not uniform. Even in connections stiffened by continuity plates across the panel zone, significantly higher stresses tend to occur at the center of the flange, where the column web produces a local stiffness concentration. Large secondary stresses are also induced into the girder flange to column flange joint by kinking of the column flanges resulting from shear deformation of the column panel zone. The dynamic loading experienced by the moment-resisting connections in earthquakes is characterized by high strain tension-compression cycling. Bridge engineers have long recognized that the dynamic loading associated with bridges necessitates different connection details in order to provide improved fatigue resistance, as compared to traditional building design that is subject to “static” loading due to gravity and wind loads. While the nature of the dynamic loads resulting from earthquakes is somewhat different than the high cycle dynamic loads for which fatigue-prone structures are designed, similar detailing may be desirable for buildings subject to seismic loading. In design and construction practice for welded steel bridges, mechanical and metallurgical notches should be avoided because they may be the initiators of fatigue cracking. As fatigue cracks grow under repetitive loading, a critical crack size may be reached whereupon the material toughness (which is a function of temperature) may be unable to resist the onset of brittle (unstable) crack growth. The beam-to-column connections in WSMF buildings are comparable to category C or D bridge details that have a reduced allowable stress range as opposed to category B details for which special metallurgical, inspection and testing requirements are applied. The rapid rate of loading imposed by seismic events, and the complete inelastic range of tension-compression-tension loading applied to these connections is Introduction 1-3

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much more severe than typical bridge loading applications. The mechanical and metallurgical notches or stress risers created by the beam-column weld joints are a logical point for fracture problems to initiate. This, coupled with the tri-axial restraint provided by the beam web and the column flange, is a recipe for brittle fracture. During the Northridge Earthquake, oOnce fractures initiated in beam-column joints, they progressed in a number of different ways. In some cases, the fractures initiated but did not grow, and could not be detected by visual observation. In other cases, In many cases, the fractures progressed completely directly through the thickness of the weld, and if fireproofing was removed, the fractures were evident as a crack through exposed faces of the weld, or the metal just behind the weld (Figure 12a). Other fracture patterns also developed. In some cases, the fracture developed into a surface that resembled a through-thickness failure of the column flange material behind the CJP weld (Figure 1-2b). In these cases, a portion of the column flange remained bonded to the beam flange, but pulled free from the remainder of the column. This fracture pattern has sometimes been termed a “divot” or “nugget” failure. A number of fractures progressed completely through the column flange, along a near horizontal plane that aligns approximately with the beam lower flange (Figure 1-3a). In some cases, these fractures extended into the column web and progressed across the panel zone Figure (1-3b). Investigators have reported some instances where columns fractured entirely across the section.

a. Fracture at Fused Zone

b. Column Flange “Divot” Fracture

Figure 1-2 - Fractures of Beam to Column Joints

a. Fractures through Column Flange

b. Fracture Progresses into Column Web

Figure 1-3 - Column Fractures Introduction 1-4

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Once these fractures have occurred, the beam - column connection has experienced a significant loss of flexural rigidity and capacity. Residual flexural strength and rigidity must be developed through a couple consisting of forces transmitted through the remaining top flange connection and the web bolts. Initial rResearch suggests that residual stiffness is approximately 20% of that of the undamaged connection and that residual strength varies from 10% to 40% of the undamaged capacity, when loading results in tensile stress normal to the fracture plane. When loading produces compression across the fracture plane, much of the original strength and stiffness remain. However, in providing this residual strength and stiffness, the beam shear connections can themselves be subject to failures, consisting of fracturing of the welds of the shear plate to the column, fracturing of supplemental welds to the beam web or fracturing through the weak section of shear plate aligning with the bolt holes (Figure 1-4).

Figure 1-4 - Vertical Fracture through Beam Shear Plate Connection It is now known that these fractures were the result of a number of complex factors that were not well understood either when these connections were first adopted as a standard design approach, or when the damage was discovered immediately following the Northridge earthquake. Engineers had commonly assumed that when these connections were loaded to yield levels, flexural stresses in the beam would be transferred to the column through a force couple comprised of nearly uniform yield level tensile and compressive stresses in the beam flanges. It was similarly assumed that nearly all of the shear stress in the beam was transferred to the column through the shear tab connection to the beam web. In fact, the actual behavior is quite different from this. As a result of local deformations that occur in the column at the location of the beam connection, a significant portion of the shear stress in the beam is actually transferred to the column through the beam flanges. This causes large localized secondary stresses in the beam flanges, both at the toe of the weld access hole and also in the complete joint penetration weld at the face of the column. The presence of the column web behind the column flange tends to locally stiffen the joint of the beam flange to the column flange, further concentrating the distribution of connection stresses and strains. Finally, the presence of the heavy beam and column flange plates, arranged in a “+” shaped pattern at the beam flange to column flange joint produces a condition of very high restraint, which retards the onset of yielding, by raising the effective yield strength of the material, and allowing the development of very large stresses. Introduction 1-5

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The most severe stresses typically occur at the root of the complete joint penetration weld of the beam bottom flange to the column flange. This is precisely the region of this welded joint that is most difficult for the welder to properly complete, as the access to the weld is restricted by the presence of the beam web and the welder often performs this weld while seated on the top flange, in the so-called “wildcat” position. The welder must therefore work from both sides of the beam web, starting and terminating the weld near the center of the joint, a practice that often results in poor fusion and the presence of slag inclusions at this location. These conditions, which are very difficult to detect when the weld backing is left in place, as was the typical practice, are ready-made crack initiators. When this region of the welded joints is subjected to the large concentrated tensile stresses, the weld defects begin to grow into cracks and these cracks can quickly become unstable and propagate as brittle fractures. Once these brittle fractures initiate, they can grow in a variety of patterns, as described above, under the influence of the stress field and the properties of the base and weld metals present at the zone of the fracture. Despite the obvious local strength impairment resulting from these fractures, many damaged buildings did not display overt signs of structural damage, such as permanent drifts or extreme damage to architectural elements. Until news of the discovery of connection fractures in some buildings began to spread through the engineering community, it was relatively common for engineers to perform cursory post-earthquake evaluations of WSMF buildings and declare that they were undamaged. In order to reliably determine if a building has sustained connection damage, it is necessary to remove architectural finishes and fireproofing and perform nondestructive examination including visual inspection and ultrasonic testing careful visual inspection of the welded joints supplemented, in some cases, by nondestructive testing. Even if no damage is found, this is a costly process. Repair of damaged connections is even more costly. A few WSMF buildings have sustained so much connection damage that it has been deemed more practical to demolish the structures rather than to repair them. In the case of one WSMF building, damaged by the Northridge earthquake, repair costs were sufficiently large that the owner elected to demolish rather than replace than building. Immediately following the Northridge Earthquake, a series of tests of beam-column subassemblies were performed at the University of Texas at Austin, under funding provided by the AISC as well as private sources. The test specimens used heavy W14 column sections and deep (W36) beam sections commonly employed in some California construction. Initial specimens were fabricated using the standard prequalified connection specified by the Uniform Building Code (UBC). Section 2211.7.1.2 of UBC-94 {NEHRP-91 Section 10.10.2.3} specified this prequalified connection as follows: “2211.7.1.2 Connection strength. The girder top column connection may be considered to be adequate to develop the flexural strength of the girder if it conforms to the following: 1. the flanges have full penetration butt welds to the columns. 2. the girder web to column connection shall be capable of resisting the girder shear determined for the combination of gravity loads and the seismic shear forces which result from compliance with Section 2211.7.2.1. This connection strength need not exceed that required to develop gravity loads plus 3(Rw/8) times the girder shear resulting from the prescribed seismic forces.

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Where the flexural strength of the girder flanges is greater than 70 percent of the flexural strength of the entire section, (i.e. btf/(d-tf)Fy>0.7ZxFy) the web connection may be made by means of welding or high-strength bolting. For girders not meeting the criteria in the paragraph above, the girder web-to-column connection shall be made by means of welding the web directly or through shear tabs to the column. That welding shall have a strength capable of developing at least 20 percent of the flexural strength of the girder web. The girder shear shall be resisted by means of additional welds or friction-type slip-critical high strength bolts or both. and: 2211.7.2.1 Strength. The panel zone of the joint shall be capable of resisting the shear induced by beam bending moments due to gravity loads plus 1.85 times the prescribed seismic forces, but the shear strength need not exceed that required to develop 0.8ΣMs of the girders framing into the column flanges at the joint...”

In order to investigate the effects that backing bars and weld tabs had on connection performance, these were removed from the specimens prior to testing. Despite these precautions, the test specimens failed at very low levels of plastic loading. Following these tests at the University of Texas at Austin, reviews of literature on historic tests of these connection types indicated a significant failure rate in past tests as well, although these had often been ascribed to poor quality in the specimen fabrication. It was concluded that the prequalified connection, specified by the building code, was fundamentally flawed and should not be used for new construction in the future. In retrospect, this conclusion may have been somewhat premature. More recent testing of connections having configurations similar to those of the prequalified connection, but incorporating tougher weld metals, having backing bars removed from the bottom flange joint, and fabricated with greater care to avoid the defects that can result in crack initiation, have performed better than those initially tested at the University of Texas. However, as a class, when fabricated using currently prevailing construction practice, these connections still do not appear to be capable of consistently developing the levels of ductility presumed by the building codes for service in moment-resisting frames that are subjected to large inelastic demands.When the first test specimens for that series were fabricated, the welder failed to follow the intended welding procedures. Further, no special precautions were taken to assure that the materials incorporated in the work had specified toughness. Some engineers, with knowledge of fracture mechanics, have suggested that if materials with adequate toughness are used, and welding procedures are carefully specified and followed, adequate reliability can be obtained from the traditional connection details. Others believe that the conditions of high triaxial restraint present in the beam flange to column flange joint (Blodgett - 1995) would prevent ductile behavior of these joints regardless of the procedure used to make the welds. Further they point to the important influence of the relative yield and tensile strengths of beam and column materials, and other variables, that can affect connection behavior. To date, there has not been sufficient research conducted to resolve this issue.

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In reaction to the University of Texas tests as well as the widespread damage discovered following the Northridge Earthquake, and the urging of the California Seismic Safety Commission, in September, 1994 the International Conference of Building Officials (ICBO) adopted an emergency code change to the 1994 edition of the Uniform Building Code (UBC-94) {1994 NEHRP Recommended Provisions Section 5.2}. This code change, jointly developed by the Structural Engineers Association of California, AISI and ICBO staff, deleted the prequalified connection and substituted the following in its place: “2211.7.1.2 Connection Strength. Connection configurations utilizing welds or high-strength bolts shall demonstrate, by approved cyclic test results or calculation, the ability to sustain inelastic rotation and develop the strength criteria in Section 2211.7.1.1 considering the effect of steel overstrength and strain hardening.” “2211.7.1.1 Required strength. The girder-to-column connection shall be adequate to develop the lesser of the following: 1. The strength of the girder in flexure. 2. The moment corresponding to development of the panel zone shear strength as determined from formula 11-1.”

Unfortunately, neither the required “inelastic rotation”, or calculation and test procedures are well defined by these code provisions. Design Advisory No. 3 (SAC-1995) included an Interim Recommendation (SEAOC-1995) that attempted to clarify the intent of this code change, and the preferred methods of design in the interim period until additional research could be performed and reliable acceptance criteria for designs re-established. The State of California similarly published a joint Interpretation of Regulations (DSA-OSHPD - 1994) indicating the interpretation of the current code requirements which would be enforced by the state for construction under its control. This applied only to the construction of schools and hospitals in the State of California. The intent of these Interim Guidelines is to supplement these previously published documents and to provide updated recommendations based on the results of the limited directed research performed to date. 1.4 The SAC Joint Venture There are no modifications to the Guidelines or Commentary of Section 1.4 at this time. 1.5 Sponsors There are no modifications to the Guidelines or Commentary of Section 1.5 at this time. 1.6 Summary of Phase 1 Research There are no modifications to the Guidelines or Commentary of Section 1.6 at this time. 1.7 Intent Introduction 1-8

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There are no modifications to the Guidelines or Commentary of Section 1.7 at this time. 1.8 Limitations There are no modifications to the Guidelines or Commentary of Section 1.8 at this time. 1.9 Use of the Guidelines There are no modifications to the Guidelines or Commentary of Section 1.9 at this time.

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12. REFERENCES ATLSS, Fractographic Analysis of Specimens from Failed Moment Connections, (publication pending, title not exact)Fracture Analysis of Failed Moment Frame Weld Joints Produced in FullScale Laboratory Tests and Buildings Damaged in the Northridge Earthquake, SAC95-08, 1995. ATLSS, Testing of Welded “T” Specimens, (publication pending, title not exact), SAC, 1995 A Study of the Effects of Material and Welding Factors on Moment-Frame Weld Joint Performance Using a Small-Scale Tension Specimen. Kauffman, E.J., and Fisher, J.W., SAC9508 1995. Allen J., Personal Correspondence, Test Reports for New Detail, July 30, 1995. Allen J., Partridge, J.E., and Richard, R.M., Stress Distribution in Welded/Bolted Beam to Column Moment Connections. The Allen Company, March, 1995. American Association of State Highway and Transportation Officials, Bridge Welding Code AASHTO/AWS D1.5, 1995. American Institute of Steel Construction, Seismic Provisions for Structural Steel Buildings, April, 1997 American Institute of Steel Construction, Statistical Analysis of Charpy V-notch Toughness For Steel Wide Flange Structural Shapes, July, 1995. American Institute of Steel Construction, Manual of Steel Construction, ASD, Ninth Edition, 1989. American Institute of Steel Construction, Manual of Steel Construction, LRFD, Second Edition, 1998. American Institute of Steel Construction, Load and Resistance Factor Design Specification for Structural Steel Buildings, December 1, 1993. American Institute of Steel Construction, Specification for Structural Joints using ASTM A325 or A490 Bolts. 1985. American Institute of Steel Construction, AISC Northridge Steel Update I, October, 1994. American Welding Society, Guide for Nondestructive Inspection of Welds, AWS B1.10-86, 1986. American Welding Society, Guide for Visual Inspection of Welds, AWS B1.11-88, 1988. American Welding Society, Surface Roughness Guide for Oxygen Cutting, AWS C4.1-77, 1977. American Welding Society, Structural Welding Code - Steel AWS D1.1-94, 1994. References 12-1

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American Welding Society, Structural Welding Code – Steel AWS D1.1-98, 1998 Anderson, J.C., Johnson, R.G., Partridge, J.E., “Post Earthquake Studies of A Damaged Low Rise Office Building” Technical Report: Case Studies of Steel Moment Frame Building Performance in the Northridge Earthquake of January 17, 1994 SAC 95-07. SAC, December, 1995. Anderson, J.C., Filippou, F.C., Dynamic Response Analysis of the 18 Story Canoga Building, SAC, March, 1995. Anderson, J.C., Test Results for Repaired Specimen NSF#1, Report to AISC Steel Advisory Committee, June, 1995. Applied Technology Council, Earthquake Damage Evaluation Data for California ATC-13, Redwood City, CA 1985. Applied Technology Counicl, Procedures for Post Earthquake Safety Evaluations of Buildings ATC-20, Redwood City, CA, 1989. Applied Technology Council, Guidelines for Cyclic Seismic Testing of Components of Steel Structures, ATC-24, Redwood City, CA, 1992. Astaneh-Asl, A. Post-Earthquake Stability of Steel Moment Frames with Damaged Connections. Proceedings of the Third International Workshop on Connections in Steel Structures, University of Trento, Trento, Italy, 1995. Avent, R., “Designing Heat-Straightening Repairs,” National Steel Construction Conference Proceedings, Las Vegas, NV, AISC, 1992. Avent, R., “Engineered Heat Straightening,” National Steel Construction Conference Proceedings, San Antonio, TX, AISC, 1995. Barsom, J. M. and Korvink, S. A. “Through-thickness Properties of Structural Steels”, manuscript submitted to ASCE Journal of Structural Engineering, 1997. Beck, J.L., May, B.S., Polidori, D.C., Vanik, M.W., “Ambient Vibration Surveys of Three SteelFrame Buildings Strongly Shaken by the 1994 Northridge Earthquake”, Analytical and Field Investigations of Buildings Affected by the Northridge Earthquake of January 17, 1994, SAC 9504 Part 2, SAC, December, 1995. Bertero, V.V., and Whittaker, A. and Gilani, A., Testing of Repaired Welded Beam Column AssembliesSeismic Tesing of Full-Scale Steel Beam-Column Assemblies, SAC96-01, publication pending (title not exact), 1995X1996. Blodgett, O., “Evaluation of Beam to Column Connections”, SAC Steel Moment Frame Connection Advisory No. 3, Feb. 1995. References 12-2

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Bonowitz, D, and Youssef, N. “SAC Survey of Steel-Moment Frames Affected by the 1994 Northridge Earthquake”, Surveys and Assessment of Damage to Buildings Affected by the Northridge Earthquake of January 17, 1994 SAC 95-06, SAC, 1995. Building Seismic Safety Council. NEHRP Recommended Provisions for Seismic Regulations for New Buildings -1991 Edition FEMA 222, (Commentary FEMA 223), Washington D.C., January, 1992. Building Seismic Safety Council. NEHRP Recommended Provisions for Seismic Regulations for New Buildings -1994 Edition FEMA 222A, (Commentary FEMA223A), Washington D.C., July, 1995. Building Seismic Safety Council. NEHRP Recommended Provisions for Seismic Regulations for New Buildings and Other Structures. – 1997 Edition, FEMA 302, (Commentary FEMA303), Washington, D.C., February, 1998 Campbell, K.W. and Bazorgnia, Y., “Near Source Attentuation of Peak Horizontal Acceleration from World Wide Accelerogram Records from 1957 - 1993,” Proceedings of the Fifth National Conference on Earthquake Engineering, Chicago, Ill, 1994. Campbell, S., “Modeling of Weld Fractures Using the Drain Programs”, Technical Report: Parametric Analytical Investigations of Ground Motion and Structural Response, Northridge Earthquake of January 17, 1994 SAC95-05. SAC, 1995. Chen, S.J. and Yeh, C.H., Enhancement of Ductility of Steel Beam-to-Column Connections for Seismic Resistance, Department of Construction Engineering, National Taiwan University, May, 1995. Diererlein, G. “Summary of Building Analysis Studies” Analytical and Field Investigations of Buildings Affected by the Northridge Earthquake of January 17, 1994, SAC 95-04 Part 1, SAC, December, 1995 Durkin, M. E., “Inspection, Damage, and Repair of Steel Frame Buildings Following the Northridge Earthquake”, Technical Report: Surveys and Assessment of Damage to Buildings Affected by the Northridge Earthquake of January 17, 1994 SAC 95-06, SAC, December, 1995. Engelhardt, M.D., and Sabol, T.A. Testing of Welded Steel Moment Connections In Response to the Northridge Earthquake, Progress Report to the AISC Advisory Subcommittee on Special Moment Resisting Frame Research, October, 1994. Engelhardt, M. D., Keedong, K.M. Sabol T. A., Ho, L., Kim, H. Uzarski, J. and Abunnasar, H. “Analysis of a Six Story Steel Moment Frame Building in Santa Monica”, Analytical and Field Investigations of Buildings Affected by the Northridge Earthquake of January 17, 1994, SAC 9504 Part 1 SAC, December, 1995.

References 12-3

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Engelhardt, M. D., Keedong, K.M., Uzarski, J., Abunassar, H., Sabol, T.A., Ho, L., and Kim, H. “Parametric Studies on Inelastic Modeling of Steel Moment Frames”, Technical Report: Parametric Analytical Investigations of Ground Motion and Structural Response, Northridge Earthquake of January 17, 1994 SAC95-05. SAC, 1995. Engelhardt, M.D., Sabol, T. A., and Shuey, B.D. Testing of Repair Concepts for Damaged Steel Moment Connections.et. al. Testing of Repaired Welded Beam Column Assemblies, SAC96-01, publication pending (title not exact), 19951996. Englehardt, M.D. Fowler, T.J., and Barnes, C.A., Acoustic Emission Monitoring of Welded Steel Moment Connection Tests.et. al. Accoustic Emission Recordings for Welded Beam Column Assembly Tests, SAC95-08, publication pending (title not exact), 1995. Frank, K.H. “The Physical and Metallurgical Properties Of Structural Steels” State of Art Papers: Metallurgy, Fracture Mechanics, Welding, Moment Connections and Frame System Behavior SAC 95-09. SAC, September, 1996 Fillippou, F.C. “Nonlinear Static and Dynamic Analysis of Canoga Park Towers with FEAPSTRUC”, Analytical and Field Investigations of Buildings Affected by the Northridge Earthquake of January 17, 1994, SAC 95-04 Part 2, SAC., December, 1995. Fisher, J.W., Dexter, R.J., and Kauffman, E.J., “Fracture Mechanics of Welded Structural Steel Connections.” State of Art Papers: Metallurgy, Fracture Mechanics, Welding, Moment Connections and Frame System Behavior SAC 95-09. SAC, September, 1996 Forrel/Elsesser Engineers, Inc., Lawrence Berkeley National Labs Steel Joint Test - Technical Brief, San Francisco, CA, July 17, 1995. Gates, W.E., and Morden, M., “Lessons from Inspection, Evaluation, Repair and Construction of Welded Steel Moment Frames Following the Northridge Earthquake”, Surveys and Assessment of Damage to Buildings Affected by the Northridge Earthquake of January 17, 1994 SAC 95-06 SAC, December, 1995. Gates, W.E. “Interpretation of SAC Survey Data on Damaged Welded Steel Moment Frames Following the Northridge Earthquake”, Surveys and Assessment of Damage to Buildings Affected by the Northridge Earthquake of January 17, 1994 SAC 95-06, SAC, December, 1995. Green, M. “Santa Clarita City Hall; Northridge Earthquake Damage” Technical Report: Case Studies of Steel Moment Frame Building Performance in the Northridge Earthquake of January 17, 1994 SAC 95-07. SAC, December, 1995. Hall, J.F., “Parameter Study of the Response of Moment-Resisting Steel Frame Buildings to Near-Source Ground Motions”, Technical Report: Parametric Analytical Investigations of Ground Motion and Structural Response, Northridge Earthquake of January 17, 1994 SAC9505. SAC, 1995.

References 12-4

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Hajjar, J.F., O’Sullivan D.P., Leon, R. T., Gourley, B.C. “Evaluation of the Damage to the Borax Corporate Headquarters Building As A Result of the Northridge Earthquake”, Technical Report: Case Studies of Steel Moment Frame Building Performance in the Northridge Earthquake of January 17, 1994 SAC 95-07. SAC, December, 1995. Harrison, P.L. and Webster, S.E., Examination of Two Moment Resisting Frame Connectors Utilizing a Cover-Plate Design, British Steel Technical, Swinden Laboratories, Moorgate, Rotherham, 1995. Hart, G.C., Huang, S.C., Lobo, R.F., Van Winkle, M., Jain, A., “Earthquake Response of Strengthened Steel Special moment Resisting Frames” Analytical and Field Investigations of Buildings Affected by the Northridge Earthquake of January 17, 1994, SAC 95-04 Part 1, SAC., December, 1995 Hart, G.C., Huang, S., Lobo, R., and Stewart, J., “Elastic and Inelastic Analysis for Weld Failure Prediction of Two Adjacent Steel Buildings”, ” Analytical and Field Investigations of Buildings Affected by the Northridge Earthquake of January 17, 1994, SAC 95-04 Part 1, SAC, December, 1995. Hart, G.C., Huang, S., Lobo, R., and Stewart, J., “Influence of Vertical Ground Motion on Special Moment-Resisting Frames”, Technical Report: Parametric Analytical Investigations of Ground Motion and Structural Response, Northridge Earthquake of January 17, 1994 SAC9505. SAC, 1995. Heaton, T.H., Hall, J.F., Wald, D.J., and Halling, M.W. “Response of High-Rise and BaseIsolated Buildings to a Hypothetical Mw 7.0 Blind Thrust Earthquake” Science Vol. 26, pp 206211, January, 1995. International Conference of Building Officials, Uniform Building Code UBC-97, Whittier, CA, 1997. International Conference of Building Officials, Uniform Building Code UBC-94. Whittier, CA, 1994. Iwan, W.D., “Drift Demand Spectra for Selected Northridge Sites”, Technical Report: Parametric Analytical Investigations of Ground Motion and Structural Response, Northridge Earthquake of January 17, 1994 SAC95-05. SAC, 1995. Joyner, W.B., and Boore, D.M., “Ground Motion Parameters for Seismic Design,”Bulletin of the Sesimological Society of America, 1994. Kariotis, J. and Eimani, T.J., “Analysis of a Sixteen Story Steel Frame Building at Site 5, for the Northridge Earthquake”, Analytical and Field Investigations of Buildings Affected by the Northridge Earthquake of January 17, 1994, SAC 95-04 Part 2, SAC, December, 1995.

References 12-5

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Krawinkler, H.K., “Systems Behavior of Structural Steel Frames Subjected to Earthquake Ground Motions” State of Art Papers: Metallurgy, Fracture Mechanics, Welding, Moment Connections and Frame System Behavior SAC 95-09. SAC, September, 1996 Krawinkler, H.K., Ali, A.A., Thiel, C.C., Dunlea, J.M., “Analysis of a Damaged 4-Story Building and an Undamaged 2- Story Building”, Analytical and Field Investigations of Buildings Affected by the Northridge Earthquake of January 17, 1994, SAC 95-04 Part 1, SAC, December, 1995. Ksai, K. , and Bleiman, D. “Bolted Brackets for Repair of Damaged Steel Moment Frame Connections,” 7th U.S.-Japan Workshop on the Improvement of Structural Design and Construction Practices: Lessons Learned from Northridge and Kobe, Kobe, Japan, January, 1996 Leon, R. T., “Seismic Performance of Bolted and Riveted Connections” State of Art Papers: Metallurgy, Fracture Mechanics, Welding, Moment Connections and Frame System Behavior SAC 95-09. SAC, September, 1996 Miller, D.K. “Welding of Seismically Resistant Steel Structures” State of Art Papers: Metallurgy, Fracture Mechanics, Welding, Moment Connections and Frame System Behavior SAC 95-09. SAC, September, 1996 Naeim F., DiJulio, R., Benuska, K., Reinhorn, A. M., and Chen, L. “Evaluation of Seismic Performance of an 11 Story Steel Moment Frame Building During the 1994 Northridge Earthquake”, ” Analytical and Field Investigations of Buildings Affected by the Northridge Earthquake of January 17, 1994, SAC 95-04 Part 2 SAC, December, 1995. Newmark, N.M. and Hall W.J., Earthquake Spectra and Design. Earthquake Engineering Research Institute, 1982. NIST and AISC. Modification of Existing Welded Steel Moment Frame Connections for Seismic Resistance. National Institute of Standards and Technology and American Institute of Steel Construction. 1999 Paret, T.F., Sasaki, K.K., “Analysis of a 17 Story Steel Moment Frame Building Damaged by the Northridge Earthquake”, Analytical and Field Investigations of Buildings Affected by the Northridge Earthquake of January 17, 1994, SAC 95-04 Part 2, SAC, December, 1995. Popov, E.P. and Yang, T.S. Steel Seismic Moment Resisting Connections. University of California at Berkeley, May, 1995. Popov, E.P. Blondet, M., Stepanov, L, and Stodjadinovic, B. Full-Scale Beam-Column Connection Tests. et. al. Testing of Repaired Welded Beam Column Assemblies, SAC, publication pending (title not exact), 1995 SAC 96-01. 1996.. SAC, Proceedings of the International Workshop on Steel Moment Frames, October 23-24, 1994 SAC-94-01. Sacramento, CA, December, 1994.

References 12-6

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SAC . Steel Moment Frame Advisory No. 1. September, Sacramento, CA, 1994. SAC . Steel Moment Frame Advisory No. 2. October, Sacramento, CA, 1994. SAC . Steel Moment Frame Advisory No. 3 SAC-95-01, February, Sacramento, CA, 1995. Shonafelt, G.O., and Horn, W.B.. Guidelines for Evaluation and Repair of Damaged Steel Bridge Members, NCHRP Report 271, Transportation Research Board, 1984. Skiles, J.L. and Campbell, H.H., “Why Steel Fractured in the Northridge Earthquake” SAC Advisory No. 1, October, 1994. Seismic Safety Commission, Northridge Earthquake Turning Loss to Gain, Report to the Governor, Sacramento, CA, 1995. Smith Emery Company. Report of Test, July, 1995. Sommerville, P, Graves, R., Chandan, S. Technical Report: Characterization of Ground Motion During the Northridge Earthquake of January 17, 1994, SAC 95-03, SAC, December, 1995. State of California. Division of the State Architect (DSA) and Office of Statewide Health Planning and Development (OSHPD). Interpretation of Regulations Steel Moment Resisting Frames, Sacramento, CA, 1994. Structural Engineers Association of California (SEAOC), Seismology Committee, Recommended Lateral Force Requirements and Commentary, Sacramento, CA. 1990. Structural Engineers Association of California (SEAOC), Seismology Committee, Interim Recommendations for Design of Steel Moment Resisting Connection,. Sacramento, CA, January, 1995. Structural Engineers Association of California (SEAOC), Vision 2000: A Framework for Performance Based Engineering of Buildings, Sacramento, CA, April, 1995. Structural Shape Producers Council, Statistical Analysis of Tensile Data for Wide Flange Structural Shapes, 1994. Thiel, C.C., and Zsutty, T.C., “Earthquake Characteristics and Damage Statistics,” Earthquake Spectra, Volume 3, No. 4., Earthquake Engineering Research Institute, Oakland, Ca. 1987. Tremblay, R., Tchebotarev, N., and Filiatrault, A., “Seismic Performance of RBS Connections for Steel Moment Resisting Frames: Influence of Loading Rate and Floor Slab,” Proceedings of the Second International Conference on the Behavior of Steel Structures in Seismic Area, Kyoto, Japan, August, 1997

References 12-7

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Tsai, K.C. and Popov, E. P. “Seismic Steel Beam-Column Moment Connections” State of Art Papers: Metallurgy, Fracture Mechanics, Welding, Moment Connections and Frame System Behavior SAC 95-09. SAC, September, 1996 Uang, C.M. and Latham, C.T. Cyclic Testing of Full-Scale MNH-SMRF Moment Connections, Structural Systems Research, University of California, San Diego, March, 1995. Tsai, K.C. and Popov, E.P., Steel Beam - Column Joints In Seismic Moment Resisting Frames, Report No. UCB/EERC-88/19, Earthquake Engineering Research Center, University of California, Berkeley, Nov., 1988. Uang, C.M., Yu, Q.S., Sadre, A., Bonowitz, D., Youssef, N. “Performance of a 13 Story Steel Moment-Resisting Frame Damaged in the 1994 Northridge Earthquake”, ” Analytical and Field Investigations of Buildings Affected by the Northridge Earthquake of January 17, 1994, SAC 9504 Part 2 SAC, December, 1995. Uang, C.M. and Bondad, D. Progress Report on Cyclic Testing of Three Repaired UCSD Specimens, SAC, 1995. Uang, C.M. and Lee, C.H. “Seismic Response of Haunch Repaired Steel SMRFs: Analytical Modelling and Case Studies” ” Analytical and Field Investigations of Buildings Affected by the Northridge Earthquake of January 17, 1994, SAC 95-04 Part 2, SAC., December, 1995 Wald, D.J., Heaton, T.H., and Hudnut, K.W., The Slip History of the 1994 Northridge, California, Earthquake Determined from Strong-Motion, Teleseismic, GPS, and Leveling Data, United Sates Geologic Survey, 1995. Watabe, M. Peformance of Wooden Houses and Steel Buildings during the Great Hanshin Earthquake, Architectural Institute of Japan, May, 1995. Youssef, N.F.G, Bonowitz, D., and Gross, J.L., A Survey of Steel Moment-Resisting Frame Buildings Affected by the 1994 Northridge Earthquake, NISTR 5625, Gaithersburg Md, April, 1995.

