Franki Guide

THIRD EDITION

A Guide to Practical

GEOTECHNICAL ENGINEERING in Southern Africa First Edition January 1976 written and compiled by IH Braatvedt Pr Eng, BSc (Eng), MICE, FSAICE Second Edition December 1986 revised and updated by JP Everett Pr Eng, BSc (Eng), FSAICE Third Edition July 1995 re-written and updated by G Byrne Pr Eng, BSc (Eng), MSAICE JP Everett Pr Eng, BSc (Eng), FSAICE K Schwartz Pr Eng, BSc (Eng), GDE, FSAICE assisted by EA Friedlaender Pr Eng, BSc (Eng), MSAICE N Mackintosh NH Dip (Civ Eng), MSAICE C Wetter BSc (Eng)

THE PURPOSE OF THIS BOOK

When Frankipile South Africa first published "The Guide" in 1976 the main purpose was to create a practical reference on all aspects of soil investigation and piling as carried out by the company in Southern Africa at that time. Judging from the popularity of the first edition this objective was achieved and most design engineers in Southern Africa have a copy on their bookshelves. The second edition was published in 1986 as an update of the first and it was equally popular. This, the third edition, is in fact a re-write of the book as Frankipile has expanded its activities into soil improvement and lateral support as well as environmental engineering. The name of the Guide has thus changed to include all aspects of Geotechnical Engineering as carried out by the Company in Southern Africa. The purpose of the book, however, remains the same; it is a reference with a wealth of practical information on geotechnical topics which we are confident all those who receive a copy will find extremely useful. The contents of this book are presented in good faith. As in all geotechnical design the methods and data presented in the book must be interpreted and used with a degree of knowledge, experience and judgement. Frankipile South Africa (Pty) Ltd does not hold itself in any way responsible for any inaccuracies or errors in the book or for any interpretation thereof by persons other than its own employees.

The company acknowledges, with appreciation, the contribution by Messrs. OVE ARUP & PARTNERS to the section on pilecap design.

FOREWORD by PROFESSOR KEN KNIGHT

Over the twenty years since Ian Braatvedt wrote the original Guide to Piling and Foundation Systems the book has become a standard text for all those in the industry in Southern Africa. It is also widely used by many outside the industry and don't be surprised i/you come across a copy in any country of the world. What has made the Guide such a valuable text is the wealth of practical information it contains on piling as well as a number of other geotechnical topics. Franki's product diversification has been dramatic since the publishing of the second edition in 1986. Through its subsidiary GeoFranki the company has entered the lateral support market and has considerably increased its market share in soil improvement. There have been other product improvements which have been developed and the company is also involved in environmental engineering. These developments are all part of Franki's ongoing drive for improvement which is backed up by some of the most experienced geotechnical engineers in the country. With all the changes it is not surprising that John Everett and his editorial team decided that the third edition of the Guide had to involve a change in name which in turn signifies that the book now covers a much wider cross section of geotechnical engineering. As such this edition will no doubt prove an even more valuable reference. PROFESSOR K. KNIGHT PrEng

Durban, July 1995.

CONTENTS Page 1.0

FRANKIPILE SOUTH AFRICA (PTY) LIMITED

1

2.0

GEOTECHNICAL INVESTIGATION 2.2 FIELD INVESTIGATION TECHNIQUES 2.3 GEOTECHNICAL ENGINEERING LABORATORY SERVICES

4 9 27

3.0

SOIL AND ROCK CLASSIFICATION AND DESIGN PARAMETERS 3.1 NOTES ON SOIL PROFILING 3.2 NOTES ON ROCK MASS DESCRIPTION 3.3 INTERPRETATION OF GEOTECHNICAL INVESTIGATION AND LABORATORY / IN-SITU TESTING DATA

30 30 37

4.0

FACTORS INFLUENCING THE SELECTION OF A PILE TYPE

61

5.0

CLASSIFICATION OF PILING SYSTEMS

63

6.0

SUMMARY DETAILS OF PILING SYSTEMS

64

7.0

TECHNICAL DETAILS OF PILING SYSTEMS 7.1 FRANKI DRIVEN CAST-IN-SITU PILES 7.2 DRIVEN TUBE PILES 7.3 PRECAST PILES 7.4 STEEL H-PILE 7.5 TIMBER PILES 7.6 AUGER PILES 7.7 UNDERSLURRY PILES 7.8 CONTINUOUS FLIGHT AUGER (CFA