References 12-8

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3. CLASSIFICATION AND IMPLICATIONS OF DAMAGE 3.1 Summary of Earthquake Damage There are no modifications to the Guidelines or Commentary of Section 3.1 at this time. 3.2 Damage Types There are no modifications to the Guidelines or Commentary of Section 3.2 at this time. 3.2.1 Girder Damage

There are no modifications to the Guidelines or Commentary of Section 3.2.1 at this time. 3.2.2 Column Flange Damage

There are no modifications to the Guidelines or Commentary of Section 3.2.2 at this time. 3.2.3 Weld Damage, Defects and Discontinuities

Six types of weld discontinuities, defects and damage are defined in Table 3-3 and illustrated in Figure 3-4. All apply to the complete joint penetration (CJP) welds between the girder flanges and the column flanges. This category of damage was the most commonly reported type fFollowing the Northridge Earthquake, many instances of W1a and W1b conditions were reported as damage. These conditions, which are detectable only by ultrasonic testing or by removal of weld backing, are now thought more likely to be construction defects than damage. Table 3-3 - Types of Weld Damage, Defects and Discontinuities Type W1 W1a W1b W2 W3 W4 W5

Description Weld root indications Incipient indications -– depth bf/4 Crack through weld metal thickness Fracture at girder interface Fracture at column interface Root indication— non-rejectable Partial crack at weld to column (beam flanges sound) Partial crack at weld to column (beam flange cracked) Crack in Supplemental Weld (beam flanges sound) Crack in Supplemental Weld (beam flange cracked) Fracture through tab at bolt holes Yielding or buckling of tab Damaged, or missing bolts4 Full length fracture of weld to column Fracture, buckle, or yield of continuity plate3 Fracture of continuity plate welds3 Yielding or ductile deformation of web3 Fracture of doubler plate welds3 Partial depth fracture in doubler plate3 Partial depth fracture in web3 Full (or near full) depth fracture in web or doubler plate3 Web buckling3 Fully severed column

Index2dj 4 1 8 8 10 4 10 8 4 8 8 8 6 8 8 01 04 8 8 8 0 4 8 1 8 10 6 6 10 4 4 1 4 4 8 8 6 10

Notes To Table 4-3a: 1. See Figures 3-2 through 3-6 for illustrations of these types of damage. 2. Where multiple damage types have occurred in a single connection, then: a. Sum the damage indices for all types of damage with d=1 and treat as one type. If multiple types still exist; then:

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b. For two types of damage refer to Table 4-3b. If the combination is not present in Table 4-3b and the damage indices for both types are greater than or equal to 4, use 10 as the damage index for the connection. If one is less than 4, use the greater value as the damage index for the connection. c. If three or more types of damage apply and at least one is greater than 4, use an index value of 10, otherwise use the greatest of the applicable individual indices. 3. Panel zone damage should be reflected in the damage index for all moment connections attached to the damaged panel zone within the assembly. 4. Missing or loose bolts may be a result of construction error rather than damage. The condition of the metal around the bolt holes, and the presence of fireproofing or other material in the holes can provide clues to this. Where it is determined that construction error is the cause, the condition should be corrected and a damage index of “0” assigned.

Table 4-3b - Connection Damage Indices for Common Damage Combinations1 Girder, Column or Weld Damage

Shear Tab Damage

Damage Index

Girder, Column or Weld Damage

Shear Tab Damage

Damage Index

G3 or G4

S1a S1b S2a S2b S3 S4 S5 S6 S1a S1b S2a S2b S3 S4 S5 S6 S1a S1b S2a S2b S3 S4 S5 S6

8 10 8 10 10 10 10 10 8 10 8 10 10 10 10 10 8 10 8 10 10 10 10 10

C5

S1a S1b S2a S2b S3 S4 S5 S6 S1a S1b S2a S2b S3 S4 S5 S6

6 10 6 10 10 10 10 10 8 10 8 10 10 10 10 10

C2

C3 or C4

1.

W2, W3, or W4

See Table 4-3a, footnote 2 for combinations other than those contained in this table.

More complete descriptions (including sketches) of the various types of damage are provided in Section 3.1. When the engineer can show by rational analysis that other values for the relative severities of damage are appropriate, these may be substituted for the damage indices provided in Post Earthquake Evaluation 4-7

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the tables. A full reporting of the basis for these different values should be provided to the building official, upon request. Commentary: The connection damage indices provided in Table 4-3 (ranging from 0 to 10) represent judgmental estimates of the relative severities of this damage. An index of 0 indicates no damage and an index of 10 indicates very severe damage. When initially developed, these connection damage indices were conceptualized as estimates of the connection’s lost capacity to reliably participate in the building’s lateral-force-resisting system in future earthquakes (with 0 indicating no loss of capacity and 10 indicating complete loss of capacity). However, due to the limited data available, no direct correlation between these damage indices and the actual residual strength and stiffness of a damaged connection was ever made. They do provide a convenient measure, however, of the extent of damage that various connections in a building have experienced. When FEMA-267 was first published, weld root discontinuities, Type W1a and defects, type W1b, were classified as damage in Table 4-3a with damage indices of 1 and 4, respectively assigned. Recent evidence and investigations, however, suggest strongly that these W1 conditions are not likely to be damage, and also are difficult to reliably detect. As a result, with the publication of Interim Guidelines Advisory No. 2, the damage indices for these conditions has been reduced to a null value, consistent with classifying them as pre-existing conditions, rather than damage. It should be noted that the reduced damage index associated with these conditions is not intended to indicate that these are not a concern with regard to future performance of the building. In particular, type W1b conditions can serve as ready initiators for the types of brittle fractures associated with the other damage types and connections having such conditions are more susceptible to future earthquake-induced damage than connections that do not have these conditions. Correction of these conditions should generally be considered an upgrade or modification, rather than a damage repair. 4.3.5 Step 4— Inspect Connections Adjacent to Damaged Connections

There are no modifications to the Guidelines or Commentary of Section 4.3.5 at this time. 4.3.6 Step 5— Determine Average Damage Index for Each Group

There are no modifications to the Guidelines or Commentary of Section 4.3.6 at this time.

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4.3.7 Step 6— Determine the Probability that the Connections in a Group at a Floor Level Sustained Excessive Damage

There are no modifications to the Guidelines or Commentary of Section 4.3.7 at this time. 4.3.7.1 Some Connections in Group Not Inspected

There are no modifications to the Guidelines or Commentary of Section 4.3.7.1 at this time. 4.3.7.2 All Connections in Group Inspected

There are no modifications to the Guidelines or Commentary of Section 4.3.7.2 at this time. 4.3.8 Step 7— Determine Recommended Recovery Strategies for the Building

There are no modifications to the Guidelines or Commentary of Section 4.3.8 at this time. 4.3.9 Step 8 - Evaluation Report

There are no modifications to the Guidelines or Commentary of Section 4.3.9 at this time. 4.4 Alternative Group Selection for Torsional Response There are no modifications to the Guidelines or Commentary of Section 4.4 at this time. 4.5 Qualified Independent Engineering Review There are no modifications to the Guidelines or Commentary of Section 4.5 at this time. 4.5.1 Timing of Independent Review

There are no modifications to the Guidelines or Commentary of Section 4.5.1 at this time. 4.5.2 Qualifications and Terms of Employment

There are no modifications to the Guidelines or Commentary of Section 4.5.2 at this time. 4.5.3 Scope of Review

There are no modifications to the Guidelines or Commentary of Section 4.5.3 at this time. 4.5.4 Reports

There are no modifications to the Guidelines or Commentary of Section 4.5.4 at this time.

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4.5.5 Responses and Corrective Actions

There are no modifications to the Guidelines or Commentary of Section 4.5.5 at this time. 4.5.6 Distribution of Reports

There are no modifications to the Guidelines or Commentary of Section 4.5.6 at this time. 4.5.7 Engineer of Record

There are no modifications to the Guidelines or Commentary of Section 4.5.7 at this time. 4.5.8 Resolution of Differences

There are no modifications to the Guidelines or Commentary of Section 4.5.8 at this time.

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5. POST-EARTHQUAKE INSPECTION When required by the building official, or recommended by the Interim Guidelines in Chapter 4, post-earthquake inspections of buildings may be conducted in accordance with the Interim Guidelines of this Chapter. In order to determine, with certainty, the actual post-earthquake condition of a building, it is necessary to inspect all elements and their connections. However, it is permissible to select An an appropriate sample (or samples) of WSMF connections should be selected for inspection in accordance with the Chapter 4 Guidelines. These connections, and others deemed appropriate by the engineer, should be subjected to visual inspection (VI) and supplemented by non-destructive testing (NDT) as required by this Chapter. Commentary: The only way to be certain that all damage sustained by a building is detected is to perform complete inspections of every structural element and connection. In most cases, such exhaustive post-earthquake inspections would be both economically impractical and also unnecessary. As recommended by these guidelines, the purpose of post-earthquake inspections is not to detect all damage that has been sustained by a building, but rather, to detect with reasonable certainty, that damage likely to result in a significant degradation in the building’s ability to resist future loading. The connection sampling process, suggested by Chapter 4 of these Interim Guidelines was developed to provide a low probability that damage in buildings that had sustained a substantial reduction in load carrying capacity would be overlooked while avoiding the performance of exhaustive investigations of buildings that have sustained relatively insignificant damage. Where greater certainty in the detection of damage is desired for a building, a more extensive program of inspection can be conducted. For those cases in which it is desired to perform an analytical determination of the residual load carrying capacity of the structure, complete inspections of elements and connections should be performed so that an analytical model of the building can be developed that reasonably represents its post-earthquake condition. 5.1 Connection Types Requiring Inspection 5.1.1 Welded Steel Moment Frame (WSMF) Connections

The inspection of a WSMF connection should start with visual inspection of the welded bottom beam flange to column flange joint and the base materials immediately adjacent to this joint. If damage to this joint is apparent, or suspected, then inspections of that connection should be extended to include the complete joint penetration (CJP) groove welds connecting both top and bottom beam flanges to the column flange, including the backing bar and the weld access holes in the beam web; the shear tab connection, including the bolts, supplemental welds and Post-Earthquake Inspection 5-1

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beam web; the column's web panel zone, including doubler plates; and the continuity plates and continuity plate welds (See Figure 3-1). In addition, where visual inspection indicates potential concealed damage, visual inspection should be supplemented with other methods of nondestructive testing. Commentary: The largest concentration of reported damage following the Northridge Earthquake occurred at the welded joint between the bottom girder flange and column, or in the immediate vicinity of this joint. To a much lesser extent, damage was also observed in some buildings at the joint between the top girder flange and column. If damage at either of these locations is substantial (dj per Chapter 4 greater than 5), then damage is also commonly found in the panel zone or shear tab areas. When originally published,These these Interim Guidelines recommended complete inspection, by visual and NDT assisted means, of all of these potential damage areas for a small representative sample of connections. This practice is was consistent with that followed by most engineers in the Los Angeles area, following the Northridge Earthquake. It requires removal of fireproofing from a relatively large surface of the steel framing, which at most connections will be undamaged. In the time since the Interim Guidelines were first published, extensive investigations have been conducted of the statistical distribution of damage sustained by buildings in the Northridge earthquake, the nature of this damage and the effect of this damage on the future load-carrying capacity of the buildings. These investigations strongly suggest that the W1a and W1b conditions at the weld root are unlikely to be earthquake damage, but rather, conditions of discontinuity and defects from the original construction. Further, studies have shown that NDT methods are generally unreliable in the detection of these conditions. As a result, the current recommendation is not to conduct exhaustive NDT investigations of connections in order to discover hidden damage, as was originally recommended. In a series of analytical investigations of the effect of moment-resisting connection damage on building behavior, it was determined that even if a large number of connections experience fracture at one beam flange to column joint, there is relatively little increase in the probability of global collapse in a future earthquake. Similarly, these investigations indicate that if both the top and bottom beam flange to column joints fracture in a large a number of connections, a very significant increase in the probability of global building collapse occurs. Therefore, to reduce the costs associated with post-earthquake inspections, with the publication of Interim Guidelines Advisory No.2 it is recommended that postearthquake inspections initially be limited to visual inspection of the beam bottom Post-Earthquake Inspection 5-2

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flange to column joint region. If there is evidence of potential damage in this region that is not directly observable by visual means, for example, a gap between the weld backing and column flange, then supplemental investigations of this joint should be conducted using NDT. Similarly, if it is determined that fractures have occurred at the beam bottom flange joint, then inspections of that connection should be extended to encompass the entire connection including the top beam flange joint, the shear tab and column panel zone. This approach was permitted as an alternate, in the original publication of the Interim Guidelines. Some engineers have suggested an alternative approach consisting of visual only inspections, limited to the girder bottom flange to column joint, but for a very large percentage of the total connections in the building. These bottom flange joint connections can be visually inspected with much less fireproofing removed from the framing surfaces. When significant damage is found at the exposed bottom connection, then additional fireproofing is removed to allow full exposure of the connection and inspection of the remaining surfaces. These engineers feel that by inspecting more connections, albeit to a lesser scope than recommended in these Interim Guidelines, their ability to locate the most severe occurrences of damage in a building is enhanced. These engineers use NDT assisted inspection on a very small sample of the total connections exposed to obtain an indication of the likelihood of hidden problems including damage types. If properly executed, such an approach can provide sufficient information to evaluate the post-earthquake condition of a building and to make appropriate occupancy, structural repair and/or modification decisions. It is important that the visual inspector be highly trained and that visual inspections be carefully performed, preferably by a structural engineer. Casual observation may miss clues that hidden damage exists. If, as a result of the partial visual inspection, there is any reason to believe that damage exists at a connection (such as small gaps between the CJP weld backing and column face), then complete inspection of the suspected connection, in accordance with the recommendations of these Interim Guidelines should be performed. If this approach is followed, it is recommended that a significantly larger sample of connections than otherwise recommended by these Interim Guidelines, perhaps nearly all of the connections, be inspected. 5.1.2 Gravity Connections

There are no modifications to the Guidelines or Commentary of Section 5.1.2 at this time. 5.1.3 Other Connection Types

There are no modifications to the Guidelines or Commentary of Section 5.1.3 at this time. Post-Earthquake Inspection 5-3

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5.2 Preparation 5.2.1 Preliminary Document Review and Evaluation 5.2.1.1 Document Collection and Review

There are no modifications to the Guidelines or Commentary of Section 5.2.1.1 at this time. 5.2.1.2 Preliminary Building Walk-Through.

There are no modifications to the Guidelines or Commentary of Section 5.2.1.2 at this time. 5.2.1.3 Structural Analysis

There are no modifications to the Guidelines or Commentary of Section 5.2.1.3 at this time. 5.2.1.4 Vertical Plumbness Check

There are no modifications to the Guidelines or Commentary of Section 5.2.1.4 at this time. 5.2.2 Connection Exposure

Pre-inspection activities to expose and prepare a connection for inspection should include the local removal of suspended ceiling panels or (as applicable) local demolition of permanent ceiling finish to access the connection; and cleaning of sufficient fireproofing from the beam and column surfaces to allow visual observation of the area to be inspected. If initial inspections are to be limited to the beam bottom flange to column joint and the surrounding material, fireproofing should be removed from the connection as indicated in Figure 5-1a. Removal of fireproofing need only be sufficient to permit observation of the surfaces of base and weld metals. Wire brushing and cleaning to remove all particles of fireproofing material is not necessary unless ultrasonic testing of the joint area is to be conducted. In the event that damage is found at the bottom beam flange to column joint, then additional fireproofing should be removed, as indicated in Figure 51b, to expose the column panel zone, the column flange, continuity plates, beam web and flanges. The extent of the removal of fireproofing should be sufficient to allow adequate inspection of the surfaces to be inspected. Figure 5-1b suggests a pattern that will allow both visual and NDT inspection of the top and bottom beam flange to column joints, the beam web and shear connection, column panel zone and continuity plates, and column flanges in the areas of highest expected demands. The maximum extent of the removal of fireproofing need not be greater than a distance equal to the beam depth "d" into the beam span to expose evidence of any yielding.

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Exposed surfaces

6”

6” 6” Fireproofing

Figure 5-1a Recommended Zone for Fireproofing Removal for Initial Inspections

6”

12” 6” Fireproofing

Figure 5-1b Recommended Zone for Removal of Fireproofing for Complete Inspections Commentary: If inspection is to be limited to visual observation of the surfaces of the base metal and welds, cleaning of fireproofing need only be sufficient to expose these surfaces. However, if ultrasonic testing is to be performed, the surface over which the scanning will be performed must be free Cleaning of weld areas and removal of mill scale and weld spatter. Such cleaning should be done with care, preferably using a power wire brush, to ensure a clean surface that does not affect the accuracy of ultrasonic testing. The resulting surface finish should be clean, free of mill scale, rust and foreign matter. The use of a chisel should be avoided to preclude scratching the steel surfaces which could be mistaken for yield lines. Sprayed-on fireproofing on WSMFs erected prior to about 19801970 is likely to contain asbestos and should be handled according to Post-Earthquake Inspection 5-5

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applicable standards for the removal of hazardous materials. Health hazards associated with asbestos were recognized by industry in the late 1960s and by 1969, most commercial production of asbestos containing materials had ceased. In April, 1973, the federal government formally prohibited the production of asbestos containing materials with the adoption of the National Emission Standards for Hazardous Air Pollutants. Allowing for shelf life of materials produced prior to that date, it should be considered possible that buildings constructed prior to 1975 contain some asbestos hazards. To preclude physical exposure to hazardous materials and working conditions in such buildings, the structural engineer should require by contractual agreement with the building owner, prior to the start of the inspection program, that the building owner deliver to the structural engineer for his/her review and files a laboratory certificate that confirms the absence of asbestos in structural steel fireproofing, local pipe insulation, ceiling tiles, and drywall joint compound. The pattern of fireproofing removal indicated in Figure 5-1 is adequate to allow visual and UT inspection of the top and bottom girder flange to column joints, the beam web and shear connection and the column panel zone. As discussed in the commentary to Section 5.1.1, some engineers prefer to initially inspect only the bottom beam flange to column joint. In such cases, the initial removal of fireproofing can be more limited than indicated in the figure. If after initial inspection, damage at a connection is suspected, then full removal, as indicated in the figure, should be performed to allow inspection of all areas of the connection. 5.3 Inspection Program 5.3.1 Visual Inspection (VI)

There are no modifications to the Guidelines or Commentary of Section 5.3.1 at this time. 5.3.1.1 Top Flange

There are no modifications to the Guidelines or Commentary of Section 5.3.1.1 at this time. 5.3.1.2 Bottom Flange

There are no modifications to the Guidelines or Commentary of Section 5.3.1.2 at this time. 5.3.1.3 Column and Continuity Plates

There are no modifications to the Guidelines or Commentary of Section 5.3.1.3 at this time.

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5.3.1.4 Beam Web Shear Connection

There are no modifications to the Guidelines or Commentary of Section 5.3.1.4 at this time. 5.3.2 Nondestructive Testing (NDT)

NDT should may be used to supplement the visual inspection of connections selected in accordance with the Interim Guidelines of Chapter 4. The testing agency and NDT personnel performing this work should conform to the qualifications indicated in Chapter 11 of these Interim Guidelines. The following NDT techniques should may be used at the top and bottom of each connection, where accessible, to supplement visual inspection: These techniques should be used whenever visual inspection indicates the potential for damage that is not directly observable. a) Magnetic particle testing (MT) of the beam flange to column flange weld surfaces may be used to confirm the presence of suspected surface cracks based on visual evidence. Where fractures are evident from visual inspection, MT should be used to confirm the lateral extent of the fracture.All surfaces which were visually inspected should be tested using the magnetic particle technique. Commentary: The color of powder should be selected to achieve maximum contrast to the base and weld metal under examination. The test may be further enhanced by applying a white coating made specifically for MT or by applying penetrant developer prior to the MT examination. This background coating should be allowed to thoroughly dry before performing the MT. b) Ultrasonic testing (UT) may be used to detect the presence of hidden fractures, where visual inspection reveals the potential for such fractures. of all faces at the beam flange welds and adjacent column flanges (extending at least 3 inches above and below the location of the CJP weld, along the face of the column, but not less than 1-1/2 times the column flange thickness). Commentary: The purpose of UT is to 1) locate and describe the extent of internal defects not visible on the surface and 2) to determine the extent of cracks observed visually and by MT. These guidelines recommend the use of visual inspection as the primary tool for detecting earthquake damage (See commentary to Sec. 5..1.1). UT can be a useful technique for confirmation of the presence of suspected fractures at the beam flange to column flange joints. Visual evidence that may suggest the need for such testing could include apparent separation of the base of the weld backing from the face of the column. Requirements and acceptance criteria for NDT should be as given in AWS D1.1-98 Sections 6 and 8. Acceptance or rejection of planar weld discontinuity (cracks, slag inclusion, or lack of fusion), including root indications, should, as a minimum, be consistent with AWS Discontinuities Severity Class designations of cracks and defects per Table 8.26.2 of AWS D1.1-98 for Static Post-Earthquake Inspection 5-7

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Structures. Beam flange welds should be tested as "tension welds" per AWS D1.1 Table 8.15.3, Note 3. Backing bars need not be removed prior to performing UT. Commentary: The value of UT for locating small discontinuities at the root of beam flange to column flange welds when the backing is left in place is not universally accepted. The reliability of this technique is particularly questionable at the center of the joint, where the beam web obscures the signal. There have been a number of reported instances of UT detected indications which were not found upon removal of the backing, and similarly, there have been reported instances of defects which were missed by UT examination but were evident upon removal of the backing. The smaller the defect, the less likely it is that UT alone will reliably detect its presence. Despite the potential inaccuracies of this technique, it is the only method currently available, short of removal of the backing, to find subsurface damage in the welds. It is also the most reliable method for finding lamellar problems in the column flange (type C4 and C5 damage) opposite the girder flange. Removal of weld backing at these connections results in a significant cost increase that is probably not warranted unless UT indicates widespread, significant defects and/or damage in the building. The proper scanning techniques, beam angle(s) and transducer sizes should be used as specified in the written UT procedure contained in the Written Practice, prepared in accordance with Section 5.3.3 of these Interim Guidelines. The acceptance standard should be that specified in the original contract documents, but in no case should it be less than the acceptance criteria of AWS D1.1, Chapter 8, for Statically Loaded Structures. The base metal should be scanned with UT for cracks. Cracks which have propagated to the surface of the weld or beam and column base metal will probably have been detected by visual inspection and magnetic particle tests performed earlier. The purpose of ultrasonic testing of the base metal is to: 1. Locate and describe the extent of internal indications not apparent on the surface and, 2. Determine the extent of cracks found visually and by magnetic particle test. Commentary: Liquid dye penetrant testing (PT) may be used where MT is precluded due to geometrical conditions or restricted access. Note that more stringent requirements for surface preparation are required for PT than MT, per AWS D1.1. If practical, NDT should be performed across the full width of the bottom beam flange joint. However, if there are no discontinuity signals from UT of Post-Earthquake Inspection 5-8

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accessible faces on one side of the bottom flange weld, obstructions on the other side of the connection need not be removed for testing of the bottom flange weld. Slabs, flooring and roofing need not be removed to permit NDT of the top flange joint unless there is significant visible damage at the bottom beam flange, adjacent column flange, column web, or shear connection. Unless such damage is present, NDT of the top flange should be performed as permitted, without local removal of the diaphragms or perimeter wall obstructions. It should be noted that UT is not 100% effective in locating discontinuities and defects in CJP beam flange to column flange welds. The ability of UT to reliably detect such defects is very dependent on the skill of the operator and the care taken in the inspection. Even under perfect conditions, it is difficult to obtain reliable readings of conditions at the center of the beam flange to column flange connection as return signals are obscured by the presence of the beam web. If backing is left in place on the welds, UT becomes even less reliable. There have been a number of reported instances in which UT indicated apparent defects, that were found not to exist upon removal of the backing. Similarly, UT has failed in some cases to locate defects that were later discovered upon removal of the backing. Additional information on UT may be found in AWS B1.10. 5.3.3 Inspector Qualification 5.3.4 Post-Earthquake Field Inspection Report

There are no modifications to the Guidelines or Commentary of Section 5.3.4 at this time. 5.3.5 Written Report

There are no modifications to the Guidelines or Commentary of Section 5.3.5 at this time.

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6. POST-EARTHQUAKE REPAIR AND MODIFICATION 6.1 Scope There are no modifications to the Guidelines or Commentary of Section 6.1 at this time. 6.2 Shoring There are no modifications to the Guidelines or Commentary of Section 6.2 at this time. 6.3 Repair Details There are no modifications to the Guidelines or Commentary of Section 6.3 at this time. 6.4 Preparation There are no modifications to the Guidelines or Commentary of Section 6.4 at this time. 6.5 Execution There are no modifications to the Guidelines or Commentary of Section 6.5 at this time. 6.6 STRUCTURAL MODIFICATION 6.6.1 Definition of Modification

There are no modifications to the Guidelines or Commentary of Section 6.6.1 at this time. 6.6.2 Damaged vs. Undamaged Connections

There are no modifications to the Guidelines or Commentary of Section 6.6.2 at this time. 6.6.3 Criteria

Connection modification intended to permit inelastic frame behavior should be proportioned so that the required plastic deformation of the frame may be accommodated through the development of plastic hinges at pre-determined locations within the girder spans, as indicated in Figure 6-12 Figure 6.6.3-1. Beam-column connections should be designed with sufficient strength (through the use of cover plates, haunches, side plates, etc.) to force development of the plastic hinge away from the column face. This condition may also be attained through local weakening of the beam section, at the desired location for plastic hinge formation. All elements of the connection should have adequate strength to develop the forces resulting from the

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formation of the plastic hinge at the predetermined location, together with forces resulting from gravity loads.

h

Undeformed frame

Deformed frame shape

Plastic Hinges

drift angle - θ

L’ L

Figure 6-12 Figure 6.6.3-1 - Desired Plastic Frame Behavior Commentary: Nonlinear deformation of frame structures is typically accommodated through the development of inelastic flexural or shear strains within discrete regions of the structure. At large inelastic strains these regions can develop into plastic hinges, which can accommodate significant concentrated rotations at constant (or nearly constant) load through yielding at tensile fibers and buckling at compressive fibers. If a sufficient number of plastic hinges develop in a frame, a mechanism is formed and the frame can deform laterally in a plastic manner. This behavior is accompanied by significant energy dissipation, particularly if a number of members are involved in the plastic behavior, as well as substantial local damage to the highly strained elements. The formation of hinges in columns, as opposed to beams, is undesirable, as this results in the formation of weak story mechanisms with relatively few elements participating, and consequently little energy dissipation occurring. In addition, such mechanisms also result in local damage to critical gravity load bearing elements. The prescriptive connection contained in the UBC and NEHRP Recommended Provisions prior to the Northridge Earthquake was based on the assumed development of plastic hinge zones within the beams at adjacent to the face of the column, or within the column panel zone itself. If the plastic hinge develops in the column panel zone, the resulting column deformation results in very large secondary stresses on the beam flange to column flange joint, a condition which can contribute to brittle failure. If the plastic hinge forms in the beam, at the face of the column, this can result in very large through-thickness strain demands on Post-earthquake Repair and Modification 6-2

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the column flange material and large inelastic strain demands on the weld metal and surrounding heat affected zones stress and strain demands on the welded beam flange to column flange joint. These conditions can also lead to brittle joint failure. Although ongoing research may reveal conditions of material properties, design and detailing configurations that permit connections with yielding occurring at the column face to perform reliably, for the present, it is recommended In order to achieve more reliable performance, it is recommended that the connection of the beam to the column be modified to be sufficiently strong to force the inelastic action (plastic hinge) away from the column face. Plastic hinges in steel beams have finite length, typically on the order of half the beam depth. Therefore, the location for the plastic hinge should be shifted at least that distance away from the face of the column. When this is done, the flexural demands on the columns are increased. Care must be taken to assure that weak column conditions are not inadvertently created by local strengthening of the connections. It should be noted that connection modifications of the type described above, while believed to be effective in preventing brittle connection fractures, will not prevent structural damage from occurring. Brittle connection fractures are undesirable because they result in a substantial reduction in the lateral-forceresisting strength of the structure which, in extreme cases, can result in instability and collapse. Connections modified as described in these Interim Guidelines should experience many fewer such brittle fractures than unmodified connections. However, the formation of a plastic hinge within the span of a beam is not a completely benign event. Beams which have formed such hinges may exhibit large buckling and yielding deformation, damage which typically must be repaired. The cost of such repairs could be comparable to the costs incurred in repairing fracture damage experienced in the Northridge Earthquake. The primary difference is that life safety protection will be significantly enhanced and most structures that have experienced such plastic deformation damage should continue to be safe for occupancy while repairs are made. If the types of damage described above are unacceptable for a given building, then alternative methods of structural modification should be considered that will reduce the plastic deformation demands on the structure during a strong earthquake. Appropriate methods of achieving such goals include the installation of supplemental braced frames, energy dissipation systems, and similar systematic modifications of the building’s basic lateral force resisting system. It is important to recognize that in frames with relatively short bays, the flexural hinging indicated in Figure 6.6.3-1 may not be able to form. If the effective flexural length (L’in the figure) of beams in a frame becomes too short, then the beams or girders will yield in shear before zones of flexural plasticity

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can form, resulting in an inelastic behavior that is more like that of an eccentrically braced frame than that of a moment frame. This behavior may inadvertently occur in frames in which relatively large strengthened connections, such as haunches, cover plates or side plates have been used on beams with relatively short spans. This behavior is illustrated in Figure 6.6.3-2. The guidelines contained in this section are intended to address the design of flexurally dominated moment resisting frames. When utilizing these guidelines, it is important to confirm that the configuration of the structure is such that the presumed flexural hinging can actually occur. It is possible that shear yielding of frame beams, such as that schematically illustrated in Figure 6.6.3-2 may be a desirable behavior mode. However, to date, there has not been enough research conducted into the behavior of such frames to develop recommended design guidelines. If modifications to an existing frame result in such a configuration designers should consider referring to the code requirements for eccentrically braced frames. Particular care should be taken to brace the shear link of such beams against lateral-torsional buckling and also to adequately stiffen the webs to avoid local buckling following shear plastification. Shear Link

Shear Link

Figure 6.6.3-2 Shear Yielding Dominated Behavior of Short Bay Frames 6.6.4 Strength and Stiffness 6.6.4.1 Strength

When these Interim Guidelines require determination of the strength of a framing element or component, this shall be calculated in accordance with the criteria contained in UBC-94, Section

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2211.4.2 {NEHRP-91 Section 10.2, except that the factor φshould be taken as 1.0}, restated as follows: 2211.4.1 Member strength. Where this section requires that the strength of the member be developed, the following shall be used: Flexure Shear Axial compression Axial tension Connectors Full Penetration welds Partial Penetration welds Bolts and fillet welds

Ms = Z F y Vs = 0.55 Fy d t Psc = 1.7 Fa A Pst = Fy A Fy A 1.7 allowable (see commentary) 1.7 allowable

Alternatively, the criteria contained in the 1997 edition of the AISC Seismic Provisions for Structural Steel Buildings (AISC, 1997) may be used. Commentary: At the time the Interim Guidelines were first published, they were based on the 1994 edition of the Uniform Building Code and the 1994 edition of the NEHRP Provisions. In the time since that initial publication, more recent editions of both documents have been published, and codes based on these documents have been adopted by some jurisdictions. In addition, the American Institute of Steel Construction has adopted a major revision to its Seismic Provisions for Structural Steel Buildings (AISC Seismic Provisions), largely incorporating, with some modification, the recommendations contained in the Interim Guidelines. This updated edition of the AISC Seismic Provisions has been incorporated by reference into the 1997 edition of the NEHRP Provisions and has also been adopted by some jurisdictions as an amendment to the model building codes. Structural upgrades designed to comply with the requirements of the 1997 AISC Seismic Provisions may be deemed to comply with the intent of these Interim Guidelines. Where reference is made herein to the requirements of the 1994 Uniform Building Code or 1994 NERHP Provisions, the parallel provisions of the 1997 editions may be used instead, and should be used in those jurisdictions that have adopted codes based on these updated standards. Partial penetration welds are not recommended for tension applications in critical connections resisting seismic induced stresses. The geometry of partial penetration welds creates a notch-like condition that can initiate brittle fracture under conditions of high tensile strain. Many WSMF structures are constructed with concrete floor slabs that are provided with positive shear attachment between the slab and the top flanges of the girders of the moment-resisting frames. Although not generally accounted for in the design of the frames, the resulting composite action can increase the

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effective strength of the girder significantly, particularly at sections where curvature of the girder places the top flange into compression. Although this effect is directly accounted for in the design of composite systems, it is typically neglected in the design of systems classified as moment resisting steel frames. The increased girder flexural strength caused by this composite action can result in a number of effects including the unintentional creation of weak column strong beam and weak panel zone conditions. In addition, this composite effect has the potential to reduce the effectiveness of reduced section or “dog-bone” type connection assemblies. Unfortunately, very little laboratory testing of large scale connection assemblies with slabs in place has been performed to date and as a result, these effects are not well quantified. In keeping with typical contemporary design practice, the design formulae provided in these Guidelines neglect the strengthening effects of composite action. Designers should, however, be alert to the fact that these composite effects do exist. Similar, and perhaps more severe, effects may also exist where steel beams support masonry or concrete walls. 6.6.4.2 Stiffness

Calculation of frame stiffness for the purpose of determining interstory drift under the influence of the design lateral forces should be based on the properties of the bare steel frame, neglecting the effects of composite action with floor slabs. The stiffening effects of connection reinforcements (e.g.: haunches, side plates, etc.) may be considered in the calculation of overall frame stiffness and drift demands. When reduced beam section connections are utilized, the reduction in overall frame stiffness, due to local reductions in girder cross section, should be considered. Commentary: For design purposes, frame stiffness is typically calculated considering only the behavior of the bare frame, neglecting the stiffening effects of slabs, gravity framing, and architectural elements such as walls. The resulting calculation of building stiffness and period typically underestimates the actual properties, substantially. Although this approach can result in unconservative estimates of design force levels, it typically produces conservative estimates of interstory drift demands. Since the design of most moment-resisting frames are controlled by considerations of drift, this approach is considered preferable to methods that would have the potential to over-estimate building stiffness. Also, many of the elements that provide additional stiffness may be subject to rapid degradation under severe cyclic lateral deformation, so that the bare frame stiffness may provide a reasonable estimate of the effective stiffness under long duration ground shaking response. Notwithstanding the above, designers should be alert to the fact that unintentional stiffness introduced by walls and other non-structural elements can

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significantly alter the behavior of the structure in response to ground shaking. Of particular concern, if these elements are not uniformly distributed throughout the structure, or isolated from its response, they can cause soft stories and torsional irregularities, conditions known to result in poor behavior. 6.6.5 Plastic Rotation Capacity

The plastic rotation capacity of modified connections should reflect realistic estimates of the required level of plastic rotation demand. In the absence of detailed calculations of rotation demand, connections should be shown to be capable of developing a minimum plastic rotation capacity on the order of 0.025 to 0.030 radian. The demand may be lower when braced frames, supplemental damping, base isolation, or other elements are introduced into the moment frame system, to control its lateral deformation; when the design ground motion is relatively low in the range of predominant periods for the structure; and when the frame is sufficiently strong and stiff. As used in these Guidelines, plastic rotation is defined as the plastic chord rotation angle. The plastic chord rotation angle is calculated using the rotated coordinate system shown in Fig. 6.6.51 as the plastic deflection of the beam or girder, at the point of inflection (usually at the center of its span,) ∆CI, divided by the distance between the center of the beam span and the centerline of the panel zone of the beam column connection, LCL. This convention is illustrated in Figure 6.6.51. It is important to note that this definition of plastic rotation is somewhat different than the plastic rotation that would actually occur within a discrete plastic hinge in a frame model similar to that shown in Figure 6.6.3-1. These two quantities are related to each other, however, and if one of them is known, the other may be calculated from Eq. 6.6.5-1. cL

LCL

Plastic hinge

Beam span center line

cL

∆CL

lh

θp =

∆ CL

LCL

Figure 6.6.5-1 Calculation of Plastic Rotation Angle

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θ p = θ ph where: θp θph LCL lh

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( LCL − lh ) LCL

(6.6.5-1)

is the plastic chord angle rotation, as used in these Guidelines is the plastic rotation, at the location of a discrete hinge is the distance from the center of the beam span to the center of the beam-column assembly panel zone is the assumed location of the discrete plastic hinge relative to the center of the beam-column assembly panel zone

If calculations are performed to determine the required connection plastic rotation capacity, the capacity should be taken somewhat greater than the calculated deformation demand, due to the high variability and uncertainty inherent in predictions of inelastic seismic response. Until better guidelines become available, a required plastic rotation capacity on the order of 0.005 radians greater than the demand calculated for the design basis earthquake (or if greater conservatism is desired - the maximum capable considered earthquake) is recommended. Rotation demand calculations should consider the effect of plastic hinge location within the beam span, as indicated in Figure 6-12 Figure 6.6.3-1, on plastic rotation demand. Calculations should be performed to the same level of detail specified for nonlinear dynamic analysis for base isolated structures in UBC-94 Section 1655 {NEHRP-94 Section 2.6.4.4}. Ground motion time histories utilized for these nonlinear analyses should satisfy the scaling requirements of UBC-94 Section 1655.4.2 {NEHRP-94 Section 2.6.4.4} except that instead of the base isolated period, TI, the structure period, T, calculated in accordance with UBC-94 Section 1628 {NEHRP-94 Section 2.3.3.1} should be used. Commentary. When the Interim Guidelines were first published, the plastic rotation was defined as that rotation that would occur at a discrete plastic hinge, similar to the definition of θph. in Eq. 6.6.5-1, above. In subsequent testing of prototype connection assemblies, it was found that it is often very difficult to determine the value of this rotation parameter from test data, since actual plastic hinges do not occur at discrete points in the assembly and because some amount of plasticity also occurs in the panel zone of many assemblies. The plastic chord angle rotation, introduced in Interim Guidelines Advisory No. 1, may more readily be obtained from test data and also more closely relates to the drift experienced by a frame during earthquake response. Traditionally, structural engineers have calculated demand in moment frames by sizing the members for strength and drift using code forces (either equivalent static or reduced dynamic forces) and then "developing the strength of the members." Since 1988, "developing the strength" has been accomplished by prescriptive means. It was assumed that the prescribed connections would be strong enough so that the girder would yield (in bending), or the panel zone

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would yield (in shear) in a nearly perfectly plastic manner producing the plastic rotations necessary to dissipate the energy of the earthquake. It is now known that the prescriptive connection is often incapable of behaving in this manner. In the 1994 Northridge earthquake, many moment-frame connections fractured with little evidence of plastic hinging of the girders or yielding of the column panel zones. Testing of moment frame connections both prior to and subsequent to the earthquake suggests that the standard welded flange-bolted web connection is unable to reliably provide plastic rotations beyond about 0.005 radian for all ranges of girder depths and often fails below that level. Thus, for frames designed for code forces and for the code drift limits, new connection configurations must be developed to reliably accommodate such rotation without brittle fracture. In order to develop reasonable estimates of the plastic rotation demands on a frame’s connections, it is necessary to perform inelastic time history analyses. For regular structures, approximations of the plastic rotation demands can be obtained from linear elastic analyses. Analytical research (Newmark and Hall 1982) suggests that for structures having the dynamic characteristics of most WSMF buildings, and for the ground motions typical of western US earthquakes, the total frame deflections obtained from an unreduced (no R or Rw factor) dynamic analysis provide an approximate estimate of those which would be experienced by the inelastic structure. For the typical spectra contained in the building code, this would indicate expected drift ratios on the order of 1%. The drift demands in a real structure, responding inelastically, tend to concentrate in a few stories, rather than being uniformly distributed throughout the structure’s height. Therefore, it is reasonable to expect typical drift demands in individual stories on the order of 1.5% to 2% of the story height. As a rough approximation, the drift demand may be equated to the joint rotation demand, yielding expected rotation demands on the order of perhaps 2%. Since there is considerable variation in ground motion intensity and spectra, as well as the inelastic response of buildings to these ground motions, conservatism in selection of an appropriate connection rotation demand is warranted. In recent testing of large scale subassemblies incorporating modified connection details, conducted by SAC and others, when the connection design was able to achieve a plastic rotation demand of 0.025 radians or more for several cycles, the ultimate failure of the subassembly generally did not occur in the connection, but rather in the members themselves. Therefore, the stated connection capacity criteria would appear to result in connections capable of providing reliable performance. It should be noted that the connection assembly capacity criteria for the modification of existing buildings, recommended by these Interim Guidelines, is Post-earthquake Repair and Modification 6-9

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somewhat reduced compared to that recommended for new buildings (Chapter 7). This is typical of approaches normally taken for existing structures. For new buildings, these Interim Guidelines discourage building-specific calculation of required plastic rotation capacity for connections and instead, encourage the development of highly ductile connection designs. For existing buildings, such an approach may lead to modification designs that are excessively costly, as well as the modification of structures which do not require such modification. Consequently, an approach which permits the development of semi-ductile connection designs, with sufficient plastic rotation capacity to withstand the expected demands from a design earthquake is adopted. It should be understood that buildings modified to this reduced criteria will not have the same reliability as new buildings, designed in accordance with the recommendations of Chapter 7. The criteria of Chapter 7 could be applied to existing buildings, if superior reliability is desired. When performing inelastic frame analysis, in order to determine the required connection plastic rotation capacity, it is important to accurately account for the locations at which the plastic hinges will occur. Simplified models, which represent the hinge as occurring at the face of the column, maywill underestimate the plastic rotation demand. This problem becomes more severe as the column spacing, L, becomes shorter and the distance between plastic hinges, L’, a greater portion of the total beam span. Eq. 6.6.5-1 may be used to convert calculated values of plastic rotation at a hinge remotely located from the column, to the chord angle rotation, used for the definition of acceptance criteria contained in these Guidelines. In extreme cases, the girder will not form plastic hinges at all, but instead, will develop a shear yield, similar to an eccentric braced frame. 6.6.6

Connection Qualification and Design

Modified girder-column connections may be qualified by testing or designed using calculations. Qualification by testing is the preferred approach. Preliminary designs of connections to be qualified by test may be obtained using the calculation procedures of Section 6.6.6.3. The procedures of that section may also be used to calibrate previous tests of similar connection configurations to slightly different applications, by extrapolation. Extrapolation of test results should be limited to connections of elements having similar geometries and material specifications as the tested connections. Designs based on calculation alone should be subject to qualified independent third party review. Commentary: Because of the cost of testing, use of calculations for interpolation or extrapolation of test results is desirable. How much extrapolation should be accepted is a difficult decision. As additional testing is done, more information may be available on what constitutes "conservative" testing conditions, thereby

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allowing easier decisions relative to extrapolating tests to actual conditions which are likely to be less demanding than the tests. For example, it is hypothesized that connections of shallower, thinner flanged members are likely to be more reliable than similar connections consisting of deeper, thicker flanged members. Thus, it may be possible to test the largest assemblages of similar details and extrapolate to the smaller member sizes? - at least within comparable member group families. However, there is evidence to suggest that extrapolation of test results to assemblies using members of reduced size is not always conservative. In a recent series of tests of cover plated connections, conducted at the University of California at San Diego, a connection assembly that produced acceptable results for one family of beam sizes, W24, did not behave acceptably when the beam depth was reduced significantly to W18. In that project, the change in relative flexibilities of the members and connection elements resulted in a shift in the basic behavior of the assembly and initiation of a failure mode that was not observed in the specimens with larger member sizes. In order to minimize the possibility of such occurrences, when extrapolation of test results is performed, it should be done with a basic understanding of the behavior of the assembly, and the likely effects of changes to the assembly configuration on this behavior. Test results should not be extrapolated to assembly configurations that are expected to behave differently than the tested configuration. Extrapolation or interpolation of results with differences in welding procedures, details or material properties is even more difficult. 6.6.6.1 Qualification Test Protocol

There are no modifications to the Guidelines or Commentary of Section 6.6.6.1 at this time. 6.6.6.2 Acceptance Criteria

The minimum acceptance criteria for connection qualification for specimens tested in accordance with these Interim Guidelines should be as follows: a) The connection should develop beam plastic rotations as indicated in Section 6.6.5, for at least one complete cycle. b) The connection should develop a minimum strength equal to 80% of the plastic strength of the girder, calculated using minimum specified yield strength Fy, throughout the loading history required to achieve the required plastic rotation capacity, as indicated in a), above. c) The connection should exhibit ductile behavior throughout the loading history. A specimen that exhibits a brittle limit state (e.g. complete flange fracture, column cracking, through-thickness failures of the column flange, fractures in welds subject to

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tension, shear tab cracking, etc. ) prior to reaching the required plastic rotation shall be considered unsuccessful. d) Throughout the loading history, until the required plastic rotation is achieved, the connection should be judged capable of supporting dead and live loads required by the building code. In those specimens where axial load is applied during the testing, the specimen should be capable of supporting the applied load throughout the loading history. The evaluation of the test specimen’s performance should consistently reflect the relevant limit states. For example, the maximum reported moment and the moment at the maximum plastic rotation are unlikely to be the same. It would be inappropriate to evaluate the connection using the maximum moment and the maximum plastic rotation in a way that implies that they occurred simultaneously. In a similar fashion, the maximum demand on the connection should be evaluated using the maximum moment, not the moment at the maximum plastic rotation unless the behavior of the connection indicated that this limit state produced a more critical condition in the connection. Commentary: Many connection configurations will be able to withstand plastic rotations on the order of 0.025 radians or more, but will have sustained significant damage and degradation of stiffness and strength in achieving this deformation. The intent of the acceptance criteria presented in this Section is to assure that when connections experience the required plastic rotation demand, they will still have significant remaining ability to participate in the structure’s lateral load resisting system. In evaluating the performance of specimens during testing, it is important to distinguish between brittle behavior and ductile behavior. It is not uncommon for small cracks to develop in specimens after relatively few cycles of inelastic deformation. In some cases these initial cracks will rapidly lead to ultimate failure of the specimen and in other cases they will remain stable, perhaps growing slowly with repeated cycles, and may or may not participate in the ultimate failure mode. The development of minor cracks in a specimen, prior to achievement of the target plastic rotation demand should not be cause for rejection of the design if the cracks remain stable during repeated cycling. Similarly, the occurrence of brittle fracture at inelastic rotations significantly in excess of the target plastic rotation should not be cause for rejection of the design. 6.6.6.3 Calculations

There are no modifications to the Guidelines or Commentary of Section 6.6.6.3 at this time.

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6.6.6.3.1 Material Strength Properties In the absence of project specific material property information (for example, mill test reports), the values listed in Table 6-3 Table 6.6.6.3.1-1 should be used to determine the strength of steel shape and plate for purposes of calculation. The permissible strength for weld metal should be taken in accordance with the building code. Table 6-3Table 6.6.6.3-1 - Properties for Use in Connection Modification Design Material Fy (ksi) Fy m (ksi) Fu (ksi) 1 1 A36 Beam 36 Dual Certified Beam Axial, Flexural 50 65 min. Shape Group 1 552 Shape Group 2 582 Shape Group 3 572 Shape Group 4 542 Through-Thickness Note 3 A572 Column/Beam Axial, Flexural 50 65 min. Shape Group 1 582 Shape Group 2 582 Shape Group 3 572 Shape Group 4 572 Shape Group 5 552 Through-Thickness Note 3 A992 Structural Shape1 Use same values as for A572, Gr. 50 Notes: 1. See Commentary 2. Based on coupons from web. For thick flanges, the Fy flange is approximately 0.95 Fy web. 3. See Commentary

Commentary: Table 6-3, Note 1 - The material properties for steel nominally designated on the construction documents as ASTM A36 can be highly variable and in recent years, steel meeting the specified requirements for both ASTM A36 and A572 has routinely been incorporated in projects calling for A36 steel. Consequently, unless project specific data is available to indicate the actual strength of material incorporated into the project, the properties for ASTM A572 steel should be assumed when ASTM A36 is indicated on the drawings, and the assumption of a higher yield stress results in a more severe design condition. The ASTM A992 specification was specifically developed by the steel industry in response to expressed concerns of the design community with regard to the permissible variation in chemistry and mechanical properties of structural steel under the A36 and A572 specifications. This new specification, which was adopted in late 1998, is very similar to ASTM A572, except that it includes somewhat more restrictive limits on chemistry and on the permissible variation in Post-earthquake Repair and Modification 6-13

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yield and ultimate tensile stress, as well as the ratio of yield to tensile strength. At this time, no statistical data base is available to estimate the actual distribution of properties of material produced to this specification. However, the properties are likely to be very similar, albeit with less statistical scatter, to those of material recently produced under ASTM A572, Grade 50. Table 6-3Table 6.6.6.3-1, Note 3 - In the period immediately following the Northridge earthquake, the Seismology Committee of the Structural Engineers Association of California and the International Conference of Building Officials issued Interim Recommendation No. 2 (SEAOC-1995) to provide guidance on the design of moment resisting steel frame connections. Interim Recommendation No. 2 included a recommendation that the through-thickness stress demand on column flanges be limited to a value of 40 ksi, applied to the projected area of beam flange attachment. This value was selected somewhat arbitrarily, to ensure that through-thickness yielding did not initiate in the column flanges of momentresisting connections and because it was consistent with the successful tests of assemblies with cover plates conducted at the University of Texas at Austin (Engelhardt and Sabol - 1994), rather than being the result of a demonstrated through-thickness capacity of typical column flange material. Despite the somewhat arbitrary nature of the selection of this value, its use often controls the overall design of a connection assembly including the selection of cover plate thickness, haunch depth, and similar parameters. It would seem to be important to prevent the inelastic behavior of connections from being controlled by through-thickness yielding of the column flanges. This is because it would be necessary to develop very large local ductilities in the column flange material in order to accommodate even modest plastic rotation demands on the assembly. However, extensive investigation of the throughthickness behavior of column flanges in a “T” joint configuration reveals that neither yielding, nor through-thickness failure are likely to occur in these connections. Barsom and Korvink (1997) conducted a statistical survey of available data on the tensile strength of rolled shape material in the throughthickness direction. These tests were generally conducted on small diameter coupons, extracted from flange material of heavy shapes. The data indicates that both the yield stress and ultimate tensile strength of this material in the throughthickness direction is comparable to that of the material in the direction parallel to rolling. However, it does indicate somewhat greater scatter, with a number of reported values where the through-thickness strength was higher, as well as lower than that in the longitudinal direction. Review of this data indicates with high confidence that for small diameter coupons, the yield and ultimate tensile values of the material in a through-thickness direction will exceed 90% and 80% respectively of the comparable values in the longitudinal direction. theThe causes for through-thickness failures of column flanges (types C2, C4, and C5), observed

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both in buildings damaged by the Northridge Earthquake and in some test specimens, are not well understood. They are thought to be a function of the metallurgy and “purity” of the steel; conditions of loading including the presence of axial load and rate of loading application; conditions of tri-axial restraint; conditions of local hardening and embrittlement within the weld’s heat affected zone; stress concentrations induced by the presence of backing bars and defects at the root of beam flange to column flange welds; and by the relationship of the connection components as they may affect flange bending stresses and flange curvature induced by panel zone yielding. Given the many complex factors which can affect the through-thickness strength of the column flange, determination of a reliable basis upon which to set permissible design stresses will require significant research. Such research is currently being conducted under the SAC phase II program. While this statistical distribution suggests the likelihood that the throughthickness strength of column flanges could be less than the flexural strength of attached beam elements, testing of more than 40 specimens at Lehigh University indicates that this is not the case. In these tests, high strength plates, representing beam flanges and having a yield strength of 100 ksi were welded to the face of A572, Grade 50 and A913, Grade 50 column shapes, to simulate the portion of a beam-column assembly at the beam flange. These specimens were placed in a universal testing machine and loaded to produce high throughthickness tensile stresses in the column flange material. The tests simulated a wide range of conditions, representing different weld metals as well and also to include eccentrically applied loading. In 40 of 41 specimens tested, the assembly strength was limited by tensile failure of the high strength beam flange plate as opposed to the column flange material. In the one failure that occurred within the column flange material, fracture initiated at the root of a low-toughness weld, at root defects that were intentionally introduced to initiate such a fracture. The behavior illustrated by this test series is consistent with mechanics of materials theory. At the joint of the beam flange to column flange, the material is very highly restrained. As a result of this, both the yield strength and ultimate tensile strength of the material in this region is significantly elevated. Under these conditions, failure is unlikely to occur unless a large flaw is present that can lead to unstable crack propagation and brittle fracture. In light of this evidence, Interim Guidelines Advisory No. 2 deletes any requirement for evaluation of through-thickness flange stress in columns. Interim Recommendation No. 2 (SEAOC-1995) included a value of 40 ksi, applied to the projected area of beam flange attachment, for the throughthickness strength to be used in calculations. This value was selected because it was consistent with the successful tests of cover plated assemblies conducted at

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the University of Texas at Austin (Engelhardt and Sabol - 1994). However, because of the probable influence of all the factors noted above, this value can only be considered to reflect the specific conditions of those tests and specimens. Although reduced stresses at the column face produced acceptable results in the University of Texas tests, the key to that success was more likely the result of forcing the plastic hinge away from the column than reduction of the throughthickness stress by the cover plates. Reduction of through-thickness column flange stress to ever lower levels by the use of thicker cover plates is not recommended, since such cover plates will result in ever higher forces on the face of the column flange as well as larger weldments with potential for enlarged heat affected zones, higher residual stresses and conditions of restraint. Since the initial publication of the Interim Guidelines, a significant number of tests have been performed on reduced beam section connections (See section 7.5.3), most of which employed beam flanges which were welded directly to the column flanges using improved welding techniques, but without reinforcement plates. No through-thickness failures occurred in these tests despite the fact that calculated through-thickness stresses at the root of the beam flange to column flange joint ranged as high as 58 ksi. The successful performance of these welded joints is most probably due to the shifting of the yield area of the assembly away from the column flange and into the beam span. Based on the indications of the above described tests, and noting the undesirability of over reinforcing connections, it is now suggested that a maximum through-thickness stress of 0.9Fyc may be appropriate for use with connections that shift the hinging away from the column face. Notwithstanding this recommendation, engineers are still cautioned to carefully consider the through-thickness issue when these other previously listed conditions which are thought to be involved in this type of failure are prevalent. Notwithstanding all of the above, successful tests using cover plates and other measures of moving hinges (and coincidentally reducing through-thickness stress) continue to be performed. In the interim, structural engineers choosing to utilize connections relying on through-thickness strength should recognize that despite the successful testing, connections relying on through-thickness strength can not be considered to be fully reliable until the influence of the other parameters discussed above can be fully understood. A high amount of structural redundancy is recommended for frames employing connections which rely on through-thickness strength of the column flange. 6.6.6.3.2 Determine Plastic Hinge Location The desired location for the formation of plastic hinges should be determined as a basic parameter for the calculations. For beams with gravity loads representing a small portion of the Post-earthquake Repair and Modification 6-16

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total flexural demand, the location of the plastic hinge may be assumed to occur as indicated in Table 6.6.6.3.2-1 and illustrated in Figure 6.6.6.3.2-1, at a distance equal to 1/3 of the beam depth from the edge of the reinforced connection (or start of the weakened beam section), unless specific test data for the connection indicates that a different value is appropriate. Refer to Figure 6-13. Table 6.6.6.3.2-1 Plastic Hinge Location - Strengthened Connections Connection Type

Reference Section

Hinge Location “sh”

Sect. 7.9.1

d/4 beyond end of cover plates

Haunches

Sect. 7.9.3, 7.9.4

d/3 beyond toe of haunch

Vertical Ribs

Sect. 7.9.2

d/3 beyond toe of ribs

Plastic hinge s h=

Edge of reinforced connection

d/4

Connection reinforcement

sh = d/3

Edge of reinforced connection

Beam depth - d

Cover plates

L’

L

Figure 6-13 Figure 6.6.6.3.2-1 - Location of Plastic Hinge Commentary: The suggested locations for the plastic hinge, at a distance d/3 away from the end of the reinforced section indicated in Table 6.6.6.3.2-1 and Figure 6.6.6.3.2-1 are is based on the observed behavior of test specimens, with no significant gravity load present. If significant gravity load is present, this can shift the locations of the plastic hinges, and in the extreme case, even change the form of the collapse mechanism. If flexural demand on the girder due to gravity load is less than about 30% of the girder plastic capacity, this effect can safely be neglected, and the plastic hinge locations taken as indicated. If gravity demands significantly exceed this level then plastic analysis of the girder should be

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performed to determine the appropriate hinge locations. Note that in zones of high seismicity (UBC Zones 3 and 4, and NEHRP Map Areas 6 and 7) gravity loading on the girders of earthquake resisting frames typically has a very small effect. 6.6.6.3.3 Determine Probable Plastic Moment at Hinges The probable value of the plastic moment, Mpr, at the location of the plastic hinges should be determined from the equation: M pr = 0.95aZ b Fya M pr = 1.1Z b Fya where: α

(6-1) (6.6.6.3.3-1)

is a coefficient that accounts for the effects of strain hardening and modeling uncertainty, taken as: 1.1

when qualification testing is performed or calculations are correlated with previous qualification testing

1.3

when design is based on calculations, alone.

Fya

is the actual yield stress of the material, as identified from mill test reports. Where mill test data for the project is not traceable to specific framing elements, the average of mill test data for the project for the given shape may be used. When mill test data for the project is not available, the value of Fym, from table 6-3Table 6.6.6.3-1 may be used.

Zb

is the plastic modulus of the section

Commentary: The 1.10.95 factor, in equation 6.6.6.3.3-1, is used to adjust account for two effects. First, it is intended to account for the typical difference between the yield stress in the beam web, where coupons for mill certification tests are normally extracted, andto the value in the beam flange. Beam flanges, being comprised of thicker material, typically have somewhat lower yield strengths than do beam web material. Second, it is intended toThe factor of 1.1 recommended to account for strain hardening, or other sources of strength above yield, and agrees fairly well with available test results. It should be noted that the 1.1 factor could underestimate the over-strength where significant flange buckling does not act as the gradual limit on the connection. Nevertheless, the 1.1 factor seems a reasonable expectation of over-strength considering the complexities involved.

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Connection designs that result in excessive strength in the girder connection relative to the column or excessive demands on the column panel zone are not expected to produce superior performance. There is a careful balance that must be maintained between developing connections that provide for an appropriate allowance for girder overstrength and those that arbitrarily increase connection demand in the quest for a “conservative” connection design. The factors suggested above were chosen in an attempt to achieve this balance, and arbitrary increases in these values are not recommended. When the Interim Guidelines were first published, Eq. 6.6.6.3.3-1 included a coefficient, α, intended to account both for the effects of strain hardening and also for modeling uncertainty when connection designs were based on calculations as opposed to a specific program of qualification testing. The intent of this modeling uncertainty factor was twofold: to provide additional conservatism in the design when specific test data for a representative connection was not available, and also as an inducement to encourage projects to undertake connection qualification testing programs. After the Interim Guidelines had been in use for some time, it became apparent that this approach was not an effective inducement for projects to perform qualification testing, and also that the use of an overly large value for the α coefficient often resulted in excessively large connection reinforcing elements (cover plates, e.g.) and other design features that did not appear conducive to good connection behavior. Consequently, it was decided to remove this modeling uncertainty factor from the calculation of the probable strength of an assembly. 6.6.6.3.4 Determine Beam Shear The shear in the beam at the location of the plastic hinge should be determined. A free body diagram of that portion of the beam located between plastic hinges is a useful tool for obtaining the shear at each plastic hinge. Figure 6-14Figure 6.6.3.4-1 provides an example of such a calculation.

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Plastic hinge

P

Note: if 2Mpr /L’is less then the gravity shear in the free body (in this case P/2 + wL’/2), then the plastic hinge location will shift and L’ must be adjusted, accordingly

L’ sh L

P

VA

w Mpr

“A”

Vp

Mpr

L’

taking the sum of moments about “A” = 0 Vp ={Mpr + Mpr + P L’/2 + wL’2/2}/L’

Figure 6-14 Figure 6.6.3.4-1 - Sample Calculation of Shear at Plastic Hinge 6.6.6.3.5 Determine Strength Demands on Connection In order to complete the design of the connection, including sizing the various plates and joining welds which make up the connection, it is necessary to determine the shear and flexural strength demands at each critical section. These demands may be calculated by taking a free body of that portion of the connection assembly located between the critical section and the plastic hinge. Figure 6-15 Figure 6.6.3.5-1 demonstrates this procedure for two critical sections, for the beam shown in Figure 6-14Figure 6.6.3.4-1.

Plastic hinge

Plastic hinge

Mpr

Mf

Vp

dc

x

Vp x+dc/2

Mf=Mpr +Vpx Critical Section at Column Face

Mpr

Mc

Mc=Mpr +Vp(x+dc/2) Critical Section at Column Centerline

Figure 6-15 Figure 6.6.3.5-1 - Calculation of Demands at Critical Sections Post-earthquake Repair and Modification 6-20

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Commentary: Each unique connection configuration may have different critical sections. The vertical plane that passes through the joint between the beam flanges and column (if such joining occurs) will typically define at least one such critical section, used for designing the joint of the beam flanges to the column, as well as evaluating shear demands on the column panel zone. A second critical section occurs at the center line of the column. Moments calculated at this point are used to check weak beam - strong column conditions. Other critical sections should be selected as appropriate to the connection configuration. 6.6.6.3.6 Check for Strong Column - Weak Beam Condition Buildings which form sidesway mechanisms through the formation of plastic hinges in the beams can dissipate more energy than buildings that develop mechanisms consisting primarily of plastic hinges in the columns. Therefore, if an existing building’s original design was such that hinging would occur in the beams rather than the columns, care should be taken not to alter this behavior with the addition of connection reinforcement. To determine if the desired strong column - weak beam condition exists, the connection assembly should be checked to determine if the following equation is satisfied:

∑Z where:

c

(Fyc − f a )

∑M

c

> 1.0

(6.6.6.3.6-12)

Zc is the plastic modulus of the column section above and below the connection Fyc is the minimum specified yield stress for the column above and below fa is the axial load in the column above and below ΣMc is the moment calculated at the center of the column in accordance with Section 6.6.6.3.5 sum of the column moments at the top and bottom of the panel zone, respectively, resulting from the development of the probable beam plastic moments, Mpr, within each beam in the connection. Commentary: Equation 6.6.6.3.6-12 is based on the building code provisions for strong column - weak beam design. The building code provisions for evaluating strong column - weak beam conditions presume that the flexural stiffness of the columns above and below the beam are approximately equal, that the beams will yield at the face of the column, and that the depth of the columns and beams are small relative to their respective span lengths. This permits the code to use a relatively simple equation to evaluate strong column - weak beam conditions in which the sum of the flexural capacities of columns at a connection are compared to the sums of the flexural capacities in the beams. The first publication of the Interim Guidelines took this same approach, except that the definition of ΣMc was modified to explicitly recognize that because flexural hinging of the beams would occur at a location removed from the face of the column, the moments delivered by the beams to the connection would be larger than the plastic moment strength of the beam. In this equation, ΣMc was taken as the sum of the moments at the Post-earthquake Repair and Modification 6-21

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center of the column, calculated in accordance with the procedures of Sect. 6.6.3.5. This simplified approach is not always appropriate. If non-symmetrical connection configurations are used, such as a haunch on only the bottom side of the beam, this can result in an uneven distribution of stiffness between the two column segments, and premature yielding of the column, either above, or below, the beam-column connection. Also, it was determined that for connection configurations in which the panel zone depth represents a significant fraction of the total column height, such as can occur in some haunched and side-plated connections, the definition of ΣMc contained in the initial printing of the Guidelines could lead to excessive conservatism in determining whether or not a strong column - weak beam condition exists in a structure. Consequently, Interim Guidelines Advisory No. 1 adopted the current definition of ΣMc for use in this evaluation. This definition requires that the moments in the column, at the top and bottom of the panel zone be determined for the condition when a plastic hinge has formed at all beams in the connection. Figure 6.6.6.3.6-1 illustrates a method for determining this quantity. In such cases, When evaluation indicates that a strong column - weak beam condition does not exist, a plastic analysis should be considered to determine if an undesirable story mechanism is likely to form in the building. assumed point of zero moment

ht

Vc

Vp

Vc =

∑ [M

M ct = Vc ht

Mct

dp

Mpr Vf

(

pr

]

(

+ V p ( L − L ' ) / 2) − V f hb + d p / 2

)

hb + d p + ht

)

M cb = Vc + V f hb

∑

M c = M ct + M cb

Mpr Mcb

hb

Vp

Vc+Vf (L-L’)/2

Note: The quantities Mpr, Vp, L, and L’are as previously identified. Vf is the incremental shear distributed to the column at the floor level. Other quantities are as shown.

Figure 6.6.6.3.6-1 Calculation of Column Moment for Strong Column Evaluation

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6.6.6.3.7 Check Column Panel Zone The adequacy of the shear strength of the column panel zone should be checked. For this purpose, the term 0.8ΣMf should be substituted for the term 0.8ΣMs in UBC-94 Section 2211.7.2.1 {0.9ΣφbMp in NEHRP-91 Section 10.10.3.1} repeated below for convenience of reference. Mf is the calculated moment at the face of the column, when the beam mechanism forms, calculated as indicated in Section 6.6.6.3.5, above. In addition, it is recommended not to use the alternative design criteria indicated in UBC-94 Section 2211.7.2.1 (NEHRP-91 Sect. 10.10.3.1), permitting panel zone shear strength to be proportioned for the shear induced by bending moments from gravity loads plus 1.85 times the prescribed seismic forces. For convenience of reference, UBC-94 Section 2211.7.2.1 is reproduced below, edited, to indicate the recommended application: 2211.7.2.1 Strength (edited). The panel zone of the joint shall be capable of resisting the shear induced by beam bending moments due to gravity loads plus 1.85 times the prescribed seismic forces, but the shear strength need not exceed that required to develop 0.8ΣMs 0.8ΣMf of the girders framing into the column flanges at the joint. The joint panel zone shear strength may be obtained from the following formula:

 3b c t c f 2  V = 0.55Fy d c t 1 +  dbdct  

(11-1)

where: bc = width of column flange db = the depth of the beam (including haunches or cover plates) dc = the depth of the column t = the total thickness of the panel zone including doubler plates tcf = the thickness of the column flange

Commentary: The effect of panel zone shear yielding on connection behavior is not well understood. In the past, panel zone shear yielding has been viewed as a benign mechanism that permits overall frame ductility demands to be accommodated while minimizing the extent of inelastic behavior required of the beam and beam flange to column flange joint. The criteria permitting panel zone shear strength to be proportioned for the shears resulting from moments due to gravity loads plus 1.85 times the design seismic forces was adopted by the code specifically to encourage designs with weak panel zones. However, during recent testing of large scale connection assemblies with weak panel zones, it has been noted that in order to accommodate the large shear deformations that occur in the panel zone, extreme “kinking” deformations were induced into the column flanges at the beam flange to column flange welded joint. While this did not lead to premature joint failure in all cases, it is believed to have contributed to such premature failures in at least some of the specimens. The recommendations of this section are intended to result in stronger panel zones than previously Post-earthquake Repair and Modification 6-23

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permitted by the code, thereby avoiding potential failures due to this kinking action on the column flanges. 6.6.7

Modification Details

There are no modifications to the Guidelines or Commentary of Section 6.6.7 at this time. 6.6.7.1 Haunch at Bottom Flange

Figure 6-166.6.7.1-1 illustrates the basic configuration for a connection modification consisting of the addition of a welded haunch at the bottom beam flange. Several tests of such a modification were conducted by Uang under the SAC phase I project (Uang, 1995). Following that work, additional research on the feasibility of improving connection performance with welded haunches was conducted under a project that was jointly sponsored by NIST and AISC (NIST, 1998). As indicated in the report of that work, the haunched modification improves connection performance by altering the basic behavior of the connection. In essence, the haunch creates a prop type support, beneath the beam bottom flange. This both reduces the effective flexural stresses in the beam at the face of the support, and also greatly reduces the shear that must be transmitted to the column through the beam. Laboratory tests indicate this modification can be effective when the existing low-toughness welds between the beam bottom flange and column are left in place, however, more reliable performance is obtained when the top welds are modified. A complete procedure for the design of this modification may be found in NIST, 1998. two alternative configurations of this detail that have been tested (Uang - 1995). The basic concept is to reinforce the connection with the provision of a triangular haunch at the bottom flange. The intended behavior of both configurations is to shift the plastic hinge from the face of the column and to reduce the demand on the CJP weld by increasing the effective depth of the section. In one test, shown on the left of Figure 6-16, the joint between the girder bottom flange and column was cut free, to simulate a condition which might occur if the bottom joint had been damaged, but not repaired. In a second tested configuration, the bottom flange joint was repaired and the top flange was replaced with a locally thickened plate, similar to the detail shown in Figure 6-9. Design Issues: This approach developed acceptable levels of plastic rotation. Acceptable levels of connection strength were also maintained during large inelastic deformations of the plastic hinge. This approach does not require that the top flange be modified, or slab disturbed, unless other conditions require repair of the top flange, as in the detail on the left of Figure 6-16. The bottom flange is generally far more accessible than the top flange because a slab does not have to be removed. In addition, the haunch can be installed at perimeter frames without removal of the exterior building cladding. There did not appear to be any appreciable degradation in performance when the bottom beam flange was not re-welded to the face of the column. Eliminating this additional welding should help reduce the cost of the repair. Performance is dependent on properly executed complete joint penetration welds at the column face and at the attachment of the haunch to the girder bottom flange. The joint can be subject to through-thickness flaws in the column flange; however, this connection may not be as sensitive Post-earthquake Repair and Modification 6-24

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to this potential problem because of the significant increase in the effective depth of the beam section which can be achieved. Welding of the bottom haunch requires overhead welding. The skewed groove welds of the haunch flanges to the girder and column flanges may be difficult to execute. Experimental Results: This approach developed excellent levels of plastic rotation. In Specimen 1, the bottom flange CJP weld was damaged in a prior test but was not repaired: only the bottom haunch was added. During the test of specimen 1, a slowly growing crack developed at the underside of the top flange-web intersection, perhaps exacerbated by significant local buckling of the top flange. Some of the buckling may be attributed to lateral torsional buckling that occurred because the bottom flange was not restrained by a CJP weld. A significant portion of the flexural strength was lost during the cycles of large plastic rotation. In the second specimen, the bottom girder flange weld was intact during the haunch testing, and its performance was significantly improved compared with the first specimen. The test was stopped when significant local buckling led to a slowly growing crack at the beam flange and web intersection. At this time, it appears that repairing damaged bottom flange welds in this configuration can produce better performance. Acceptable levels of flexural strength were maintained during large inelastic deformations of the plastic hinge for both specimens. As reported in NIST, 1998, a total of 9 beam-column connection tests incorporating bottom haunch modifications of preNorthridge connections have been tested in the laboratory, including two dynamic tests. Most of the connection assemblies tested resisted in excess of 0.02 radians of imposed plastic rotation. However, for those specimens in which the existing low-toughness weld was left in place at the beam top flange, without modification, connection behavior was generally limited by fractures generating at these welds at relatively low plastic rotations. It may be expected that enhanced performance can be obtained by replacing or reinforcing these welds as part of the modification.

Figure 6-166.6.7.1-1 - Bottom Haunch Connection Modification

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Quantitative Results: No. of specimens tested: 29 Girder Size: W30 x 99 Column Size: W14 x 176 Plastic Rotation achievedSpecimen 1 UCSD-1R: 0.04 radian (w/o bottom flange weld) Specimen 2 UCSD-3R:0.05 radian (with bottom flange weld) Specimen UCSD-4R: 0.014 radian (dynamic- limited by test setup) Speciemn UCSD-5R: 0.015 radian (dynamic- limited by test setup) Girder Size: W36x150 Column Size: W14x257 Plastic Rotation achieved Specimen UCB-RN2: 0.014 radian (no modification of top weld) Specimen UTA-1R: 0.019 radian (partial modification of top weld) Specimen UTA-1RB: 0.028 radian (modified top weld) Girder Size: W36x150 Column Size: W14x455 Plastic Rotation achievedSpecment UTA-NSF4: 0.015 radian (no modification of top weld) Girder Size: W18x86 Column Size: W24x279 Plastic Rotation achievedSpecimen SFCCC-8: 0.035 radian (cover plated top flange) 6.6.7.2 Top and Bottom Haunch

There are no modifications to the Guidelines or Commentary of Section 6.6.7.2 at this time. 6.6.7.3 Cover Plate Sections

Figure 6.6.7.3-1 Figure 6-18 illustrates the basic configurations of cover plate connections. The assumption behind the cover plate is that it reduces the applied stress demand on the weld at the column flange and shifts the plastic hinge away from the column face. Only the connection with cover plates on the top of the top flange has been tested. There are no quantitative results for cover plates on the bottom side of the top flange, such as might be used in repair. It is likely that thicker plates would be required where the plates are installed on the underside of the top flange. The implications of this deviation from the tested configuration should be considered.

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d

Near and Far Sides

d/2, typical Top &Bottom

Top &Bottom

Figure 6-18 Figure 6.6.7.3-1 - Cover Plate Connection Modification Design Issues: Following the Northridge earthquake, the University of Texas at Austin conducted a program of research, under private funding, to develop a modified connection configuration for a specific project. Following a series of unsuccessful tests on various types of connections,Approximately eight connections similar to that shown in Figure 6-18Figure 6.7.3-1 have been were tested (Engelhardt & Sabol - 1994), and have demonstrated the ability to achieve acceptable levels of plastic rotation provided that the beam flange to column flange welding wasis correctly executed and through-thickness problems in the column flange wereare avoided. Due to the significant publicity that followed these successful tests, as well as the economy of these connections relative to some other alternatives, cover plated connections quickly became the predominant configuration used in the design of new buildings. As a result, a number of qualification tests have now been performed on different variations of cover plated connections, covering a wide range of member sizes ranging from W16 to W36 beams, as part of the design process for individual building projects. The results of these tests have been somewhat mixed, with a significant number of failures reported. Although this connection type appears to be significantly more reliable than the typical pre-Northridge connection, it should be expected that some connections in buildings incorporating this detail may still be subjected to earthquake initiated fracture damage. Designers should consider using alternative connection types, unless highly redundant framing systems are employed. The option with the top flange cover plate located on top of the flange can be used on perimeter frames where access to the outer side of the beam is restricted by existing building cladding. The option with the cover plate for the top flange located beneath the flange can be installed without requiring modification of the slab. In the figures shown, the bottom cover plate is rectangular, and sized slightly wider than the beam flange to allow downhand fillet welding of the joint between the two plates. Some configurations using triangular plates at the bottom flange, similar to the top flange have also been tested.

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Designers using this detail are cautioned to be mindful of not making cover plates so thick that excessively large welds of the beam flange combination to column flange result. As the cover plates increase in size, the weld size must also increase. Larger welds invariably result in greater shrinkage stresses and increased potential for cracking prior to actual loading. In addition, larger welds will lead to larger heat affected zones in the column flange, a potentially brittle area. Performance is dependent on properly executed girder flange welds. The joint can be subject to through-thickness failures in the column flange. Access to the top of the top flange requires demolition of the existing slab. Access to the bottom of the top flange requires overhead welding and may be problematic for perimeter frames. Costs are greater than those associated with approaches that concentrate modifications on the bottom flange Experimental Results: Six of eight connections tested by the University of Texas at Austin were able to achieve plastic rotations of at least 0.025 radians, or better. These tests were performed using heavy column sections which forced nearly all of the plastic deformation into the beam plastic hinge; very little column panel zone deformation occurred. Strength loss at the extreme levels of plastic rotation did not reduce the flexural capacity to less than the plastic moment capacity of the section based on minimum specified yield strength. One specimen achieved plastic rotations of 0.015 radians when a brittle fracture of the CJP weld (type W2 failure) occurred. This may partially be the result of a weld that was not executed in conformance with the specified welding procedure specification. The second unsuccessful test specimen achieved plastic rotations of 0.005 radian when a section of the column flange pulled out (type C2 failure). The successful tests were terminated either when twisting of the specimen threatened to damage the test setup or the maximum stroke of the loading ram was achieved. Since the completion of that testing, a number of additional tests have been performed. Data for 18 tests, including those performed by Engelhardt and referenced above, are in the public domain. At least 12 other tests have been performed on behalf of private parties, however, the data from these tests are not available. Some of those tests exhibited premature fractures. Quantitative Results: No. of specimens tested: 18 Girder Size: W21 x 68 to W36 x 150 Column Size: W12 x 106 to W14 x 455, and 426 Plastic Rotation achieved6 13 Specimens : >.025 radian to 0.05 radian 13 Specimens: 0.005 < θp < 0.0250.015 radian (W2 failure) 12 Specimens: 0.005 radian (C2 failure) 6.6.7.4 Upstanding Ribs

There are no modifications to the Guidelines or Commentary of Section 6.6.7.4 at this time.

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6.6.7.5 Side-Plate Connections

There are no modifications to the Guidelines or Commentary of Section 6.6.7.5 at this time. 6.6.7.6 Bolted Brackets

Heavy bolted brackets, incorporating high strength bolts, may be added to existing welded connections to provide an alternative load path for transfer of stress between the beams and columns. To be compatible with existing welded connections, the brackets must have sufficient strength and rigidity to transfer beam stresses with negligible deformation. Pre-tensioning of the bolts or threaded rods attaching the brackets to the column flanges and use of welds or slipcritical connections between the brackets and beam flanges can help to minimize deformation under load. Reinforcement of the column flanges may be required to prevent local yielding and excessive deformation of these elements. Two alternative configurations, which may be used either to repair an existing damaged, welded connection or to reinforce an existing undamaged connection are illustrated in Figure 6.6.7.6-1. The developer of these connections offers the brackets in the form of proprietary steel castings. Several tests of these alternative connections have been performed on specimens with beams ranging in size from W16 to W36 sections and with large plastic rotations successfully achieved. Under a project jointly funded by NIST and AISC, the use of a single bracket at the bottom flange of the beam was investigated. It was determined that significant improvement in connection behavior could be obtained by placing a bracket at the bottom beam flange and by replacing existing low-toughness welds at the top flange with tougher material. NIST, 1998 provides a recommended design procedure for such connection modifications. Design Issues: The concept of bolted bracket connections is similar to that of the riveted “wind connections” commonly installed in steel frame buildings in the early twentieth century. The primary difference is that the riveted wind connections were typically limited in strength either by flexural yielding of outstanding flanges of the brackets, or shear and tension on the rivets, rather than by flexural hinging of the connected framing. Since the old-style wind connections could not typically develop the flexural strength of the girders and also could be quite flexible, they would be classified either as partial strength or partially restrained connections. Following the Northridge earthquake, the concept of designing such connections with high strength bolts and heavy plates, to behave as fully restrained connections, was developed and tested by a private party who has applied for patents on the concept of using steel castings for this purpose. Bolted bracket connections can be installed in an existing building without field welding. Since this reduces the risk of construction-induced fire, brackets may be installed with somewhat less demolition of existing architectural features than with welded connections. In addition, the quality assurance issues related to field welding are eliminated. However, the fabrication of the brackets themselves does require quality assurance. Quality assurance is also required for operations related to the drilling of bolt holes for installation of bolts, surface preparation of faying surfaces and for installation and tensioning of the bolts themselves.

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Steel casting

SAC 99-01 High tensile threaded rod

Pipe Plate Bolts

Bolts

WARNING: The information presented in this figure is PROPRIETARY. US and Foreign Patents have been applied for. Use of this information is strictly prohibited except as authorized in writing by the developer. Violators shall be prosecuted in accordance with US and Foreign Patent Intellectual Property Laws.

Figure 6.6.7.6-1 Bolted Bracket Modification Bolted brackets can be used to repair damaged connections. If damage is limited to the beam flange to column flange welds, the damaged welds should be dressed out by grinding. Any existing fractures in base metal should be repaired as indicated in Section 6.3, in order to restore the strength of the damaged members and also to prevent growth of the fractures under applied stress. Since repairs to base metal typically require cutting and welding, this reduces somewhat the advantages cited above, with regard to avoidance of field welding. Experimental Results: A series of tests on several different configurations of proprietary heavy bolted bracket connections have been performed at Lehigh University (Ksai & Bleiman, 1996) to qualify these connections for use in repair and modification applications. To test repair applications, brackets were placed only on the bottom beam flange to simulate installations on a connection where the bottom flange weld in the original connection had failed. In these specimens, bottom flange welds were not installed, to approximate the condition of a fully fractured weld. The top flange welds of these specimens were made with electrodes rated for notch toughness, to preclude premature failure of the specimens at the top flange. For specimens in which brackets were placed at both the top and bottom beam flanges, both welds were omitted. Acceptable plastic rotations were achieved for each of the specimens tested. No testing has yet been performed to determine the effectiveness of bolted brackets when applied to an existing undamaged connection with full penetration beam flange to column flange welds with low toughness or significant defects or discontinuities. Quantitative Results: No. of specimens tested: 8 Girder Size: W16x40 and W36x150 Column Size: W12x65 and W14x425 Plastic Rotation achieved - 0.05 radians - 0.07 radians

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7. NEW CONSTRUCTION 7.1 Scope This Chapter presents interim design guidelines for new welded steel moment frames (WSMFs) intended to resist seismic demands through inelastic behavior. The criteria apply to all SMRF structures designed for earthquake resistance and those IMRF and OMRF structures located in Uniform Building Code (UBC) Seismic Zones 3 and 4 {National Earthquake Hazards Reduction Program (NEHRP) Map Areas 6 and 7} or assigned to 1997 NEHRP Seismic Design Categories D, E, or F. Light, single-story buildings, the design of which is governed by wind, need not consider these Interim Guidelines. Frames with bolted connections, either fully restrained (type FR) or partially restrained (type PR), are beyond the scope of this document. However, the acceptance criteria for connections may be applied to type FR bolted connections as well. Commentary: Observation of damage experienced by WSMF buildings in the Northridge Earthquake and subsequent laboratory testing of large scale beamcolumn assemblies has demonstrated that the standard details for WSMF connections commonly used in the past are not capable of providing reliable service in the post-elastic range. Therefore, structures which are expected to experience significant post-elastic demands from design earthquakes, or for which highly reliable seismic performance is desired, should be designed using the Interim Guidelines presented herein. In order to determine if a structure will experience significant inelastic behavior in a design earthquake, it is necessary to perform strength checks of the frame components for the combination of dead and live loads expected to be present, together with the full earthquake load. Except for structures with special performance goals, or structures located within the near field (within 10 kilometers) of known active earthquake faults, the full earthquake load may be taken as the minimum design earthquake load specified in the building code, but calculated using a lateral force reduction coefficient (Rw or R) of unity. If all components of the structure and its connections have adequate strength to resist these loads, or nearly so, then the structure may be considered to be able to resist the design earthquake, elastically. Design of frames to remain elastic under unreduced (Rw {R} taken as unity) earthquake forces may not be an overly oppressive requirement, particularly in more moderate seismic zones. Most frame designs are currently controlled by drift considerations and have substantially more strength than the minimum specified for design by the building code. As part of the SAC Phase 1 research, a number of modern frame buildings designed with large lateral force reduction coefficients (Rw = 12, {R = 8}) were evaluated for unreduced forces calculated New Construction 7-1

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using the standard building code spectra. It was determined that despite the nominally large lateral force reduction coefficients used in the original design, the maximum computed demands from the dynamic analyses were only on the order of 2 to 3 times those which would cause yielding of the real structures (Krawinkler, et. al. - 1995; Uang, et. al. - 1995; Engelhardt, et. al. - 1995, Hart, et. al. - 1995; Kariotis and Eimani - 1995). Therefore, it is not unreasonable to expect that OMRF structures (nominally designed with a lateral force reduction coefficient Rw = 6 {R = 4.5}) could resist the design earthquakes with near elastic behavior. Regardless of these considerations, better seismic performance can be expected by designing structures with greater ductility rather than less, and engineers are not encouraged to design structures for elastic behavior using brittle or unreliable details.. For structures designed to meet special performance goals, and buildings located within the near field of major active faults, full earthquake loads calculated in accordance with the above procedure may not be adequate. For such structures, the full earthquake load should be determined using a site specific ground motion characterization and a suitable analysis procedure. Recent research (Heaton, et. al. - 1995) suggests that the elastic response spectrum technique, typically used for determining seismic forces for structural design, may not provide an adequate indication of the true earthquake demands produced by the large impulsive ground motions common in the near field of large earthquake events. Further, this research indicates that frame structures, subjected to such impulsive ground motions can experience very large drifts, and potential collapse. In an attempt to address this, both the 1997 edition of the Uniform Building Code and the 1997 edition of the NEHRP Provisions specify design ground motions for structures located close to major active faults that are substantially more severe than those contained in earlier codes. While the more severe ground motion criteria contained in these newer provisions are probably adequate for the design of most structures, analytical studies conducted by SAC confirm that even structures designed to these criteria can experience very large drift demands, and potentially collapse, if the dynamic characteristics of the impulsive loading and those of the structure are matched. Direct nonlinear time history analysis, using an appropriate ground motion representation would be one method of more accurately determining the demands on structures located in the near field. Additional research on these effects is required. As an alternative to use of the criteria contained in these Interim Guidelines, OMRF structures in zones of high seismicity (UBC seismic zones 3 and 4 and NEHRP map areas 6 and 7) and OMRF structures assigned to 1997 NEHRP Seismic Design Categories D, E or F, may be designed for the connections to remain elastic (Rw or R taken as 1.0) while the beams and columns are designed using the standard lateral force reduction coefficients specified by the building

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code. Although this is an acceptable approach, it may result in much larger connections than would be obtained by following these Interim Guidelines. The use of partially restrained connections may be an attractive and economical alternative to the design of frames with fully restrained connections. However, the design of frames with partially restrained connections is beyond the scope of this document. The AISC is currently working on development of practical design guidelines for frames with partially restrained connections. 7.2 General - Welded Steel Frame Design Criteria 7.2.1 Criteria

Welded Steel Moment Frame (WSMF) systems should, as a minimum, be designed for the provisions of the prevailing building code and these Interim Guidelines. Special MomentResisting Frames (SMRF)s and Ordinary Moment-Resisting Frames (OMRF)s with FR connections, should additionally be designed in accordance with either the 1997 edition of the AISC Seismic Provisions for Structural Steel Buildings (AISC, 1997) or the emergency code change to the 1994 UBC {NEHRP-1994}, restated as follows: 2211.7.1.1. Required Strength {NEHRP-1994 Section 5.2, revision to Ref. 8.2c of Ref. 5.3} The girder-to-column connections shall be adequate to develop the lesser of the following: 1.

The strength of the girder in flexure.

2.

The moment corresponding to development of the panel zone shear strength as determined by Formula (11-1).

2211.7.1.3-2 Connection Strength Connection configurations utilizing welds and high strength bolts shall demonstrate, by approved cyclic test results or calculation, the ability to sustain inelastic rotations and to develop the strength criteria in Section 2211.7.1.1 considering the effects of steel overstrength and strain hardening.

Commentary: At the time the Interim Guidelines were first published, they were based on the 1994 edition of the Uniform Building Code and the 1994 edition of the NEHRP Provisions. In the time since that initial publication, more recent editions of both documents have been published, and codes based on these documents have been adopted by some jurisdictions. In addition, the American Institute of Steel Construction has adopted a major revision to its Seismic Provisions for Structural Steel Buildings (AISC Seismic Provisions), largely incorporating, with some modification, the recommendations contained in the Interim Guidelines. This updated edition of the AISC Seismic Provisions has been incorporated by reference into the 1997 edition of the NEHRP Provisions and has also been adopted by some jurisdictions as an amendment to the model building codes. SMRF and OMRF systems that are designed to comply with the requirements of the 1997 AISC Seismic Provisions may be deemed to comply with the intent of these Interim Guidelines. Where reference is made herein to the New Construction 7-3

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requirements of the 1994 Uniform Building Code or 1994 NERHP Provisions, the parallel provisions of the 1997 editions may be used instead, and should be used in those jurisdictions that have adopted codes based on these updated standards. The 1997 edition NEHRP Provisions and AISC Seismic Provisions introduce a new structural system termed an Intermediate Moment Resisting Frame (IMRF). Provisions for IMRF structures include somewhat more restrictive detailing and design requirements than those for OMRF structures, and less than those for SMRF structures. The intent was to provide a system that would be more economical than SMRF structures yet have better inelastic response capability than OMRF structures. The SAC project is currently conducting research to determine if the provisions for the new IMRF system are adequate, but has not developed a position on this at this time. At this time, no recommendations are made to change the minimum lateral forces, drift limitations or strength calculations which determine member sizing and overall performance of moment frame systems, except as recommended in Sections 7.2.2, 7.2.3 and 7.2.4. The design of joints and connections is discussed in Section 7.3. The UBC permits OMRF structures with FR connections, designed for 3/8Rw times the earthquake forces otherwise required, to be designed without conforming to Section 2211.7.1. However, this is not recommended. 7.2.2 Strength and Stiffness 7.2.2.1 Strength

When these Interim Guidelines require determination of the strength of a framing element or component, this shall be calculated in accordance with the criteria contained in UBC-94, Section 2211.4.2 {NEHRP-91 Section 10.2, except that the factor φshould be taken as 1.0}, restated as follows: 2211.4.1 Member strength. Where this section requires that the strength of the member be developed, the following shall be used: Flexure Shear Axial compression Axial tension Connectors Full Penetration welds Partial Penetration welds Bolts and fillet welds

Ms = Z F y Vs = 0.55 Fy d t Psc = 1.7 Fa A Pst = Fy A Fy A 1.7 allowable (see commentary) 1.7 allowable

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Commentary: Partial penetration welds are not recommended for tension applications in critical connections resisting seismic-induced stresses. The geometry of partial penetration welds creates a notch-like condition that can initiate brittle fracture under conditions of high tensile strain. Many WSMF structures are constructed with concrete floor slabs that are provided with positive shear attachment between the slab and the top flanges of the girders of the moment-resisting frames. Although not generally accounted for in the design of the frames, the resulting composite action can increase the effective strength of the girder significantly, particularly at sections where curvature of the girder places the top flange into compression. Although this effect is directly accounted for in the design of composite systems, it is typically neglected in the design of systems classified as moment resisting steel frames. The increased girder flexural strength caused by this composite action can result in a number of effects including the unintentional creation of weak column strong beam and weak panel zone conditions. In addition, this composite effect has the potential to reduce the effectiveness of reduced section or “dog-bone” type connection assemblies. Unfortunately, very little laboratory testing of large scale connection assemblies with slabs in place has been performed to date and as a result, these effects are not well quantified. In keeping with typical contemporary design practice, the design formulae provided in these Guidelines neglect the strengthening effects of composite action. Designers should, however, be alert to the fact that these composite effects do exist. 7.2.2.2 Stiffness

Calculation of frame stiffness for the purpose of determining interstory drift under the influence of the design lateral forces should be based on the properties of the bare steel frame, neglecting the effects of composite action with floor slabs. The stiffening effects of connection reinforcements (e.g.: haunches, side plates, etc.) may be considered in the calculation of overall frame stiffness and drift demands. When reduced beam section connections are utilized, the reduction in overall frame stiffness, due to local reductions in girder cross section, should be considered. Commentary: For design purposes, frame stiffness is typically calculated considering only the behavior of the bare frame, neglecting the stiffening effects of slabs, gravity framing, and architectural elements. The resulting calculation of building stiffness and period typically underestimates the actual properties, substantially. Although this approach can result in unconservative estimates of design force levels, it typically produces conservative estimates of interstory drift demands. Since the design of most moment-resisting frames are controlled by considerations of drift, this approach is considered preferable to methods that would have the potential to over-estimate building stiffness. Also, many of the

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elements that provide additional stiffness may be subject to rapid degradation under severe cyclic lateral deformation, so that the bare frame stiffness may provide a reasonable estimate of the effective stiffness under long duration ground shaking response. Notwithstanding the above, designers should be alert to the fact that unintentional stiffness introduced by walls and other non-structural elements can significantly alter the behavior of the structure in response to ground shaking. Of particular concern, if these elements are not uniformly distributed throughout the structure, or isolated from its response, they can cause soft stories and torsional irregularities, conditions known to result in poor behavior. 7.2.3 Configuration

Frames should be proportioned so that the required plastic deformation of the frame can may be accommodated through the development of plastic hinges at pre-determined locations within the girder spans, as indicated in Figure 7-1Figure 7.2.3-1. Beam-column connections should be designed with sufficient strength (through the use of cover plates, haunches, side plates, etc.) to force development of the plastic hinge away from the column face. This condition may also be attained through local weakening of the beam section at the desired location for plastic hinge formation.

h

Undeformed frame

Deformed frame shape

Plastic Hinges

drift angle - θ

L’ L

Figure 7-1 Figure 7.2.3-1 - Desired Plastic Frame Behavior Commentary: Nonlinear deformation of frame structures is typically accommodated through the development of inelastic flexural or shear strains within discrete regions of the structure. At large inelastic strains these regions can develop into plastic hinges, which can accommodate significant concentrated rotations at constant (or nearly constant) load through yielding at tensile and compressive fibers and by buckling at compressive fibers. If a sufficient number of plastic hinges develop in a frame, a mechanism is formed and the frame can New Construction 7-6

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deform laterally in a plastic manner. This behavior is accompanied by significant energy dissipation, particularly if a number of members are involved in the plastic behavior, as well as substantial local damage to the highly strained elements. The formation of hinges in columns, as opposed to beams, is undesirable, as this results in the formation of weak story mechanisms with relatively few elements participating, so called “story mechanisms” and consequently little energy dissipation occurring. In addition, such mechanisms also result in local damage to critical gravity load bearing elements. The prescriptive connection contained in the UBC and NEHRP Recommended Provisions prior to the Northridge Earthquake was based on the assumed development of plastic hinge zones within the beams at adjacent to the face of the column, or within the column panel zone itself. If the plastic hinge develops in the column panel zone, the resulting column deformation results in very large secondary stresses on the beam flange to column flange joint, a condition which can contribute to brittle failure. If the plastic hinge forms in the beam, at the face of the column, this can result in very large through-thickness strain demands on the column flange material and large inelastic strain demands on the weld metal and surrounding heat affected zones. These conditions can also lead to brittle joint failure. Although ongoing research may reveal conditions of material properties, design and detailing configurations that permit connections with yielding occurring at the column face to perform reliably, for the present it is recommended In order to achieve more reliable performance, it is recommended that the connection of the beam to the column be configured to force the inelastic action (plastic hinge) away from the column face. This can be done either by local reinforcement of the connection, or locally reducing the cross section of the beam at a distance away from the connection. Plastic hinges in steel beams have finite length, typically on the order of half the beam depth. Therefore, the location for the plastic hinge should be shifted at least that distance away from the face of the column. When this is done through reinforcement of the connection, the flexural demands on the columns, for a given beam size, are increased. Care must be taken to assure that weak column conditions are not inadvertently created by local strengthening of the connections. It should be noted that some professionals and researchers believe that configurations which permit plastic hinging to occur adjacent to the column face may still provide reliable service under some conditions. These conditions may include limitations on the size of the connected sections, the use of base and weld metals with adequate notch toughness, joint detailing that minimizes notch effects, and appropriate control of the relative strength of the beam and column materials. Sufficient research has not been performed to date either to confirm these suggestions or define the conditions in which they are valid. Research however does indicate that reliable performance can be attained if the plastic

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hinge is shifted away from the column face, as suggested above. Consequently, these Interim Guidelines make a general recommendation that this approach be taken. Additional research should be performed to determine the acceptability of other approaches. It should also be noted that reinforced connection (or reduced beam section) configurations of the type described above, while believed to be effective in preventing brittle connection fractures, will not prevent structural damage from occurring. Brittle connection fractures are undesirable because they result in a substantial reduction in the lateral-force-resisting strength of the structure which, in extreme cases, can result in instability and collapse. Connections configured as described in these Interim Guidelines should experience many fewer such brittle fractures than unmodified connections. However, the formation of a plastic hinge within the span of a beam is not a completely benign event. Beams which have formed such hinges may, if plastic rotations are large, exhibit significantlarge buckling and yielding deformation, damage which typically must be repaired. The cost of such repairs could be comparable to the costs incurred in repairing fracture damage experienced in the Northridge Earthquake. The primary difference is that life safety protection will be significantly enhanced and most structures that have experienced such plastic deformation damage should continue to be safe for occupancy while repairs are made. If the types of damage described above are unacceptable for a given building, then alternative structural systems should be considered that will reduce the plastic deformation demands on the structure during a strong earthquake. Appropriate methods of achieving such goals include the installation of supplemental braced frames, energy dissipation systems, base isolation systems and similar structural systems. Framing systems incorporating partially restrained connections may also be quite effective in resisting large earthquake induced deformation with limited damage. It is important to recognize that in frames with relatively short bays, the flexural hinging indicated in Figure 7.2.3-1 may not be able to form. If the effective flexural length (L’in the figure) of beams in a frame becomes too short, then the beams or girders will yield in shear before zones of flexural plasticity can form, resulting in an inelastic behavior that is more like that of an eccentrically braced frame than that of a moment frame. This behavior may inadvertently occur in frames in which relatively large strengthened connections, such as haunches, cover plates or side plates have been used on beams with relatively short spans. This behavior is illustrated in Figure 7.2.3-2. The guidelines contained in this section are intended to address the design of flexurally dominated moment resisting frames. When utilizing these guidelines, it is important to confirm that the configuration of the structure is such that the New Construction 7-8

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presumed flexural hinging can actually occur. It is possible that shear yielding of frame beams, such as that schematically illustrated in Figure 7.2.3-2 may be a desirable behavior mode. However, to date, there has not been enough research conducted into the behavior of such frames to develop recommended design guidelines. Designers wishing to utilize such configurations should refer to the code requirements for eccentrically braced frames. Particular care should be taken to brace the shear link of such beams against lateral-torsional buckling and also to adequately stiffen the webs to avoid local buckling following shear plastification. Shear Link

Shear Link

Figure 7.2.3-2 Shear Yielding Dominated Behavior of Short Bay Frames 7.2.4 Plastic Rotation Capacity

The plastic rotation capacity of tested connection assemblies should reflect realistic estimates of the total (elastic and plastic) drift likely to be induced in the frame by earthquake ground shaking, and the geometric configuration of the frame. For frames of typical configuration, and for ground shaking of the levels anticipated by the building code, a minimum plastic rotation capacity of 0.03 radian is recommended. As used in these Guidelines, plastic rotation is defined as the plastic chord rotation angle. The plastic chord rotation angle is calculated using the rotated coordinate system shown in Fig. 7.2.4-1 as the plastic deflection of the beam or girder, at its point of inflection (usually the mid-span,) ∆CI, divided by the distance between this mid-span point and the centerline of the panel zone of the beam column connection, LCL. This convention is illustrated in Figure 7.2.4-1.

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LCL

Plastic hinge

Beam span center line

cL

∆CL

lh

θp =

∆ CL

LCL

Figure 7.2.4-1 Calculation of Plastic Rotation Angle It is important to note that this definition of plastic rotation is somewhat different than the plastic rotation that would actually occur within a discrete plastic hinge in a frame model similar to that shown in Figure 7.2.3-1. These two quantities are related to each other, however, and if one of them is known, the other may be calculated from Eq. 7.2.4-1.When the configuration of a frame is such that the ratio L/L’is greater than 1.25, the plastic rotation demand should be taken as follows:

θ p = θ ph where: θp θph LCL lh

( LCL − lh ) LCL

(7.2.4-1)

is the plastic chord angle rotation, as used in these Guidelines is the plastic rotation, at the location of a discrete hinge is the distance from the center of the beam-column assembly panel zone to the center of the beam span is the location of the discrete plastic hinge relative to the center of the beam-column assembly panel zone θ = 0.025(1 + ( L − L' ) L' )

(7-1)

where: L is the center to center spacing of columns, and L’is the center to center spacing of plastic hinges in the bay under consideration The indicated rotation demands may be reduced when positive means, such as the use of base isolation or energy dissipation devices, are introduced into the design to control the building’s response. When such measures are taken, nonlinear dynamic analyses should be performed and the connection demands taken as 0.005 radians greater than the plastic rotation demands calculated in the analyses. The nonlinear analyses should conform to the criteria specified in UBC-94 Section 1655 {NEHRP-94 Section 2.6.4.2} for nonlinear dynamic analysis of base New Construction 7-10

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isolated structures. Ground motion time histories utilized for these nonlinear analyses should satisfy the scaling requirements of UBC-94 Section 1655.4.2 {NEHRP-94 Section 2.6.4.4}, except that if the building is not base isolated, the structure period T, calculated in accordance with UBC-94 Section 1628 {NEHRP-94 Section 2.3.3.1} should be substituted for TI. When using methods of nonlinear analysis to establish the plastic rotation demands on frame connections, the analysis results should not be scaled by the factor Rw (R) or RWi (Ri ), as otherwise permitted by the building code. Commentary: When the Interim Guidelines were first published, the plastic rotation was defined as that rotation that would occur at a discrete plastic hinge, similar to the definition of θph. in Eq. 7.2.4-1, above. In subsequent testing of prototype connection assemblies, it was found that it is often very difficult to determine the value of this rotation parameter from test data, since actual plastic hinges do not occur at discrete points in the assembly and because some amount of plasticity also occurs in the panel zone of many assemblies. The plastic chord angle rotation, introduced in this advisory, may more readily be obtained from test data and also more closely relates to the drift experienced by a frame during earthquake response. This change in the definition of plastic rotation does not result in any significant change in the acceptance criteria for beam-column assembly qualification testing. When the Interim Guidelines were first published, they recommended an acceptance criteria given by Eq. 7.2.4-2, below:

 L − L'  θ p = 0.0251 +   L' 

(7.2.4-2)

For typical beam-column assemblies in which the plastic hinge forms relatively close to the face of the column, perhaps within a length of 1/2 the beam depth, this typically resulted in a plastic rotation demand of 0.03 radians, as currently measured. Traditionally, engineers have calculated demand in moment frames by sizing the members for strength and drift using code forces (either equivalent static or reduced dynamic forces) and then "developing the strength of the members." Since 1988, "developing the strength" has been accomplished by prescriptive means based on a review of testing of moment frame connections to that date. It was assumed that the prescribed connections would be strong enough that the beam or girder would yield (in bending), or the panel zone would yield (in shear) in a nearly perfectly plastic manner producing the plastic rotations necessary to dissipate the energy of the earthquake.

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A realistic estimate of the interstory drift demand for most structures and most earthquakes is on the order of 0.015 to 0.025 times the story height for WSMF structures designed to code allowable drift limits. In such frames, a portion of the drift will be due to elastic deformations of the frame, while the balance must be provided by inelastic rotations of the beam plastic hinges, by yielding of the column panel zone, or by a combination of the two. In the 1994 Northridge Earthquake, many moment-frame connections fractured with little evidence of plastic hinging of the beams or yielding of the column panel zones. Testing of moment frame connections both prior to and subsequent to the earthquake suggests that the standard, pre-Northridge, welded flange-bolted web connection is unable to reliably provide plastic rotations beyond about 0.005 radian for all ranges of beam depths and often fails below that level. Since the elastic contribution to drift may approach 0.01 radian, the necessary inelastic contributions will exceed the capability of the standard connection in many cases. For frames designed for code forces and for the code drift, the necessary plastic rotational demand may be expected to be on the order of 0.02 radian or more and new connection configurations should be developed to accommodate such rotation without brittle fracture. The recommended plastic rotation connection demand of 0.03 radians was selected both to provide a comfortable margin against the demands actually expected in most cases and because in recent testing of connection assemblies, specimens capable of achieving this demand behaved in a ductile manner through the formation of plastic hinges. For a given building design, and known earthquake hazard, it is possible to more accurately estimate plastic rotation demands on frame connections. This requires the use of nonlinear analysis techniques. Analysis software capable of performing such analyses is becoming more available and many design offices will have the ability to perform such analyses and develop more accurate estimates of inelastic demands for specific building designs. However, when performing such analyses, care should be taken to evaluate building response for multiple earthquake time histories, representative of realistic ground motions for sites having similar geologic characteristics and proximity to faults as the actual building site. Relatively minor differences in the ground motion time history used as input in such an analysis can significantly alter the results. Since there is significant uncertainty involved in any ground motion estimate, it is recommended that analysis not be used to justify the design of structures with non-ductile connections, unless positive measures such as the use of base isolation or energy dissipation devices are taken to provide reliable behavior of the structure.

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It has been pointed out that it is not only the total plastic rotation demand that is important to connection and frame performance, but also the connection mechanism (for example - panel zone yielding, girder flange yielding/buckling, etc.) and hysteretic loading history. These are matters for further study in the continuing research on connection and joint performance. 7.2.5 Redundancy

The frame system should be designed and arranged to incorporate as many moment-resisting connections as is reasonable into the moment frame. Commentary: Early moment frame designs were highly redundant and nearly every column was designed to participate in the lateral-force-resisting system. In an attempt to produce economical designs, recent practice often yieldedproduced designs which utilized only a few large columns and beams in a small proportion of the building’s frames for lateral resistance, with the balance of the building columns designed not considered or designed to participate in lateral resistance. This practice led to the need for large welds at the connections and to reliance on only a few connections for the lateral stability of the building. The resulting large framing elements and connections are believed to have exacerbated the poor performance of the pre-Northridge connection. Further, if only a few framing elements are available to resist lateral demands, then failure of only a few connections has the potential to result in a significant loss of earthquakeresisting strength. Together, these effects are not beneficial to building performance. The importance of redundancy to building performance can not be overemphasized. Even connections designed and constructed according to the improved procedures recommended by these Interim Guidelines will have some potential, albeit greatly reduced, for brittle failures. As the number of individual beams and columns incorporated into the lateral-force-resisting system is increased, the consequences of isolated connection failures are significantly reduceds. Further, as more framing elements are activated in the building’s response to earthquake ground motion, the building develops greater potential for energy absorption and dissipation, and greater ability to limit control earthquake-induced deformations to acceptable levels. Incorporation of more of the building framing into the lateral-force-resisting system will lead to smaller members and therefore an anticipated increase in the reliability of individual connections. It will almost certainly lead to improved overall system reliability. Further, recent studies conducted by designers indicate that under some conditions, redundant framing systems can be constructed as economically as non-redundant systems. In these studies, the additional costs incurred in making a greater number of field-welded moment-resisting New Construction 7-13

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connections in the more redundant frame were balanced by a reduced total tonnage of steel in the lateral-force-resisting systems, and sometimes reduced foundation costs as well. In order to codify the need for more redundant structural systems, the 1997 Uniform Building Code has specifically adopted a reliability coefficient, ρx, tied to the redundancy of framing present in the building. This coefficient, with values varying from 1.0 for highly redundant structures to 1.5 for non-redundant structures, is applied to the design earthquake forces, E, in the load combination equations, and has the effect of requiring more conservative design force levels for structures with nonredundant systems. The Building Seismic Safety Council’s Provisions Update Committee has also approved a proposal to include such a coefficient in the1997 NEHRP Provisions also includes such a coefficient. The formulation of this coefficient and its application are very similar in both the 1997 Uniform Building Code and 1997 NEHRP Provisions. As proposed contained in the 1997 NEHRP Provisions, the reliability coefficient is given by the equation: ρ = 2−

20 rmax Ax

(7.25-1)

where:

rmax x =

the ratio of the design story shear resisted by the single element

carrying the most shear force in the story to the total story shear, for a given direction of loading. For moment frames,

rmax x

is taken as the

maximum of the sum of the shears in any two adjacent columns in a moment frame divided by the story shear. For columns common to two bays with moment resisting connections on opposite sides at the level under consideration, 70% of the shear in that column may be used in the column shear summation. Ax = the floor area in square feet of the diaphragm level immediately above the story. The 1997 UBC and NEHRP Provisions also require that structures utilizing moment resisting frames as the primary lateral force resisting system be proportioned such that they qualify for a maximum value of ρx of 1.25. Structures located within a few kilometers of major active faults must be configured so as to qualify for a maximum value of ρx of 1.1.

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The most redundant moment-resisting frame systems are distributed frames in which all beam-column connections are detailed to be moment resisting. In these types of structures, half of the moment-resisting connections will be to the minor axis of the column which will typically result in weak column/strong beam framing. The 1994 UBC requirements limit the portion of the building design lateral forces that can be resisted by relative number of weak column/strong beam connections in the moment frame system. This limitation was adopted to avoid the design of frames likely to develop story mechanisms as opposed to concern about the adequacy of moment-resisting connections to the minor axis of columns. However, the limited research data available on such connections suggests that they do not behave well. There is a divergence of opinion among structural engineers on the desirability of frames in which all beam-column connections are made momentresisting, including those of beams framing to the minor axis of columns. Use of such systems as a means of satisfying the redundancy recommendations of these Interim Guidelines requires careful consideration by the structural engineer. Limited testing in the past has indicated that moment connections made to the minor axis of wide flange columns are subject to the same types of fracture damage experienced by major axis connections. As of this time, there has not been sufficient research to suggest methods of making reliable connections to the column minor axis. 7.2.6 System Performance

There are no modifications to the Guidelines or Commentary of Section 7.2.6 at this time. 7.2.7 Special Systems

There are no modifications to the Guidelines or Commentary of Section 7.2.7 at this time. 7.3 Connection Design & Qualification Procedures - General 7.3.1 Connection Performance Intent

The intent of connection design should be to force the plastic hinge away from the face of the column to a pre-determined location within the beam span. This may be accomplished by local reinforcement of the connection itself (cover plates, haunches, side plates, etc.) or by local reductions of the beam section (drilled holes, trimmed flanges, etc.). All elements of the connection should have adequate strength to develop the forces resulting from the formation of the plastic hinge at the predetermined location, together with forces resulting from gravity loads. Section 7.5.2 outlines a design procedure for reinforced connection designs. Section 7.5.3 provides a design procedure for reduced section connections.

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7.3.2 Qualification by Testing

There are no modifications to the Guidelines or Commentary of Section 7.3.2 at this time. 7.3.3 Design by Calculation

There are no modifications to the Guidelines or Commentary of Section 7.3.3 at this time. 7.4 Guidelines for Connection Qualification by Testing 7.4.1 Testing Protocol

There are no modifications to the Guidelines or Commentary of Section 7.4.1 at this time. 7.4.2 Acceptance Criteria

The minimum acceptance criteria for connection qualification for specimens tested in accordance with these Interim Guidelines should be as follows: a)

The connection should develop beam plastic rotations as indicated in Section 7.2.4, for at least one complete cycle.

b)

The connection should develop a minimum strength equal to the plastic strength of the girder, calculated using minimum specified yield strength Fy, tThroughout the loading history required to achieve the required plastic rotation capacity, as indicated in a), above, the connection should develop a minimum moment at the column face as follows: i)

For strengthened connections, the minimum moment at the column face should be equal to the plastic moment of the girder, calculated using the minimum specified yield strength, Fy of the girder. If the load limiting mechanism in the test is buckling of the girder flanges, the engineer, upon consideration of the effect of strength degradation on the structure, may consider a minimum of 80% of the nominal strength as acceptable.

ii)

For reduced section connection designs, the minimum moment at the column face should be equal to the moment corresponding to development of the nominal plastic moment of the reduced section at the reduced section, calculated using the minimum specified yield strength, Fy of the girder, and the plastic section modulus for the reduced section. The moment at the column face should not be less than 80% of the nominal plastic moment capacity of the unreduced girder section.

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c)

The connection should exhibit ductile behavior throughout the loading history. A specimen that exhibits a brittle limit state (e.g. complete flange fracture, column cracking, through-thickness failures of the column flange, fractures in welds subject to tension, shear tab cracking, etc. ) prior to reaching the required plastic rotation should be considered unsuccessful.

d)

Throughout the loading history, until the required plastic rotation is achieved, the connection should be judged capable of supporting dead and live loads required by the building code. In those specimens where axial load is applied during the testing, the specimen should be capable of supporting the applied load throughout the loading history.

The evaluation of the test specimen’s performance should consistently reflect the relevant limit states. For example, the maximum reported moment and the moment at the maximum plastic rotation are unlikely to be the same. It would be inappropriate to evaluate the connection using the maximum moment and the maximum plastic rotation in a way that implies that they occurred simultaneously. In a similar fashion, the maximum demand on the connection should be evaluated using the maximum moment, not the moment at the maximum plastic rotation unless the behavior of the connection indicated that this limit state produced a more critical condition in the connection. Commentary: While the testing of all connection geometries and member combinations in any given building might be desirable, it would not be very practical nor necessary. Test specimens should replicate, within the limitations associated with test specimen simplification, the fabrication and welding procedures, connection geometry and member size, and potential modes of failure. If the testing is done in a manner consistent with other testing programs, reasonable comparisons can be made. On the other hand, testing is expensive and it is difficult to realistically test the beam-column connection using actual boundary conditions and earthquake loading histories and rates. It was suggested in Interim Recommendation No. 2 by the SEAOC Seismology Committee that three tested specimens be the minimum for qualification of a connection. Further consideration has led to the recognition that while three tests may be desirable, the actual testing program selected should consider the conditions of the project. Since the purpose of the testing program is to "qualify the connection", and since it is not practical for a given project to do enough tests to be statistically meaningful considering random factors such as material, welder skills, and other variables, arguments can be made for fewer tests of identical specimens, and concentration on testing specimens which represent the range of different properties which may occur in the project. Once a connection is qualified, that is, once it has been confirmed that the connection can work, monitoring of actual materials and quality control to assure emulation of the tested design becomes most important. New Construction 7-17

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Because of the cost of testing, use of calculations for interpolation or extrapolation of test results is desirable. How much extrapolation should be accepted is a difficult decision. As additional testing is done, more information may be available on what constitutes "conservative" testing conditions, thereby allowing easier decisions relative to extrapolating tests to actual conditions which are likely to be less demanding than the tests. For example, it is hypothesized that connections of shallower, thinner flanged members are likely to be more reliable than similar connections consisting of deeper, thicker flanged members. Thus, it may be possible to test the largest assemblages of similar details and extrapolate to the smaller member sizes - at least within comparable member group families. However, there is evidence to suggest that extrapolation of test results to assemblies using members of reduced size is not always conservative. In a recent series of tests of cover plated connections, conducted at the University of California at San Diego, a connection assembly that produced acceptable results for one family of beam sizes, W24, did not behave acceptably when the beam depth was reduced significantly, to W18. In that project, the change in relative flexibilities of the members and connection elements resulted in a shift in the basic behavior of the assembly and initiation of a failure mode that was not observed in the specimens with larger member sizes. In order to minimize the possibility of such occurrences, when extrapolation of test results is performed, it should be done with a basic understanding of the behavior of the assembly, and the likely effects of changes to the assembly configuration on this behavior. Test results should not be extrapolated to assembly configurations that are expected to behave differently than the tested configuration. Extrapolation or interpolation of results with differences in welding procedures, details or material properties is even more difficult. 7.5 Guidelines for Connection Design by Calculation In conditions where it has been determined that design of connections by calculation is sufficient, or when calculations are used for interpolation or extrapolation, the following guidelines should be used. 7.5.1 Material Strength Properties

In the absence of project specific material property information, the values listed in Table 7-1 Table 7.5.1-1 should be used to determine the strength of steel shape and plate for purposes of calculation. The permissible strength for weld metal should be taken in accordance with the building code. Additional information on material properties may be found in the Interim Guidelines of Chapter 8.

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Table 7-1Table 7.5.1-1 - Properties for Use in Connection Design Material A36

Fy (ksi) 36

Fy m (ksi) use values for Dual Certified

Fu (ksi) 58

Dual Certified Beam Axial, Flexural3 50 65 min. Shape Group 1 551 Shape Group 2 581 Shape Group 3 571 Shape Group 4 541 Through-Thickness Note 5 A572 Column/Beam Axial, Flexural3 50 65 min. Shape Group 1 581 Shape Group 2 581 Shape Group 3 571 Shape Group 4 571 Shape Group 5 551 , Through-Thickness Note 5 A9922 Use same values as ASTM A572 A913-50 Axial, Flexural 50 581 65 min. , Through-thickness Note 5 A913--— 65 Axial, Flexural 65 751(4) 80 min. Through-thickness Note 5 Notes: 1. Based on coupons from web. For thick flanges, the Fy flange is approximately 0.95 Fy web. 2. See Commentary 3. Values based on (SSPC-1994) 4. ASTM A913, Grade 65 material is not recommended for use in the beams of moment resisting frames 5. See Commentary

Commentary: Table7.5.1-1 Note 2 - The ASTM A992 specification was specifically developed by the steel industry in response to expressed concerns of the design community with regard to the permissible variation in chemistry and mechanical properties of structural steel under the A36 and A572 specifications. This new specification, which was adopted in late 1998, is very similar to ASTM A572, except that it includes somewhat more restrictive limits on chemistry and on the permissible variation in yield and ultimate tensile stress, as well as the ratio of yield to tensile strength. At this time, no statistical data base is available to estimate the actual distribution of properties of material produced to this specification. However, the properties are likely to be very similar, albeit with less statistical scatter, to those of material recently produced under ASTM A572, Grade 50.

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Table 7.5.1-1 Note 5 -In the period immediately following the Northridge earthquake, the Seismology Committee of the Structural Engineers Association of California and the International Conference of Building Officials issued Interim Recommendation No. 2 (SEAOC-1995) to provide guidance on the design of moment resisting steel frame connections. Interim Recommendation No. 2 included a recommendation that the through-thickness stress demand on column flanges be limited to a value of 40 ksi, applied to the projected area of beam flange attachment. This value was selected somewhat arbitrarily, to ensure that through-thickness yielding did not initiate in the column flanges of momentresisting connections and because it was consistent with the successful tests of assemblies with cover plates conducted at the University of Texas at Austin (Engelhardt and Sabol - 1994), rather than being the result of a demonstrated through-thickness capacity of typical column flange material. Despite the somewhat arbitrary nature of the selection of this value, its use often controls the overall design of a connection assembly including the selection of cover plate thickness, haunch depth, and similar parameters. It would seem to be important to prevent the inelastic behavior of connections from being controlled by through-thickness yielding of the column flanges. This is because it would be necessary to develop very large local ductilities in the column flange material in order to accommodate even modest plastic rotation demands on the assembly. However, extensive investigation of the throughthickness behavior of column flanges in a “T” joint configuration reveals that neither yielding, nor through-thickness failure are likely to occur in these connections. Barsom and Korvink (1997) conducted a statistical survey of available data on the tensile strength of rolled shape material in the throughthickness direction. These tests were generally conducted on small diameter coupons, extracted from flange material of heavy shapes. The data indicates that both the yield stress and ultimate tensile strength of this material in the throughthickness direction is comparable to that of the material in the direction parallel to rolling. However, it does indicate somewhat greater scatter, with a number of reported values where the through-thickness strength was higher, as well as lower than that in the longitudinal direction. Review of this data indicates with high confidence that for small diameter coupons, the yield and ultimate tensile values of the material in a through-thickness direction will exceed 90% and 80% respectively of the comparable values in the longitudinal direction. the actual The causes for through-thickness failures of column flanges (types C2, C4, and C5), observed both in buildings damaged by the Northridge Earthquake and in some test specimens, are not well understood. They are thought to be a function of the metallurgy and “purity” of the steel; conditions of loading including the presence of axial load and rate of loading application; conditions of tri-axial restraint; conditions of local hardening and embrittlement within the weld’s heat affected zone; stress concentrations induced by the presence of backing bars and

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defects at the root of beam flange to column flange welds; and by the relationship of the connection components as they may affect flange bending stresses and flange curvature induced by panel zone yielding. Given the many complex factors which can affect the through-thickness strength of the column flange, determination of a reliable basis upon which to set permissible design stresses will require significant research. Such research is currently being conducted under the SAC phase II program. While this statistical distribution suggests the likelihood that the throughthickness strength of column flanges could be less than the flexural strength of attached beam elements, testing of more than 40 specimens at Lehigh University indicates that this is not the case. In these tests, high strength plates, representing beam flanges and having a yield strength of 100 ksi were welded to the face of A572, Grade 50 and A913, Grade 50 and 65 column shapes, to simulate the portion of a beam-column assembly at the beam flange. These specimens were placed in a universal testing machine and loaded to produce high through-thickness tensile stresses in the column flange material. The tests simulated a wide range of conditions, representing different weld metals as well and also to include eccentrically applied loading. In 40 of 41 specimens tested, the assembly strength was limited by tensile failure of the high strength beam flange plate as opposed to the column flange material. In the one failure that occurred within the column flange material, fracture initiated at the root of a lowtoughness weld, at root defects that were intentionally introduced to initiate such a fracture. The behavior illustrated by this test series is consistent with mechanics of materials theory. At the joint of the beam flange to column flange, the material is very highly restrained. As a result of this, both the yield strength and ultimate tensile strength of the material in this region is significantly elevated. Under these conditions, failure is unlikely to occur unless a large flaw is present that can lead to unstable crack propagation and brittle fracture. In light of this evidence, Interim Guidelines Advisory No. 2 deletes any requirement for evaluation of through-thickness flange stress in columns. Interim Recommendation No. 2 (SEAOC-1995) included a value of 40 ksi, applied to the projected area of beam flange attachment, for the throughthickness strength to be used in calculations. This value was selected because it was consistent with the successful tests of assemblies with cover plates conducted at the University of Texas at Austin (Engelhardt and Sabol - 1994). However, because of the probable influence of all the factors noted above, this value can only be considered to reflect the specific conditions of those tests and specimens. Although reduced stresses at the column face produced acceptable results in the University of Texas tests, the key to that success was more likely the result of New Construction 7-21

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forcing the plastic hinge away from the column than reduction of the throughthickness stress by the cover plates. Reduction of through-thickness column flange stress to ever lower levels by the use of thicker cover plates is not recommended, since such cover plates will result in ever higher forces on the face of the column flange as well as larger weldments with potential for enlarged heat affected zones, higher residual stresses and conditions of restraint. Since the initial publication of the Interim Guidelines, a significant number of tests have been performed on reduced beam section connections (See section 7.5.3), most of which employed beam flanges which were welded directly to the column flanges using improved welding techniques, but without reinforcement plates. No through-thickness failures occurred in these tests despite the fact that calculated through-thickness stresses at the root of the beam flange to column flange joint ranged as high as 58 ksi. The successful performance of these welded joints is most probably due to the shifting of the yield area of the assembly away from the column flange and into the beam span. Based on the indications of the above described tests, and noting the undesirability of over reinforcing connections, it is now suggested that a maximum through-thickness stress of 0.9Fyc may be appropriate for use with connections that shift the hinging away from the column face. Notwithstanding this recommendation, engineers are still cautioned to carefully consider the through-thickness issue when these other previously listed conditions which are thought to be involved in this type of failure are prevalent. Connections relying on through-thickness strength can not be considered to be fully reliable until the influence of the other parameters discussed above can be fully understood. A high amount of structural redundancy is recommended for frames employing connections which rely on through-thickness strength of the column flange. Notwithstanding all of the above, successful tests using cover plates and other measures of moving hinges (and coincidentally reducing through-thickness stress) continue to be performed. In the interim, engineers choosing to utilize connections relying on through-thickness strength should recognize that despite the successful testing, connections relying on through-thickness strength can not be considered to be fully reliable until the influence of the other parameters discussed above can be fully understood. A high amount of structural redundancy is recommended for frames employing connections which rely on through-thickness strength of the column flange. 7.5.2 Design Procedure - Strengthened Connections

The following procedure may be followed to size the various elements of strengthened connection assemblies that are intended to promote formation of plastic hinges within the beam span by providing a reinforced beam section at the face of the column. Section 7.5.3 provides a

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modified procedure recommended for use in the design of connection assemblies using reduced beam sections to promote similar inelastic behavior. Begin by selecting Select a connection configuration, such as one of those indicated in Sections 7.9.1, 7.9.2, 7.9.3, 7.9.4, or 7.9.5, that will permit the formation of a plastic hinge within the beam span, away from the face of the column, when the frame is subjected to gravity and lateral loads. Then proceed as described in the following sections. The following procedure should be followed to size the various elements of the connection assembly: 7.5.2.1 Determine Plastic Hinge Locations

For beams with gravity loads representing a small portion of the total flexural demand, the location of the plastic hinge may be assumed to occur as indicated in Table 7.5.2.1-1 at a distance equal to 1/3 of the beam depth from the edge of the reinforced connection (or start of the reduced beam section), unless specific test data for the connection indicates that a different location value is more appropriate. Refer to Figure 7-2Figure 7.5.2.1-1. Table 7.5.2.1-1 Plastic Hinge Location - Strengthened Connections Connection Type

Reference Section

Hinge Location “sh”

Sect. 7.9.1

d/4 beyond end of cover plates

Haunches

Sect. 7.9.3, 7.9.4

d/3 beyond toe of haunch

Vertical Ribs

Sect. 7.9.2

d/3 beyond toe of ribs

Plastic hinge sh=

Edge of reinforced connection

d/4

Connection reinforcement

sh= d/3

Edge of reinforced connection

Beam depth - d

Cover plates

L’

L

Figure 7-2 Figure 7.5.2.1-1 - Location of Plastic Hinge Commentary: The suggested locations for the plastic hinge, at a distance d/3 away from the end of the reinforced section (or beginning of reduced section) indicated in Table 7.5.2.1-1 and Figure 7.5.2.1-1 are is based on the observed behavior of test specimens, with no significant gravity load present. If significant

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gravity load is present, this can shift the locations of the plastic hinges, and in the extreme case, even change the form of the collapse mechanism. If flexural demand on the girder due to gravity load is less than about 30% of the girder plastic capacity, this effect can safely be neglected, and the plastic hinge locations taken as indicated. If gravity demands significantly exceed this level, then plastic analysis of the girder should be performed to determine the appropriate hinge locations. In zones of high seismicity (UBC Zones 3 and 4, and NEHRP Map Areas 6 and 7), gravity loading on the girders of earthquake resisting frames typically has a very small effect, unless tributary areas for gravity loads are large. 7.5.2.2 Determine Probable Plastic Moment at Hinges

Determine the probable value of the plastic moment, Mpr, at the location of the plastic hinges as: M pr = βM p = βZ b Fy where: ß

Zb

(7.5.2.2-12)

is a coefficient that adjusts the nominal plastic moment to the estimated hinge moment based on the mean yield stress of the beam material and the estimated strain hardening. A value of 1.2 should be taken for β for ASTM A572, A992 and A913 steels. When designs are based upon calculations alone, an additional factor is recommended to account for uncertainty. In the absence of adequate testing of the type described above, ß should be taken as 1.4 for ASTM A572 and for A913, Grades 50 and 65 steels. Where adequate testing has been performed ß should be permitted to be taken as 1.2 for these materials. is the plastic modulus of the section

Commentary: In order to compute β, the expected yield strength, strain hardening and an appropriate uncertainty factor need to be determined. The following assumed strengths are recommended: Expected Yield:

The expected yield strength, for purposes of computing (Mpr) may be taken as: Fye = 0.95 Fym

(7.5.2.2-2-3)

The 0.95 factor is used to adjust the yield stress in the beam web, where coupons for mill certification tests are normally extracted, to the value in the beam flange. Beam flanges, being comprised of thicker material, typically have somewhat lower yield strengths than do beam web material.

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Fy m for various steels are as shown in Table 7-1 Table 7.5.1-1, based on a survey of web coupon tensile tests (Steel Shape Producers Council - 1994). The engineer is cautioned that there is no upper limit on the yield point for ASTM A36 steel and consequently, dual-certification steel having properties consistent with ASTM A572, Grade 50 is routinely supplied when ASTM A36 is specified. Consequently, it is the recommendation here that the design of connections be based on an assumption of Grade 50 properties, even when A36 steel is specified for beams. It should be noted that at least one producer offers A36 steel with a maximum yield point of 50 ksi in shape sizes ranging up to W 24x62. Refer to the commentary to Section 8.1.3 for further discussion of steel strength issues. Strain Hardening: A factor of 1.1 is recommended for use with the mean yield stress in the foregoing table when calculating the probable plastic moment capacity Mpr.. The 1.1 factor for strain hardening, or other sources of strength above yield, agrees fairly well with available test results. The 1.1 factor could underestimate the over-strength where significant flange buckling does not act as a gradual limit on the beam strength. Nevertheless, the 1.1 factor seems a reasonable expectation of over-strength considering the complexities involved. Modeling Uncertainty: Where a design is based on approved cyclic testing, the modeling uncertainty may be taken as 1.0, otherwise the recommended value is 1.2. When the Interim Guidelines were first published, the β coefficient included a 1.2 factor to account for modeling uncertainty when connection designs were based on calculations as opposed to a specific program of qualification testing. The intent of this factor was twofold: to provide additional conservatism in the design when specific test data for a representative connection was not available and also as an inducement to encourage projects to undertake connection qualification testing programs. After the Interim Guidelines had been in use for some time, it became apparent that this approach was not an effective inducement for projects to perform qualification testing, and also that the use of an overly large value for the β coefficient often resulted in excessively large connection reinforcing elements (cover plates, e.g.) and other design features that did not appear conducive to good connection behavior. Consequently, it was decided to remove this modeling uncertainty factor from the calculation of β. In summary, for Grade 50 steel, we have: β = [0.95 (54 ksi to 58 ksi)/50 ksi] (1.1) 1.2) = 1.35 t0 1.45, say 1.4 β = [0.95 (54 ksi to 59 ksi)/50 ksi] (1.1) = 1.13 to 1.21, say 1.2

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7.5.2.3 Determine Shear at the Plastic Hinge

The shear at the plastic hinge should be determined by statics, considering gravity loads acting on the beam. A free body diagram of that portion of the beam between plastic hinges, is a useful tool for obtaining the shear at each plastic hinge. Figure 7-3 Figure 7.5.2.3-1 provides an example of such a calculation. For the purposes of such calculations, gravity load should be based on the load combinations required by the building code in use. L/2

Plastic hinge

P

Note: Gravity loads can effect the location of the plastic hinges. If 2Mpr /L’ is less then the gravity shear in the free body (in this case P/2 + wL’/2), then the plastic hinge location will shift significantly and L’must be adjusted, accordingly

w

L’ sh L

P w

VA

Mpr “A”

Vp

Mpr

L’

taking the sum of moments about “A” = 0 Vp ={Mpr + Mpr + P L’/2 + wL’2/2}/L’

Figure 7-3 Figure 7.5.2.3-1- Sample Calculation of Shear at Plastic Hinge Commentary: The UBC gives no specific guidance on the load combinations to use with strength level calculations while the NEHRP Recommended Provisions do specify load factors for the various dead, live and earthquake components of load. For designs performed in accordance with the UBC, it is customary to use unfactored gravity loads when checking the strength of elements. 7.5.2.4 Determine Strength Demands at Each Critical Section

In order to complete the design of the connection, including sizing the various plates and joining welds which make up the connection, it is necessary to determine the shear and flexural strength demands at each critical section. These demands may be calculated by taking a free body of that portion of the connection assembly located between the critical section and the plastic hinge. Figure 7-4 Figure 7.5.2.4-1 demonstrates this procedure for two critical sections, for the beam shown in Figure 7-3 Figure 7.5.2.3-1.

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Plastic hinge

Plastic hinge

Mpr

Mf

Mpr

Mc

Vp

dc

x

Vp x+dc/2

Mf =Mpr +Vpx

Mc=Mpr +Vp(x+dc/2)

Critical Section at Column Face

Critical Section at Column Centerline

Figure 7-4 Figure 7.5.2.4-1 - Calculation of Demands at Critical Sections Commentary: Each unique connection configuration may have different critical sections. The vertical plane that passes through the joint between the beam flanges and column (if such joining occurs) will typically define at least one such critical section, used for designing the joint of the beam flanges to the column, as well as evaluating shear demands on the column panel zone. A second critical section occurs at the center line of the column. Moments calculated at this point are used to check strong column - weak beam conditions. Other critical sections should be selected as appropriate. 7.5.2.5 Check for Strong Column - Weak Beam Condition

When required by the building code, the connection assembly should be checked to determine if strong column - weak beam conditions are satisfied. In lieu of UBC-94 equation 11-3.1 {NEHRP-91 equation 10-3}, the following equation should be used:

∑Z where:

c

(Fyc − f a )

∑M

c

> 1.0

(7.5.2.5-1-4)

Zc is the plastic modulus of the column section above and below the connection Fyc is the minimum specified yield stress for the column above and below fa is the axial load in the column above and below ΣMc is the moment calculated at the center of the column in accordance with Section 7.5.2.4 sum of the column moments at the top and bottom of the panel zone, respectively, resulting from the development of the probable beam plastic moments, Mpr, within each beam in the connection.

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Commentary: The building code provisions for evaluating strong column - weak beam conditions presume that the flexural stiffness of the columns above and below the beam are approximately equal, that the beams will yield at the face of the column, and that the depth of the columns and beams are small relative to their respective span lengths. This permits the code to use a relatively simple equation to evaluate strong column - weak beam conditions in which the sum of the flexural capacities of columns at a connection are compared against the sums of the flexural capacities in the beams. The first publication of the Interim Guidelines took this same approach, except that the definition of ΣMc was modified to explicitly recognize that because flexural hinging of the beams would occur at a location removed from the face of the column, the moments delivered by the beams to the connection would be larger than the plastic moment strength of the beam. In this equation, ΣMc was taken as the sum of the moments at the center of the column, calculated in accordance with the procedures of Sect. 7.5.2.4. assumed point of zero moment

ht

Vc

Vp

Vc =

∑ [M

M ct = Vc ht

Mct

dp

Mpr Vf

(

pr

]

(

)

+ V p ( L − L ' ) / 2) − V f hb + d p / 2 hb + d p + ht

)

M cb = Vc + V f hb

∑

M c = M ct + M cb

Mpr Mcb

hb

Vp

Vc+Vf (L-L’)/2

Note: The quantities Mpr, Vp, L, and L’are as previously identified. Vf is the incremental shear distributed to the column at the floor level. Other quantities are as shown.

Figure 7.5.2.5-1 Calculation of Column Moment for Strong Column Evaluation This simplified approach is not always appropriate. If non-symmetrical connection configurations are used, such as a haunch on only the bottom side of the beam, this can result in an uneven distribution of stiffness between the two column segments, and premature yielding of the column, either above, or below, the beam-column connection. Also, it was determined that for connection configurations in which the panel zone depth represents a significant fraction of New Construction 7-28

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the total column height, such as can occur in some haunched and side-plated connections, the definition of ΣMc contained in the initial printing of the Guidelines could lead to excessive conservatism in determining whether or not a strong column - weak beam condition exists in a structure. Consequently, Interim Guidelines Advisory No. 1 adopted the current definition of ΣMc for use in this evaluation. This definition requires that the moments in the column, at the top and bottom of the panel zone be determined for the condition when a plastic hinge has formed at all beams in the connection. Figure 7.5.2.5-1 illustrates a method for estimating this quantity. 7.5.2.6 Check Column Panel Zone

The adequacy of the shear strength of the column panel zone should be checked. For this purpose, the term 0.8ΣMf should be substituted for the term 0.8ΣMs in UBC-94 Section 2211.7.2.1 {0.9ΣφbMp in NEHRP-91 Section 10.10.3.1}, repeated below for convenience of reference. Mf is the calculated moment at the face of the column, when the beam mechanism forms, calculated as indicated in Section 7.5.2.4 above. In addition, it is recommended that the alternative design criteria indicated in UBC-94 Section 2211.7.2.1 (NEHRP-91 Sect. 10.10.3.1), permitting panel zone shear strength to be proportioned for the shear induced by bending moments from gravity loads plus 1.85 times the prescribed seismic forces, not be used. For convenience of reference, UBC-94 Section 2211.7.2.1 is reproduced below, edited, to indicate the recommended application. 2211.7.2.1 Strength (edited). The panel zone of the joint shall be capable of resisting the shear induced by beam bending moments due to gravity loads plus 1.85 times the prescribed seismic forces, but the shear strength need not exceed that required to develop 0.8ΣMs 0.8ΣMf of the girders framing into the column flanges at the joint. The joint panel zone shear strength may be obtained from the following formula:

 3b c t c f 2  V = 0.55Fy d c t 1 +  dbdct  

(11-1)

where: bc = width of column flange db = the depth of the beam (including any haunches or cover plates) dc = the depth of the column t = the total thickness of the panel zone including doubler plates tcf = the thickness of the column flange

Commentary: The effect of panel zone shear yielding on connection behavior is not well understood. In the past, panel zone shear yielding has been viewed as a benign, or even beneficial mechanism that permits overall frame ductility demands to be accommodated while minimizing the extent of inelastic behavior required of the beam and beam flange to column flange joint. The criteria

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permitting panel zone shear strength to be proportioned for the shears resulting from moments due to gravity loads plus 1.85 times the design seismic forces was adopted by the code specifically to permit designs with somewhat weak panel zones. However, during recent testing of large scale connection assemblies with weak panel zones, it has been noted that in order to accommodate the large shear deformations that occur in the panel zone, extreme “kinking” deformations were induced into the column flanges at the beam flange to column flange welded joint. While this did not lead to premature joint failure in all cases, it is believed to have contributed to such premature failures in at least some of the specimens. The recommendations of this section are intended to result in stronger panel zones than previously permitted by the code, thereby avoiding potential failures due to this kinking action on the column flanges. 7.5.3 Design Procedure - Reduced Beam Section Connections

The following procedure may be followed to size the various elements of reduced beam section (RBS) assemblies with circular curved reductions in beam flanges, such as shown in Figure 7.5.3-1., such as those indicated in Section 7.9.6 indicates other configurations for such connections, however, the circular curved configuration shown in Figure 7.5.3-1 is currently preferred. RBS assemblies are intended to promote the formation of plastic hinges within the beam span by developing a segment of the beam with locally reduced section properties and strength. Begin by selecting an RBS configuration, such as one of those indicated in Figure 7.5.31, that will permit the formation of a plastic hinge within the reduced section of the beam. Of the configurations shown in the figure, the circular curved configuration is preferred. 2

2

4a + l R = radius of cut = 8a

a b f a c l Figure 7.5.3-1 Geometry of Reduced Beam Section

Commentary: Connection assemblies in which inelastic behavior is shifted away from the column face through development of a segment of the beam with intentionally reduced properties, so-called reduced beam section (RBS) or “dogbone” connections, appear to have the potential to provide an economical New Construction 7-30

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solution to the WMSF connection problem. These recommendations are based on limited design configurations that have successfully been tested ing that has been conducted of these types of connections to date. While a A large number of RBS tests have been conducted, these tests have not included the effects of floor slabs or loading rates approximating those that would be produced by a building’s response to earthquake ground motionsincluding some tests of assemblies with floor slabs present. Extensive additional testing of RBS connections, intended to explore these and other factors relevant to connection performance, are currently planned under funding provided by NIST and the SAC phase II program. In the interim, designers specifying RBS connections may wish to consider provision of details to minimize the participation of the slab in the flexural behavior of the beam at the reduced section. The criteria presented in this section are partially based on a draft procedure developed by AISC (Iwankiw, 1996).

Circular

Straight

Reduced Section

Drilled Constant

Tapered

Drilled Tapered

Figure 7.5.3-2 Alternative Reduced-Beam Section Patterns Figure 7.5.3-1 Reduced Beam Section Patterns Several alternative configurations of RBS connections have also been tested to date. As indicated in Figures 7.5.3-21 and 7.9.6-1, these include constant section, tapered section, curved section, and drilled hole patterns. It appears that several of these configurations are more desirable than others. In particular, the drilled hole section patterns have been subject to tensile failure across the reduced net section of the flange through the drill holes. A few RBS tests utilizing straight or tapered cuts have failed within the reduced section at plastic rotation demands less than recommended by these Guidelines. In all of these cases, the failure occurred at locations at which there was a change in direction of the cuts in the beam flange, resulting in a geometric stress riser or notch effect. It is also reported that one of these tests failed at the beam flange continuity plate - to column flange joint. There have been no reported failures of RBS connection New Construction 7-31

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assemblies employing the circular curved flange cuts, and therefore, this is the pattern recommended in these Guidelines. This would appear, therefore, to be a more desirable configuration, although some successful tests have been performed using the straight and tapered configurations. It is important that the pattern of any cuts made in the flange be proportioned so as to avoid sharp cut corners. All corners should be rounded to minimize notch effects and in addition, cut edges should be cut or ground in the direction of the flange length to have a surface roughness meeting the requirements of AWS C4.1-77 class 4, or smootherroughness value less than or equal to 1,000, as defined in ANSI/ASME B46.1. Concerns have been raised by some engineers over the strength reduction inherent in the RBS. Clearly, code requirements for strength, considering gravity loads and gravity loads in combination with wind, seismic and other loads must be met. For higher seismic zones, beam sizes are typically governed by elastic stiffness considerations (drift control) and this must be addressed. Also, for seismic loads, the Building Codes typically require that connections for Special Moment Resisting Frames must develop the “strength” or the “plastic bending moment” of the beam. There may be a problem of semantics where these requirements are applied to a system using RBS connections. Is the RBS part of the connection or is it part of the beam, the strength of which must be developed by the connection? Clearly, the latter interpretation should be applied. Notwithstanding the above, it must be kept in mind that, although unstated, and typically not quantified, there is inherent in design practice an implied relationship between the elastic behavior that we analyze and the inelastic behavior which the building is expected to experience. Elastic drift limitations commonly used are considered to be related to the anticipated inelastic drifts and ultimate lateral stability of the framed structure in at least an intuitively predictable manner. It can be shown that RBS’s such as those that have been tested will reduce the elastic stiffness (increase the drift) on the order of 5%. However, because of the reduction in strength, the effect on the inelastic drift may be more significant. Thus, it seems prudent to require that the RBS maintain a reasonably high proportion of the frame inelastic strength. For the connections tested to date, the inelastic strength of the RBS section has been in the range of 70% of that of the full section. However, the moment demand at the face of the column, corresponding to development of this reduced section strength, is likely to be in the range of 85% to 90% of the strength of the full beam. This seems to be quite reasonably high considering the accuracy of other seismic design assumptions. Although the use of RBS designs tends to reduce the total strength demand on the beam flange - to - column flange connection, relative to strengthened New Construction 7-32

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connections, designs utilizing RBS configurations should continue to follow the recommendations for beam flange continuity plates, weld metal and base metal notch toughness recommended by the Interim Guidelines for strengthened connections. 7.5.3.1 Determine Reduced Section and Plastic Hinge Locations

Beam depth - d

The reduced beam section should be located at a sufficient distance from the face of the column flange (dimension “c” in Figures 7.5.3-1 and 7.5.3.1-1) to avoid significant inelastic behavior of the material at the beam flange - to - column flange joint. Based on testing performed to date, it appears that a value of “c” on the order of ½ to ¾ of the beam width, bf, is sufficient. d/4 (where “d” is the beam depth) is sufficient. The total length of the reduced section of beam flange (dimension “l” in Figures 7.5.3-1 and 7.5.3.1-1) should be on the order of 0.65d to 0.85d, where d is the beam depth.3d/4 to d. The location of the plastic hinge, sh,, may be taken as ½ the length of the cut-out, l.indicated in Table 7.5.3.1-1, unless test data indicates a more appropriate value should be used. When tapered configurations are utilized, the slope of the tapered cut in the beam flange should be arranged such that the variation of the plastic section modulus, Zx, within the reduced section approximates the moment gradient in the beam during the condition when plastic hinges have formed within the reduced beam sections at both ends.

l

Plastic hinge

reduced section

L’

sh c

L

Figure 7.5.3.1-1 Critical Dimensions - RBS Assemblies 7.5.3.2 Determine Strength and Probable Plastic Moment in RBS

The RBS may be proportioned to meet the following criteria:

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1. The section at the RBS should be sufficient to satisfy the strength criteria specified by the building code for Dead, Live, Seismic, Snow, Wind, and other applicable design forces. 2. The elastic stiffness of the frame, considering the effects of the RBS, should be sufficient to meet the drift requirements specified by the code, under the design seismic and other forces. 3. The expected stress in the beam flange - to - column flange weld, under the application of gravity forces and that seismic force that results in development of the probable plastic moment of the reduced section at both ends of the beam, should be less than or equal to the strength of the weld, as indicated in Section 7.2.2 of the Interim Guidelines. 4. The expected through-thickness stress on the face of the column flange, calculated as Mf/Sc, under the application of gravity forces and that seismic force that results in development of the probable plastic moment of the reduced section at both ends of the beam, should be less than or equal to the values indicated in Section 7.5.1, where Mf is the moment at the face of the column flange, calculated as indicated in Section 7.5.2.4, and Sc is the elastic section modulus of the beam at the connection considering weld reinforcement, bolt holes, reinforcing plates, etc. The maximum moment at the face of the column should be in the range of 85 percent to 100 percent of the beam’s expected plastic moment capacity. The depth of cut-out, a, should be selected to be less than or equal to bf/4. The plastic section modulus of the RBS may be calculated from the equation:

(

Z RBS = Z x − bR t f d − t f

)

(7.5.3.2-1)

where: ZRBS Zx bR tf d

is the plastic section modulus of the reduced beam section is the plastic modulus of the unreduced section is the total width of material removed from the beam flange is the thickness of the beam flange is the depth of the beam

The probable plastic moment, Mpr, at the RBS shall be calculated from the equation:

M pr = Z RBS βFy where: ZRBS β

is the plastic section modulus of the reduced beam section is as defined in Section 7.5.2.2 New Construction 7-34

(7.5.3.2-2)

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The strength demand on the beam flange - to - column flange weld and on the face of the column may be determined by following the procedures of Section .7.5.2.3 and 7.5.2.4 of the Interim Guidelines, using the value of Mpr determined in accordance with Eq. 7.5.3.2-2. Commentary: Initial design procedures for RBS connections published by SAC recommended that sufficient reduction of the beam flange be made to maintain flexural stresses in the beam, at the column face, below the anticipated throughthickness yield strength of the column flange material. Since the publication of those recommendations, extensive testing of RBS connections has been conducted, both with and without composite slabs. The testing conducted to date on RBS specimens This testing has typically been for configurations that would result in somewhat larger strength demands at the face of the column flange than suggested by the criteria originally published by SAC. contained in this Advisory. Typically, the tested specimens had reductions in the beam flange area on the order of 35% to 45% and produced moments at the face of the column that resulted in stresses on the weld and column as large as large as 90 to 100% of the expected material strength of the beam, which is often somewhat in excess of the through-thickness yield strength of the column material. The specimens in these tests all developed acceptable levels of inelastic deformation. Recent studies conducted for SAC at Lehigh University confirm that the significant conditions of restraint that exist at the beam flange to column flange joint results in substantially elevated column through-thickness strength, negating a need to reduce flexural stresses below the anticipated column yield strength. In view of this evidence, SAC has elected to adopt design recommendations consistent with configurations that were successfully tested. The criteria contained in this Advisory suggest that these demands be reduced to a level which would maintain weld stresses within their normally specified values and through-thickness column flange stresses at the same levels recommended for strengthened connections. This may require the beam flanges to be reduced by as much as 50% or more for some frame configurations, or that supplemental reinforcement such as cover plates or vertical ribs be provided in addition to the reduced section. This approach was taken to maintain consistency with the criteria recommended for strengthened connections and with the knowledge that the factors affecting the performance of these connections are not yet fully understood. 7.5.3.3 Strong Column - Weak Beam Condition

The adequacy of the design to meet strong column - weak beam conditions should be checked in accordance with the procedures of Section 7.5.2.5

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7.5.3.4 Column Panel Zone

The adequacy of the column panel zone should be checked in accordance with the procedures of Section 7.5.2.6. 7.5.3.5 Lateral Bracing

The reduced section of the beam flanges should be provided with adequate lateral support to prevent lateral-torsional buckling of the section. Lateral braces should be located within a distance equal to 1/2 the beam depth from the expected location of plastic hinging, but should not be located within the reduced section of the flanges. Commentary: Unbraced compression flanges of beams are subject to lateraltorsional buckling, when subjected to large flexural stresses , such as occur in the plastic hinges of beams reduced sections of RBS connections during response to strong ground motion. To prevent such behavior lateral-torsional buckling, it is recommended that both flanges of beams be provided with lateral support. Section 9.8 of the 1997 AISC Seismic Specification requires such bracing in general, and specifically states as follows: “Both flanges of beams shall be laterally supported directly or indirectly. The unbraced length between lateral supports shall not exceed 2500ry/Fy. In addition, lateral supports shall be placed near concentrated forces, changes in cross section and other locations where analysis indicates that a plastic hinge will form during inelastic deformations of the SMF.” Adequate lateral support of the top flanges of beams supporting concrete filled metal deck or formed slabs can usually be obtained through the normal welded attachments of the deck to the beam or through shear studs. Lateral support of beam flanges can also be provided through the connections of transverse framing members or by provision of special lateral braces, attached directly to the flanges. Such attachments should not be made within the reduced section of the beam flange as the welding or bolting required to make such attachments can lead to premature fracturing in these regions of high plastic demands. For beams in moment-resisting frames, it has traditionally been assumed that the direct attachment of the beam flanges to the columns provided sufficient lateral support of both beam flanges to accommodate the plastic hinges anticipated to develop in these frames at the beam-column connection. However, connection configurations like the RBS, developed following the Northridge earthquake, are intended to promote formation of these plastic hinges at some distance from the beam-column interface. This brings to question the adequacy of the beam flange to column flange attachments to provide the necessary lateral

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support at the plastic hinge. While this issue is pertinent for any connection configuration that promotes plastic hinge formation remote from the beamcolumn interface, RBS connections could be more susceptible to lateral-torsion buckling at the plastic hinge because the reductions in the beam flange used to achieve plastic hinge formation also locally reduce the torsional resistance of the section. For that reason, FEMA-267a recommended provision of lateral bracing adjacent to the reduced beam section. Provision of lateral bracing does result in some additional cost. Therefore, SAC has engaged in specific investigations to evaluate the effect of lateral bracing both on the hysteretic behavior of individual connections as well as overall frame response to large lateral displacements. Until these investigations have concluded SAC continues to recommend provision of lateral bracing for RBS connections. It should be noted that Section 9.8 of the 1997 AISC Seismic Specification states: “If members with Reduced Beam Sections, tested in accordance with Appendix S are used, the placement of lateral support for the member shall be consistent with that used in the tests.” Most testing of RBS specimens performed as part of the SAC project have consisted of single beams cantilevered off a column to simulate the exterior connection in a multi-bay moment-resisting frame. The beams have generally been braced at the end of the cantilever length, typically located about 100 inches from the face of the column. For the ASTM A572, Grade 50, W36x150 sections typically tested, this results in a nominal length between lateral supports that is comparable to 2500ry/Fy. The appropriate design strength for lateral bracing of compression elements has long been a matter of debate. Most engineers have applied “rules of thumb” that suggest that the bracing element should be able to resist a small portion, perhaps on the order of 2% to 6% of the compressive force in the element being braced, applied normal to the line of action of the compression. A recent successful test of an RBS specimen conducted at the University of Texas at Austin incorporated lateral bracing with a strength equal to 6% of the nominal compressive yield force in the reduced section. 7.5.3.6 Welded Attachments

Headed studs for composite floor construction should not be placed on the beam flange between the face of the column and the extreme end of the RBS, as indicated in Figure 7.5.3.6-1. Other welded attachments should also be excluded from these regions of the beam.

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welded attachment prohibited welded attachment permitted

Reduced beam section

Figure 7.5.3.6-1 Welded Attachments to RBS Beams Commentary: There are two basic reasons for omitting headed studs in the region between the reduced beam section and the column. The first of these is that composite action of the slab and beam can effectively counteract the reduction in beam section properties achieved by the cutouts in the top beam flange. By omitting shear studs in the end region of the beam, this composite behavior is neutralized, protecting the effectiveness of the section reduction. The second reason is that the portion of the beam at the reduced section is expected to experience large cyclic inelastic strains. If welded attachments are made to the beam in this region, the potential for low-cycle fatigue of the beam, under these large cyclic inelastic strains is greatly increased. For this same reason, other welded attachments should also be excluded from this region. 7.6 Metallurgy and Welding There are no modifications to the Guidelines or Commentary of Section 7.6 at this time. 7.7 Quality Control/Quality Assurance There are no modifications to the Guidelines or Commentary of Section 7.7 at this time. 7.8 Guidelines on Other Connection Design Issues There are no modifications to the Guidelines or Commentary of Section 7.8 at this time.

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7.8.1 Design of Panel Zones

No current recommendations are made to supplement or modify the UBC-1994 {NEHRP-91} provisions for the design of panel zones, other than as indicated in Section 7.5.2.6, above. Panel zone demands should be calculated in accordance with Section 7.5.2.6. As with other elements of the connection, available panel zone strength should be computed using minimum specified yield stress for the material, except when the panel zone strength is used as a limit on the required connection strength, in which case Fym should be used. Where connection design for two-sided connection assemblies is relying on test data for onesided connection assemblies, consideration should be given to maintaining the level of panel zone deformation in the design to a level consistent with that of the test, or at least assume that the panel zone must remain elastic, under the maximum expected shear demands. Commentary: At present, no changes are recommended to the code requirements governing the design of panel zones, other than in the calculation of the demand. As indicated in Section 7.5.2.6, it is recommended that the formulation for panel zone demand contained in the UBC, based on 1.85 times the prescribed seismic forces, not be utilized. This formulation, which is not contained in either the AISC Seismic Provisions or the NEHRP Provisions, is felt to lead to the design of panel zones that are excessively flexible and weak in shear. There is evidence that panel zone yielding may contribute to the plastic rotation capability of a connection. However, there is also concern and some evidence that if the deformation is excessive, a kink will develop in the column flange at the joint with the beam flange and, if the local curvature induced in the beam and column flanges is significant, can contribute to failure of the joint. This would suggest that greater conservatism in column panel zone design may be warranted. In addition to the influence of the deformation of the panel zone on the connection performance, it should be recognized that the use of doubler plates and especially the welding associated with them is likely to be detrimental to the connection performance. It is recommended that the Engineer consider use of column sizes which will not require addition of doubler plates, where practical. 7.8.2 Design of Web Connections to Column Flanges

Specific modifications to the code requirements for design of shear connections are not made at this time. It should be noted that the emergency code change to the UBC-94 {NEHRP-94} deleted the former requirements for supplemental web welds on shear connections. This is felt to be appropriate since these welds can apparently contribute to the potential for shear tab failure at large induced rotations. When designing shear connections for moment-resisting assemblies, the designer should calculate shear demands on the web connection in accordance with Section 7.5.2.4, above. For

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connection designs based on tested configurations, the web connection design should be consistent with the conditions in the tested assemblies. Commentary: Some engineers consider that it is desirable to develop as much bending strength in the web as possible. Additionally, it has been observed in some laboratory testing that pre-mature slip of the bolted web connection can result in large secondary flexural stresses in the beam flanges and the welded joints to the column flange. However, there is some evidence to suggest that if flange connections should fail, welding of shear tabs to the beam web may promote tearing of the tab weld to the column flange or the tab itself through the bolt holes, and some have suggested that welding be avoided and that web connections should incorporate horizontally slotted holes to limit the moment which can be developed in the shear tab, thereby protecting its ability to resist gravity loads on the beam in the event of flexural connection failure. Some recent finite element studies of typical connections by Goel, Popov and others have suggested that even when the shear tab is welded, shear demands at the connections tend to be resisted by a diagonal tension type behavior in the web that tends to result in much of the shear being resisted by the flanges. Investigation of these effects is continuing. 7.8.3 Design of Continuity Plates

There are no modifications to the Guidelines or Commentary of Section 7.8.3 at this time. 7.8.4 Design of Weak Column and Weak Way Connections

There are no modifications to the Guidelines or Commentary of Section 7.8.4 at this time. 7.9 Moment Frame Connections for Consideration in New Construction There are no modifications to the Guidelines or Commentary of Section 7.9 at this time. 7.9.1 Cover Plate Connections

Figure 7-5 Figure 7.9.1-1 illustrates the basic configuration of cover plated connections. Short cover plates are added to the top and bottom flanges of the beam with fillet welds adequate to transfer the cover plate forces to the beam flanges. The bottom flange cover plate is shop welded to the column flange and the beam bottom flange is field welded to the column flange and to the cover plate. The top flange and the top flange cover plate are both field welded to the column flange with a common weld. The web connection may be either welded or high strength (slip critical) bolted. Limited testing of these connections (Engelhardt & Sabol - 1994), (Tsai & Popov -1988) has been performed. More than 30 tests of such connections have been performed, with data on at least 18 of these tests available in the public domain. New Construction 7-40

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A variation of this concept which has been tested successfully very recently (Forell/Elsesser Engineers -1995), uses cover plates sized to take the full flange force, without direct welding of the beam flanges themselves to the column. In this version of the detail, the cover plate provides a cross sectional area at the column face about 1.7 times that of the beam flange area. In the detail which has been tested, a welded shear tab is used, and is designed to resist a significant portion of the plastic bending strength of the beam web.

T&B

Figure 7-5 Figure 7.9.1-1 - Cover Plate Connection Design Issues: Following the Northridge earthquake, the University of Texas at Austin conducted a program of research, under private funding, to develop a modified connection configuration for a specific project. Following a series of unsuccessful tests on various types of connections, approximatelyApproximately eight connections similar to that shown in Figure 7-5 Figure 7.9.1-1 were have been recently tested (Engelhardt & Sabol - 1994), and they have demonstrated the ability to achieve acceptable levels of plastic rotation provided that the beam flange to column flange welding wasis correctly executed and through-thickness problems in the column flange were are avoided. This configuration is relatively economical, compared to some other reinforced configurations, and has limited architectural impact. As a result of these factors, and the significant publicity that followed the first successful tests of these connections, cover plated connections quickly became the predominant configuration used in the design of new buildings. As a result, a number of qualification tests have now been performed on different variations of cover plated connections, covering a wide range of member sizes ranging from W16 to W36 beams, as part of the design process for individual building projects. The results of these tests have been somewhat mixed, with a significant number of failures reported. Although this connection type appears to be significantly more reliable than the typical pre-Northridge connection, it should be expected that some connections in buildings incorporating this detail may still be subjected to earthquake initiated fracture damage. Designers should consider using alternative connection types, unless highly redundant framing systems are employed.

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Six of eight connections tested by the University of Texas at Austin were able to achieve plastic rotations of at least 0.025 radians, or better. Strength loss at the extreme levels of plastic rotation did not reduce the flexural capacity to less than the plastic moment capacity of the section based on minimum specified yield strength. One specimen achieved plastic rotations of 0.015 radians when a brittle fracture of the CJP weld (type W2 failure) occurred. This may partially be the result of a weld that was not executed in conformance with the specified welding procedure specification. The second unsuccessful test specimen achieved plastic rotations of 0.005 radian when a section of the column flange (type C2 failure) occurred. A similar failure occurred in recent testing by Popov of a specimen with cover plates having a somewhat modified plan shape. Quantitative Results: No. of specimens tested: 18 Girder Size: W21 x 68 to W36 x 150 Column Size: W12 x 106 to W14 x 455 Plastic Rotation achieved6 13 Specimens : >0.025 radian 1 3 Specimens: 0.015 0.005 < θp < 0.025 radian 1 2 Specimens: 0.005 radian Although apparently more reliable than the former prescriptive connection, this configuration is subject to some of the same flaws including dependence dependent on properly executed beam flange to column flange welds, and through-thickness behavior of the column flange. Further these effects are somewhat exacerbated as the added effective thickness of the beam flange results in a much larger groove weld at the joint, and therefore potentially more severe problems with brittle heat affected zones and lamellar defects in the column. Indeed, a significant percentage of connections of this configuration have failed to produce the desired amount of plastic rotation. One of the issues that must be faced by designers utilizing cover plated connections is the sequence of operations used to attach the cover plate and beam flange to the column. In one approach, the bottom cover plate is shop welded to the column, and then used as the backing for the weld of the beam bottom flange to the column flange. This approach has the advantage of providing an erection seat and also results in a somewhat reduced amount of field welding for this joint. A second approach is to attach the cover plate to the beam flange, and then weld it to the column, in the field, as an integral part of the beam flange. There are tradeoffs to both approaches. The latter approach results in a relatively large field weld at the bottom flange with large heat input required into the column and beam. If this operation is not performed with proper preheat and control of the heat input, it can potentially result in an enlarged and brittle heat affected zone in both members. The first approach results in reduced heat input and therefore, somewhat minimized potential for this effect. However, proper control of preheat and heat input remains as important in either case, as improper procedures can still result in brittle conditions in the heat affected zone. Further, the detail in which the cover plate is shop welded to the column can lead to a notch effect for the column flange at the seam between the beam

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flange and cover plate. This is effect is illustrated in Figure 7.9.1-2. At least one specimen employing this detail developed a premature fracture across the column flange that has been related to this notch effect. This effect has been confirmed by recent fracture mechanics modeling of this condition conducted by Deierlein. When developing cover plated connection details, designers should attempt to minimize the total thickness of beam flange and cover plate, so as to reduce the size of the complete joint penetration weld of these combined elements to the column flange. For some frame configurations and member sizes, this combined thickness and the resulting CJP weld size can approach or even exceed the thickness of the column flange. While there is no specific criteria in the AWS or AISC specifications that would suggest such weldments should not be made, judgementally they would not appear to be desirable from either a constructability or performance perspective. As a rough guideline, it is recommended that for connections in which both the beam flange and cover plate are welded to the column flange, the combined thickness of these elements should not exceed twice the thickness of the beam flange nor 100% of the thickness of the column flange. For cover plated connections in which only the cover plate is welded to the column flange, the same thickness limits should be applied to the cover plate.

seam acts as notch beam bottom flange column flange (in tension)

cover plate

Figure 7.9.1-2 Notch Effect at Cover Plated Connections 7.9.2 Flange Rib Connections

There are no modifications to the Guidelines or Commentary of Section 7.9.2 at this time.

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7.9.3 Bottom Haunch Connections

Figure 7.9.3-1 7-7 indicates the configuration of a connection with a haunch at the bottom beam flange.several potential configurations for single, haunched beam-column connections. As with the cover plated and ribbed connections, the intent is to shift the plastic hinge away from the column face and to reduce the demand on the CJP weld by increasing the depth of the section. To date, the configuration incorporating the triangular haunch has been subjected to limited testing. Testing of configurations incorporating the straight haunch are currently planned, but have not yet been performed. Several tests of this connection type were conducted by Uang under the SAC phase I project (Uang, 1995). Following that work, additional research on the feasibility of improving connection performance with welded haunches was conducted under a project that was jointly sponsored by NIST and AISC (NIST, 1998). That project was primarily focused on the problem of upgrading connections in existing buildings. As indicated in the report of that work, the haunched modification improves connection performance by altering the basic behavior of the connection. In essence, the haunch creates a prop type support, beneath the beam bottom flange. This both reduces the effective flexural stresses in the beam at the face of the support, and also greatly reduces the shear that must be transmitted to the column through the beam. A complete procedure for the design of this modification may be found in NIST, 1998. Figure 7-7 - Bottom Haunch Connection Modification

Figure 7.9.3-1 Bottom Haunch Connecction Two Nine tests are known to have been performed to date, both successfully all intended to replicate the condition of an existing connection that has been upgraded. Except for those specimens in which existing vulnerable welded joints were left in place at the top flange, these connections generally achieved large plastic rotations. Several dynamic tests have also been New Construction 7-44

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successfully conducted, although only moderate plastic deformation demands could be imposed due to limitations of the laboratory equipment. Both tests were conducted in a repair/modification configuration. In one test, a portion of the girder top flange, adjacent to the column, was replaced with a thicker plate. In addition, the bottom flange and haunch were both welded to the column. This specimen developed a plastic hinge within the beam span, outside the haunched area and behaved acceptably. A second specimen did not have a thickened top flange and the bottom girder flange was not welded to the column. Plastic behavior in this specimen occurred outside the haunch at the bottom flange and adjacent to the column face at the top flange. Failure initiated in the girder at the juncture between the top flange and web, possibly contributed to by buckling of the flange as well as lateral torsional buckling of the section. Fracture progressed slowly along the top fillet of the girder and eventually, traveled into the flange itself. Design Issues: The haunch can be attached to the girder in the shop, reducing field erection costs. Weld sizes are smaller than in cover plated connections. The top flange is free of obstructions. Quantitative Results: No. of specimens tested: 92 Girder Size: W30 x 99 Column Size: W14 x 176 Plastic Rotation achievedSpecimen 1 UCSD-1R:0.04 radian (w/o bottom flange weld and reinforced top flange) Specimen 2 UCSD-3R:0.05 radian (with bottom flange weld and reinforced top flange) Specimen UCSD-4R: 0.014 radian (dynamic- limited by test setup) Specimen UCSD-5R: 0.015 radian (dynamic- limited by test setup) Girder Size: W36x150 Column Size: W14x257 Plastic Rotation achieved Specimen UCB-RN2: 0.014 radian (no modification of top weld) Specimen UTA-1R: 0.019 radian (partial modification of top weld) Specimen UTA-1RB: 0.028 radian (modified top weld) Girder Size: W36x150 Column Size: W14x455 Plastic Rotation achievedSpecimen UTA-NSF4: 0.015 radian (no modification of top weld) Girder Size: W18x86 Column Size: W24x279 Plastic Rotation achievedSpecimen SFCCC-8: 0.035 radian (cover plated top flange) New Construction 7-45

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Performance is dependent on properly executed complete joint penetration welds at the column face. The joint can be subject to through-thickness flaws in the column flange; however, this connection may not be as sensitive to this potential problem because of the significant increase in the effective depth of the beam section which can be achieved. Welding of the bottom haunch requires overhead welding when relatively shallow haunches are used. The skewed groove welds of the haunch flanges to the girder and column flanges may be difficult to execute. The increased depth of the beam, resulting from the haunch may have undesirable impact on architectural design. Unless the top flange is prevented from buckling at the face of the column, performance may not be adequate. For configurations incorporating straight haunches, the haunch must be long, in order to adequately develop stress into the haunch, through the web. This tends to increase demands at the column face. Additional testing of all these configurations is recommended. 7.9.4 Top and Bottom Haunch Connections

There are no modifications to the Guidelines or Commentary of Section 7.9.4 at this time. 7.9.5 Side-Plate Connections

There are no modifications to the Guidelines or Commentary of Section 7.9.5 at this time. 7.9.6 Reduced Beam Section Connections

In this connection, the cross section of the beam is intentionally reduced within a segment, to produce an intended plastic hinge zone or fuse, located within the beam span, away from the column face. Several ways of performing this cross section reduction have been proposed. One method includes removal of a portion of the flanges, symmetrical about the beam centerline, in a so-called “dog bone” profile. Care should be taken with this approach to provide for smoothly contoured transitions to avoid the creation of stress risers which could initiate fracture. It has also been proposed to create the reduced section of beam by drilling a series of holes in the beam flanges. Figure 7-11 Figure 7.9.6-1 illustrates both concepts. The most successful configurations have used reduced sections formed with circular cuts. Configurations which taper the reduced section, through the use of unsymmetrical cut-outs, or variable size holes, to balance the cross section and the flexural demand have also been tested with success. Testing of this concept was first performed by a private party, and US patents were applied for and granted. These patents have now been released. Limited testing of both “dog-bone” and drilled hole configurations have been performed in Taiwan (Chen and Yeh - 1995). The American Institute of Steel Construction is currently performing additional tests of this configuration (Smith-Emery - 1995), however the full results of this testing are not yet available. has performed successful testing of 4 linearly tapered RBS connections. In the time since the first publication of the Interim Guidelines, a number of tests have been successfully conducted of RBS connections with circular curved cut-outs, including investigations and at the University of Texas at Austin, has successfully tested 4 circular curved RBS specimens. Others, including Popov at the New Construction 7-46

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University of California at Berkeley, and Texas A&M University., have also tested circular curved RBS connections with success. When this connection type was first proposed, There is a concern was expressed that the presence of a concrete slab at the beam top flange would tend to limit the effectiveness of the reduced section of that flange, particularly when loading places the top flange into compression. It may be possible to mitigate this effect with proper detailing of the slab. Limited testing of RBS specimens with composite slabs has recently been successfully conducted at Ecole Polytechnic, in Montreal, Canada. In these tests, shear studs were omitted from the portion of the top flange having a reduced section, in order to minimize the influence of the slab on flexural hinging. In addition, a 1 inch wide gap was placed in the slab, around the column, to reduce the influence of the slab on the connection at the column face. More recently, both the University of Texas at Austin and Texas A&M University have conducted successful tests of RBS connections with slabs and without such gaps present between the slab and column. This most recent testing suggests that the presence of the slab actually enhances connection behavior by retarding buckling of the top flange in compression and delaying strength degradation effects commonly observed in specimens tested without slabs. Design Issues: This connection type is potentially the most economical of the several types which have been suggested. The reliability of this connection type is dependent on the quality of the complete joint penetration weld of the beam to column flange, and the through-thickness behavior of the column flange. If the slab is not appropriately detailed, it may inhibit the intended “fuse” behavior of the reduced section beam segment. It is not clear at this time whether it would be necessary to use larger beams with this detail to attain the same overall system strength and stiffness obtained with other configurations. In limited testing conducted to date of the unsymmetrical “dog-bone” configuration (Smith-Emery - 1995), the plastic hinging which occurred at the reduced section was less prone to buckling of the flanges than in some of the other configurations which have been tested, due to the very compact nature of the flange in the region of the plastic hinge. However, the tendency for lateral-torsional buckling is significantly increased suggesting the need for lateral bracing of the beam flanges, near the reduced section. Experimental Results: A number of researchers have performed tests on RBS specimens to date. Most tests have utilized the ATC-24 loading protocol, which is similar to the protocol described in Section 7.4.1 of the Interim Guidelines. Testing employed at Ecole Polytechnic, in Montreal, Canada utilized a series of different testing protocols including the ATC-24 procedure and a dynamic excitation simulating the response of a connection in a building to an actual earthquake accelerogram (Tremblay, et. al., 1997). This research included two tests of connections with composite floor slabs. All of the reported tests with circular flange cuts have performed acceptably, however, the dynamic tests at Ecole Polytechnic only imposed 0.025 radians of plastic rotation on the assembly.

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Straight

Tapered

Circular Reduced Section

Drilled Constant

Drilled Tapered

Figure 7-11 7.9.6-1 - Reduced Beam Section Connection Quantitative Results: No. of specimens tested: 219 published (without slabs)2 Girder Size: W21 x 62W30 x 99 thru W 36 x 194 Column Size: W14x120W14 x 176 thru W 14 x 426, W24 x 229 Plastic Rotation achieved:- 0.03 radian Straight: - 0.02 radian Tapered - 0.027 - 0.045 radian Circular - 0.03 - 0.04 radian No. of specimens tested: 42published (with slabs) Girder Size: W21 x 44 to W36 x 150 Column Size: W14 x 90 to W14x257 Plastic Rotation achieved: 0.03-0.05 radians (ATC-24 loading protocol) 0.025 radians (earthquake simulation – limited by laboratory setup, no failure observed) 7.9.7 Slip - Friction Energy Dissipating Connection

There are no modifications to the Guidelines or Commentary of Section 7.9.7 at this time. 7.9.8 Column-Tree Connection

There are no modifications to the Guidelines or Commentary of Section 7.9.8 at this time. 7.9.9 Proprietary Slotted Web Connections

In the former prescriptive connection, in which the beam flanges were welded directly to the column flanges, beam flexural stress was transferred into the column web through the combined

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action of direct tension across the column flange, opposite the column web, and through flexure of the column flange. This stress transfer mechanism and its resulting beam flange prying moment results in a large stress concentration at the center of the beam flange, opposite the column web. Recent research (Allen, et. al. - 1995) indicates that the provision of continuity plates within the column panel zone reduces this stress concentration somewhat, but not completely. The intent of the proprietary slotted web connections is to further reduce this stress concentration and to achieve a uniform distribution of flexural stress across the beam flange at the connection, and also, to promote local buckling of the beam flanges under compressive loads to limit the amount of demand on the beam flange to column flange weld. Claimed assets for this connection include elimination of the vertical beam shear in the beam flange welds, elimination of the beam lateral torsional buckling mode, and the participation of the beam web in resisting its portion of the beam moment. A number of different configurations for this connection type have been developed and tested. Figure 7.9.9-17-14 indicates one such configuration for this connection type that has been successfully tested and which has been used in both new and retrofit steel moment-resisting frames. In this configuration, slots are cut into the beam web, extending from the weld access hole adjacent to the top and bottom flanges, and extending along the beam axis a sufficient length to alleviate the stress concentration effects at the beam flange to column flange weld. The beam web is welded to the column flange. vertical plates are placed between the column flanges, opposite the edges of the top and bottom beam flanges to stiffen the outstanding column flanges and draw flexural stress away from the center of the beam flange. Horizontal plates are placed between these vertical plates and the column web to transfer shear stresses to the panel zone. The web itself is softened with the cutting of a vertical slot in the column web, opposite the beam flange. High fidelity finite element models were utilized to confirm that a nearly uniform distribution of stress occurs across the beam flange.

Slot, typ. NOTICE OF CONFIDENTIAL INFORMATION: WARNING: The information presented in this figure is PROPRIETARY. US patents have been granted and Foreign Patents have been applied for. Use of this information is strictly prohibited except as authorized in writing by the developer. Violators shall be prosecuted in accordance with US and Foreign Patent Intellectual Property Laws.

Figure 7.9.9-17-14 - Proprietary Slotted Web Connection New Construction 7-49

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Design Issues: This detail is potentially quite economical, entailing somewhat more shop fabrication than the former prescriptive connection, but similar levels of field erection work. Contrary to the recommendations contained in these Interim Guidelines, this connection does not shift the location of plastic hinging away from the column face. However, two a number of connections employing details similar to that shown in Figure 7-147.9.9-1 have recently been tested successfully (Allen. - 1995). The connection detail is sensitive to the quality of welding employed in the critical welds, including those between the beam and column flanges., and between the vertical and horizontal plates and the column elements. It has been reported that one specimen, with a known defect in the beam flange to column flange weld was informally tested and failed at low levels of loading. The detail is also sensitive to the balance in stiffness of the various plates and flanges. For configurations other than those tested, detailed finite element analyses may be necessary to confirm that the desired uniform stress distribution is achieved. The developer of this detail indicates that for certain column profiles, it may be possible to omit the vertical slots in the column web and still achieve the desired uniform beam flange stress distribution. This detail may also be sensitive to the toughness of the column base metal at the region of the fillet between the web and flanges. In heavy shapes produced by some rolling processes the metal in this region may have substantially reduced toughness properties relative to the balance of the section. This condition, coupled with local stress concentrations induced by the slot in the web may have the potential to initiate premature fracture. The developer believes that it is essential to perform detailed analyses of the connection configuration, in order to avoid such problems. Popov tested one specimen incorporating a locally softened web, but without the vertical and horizontal stiffener plates contained in the detail shown in Figure 7-14. That specimen failed by brittle fracture through the column flange which progressed into the holes cut into the web. The stress patterns induced in that specimen, however, were significantly different than those which occur in the detail shown in the figure. Quantitative Results: Number of specimens tested: 2 Girder Size: W 27x94 Column Size: W 14x176 Plastic Rotation Achieved: Specimen 1: 0.025 radian Specimen 2: 0.030 radian Quantitative data on connection testing may be obtained from the licensor. 7.9.10 Bolted Bracket Connections

Framing connections employing bolted or riveted brackets have been used in structural steel construction since its inception. Early connections of this type were often quite flexible, and also had limited strength compared to the members they were connecting, resulting in partially restrained type framing. However, it is possible to construct heavy bolted brackets employing New Construction 7-50

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high strength bolts to develop fully restrained moment connections. Pretensioing of the bolts or threaded rods attaching the brackets to the column flanges and use of slip-critical connections between the brackets and beam flanges can help to provide the rigidity required to obtain fully restrained behavior. Reinforcement of the column flanges may be required to prevent local yielding and excessive deformation of these elements, as well. Two alternative configurations that have been tested recently are illustrated in Figure 7.9.10-1. The developer of these configurations offers the brackets in the form of proprietary steel castings. Several tests of these alternative connections have been performed on specimens with beams ranging in size from W16 to W36 sections and with large plastic rotations successfully achieved. Design Issues: The concept of bolted bracket connections is similar to that of the riveted “wind connections” commonly installed in steel frame buildings in the early twentieth century. The primary difference is that the riveted wind connections were typically limited in strength either by flexural yielding of outstanding flanges of the brackets, or shear and tension on the rivets, rather than by flexural hinging of the connected framing. Since the old-style wind connections could not typically develop the flexural strength of the girders and also could be quite flexible, they would be classified either as partial strength or partially restrained connections. Following the Northridge earthquake, the concept of designing such connections with high strength bolts and heavy plates, to behave as fully restrained connections, was developed and tested by a private party who has applied for patents on the concept of using steel castings for this purpose. Bracket

High tensile threaded rod

Pipe Plate

Bolts

WARNING: The information presented in this figure is PROPRIETARY. US and Foreign Patents have been applied for. Use of this information is strictly prohibited except as authorized in writing by the developer. Violators shall be prosecuted in accordance with US and Foreign Patent Intellectual Property Laws.

Figure 7.9.10-1 Bolted Bracket Connections Bolted connections offer a number of potential advantages over welded connections. Since no field welding is required for these connections, they are inherently less labor intensive during New Construction 7-51

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erection, and also less dependent on the technique of individual welders for successful performance. However, quality assurance should be provided for installation and tensioning of the bolts, as well as correction of any problems with fit-up due to fabrication tolerances. Experimental Results: A series of tests on several different configurations of proprietary heavy bolted bracket connections have been performed at Lehigh University (Ksai & Bleiman, 1996) to qualify these connections for use in repair and modification applications. To test repair applications, brackets were placed only on the bottom beam flange to simulate installations on a connection where the bottom flange weld in the original connection had failed. In these specimens, bottom flange welds were not installed, to approximate the condition of a fully fractured weld. The top flange welds of these specimens were made with electrodes rated for notch toughness, to preclude premature failure of the specimens at the top flange. For specimens in which brackets were placed at both the top and bottom beam flanges, both welds were omitted. Acceptable plastic rotations were achieved for each of the specimens tested. Quantitative Results: No. of specimens tested: 8 Girder Size: W16x40 and W36x150 Column Size: W12x65 and W14x425 Plastic Rotation achieved - 0.05 radians - 0.07 radians 7.10 Other Types of Welded Connection Structures There are no modifications to the Guidelines or Commentary of Section 7.10 at this time. 7.10.1 Eccentrically Braced Frames (EBF)

There are no modifications to the Guidelines or Commentary of Section 7.10.1 at this time. 7.10.2 Dual Systems

There are no modifications to the Guidelines or Commentary of Section 7.10.2 at this time. 7.10.3 Welded Base Plate Details

There are no modifications to the Guidelines or Commentary of Section 7.10.3 at this time. 7.10.4 Vierendeel Truss Systems

There are no modifications to the Guidelines or Commentary of Section 7.10.4 at this time. 7.10.5 Moment Frame Tubular Systems

There are no modifications to the Guidelines or Commentary of Section 7.10.5 at this time.

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7.10.6 Welded Connections of Collectors, Ties and Diaphragm Chords

There are no modifications to the Guidelines or Commentary of Section 7.10.6 at this time. 7.10.7 Welded Column Splices

There are no modifications to the Guidelines or Commentary of Section 7.10.7 at this time. 7.10.8 Built-up Moment Frame Members

There are no modifications to the Guidelines or Commentary of Section 7.10.8 at this time.

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8. METALLURGY & WELDING 8.1 Parent Materials 8.1.1 Steels

Designers should specify materials which are readily available for building construction and which will provide suitable ductility and weldability for seismic applications. Structural steels which may be used in the lateral-force-resisting systems for structures designed for seismic resistance without special qualification include those contained in Table 8.1.1-1. Refer to the applicable ASTM reference standard for detailed information. Table 8.1.1-1 - Structural Steel Prequalified for Use in Seismic Lateral-Force-Resisting Systems ASTM Specification ASTM A36 ASTM A283 Grade D ASTM A500 (Grades B & C) ASTM A501 ASTM A572 (Grades 42 & 50) ASTM A588 ASTM A9921 Notes: 1- See Commentary

Description Carbon Structural Steel Low and Intermediate Tensile Strength Carbon Steel Plates Cold-Formed Welded & Seamless Carbon Steel Structural Tubing in Rounds & Shapes Hot-Formed Welded & Seamless Carbon Steel Structural Tubing High-Strength Low-Alloy Columbium-Vanadium Steels of Structural Quality High-Strength Low-Alloy Structural Steel (weathering steel) Steel for Structural Shapes for Use in Building Framing

Structural steels which may be used in the lateral-force-resisting systems of structures designed for seismic resistance with special permission of the building official are those listed in Table 8.1.1-2. Steel meeting these specifications has not been demonstrated to have adequate weldability or ductility for general purpose application in seismic-force-resisting systems, although it may well possess such characteristics. In order to demonstrate the acceptability of these materials for such use in WSMF construction it is recommended that connections be qualified by test, in accordance with the guidelines of Chapter 7. The test specimens should be fabricated out of the steel using those welding procedures proposed for use in the actual work. Table 8.1.1-2 - Non-prequalified Structural Steel ASTM Specification Description ASTM A242 High-Strength Low-Alloy Structural Steel ASTM A709 Structural Steel for Bridges ASTM A913 High-Strength Low-Alloy Steel Shapes of Structural Quality, Produced by Quenching & Self-Tempering Process Metallurgy & Welding 8-1

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Commentary: Many WSMF structures designed in the last 10 years incorporated ASTM A36 steel for the beams and ASTM A572 grade 50 steel for the columns. This provided an economical way to design structures for the strong column weak beam provisions contained in the building code. Recent studies conducted by the Structural Shape Producers Council (SSPC), however, indicate that material produced to the A36 specification has wide variation in strength properties with actual yield strengths that often exceed 50 ksi. This wide variation makes prediction of connection and frame behavior difficult. Some have postulated that one of the contributing causes to damage experienced in the Northridge earthquake was inadvertent pairing of overly strong beams with average strength columns. The AISC and SSPC have been working for several years to develop a new specification for structural steel that would have both minimum and maximum yield values defined and provide for a margin between maximum yield and minimum ultimate tensile stress. AISC recently submitted such a specification, for a material with 50 ksi specified yield strength, to ASTM for development into a standard specification. ASTM formally adopted the new specification for structural shapes, with a yield strength of 50 ksi, under designation A992 in 1998 and It is anticipated that domestic mills will begin have begun producing structural wide flange shapes to this specification. within a few years and that eventually, this new material will replace A36 as the standard structural material for incorporation into lateral-force-resisting systems. Since the formal approval of the A992 specification by ASTM occurred after publication of the 1997 editions of the building codes and the AISC Seismic Specification, it is not listed in any of these documents as a prequalified material for use in lateral force resisting systems. Neither is it listed as prequalified in AWS D1.1-98. However, all steel that complies with the ASTM-992 specification will also meet the requirements of ASTM A572, Grade 50 and should therefore be permissible for any application for which the A572 material is approved. See also, the commentary to Section 8.2.2. Under certain circumstances it may be desirable to specify steels that are not recognized under the UBC for use in lateral-force-resisting systems. For instance, ASTM A709 might be specified if the designer wanted to place limits on toughness for fracture-critical applications. In addition, designers may wish to begin incorporating ASTM A913, Grade 65 steel, as well as other higher strength materials, into projects, in order to again be able to economically design for strong column - weak beam conditions. Designers should be aware, however, that these alternative steel materials may not be readily available. It is also important when using such non-prequalified steel materials, that precautions be taken to ensure adequate weldability of the material and that it has sufficient ductility to perform under the severe loadings produced by earthquakes. The Metallurgy & Welding 8-2

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cyclic test program recommended by these Interim Guidelines for qualification of connection designs, by test, is believed to be an adequate approach to qualify alternative steel material for such use as well. Note that ASTM A709 steel, although not listed in the building code as prequalified for use in lateral-force-resisting systems, actually meets all of the requirements for ASTM A36 and ASTM A572. Consequently, special qualification of the use of this steel should not be required. Although the 1994 editions of the Uniform Building Code and the NEHRP Provisions do not prequalify the use of ASTM A913 steel in lateral force resisting systems, the pending 1997 edition of the UBC does prequalify its use. Both the 1997 NEHRP Provisions and the AISC Seismic Provisions prequalify the use of this steel in elements that do not undergo significant yielding, for example, the columns of moment-resisting frames designed to meet strong column - weak beam criteria. Consequently, special approval of the Building Official should no longer be required as a pre-condition of the use of material conforming to this specification, at least for columns. 8.1.2 Chemistry

There are no modifications to the Guidelines of Section 8.1.2 at this time. Commentary: Some concern has been expressed with respect to the movement in the steel producing industry of utilizing more recycled steel in its processes. This results in added trace elements not limited by current specifications. Although these have not been shown quantitatively to be detrimental to the performance of welding on the above steels, a the new A992specification for structural steel proposed by AISC does place more control on these trace elements. Mill test reports now include elements not limited in some or all of the specifications. They include copper, columbium, chromium, nickel, molybdenum, silicon and vanadium. The analysis and reporting of an expanded set of elements should be possible, and could be beneficial in the preparation of welding procedure specifications (WPSs) by the welding engineer if critical welding parameters are required. Modern spectrographs used by the mills are capable of automated analyses. When required by the engineer, a request for special supplemental requests should be noted in the contract documents. 8.1.3 Tensile/Elongation Properties

Mechanical property test specimens are taken from rolled shapes or plates at the rolling mill in the manner and location prescribed by ASTM A6 and ASTM A370. Table 8-3 Table 8.1.3-1 gives the basic mechanical requirements for commonly used structural steels. Properties specified, and controlled by the mills, in current practice include minimum yield strength or yield point, ultimate

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tensile strength and minimum elongation. However, there can be considerable variability in the actual properties of steel meeting these specifications. SSPC, in cooperation with SEAOC, has collected statistical data on the strength characteristics of two grades (ASTM A36 and ASTM A572 Grade 50) of structural steels, based on mill test reports from selected domestic producers for the 1992 production year. Data were also collected for "Dual Grade" material that was certified by the producers as complying with both ASTM A36 and ASTM A572 Grade 50. Table 8-4 Table 8.1.3-2 summarizes these results as well as data provided by a single producer for ASTM A913 material. Unless special precautions are taken to limit the actual strength of material incorporated into the work to defined levels, new material specified as ASTM A36 should be assumed to be the dual grade for connection demand calculations, whenever the assumption of a higher strength will result in a more conservative design condition. Table 8-3 Table 8.1.3-1 - Typical Tensile Requirements for Structural Shapes Minimum Yield Ultimate Tensile Minimum Elongation Minimum Elongation Strength or Yield Strength, Ksi % % Point, Ksi in 2 inches in 8 inches A36 36 Min. 58-801 212 20 4 A242 42 Min.. 63 MIN. 213 18 A572, Gr. 42 42 Min. 60 Min. 24 20 A572, GR50 50 Min. 65 MIN. 212 18 A588 50 Min. 70 MIN. 213 18 A709, GR36 36 Min. 58-80 212 20 A709, GR50 50 Min. 65 MIN. 21 18 A913, GR50 50 Min. 65 MIN. 21 18 A913, GR65 65 Min. 80 MIN. 17 15 A992 50 Min. – 65 Max. 65 MIN 21 18 Notes: 1No maximum for shapes greater than 426 lb./ft. 2Minimum is 19% for shapes greater than 426 lb. /ft. 3No limit for Shape Groups 1, 2 and 3.Minimum is 18% for shapes greater than 426 lb./ft. 4. Minimum is 50 ksi for Shape Groups 1 and 2, 46 ksi for Shape Group 3, and 42 ksi for Shape Groups 4 and 5. ASTM

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Table 8-4 Table 8.1.3-2 - Statistics for Structural Shapes1,2 Statistic

A 36

Dual GRADE

A572 GR50

A913 GR65

Mean Minimum Maximum Standard Deviation [ s ] Mean + 1 s

Yield Point (ksi) 49.2 36.0 72.4 4.9 54.1

55.2 50.0 71.1 3.7 58.9

57.6 50.0 79.5 5.1 62.7

75.3 68.2 84.1 4.0 79.3

Mean Minimum Maximum Standard Deviation [ s ] Mean + 1 s

Tensile Strength (ksi) 68.5 58.0 88.5 4.6 73.1

73.2 65.0 80.0 3.3 76.5

75.6 65.0 104.0 6.2 81.8

89.7 83.4 99.6 3.5 93.2

Mean Minimum Maximum Standard Deviation [ s ] Mean + 1 s Mean - 1 s

Yield/Tensile Ratio 0.72 0.51 0.93 0.06 0.78 0.66

0.75 0.65 0.92 0.04 0.79 0.71

0.76 0.62 0.95 0.05 0.81 0.71

0.84 0.75 0.90 0.03 0.87 0.81

1: The data presented for ASTM A36, “Dual Grade” and ASTM A572 Grade 50 were included as part of the SSPC study (SSPC-1994). The data for ASTM A913 were derived from a single producer and may not be available from all producers. 2. Statistical Data on the distribution of strength properties for material meeting ASTM A992 are not presently available. Pending the development of such statistics, it should be assumed that A992 material will have similar properties to ASTM A572, Gr. 50 material.

Commentary: The data given in Table 8-4 Table 8.1.3-2 for A36 and A572 Grade 50 is somewhat weighted by the lighter, Group 1 shapes that will not ordinarily be used in WSMF applications. Excluding Group 1 shapes and combining the Dual Grade and A572 Grade 50 data results in a mean yield strength of 48 ksi for A36 and 57 ksi for A572 Grade 50 steel. It should also be noted that approximately 50% of the material actually incorporated in a project will have yield strengths that exceed these mean values. For the design of facilities with stringent requirements for limiting post-earthquake damage, consideration of more conservative estimates of the actual yield strength may be warranted. Until recently, In wide flange sections the tensile test coupons in wide flange sections are currently were taken from the web. The amount of reduction rolling, finish rolling temperatures and cooling conditions affect the tensile and impact Metallurgy & Welding 8-5

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properties in different areas of the member. Typically, the web exhibits about five percent higher strength than the flanges due to faster cooling. In 1998 ASTM A6 was revised to specify that coupons be taken from the flange of wide flange shapes. Design professionals should be aware of the variation in actual properties permitted by the ASTM specifications. This is especially important for yield strength. Yield strengths for ASTM A36 material have consistently increased over the last 15 years so that several grades of steel may have the same properties or reversed properties, with respect to beams and columns, from those the designer intended. Investigations of structures damaged by the Northridge earthquake found some WSMF connections in which beam yield strength exceeded column yield strength despite the opposite intent of the designer. As an example of the variations which can be found, Table 8-5 Table 8.1.3-2 presents the variation in material properties found within a single building affected by the Northridge earthquake. Properties shown include measured yield strength (Fya,), measured tensile strength (Fua ) and Charpy V-Notch energy rating (CVN). Table 8-5Table 8.1.3-2 - Sample Steel Properties from a Building Affected by the Northridge Earthquake Shape

Fya1 ksi

Fua, ksi

CVN, ft-lb.

W36 X 182

38.0

69.3

18

W36 X 230

49.3

71.7

195

Note 1 - ASTM A36 material was specified for both structures.

The practice of dual certification of A36 and A572, Grade 50 can result in mean yield strengths that are fifty percent higher than the specified yield of A36. Since there is no practical way to discern whether dual grade steel will be supplied, unless direct purchase of steel from specific suppliers is made, in the absence of such procurement practices, the prudent action for determining connection requirements, where higher strengths could be detrimental to the design, would be to assume the dual grade material whenever A36 or A572 Grade 50 is specified. In the period since the initial publication of the Interim Guidelines, several researchers and engineers engaged in connection assembly prototype testing have reported that tensile tests on coupons extracted from steel members used in the prototype tests resulted in lower yield strength than reported on the mill test report furnished with the material, and in a few cases lower yield strength than would be permitted by the applicable ASTM specification. This led to some confusion and concern, as to how mill test reports should be interpreted. Metallurgy & Welding 8-6

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The variation of the measured yield strength of coupons reported by researchers engaged in connection prototype testing, as compared to that indicated on the mill test reports, is not unusual and should be expected. These variations are the result of a series of factors including inconsistencies between the testing procedures employed as well as normal variation in the material itself. The following paragraphs describe the basis for the strengths reported by producers on mill certificates, as well as the factors that could cause independent investigators to determine different strengths for the same material. Mill tests of mechanical properties of steel are performed in accordance with the requirements of ASTM specifications A6 and A370. ASTM A6 had historically required that test specimens for rolled W shapes be taken from the webs of the shapes, but recently was revised to require testing from the flanges of wide flange shapes with 6 inch or wider flanges. A minimum of two tests must be made for each heat of steel, although additional tests are required if shapes of significantly different thickness are cast from the same heat. Coupon size and shape is specified based on the thickness of the material. The size of the coupon used to test material strength can effect the indicated value. Under ASTM A6, material that is between 3/4 inches thick and 4 inches thick can either be tested in full thickness “straps” or in smaller 1/2” diameter round specimens. In thick material, the yield strength will vary through the thickness, as a result of cooling rate effects. The material at the core of the section cools most slowly, has larger grain size and consequently lower strength. If full-thickness specimens are used, as is the practice in most mills, the recorded yield strength will be an average of the relatively stronger material at the edges of the thickness and the lower yield material at the center. Many independent laboratories will use the smaller 1/2” round specimens, and sometimes even sub-sized 1/4” round specimens for tensile testing, due to limitations of their testing equipment. Use of these smaller specimens for thick material will result in testing only of the lower yield strength material at the center of the thickness. ASTM A370 specifies the actual protocol for tensile testing including the loading rate and method of reporting test data. Strain rate can affect the strength and elongation values obtained for material. High strain rates result in elevated strength and reduced ductility. Under ASTM A370, yield values may be determined using any convenient strain rate, but not more than 1/16 inch per inch, per minute which corresponds to a maximum loading rate of approximately 30 ksi per second. Once the yield value is determined, continued testing to obtain ultimate tensile values can proceed at a more rapid rate, not to exceed 1/2 inch per inch per minute. Under ASTM A370, there are two different ways in which the yield property for structural steel can be measured and reported. These include yield point and yield strength. These are illustrated in Figure 8.1.3-1. The yield point is the peak Metallurgy & Welding 8-7

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stress that occurs at the limit of the elastic range, while the yield strength is a somewhat lower value, typically measured at a specified offset or elongation under load. Although a number of methods are available to determine yield point, the so-called “drop of the beam” method is most commonly used for structural steel. In this method the load at which a momentary drop-off in applied loading occurs is recorded, and then converted to units of stress to obtain the yield point. Yield strength may also be determined by several methods, but is most commonly determined using the offset method. In this method, the stress strain diagram for the test is drawn, as indicated in Figure 8.1.3-1. A specified offset, typically 0.2% strain for structural steel, is laid off on the abscissa of the curve and a line is drawn from this offset, parallel to the slope of the elastic portion of the test. The stress at the intersection of this offset line with the stressstrain curve is taken as the yield strength.

σ

Yield Point Yield Strength

ε

Offset

Figure 8.1.3-1 Typical Stress - Strain Curve for Structural Steel The material specifications for structural steels typically specify minimum values for yield point but do not control yield strength. The SSPC has reported that actual practice among the mills varies, with some mills reporting yield strength and others reporting yield point. This practice is permissible as yield strength will always be a somewhat lower value than yield point, resulting in a somewhat conservative demonstration that the material meets specified requirements. However, this does mean that there is inconsistency between the values reported by the various mills on certification reports. Similarly, the procedures followed by independent testing laboratories may be different than those followed by the mill, particularly with regard to strain rate and the location at which a coupon is obtained. Metallurgy & Welding 8-8

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Under ASTM A6, coupons for tensile tests had historically been obtained from the webs of structural shapes. However, most engineers and researchers engaged in connection testing have preferred to extract material specimens from the flanges of the shape, since this is more representative of the flexural strength of the section. Coupons removed from the web of a rolled shape tend to exhibit somewhat higher strength properties than do coupons removed from the flanges, due to the extra amount of working the thinner web material typically experiences during the rolling process and also because the thinner material cools more rapidly after rolling, resulting in finer grain size. Given these differences in testing practice, as well as the normal variation that can occur along the length of an individual member and between different members rolled from the same heat, the reported differences in strength obtained by independent laboratories, as compared to that reported on the mill test reports, should not be surprising. It is worth noting that following the recognition of these differences in testing procedure, the SSPC in coordination with AISC and ASTM developed and proposed a revision to the A6 specification to require test specimens to be taken from the flanges of rolled shapes when the flanges are 6 inches or more wide. It is anticipated that mills will begin to alter practice to conform to a revised specification in early 1997 This has since become the standard practice. The discovery of the somewhat varied practice for reporting material strength calls into question both the validity of statistics on the yield strength of structural steel obtained from the SSPC study, and its relevance to the determination of the expected strength of the material for use in design calculations. Although the yield point is the quantity controlled by the ASTM material specifications, it has little relevance to the plastic moment capacity of a beam section. Plastic section capacity is more closely related to the stress along the lower yield plateau of the typical stress-strain curve for structural steel. This strength may often be somewhat lower than that determined by the offset drop-of-the-beam method. Since the database of material test reports on which the SSPC study was based appears to contain test data based on both the offset and drop-of-the-beam methods, it is difficult to place great significance in the statistics derived from it and to draw a direct parallel between this data and the expected flexural strength of rolled shapes. It would appear that the statistics reported in the SSPC study provide estimates of the probable material strength that are somewhat high. Thus, the recommended design strengths presented in Tables 6.6.6.3-1 and 7.5.11 of the Interim Guidelines would appear to be conservative with regard to design of welds, panel zones and other elements with demands limited by the beam yield strength. Under the phase II program of investigation, SAC, together with the shape producers, is engaged in additional study of the statistical distribution of yield strength of various materials produced by the mills. This study is intended to provide an improved understanding of the statistical distribution of the lower Metallurgy & Welding 8-9

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yield plateau strength of material extracted from section flanges, measured in a consistent manner. In addition, it will provide correlation with yield strengths determined by other methods such that the data provided on mill test certificates can be properly interpreted and utilized. In addition, the possibility of revising the ASTM specifications to provide for more consistent reporting of strength data as well as the reporting of strength statistics that are directly useful in the design process will be evaluated. In the interim period, the data reported in Table 8-1.32, extracted from the SSPC study, remain the best currently available information. 8.1.4 Toughness Properties

There are no modifications to the Guidelines or Commentary of Section 8.1.4 at this time. 8.1.5 Lamellar Discontinuities

There are no modifications to the Guidelines or Commentary of Section 8.1.5 at this time. 8.1.6 K-Area Fractures

Recently, there have beenIn the period 1995-96 there were several reports of fractures initiating in the webs of column sections during the fabrication process, as flange continuity plates and/or doubler plates were welded into the sections. This fracturing typically initiated in the region near the fillet between the flange and web. This region has been commonly termed the “k-area” because the AISC Manual of Steel Construction indicates the dimension of the fillet between the web and flange with the symbol “k”. The k-area may be considered to extend from mid-point of the radius of the fillet into the web, approximately 1 to 1-1/2 inches beyond the point of tangency between the fillet and web. The fractures typically extended into, and sometimes across, the webs of the columns in a characteristic “half-moon” or “smiley face” pattern. Investigations of materials extracted from fractured members have indicated that the material in this region of the shapes had elevated yield strength, high yield/tensile ratio, high hardness and very low toughness, on the order of a few foot-pounds at 70oF. Material with these properties can behave in a brittle manner. Fracture can be induced by thermal stresses from the welding process or by subsequent weld shrinkage, as apparently occurred in the reported cases. There have been no reported cases of inservice k-line fracture from externally applied loading, as in beam-column connections, although such a possibility is perceived to exist under large inelastic demand. It appears that this local embrittling of sections can be attributed to the rotary straightening process used by some mills to bring the rolled shapes within the permissible tolerances under ASTM A6. The straightening process results in local cold working of the sections, which strain hardens the material. The amount of cold working that occurs depends on the initial straightness of the section and consequently, the extent that mechanical properties are effected is likely to vary along the length of a member. The actual process used to straighten the section can also affect the amount of local cold working that occurs.

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Engineers can reduce the potential for weld-induced fracture in the k-area by avoiding welding within the k-area region. This can be accomplished by detailing doubler plates and continuity plates such that they do not contact the section in this region. The use of large corner clips on beam flange continuity plates can permit this. Selection of column sections with thicker webs, to eliminate the need for doubler plates; the use of fillet welds rather than full penetration groove welds to attach doubler plates to columns, when acceptable for stress transfer; and detailing of column web doubler plates such that they are offset from the face of the column web can also help to avoid these fabrication-induced fracture problems. Commentary: It appears that detailing and fabrication practice can be adjusted to reduce the potential for k-area fracture during fabrication. However, the acceptability of having low-toughness material in the k-area region for service is a question that remains. It is not clear at this time what percentage of the material incorporated in projects is adversely affected, or even if a problem with regard to serviceability exists. SAC recently placed a public call, asking for reports of fabrication-induced fractures at the k-area, but only received limited response. However, in one of the projects that did report this problem, a significant number of columns were affected. This may have been contributed to by the detailing and fabrication practices applied on that project. Other than detailing structures to minimize the use of doubler plates, and to avoid large weldments in the potentially sensitive k-area of the shape, it is not clear at this time, what approach, if any, engineers should take with regard to this issue. There are several methods available to identify possible low notch toughness in structural carbon steels, including Charpy V-Notch testing and hardness testing of samples extracted from the members. However, both of these approaches are quite costly for application as a routine measure on projects and the need for such measures has not yet been established. Following publication of advisories on the k-line problem by AISC, and the publication of similar advisory information in FEMA-267a,reports on this problem diminished. It is not clear whether this is due to revised detailing practice on the part of engineers and fabricators, revised mill rolling practice, or a combination of both. SAC, AISC and SSPC are continuing to research this issue in order to identify if a significant problem exists, and if it does, to determine its basic causes, and to develop appropriate recommendations for mill, design, detailing, and fabrication practices to mitigate the problem. 8.2 Welding 8.2.1 Welding Process

There are no modifications to the Guidelines or Commentary of Section 8.2.1 at this time.

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8.2.2 Welding Procedures

Welding should be performed within the parameters established by the electrode manufacturer and the Welding Procedure Specification (WPS), required under AWS D1.1. Commentary: A welding procedure specification identifies all the important parameters for making a welded joint including the material specifications of the base and filler metals, joint geometry, welding process, requirements for pre- and post-weld heat treatment, welding position, electrical characteristics, voltage, amperage, and travel speed. Two types of welding procedure specifications are recognized by AWS D1.1. These are prequalified procedures and qualified-bytest procedures. Prequalified procedures are those for which the important parameters are specified within the D1.1 specification. If a prequalified procedure is to be used for a joint, all of the variables for the joint must fall within the limits indicated in the D1.1 specification for the specific procedure. If one or more variables are outside the limits specified for the prequalified procedures, then the fabricator must demonstrate the adequacy of the proposed procedure through a series of tests and submit documentation (procedure qualification records) demonstrating that acceptable properties were obtained. Regardless of whether or not a prequalified or qualified-by-test procedure is employed, the fabricator should prepare a welding procedure specification, which should be submitted to the engineer of record for review and be maintained at the work location for reference by the welders and inspectors. The following information is presented to help the engineer understand some of the issues surrounding the parameters controlled by the welding procedure specification. For example, the position (if applicable), electrode diameter, amperage or wire feed speed range, voltage range, travel speed range and electrode stickout (e.g. all passes, 0.072 in. diameter, 248 to 302 amps, 19 to 23 volts, 6 to 10 inches/minute travel speed, 170 to 245 inches/minute wire feed speed, 1/2" to 1" electrode stickout) should be established. This information is generally submitted by the fabricator as part of the Welding Procedure Specification. Its importance in producing a high quality weld is essential. The following information is presented to help the engineer understand some of the issues surrounding these parameters. The amperage, voltage, travel speed, electrical stickout and wire feed speed are functions of each electrode. If prequalified WPSs are utilized, these parameters must be in compliance with the AWS D1.1 requirements. For FCAW and SMAW, the parameters required for an individual electrode vary from manufacturer to manufacturer. Therefore, for these processes, it is essential that the fabricator/erector utilize parameters that are within the range of recommended operation published by the filler metal manufacturer. Alternately, the fabricator/erector could qualify the welding procedure by test in accordance

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with the provisions of AWS D1.1 and base the WPS parameters on the test results. For submerged arc welding, the AWS D1.1 code provides specific amperage limitations since the solid steel electrodes used by this process operate essentially the same regardless of manufacture. The filler metal manufacturer’s guideline should supply data on amperage or wire feed speed, voltage, polarity, and electrical stickout. The guidelines will not, however, include information on travel speed which is a function of the joint detail. The contractor should select a balanced combination of parameters, including travel speed, that will ensure that the code mandated weld-bead sizes (width and height) are not exceeded. Recently, ASTM approved a new material specification for structural steel shape, ASTM A992. This specification is very similar to the ASTM A572, Grade 50 specification except that it includes additional limitations on yield and tensile strengths and chemical composition. Although material conforming to A992 is expected to have very similar welding characteristics to A572 material, it was adopted too late to be included as a prequalified base material in AWS D1.1-98. Although the D1 committee has evaluated A992 and has taken measures to incorporate it as a prequalified material in AWS D1.1-2000, technically, under AWS D1.1-98, welded joints made with this material should follow qualified-bytest procedures. In reality, structural steel conforming to ASTM A992 may actually have somewhat better weldability than material conforming to the A572 specification. This is because A992 includes limits on carbon equivalent, precluding the delivery of steels where all alloys simultaneously approach the maximum specified limits. Therefore, it should be permissible to utilize prequalified procedures for joint with base metal conforming to this specification. 8.2.3 Welding Filler Metals

There are no modifications to the Guidelines of Section 8.2.3 at this time. Commentary: Currently, there are no notch toughness requirements for weld metal used in welding ASTM A 36 or A 572, Grade 50, steel in AWS D1.1. This topic has been extensively discussed by the Welding Group at the Joint SAC/AISC/AISI/NIST Invitational Workshop on September 8 and 9, 1994, and by all participants of the SAC Invitational Workshop on October 28 and 29, 1994. The topic was also considered by the AWS Presidential Task Group, which decided that additional research was required to determine the need for toughness in weld metal. There is general agreement that adding a toughness requirement for filler metal would be desirable and easily achievable. Most filler metals are fairly tough, but some will not achieve even a modest requirement such as 5 ft-lb. at + 70? F. What is not in unanimous agreement is what level of toughness should be required. The recommendation from the Joint Workshop was Metallurgy & Welding 8-13

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20 ft-lb. at -20? F per Charpy V-Notch [CVN] testing. The recommendation from the SAC Workshop was 20 ft-lb. at 30? F lower than the Lowest Ambient Service Temperature (LAST) and not above 0? F. The AWS Presidential Task Group provided an interim recommendation for different toughness values depending on the climatic zone, referenced to ASTM A709. Specifically, the recommendation was for 20 ft-lb. at temperatures of 70 degrees F for Zone 1, 40 degrees F for Zone 2, and 10 degrees F for Zone 3. The AWS also suggested toughness values for base metals used in these applications. Some fractured surfaces in the Northridge and Kobe Earthquakes revealed evidence of improper use of electrodes and welding procedures. Prominent among the misuses were high production deposition rates. Pass widths of up to 11/2 inches and pass heights of 1/2 inch were common. The kind of heat input associated with such large passes promotes grain growth in the HAZ and attendant low notch toughness. In evaluation of welds in buildings affected by the Northridge earthquake, the parameters found to be most likely to result in damage-susceptible welds included root gap, access capability, electrode diameter, stick-out, pass thickness, pass width, travel speed, wire feed rate, current and voltage were found to be the significant problems in evaluation of welds in buildings affected by the Northridge earthquake. Welding electrodes for common welding processes include: AWS A5.20: AWS A5.29: AWS A5.1: AWS A5.5: AWS A5.17: AWS A5.23: AWS A5.25:

Carbon Steel Electrodes for FCAW Low Alloy Steel Electrodes for FCAW Carbon Steel Electrodes for SMAW Low Alloy Steel Covered Arc Welding Electrodes (for SMAW) Carbon Steel Electrodes and Fluxes for SAW Low Alloy Steel Electrodes and Fluxes for SAW Carbon and Low Alloy Steel Electrodes and Fluxes for Electroslag Welding

In flux cored arc welding, one would expect the use of electrodes that meet either AWS A5.20 or AWS A5.29 provided they meet the toughness requirements specified below. Except to the extent that one requires Charpy V-Notch toughness and minimum yield strength, the filler metal classification is typically selected by the Fabricator. Compatibility between different filler metals must be confirmed by the Fabricator, particularly when SMAW and FCAW-SS processes are mixed. Generally speaking, SMAW-type filler metals may not be applied to FCAW-SS type filler metals (e.g. when a weld has been partially removed) while FCAW-type filler metals may be applied to SMAW-type filler metals. This recommendation considers the use of aluminum as a killing agent in FCAW-SS electrodes that can

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be incorporated into the SMAW filler metal with a reduction in impact toughness properties. As an aid to the engineer, the following interpretation of filler metal classifications is provided below: E1X2X3T4X5 For electrodes specified under AWS A5.20 (e.g. E71T1) 1 2 3 4 5 6 E X X T X X For electrodes specified under AWS A5.29 (e.g. E70TGK2) E1XX7X8X9X10 For electrodes specified under AWS A5.1 or AWS A5.5. (e.g. E7018) NOTES: 1.

Indicates an electrode.

2.

Indicates minimum tensile strength of deposited weld metal (in tens of ksi, e.g., 7 = 70 ksi).

3.

Indicates primary welding position for which the electrode is designed (0 = flat and horizontal and 1 = all positions).

4.

Indicates a flux cored electrode. Absence of a letter indicates a "stick" electrode for SMAW.

5.

Describes usability and performance capabilities. For our purposes, it conveys whether or not Charpy V-Notch toughness is required (1, 5, 6 and 8 have impact strength requirements while 2, 4, 7, 10 and 11 do not). A "G" signifies that the properties are not defined by AWS and are to be agreed upon between the manufacturer and the specifier. Impact strength is specified in terms of the number of foot-pounds at a given temperature (e.g., 20 ft-lb. at 0 degrees F). Note that for electrodes specified under AWS A5.20, the format for usage is "T-X".

6.

Designates the chemical composition of deposited metal for electrodes specified under AWS A5.29. Note that there is no equivalent format for chemical composition for electrodes specified under AWS A5.20.

7.

The first two digits (or three digits in a five digit number) designate the minimum tensile strength in ksi.

8.

The third digit (or fourth digit in a five digit number) indicates the primary welding position for which the electrode is designed (1 = all positions, 2 = flat position and fillet welds in the horizontal position, 4 = vertical welding with downward progression and for other positions.)

9.

The last two digits, taken together, indicate the type of current with which the electrode can be used and the type of covering on the electrode. Metallurgy & Welding 8-15

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Indicates a suffix (e.g., A1, A2, B1, etc.) designating the chemical composition of the deposited metal.

Electrode Diameter: (See AWS D1.1 Section 4.14.1.2) The issue of maximum electrode diameter has not been studied sufficiently to determine whether or not electrode diameter is a critical variable. Recent tests have produced modified frame joints with acceptable test results using the previous standard-of-practice 0.120 in. diameter wire. The use of smaller diameter electrodes will slow the rate of deposition (as measured by volume) but will not, in and of itself, produce an acceptable weld. The following lists the maximum allowable electrode diameters for prequalified FCAW WPS’s according to D1.1: • • • • •

Horizontal, complete or partial penetration welds: 1/8 inch (0.125")* Vertical, complete or partial penetration welds: 5/64 inch (0.078") Horizontal, fillet welds: 1/8 inch (0.125") Vertical, fillet welds: 5/64 inch (0.078") Overhead, reinforcing fillet welds: 5/64 inch (0.078") * This value is not part of D1.1-94, but will be part of D1.1-96.

For a given electrode diameter, there is an optimum range of weld bead sizes that may be deposited. Weld bead sizes that are outside the acceptable size range (either too large or too small) may result in unacceptable weld quality. The D1.1 code controls both maximum electrode diameters and maximum bead sizes (width and thickness). Prequalified WPS’s are required to meet these code requirements. Further restrictions on suitable electrode diameters are not recommended. Low-hydrogen electrodes. Low hydrogen electrodes should be used to minimize the risk of hydrogen assisted cracking (HAC) when conditions of high restraint and the potential for high hardness microstructures exist. Hydrogen assisted cracking can occur in the heat affected zone or weld metal whenever sufficient concentrations of diffusible hydrogen and sufficient stresses are present along with a hard microstructure at a temperature between 100 C and –100 C. Hydrogen is soluble in steel at high temperatures and is introduced into the weld pool from a variety of sources including but not limited to: moisture from coating or core ingredients, drawing lubricants, hydrogenous compounds on the base material, and moisture from the atmosphere. At the present time, the term “low hydrogen” is not well defined by AWS. The degree of hydrogen control required to reduce the risk of hydrogen assisted cracking will depend on the material being welded, level of restraint, preheat/interpass temperature, and heat input level. When a controlled level of diffusible hydrogen is required, electrodes can be purchased with a supplemental designator that indicates a diffusible hydrogen concentration below 16, 8, or 4 ml

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H2/100g in the weld metal can be maintained (H16, H8, and H4 respectively) under most welding conditions . The diffusible hydrogen potential (measured in ml/100g deposited weld metal) will depend on the type of consumable, welding process, plate/joint cleanliness, and atmospheric conditions in the area of welding. Some consumables may absorb moisture after exposure to the atmosphere. Depending on the type of consumable, this may result in a significant increase in the weld metal diffusible hydrogen concentration. In situations where control of diffusible hydrogen concentrations is important, the manufacturer should be consulted for advice on proper storage and handling conditions required to limit moisture absorption. Hydrogen assisted cracking may be avoided through the selection and maintenance of an adequate preheat /interpass temperature and/or minimum heat input. Depending on the type of steel and restraint level, a trade-off between an economic preheat/interpass temperature and the diffusible hydrogen potential of a given process exists. There have been several empirical approaches developed to determine safe preheat levels for a given application that include consideration of carbon equivalent, restraint level, electrode type, and preheat. When followed, the guidelines for preheat that have been established in AWS D1.1 and D1.5 are generally sufficient to reduce the risk of hydrogen assisted cracking in most mild steel weldments. Hydrogen assisted cracking will typically occur up to 72 hours after completion of welding. For the strength of materials currently used in moment frame construction, inspection of completed welds should be conducted no sooner than 24 hours following weld completion. 8.2.4 Preheat and Interpass Temperatures

There are no modifications to the Guidelines or Commentary of Section 8.2.4 at this time. 8.2.5 Postheat

There are no modifications to the Guidelines or Commentary of Section 8.2.5 at this time. 8.2.6 Controlled Cooling

There are no modifications to the Guidelines or Commentary of Section 8.2.6 at this time. 8.2.7 Metallurgical Stress Risers

There are no modifications to the Guidelines or Commentary of Section 8.2.7 at this time.

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8.2.8 Welding Preparation & Fit-up

There are no modifications to the Guidelines or Commentary of Section 8.2.8 at this time.

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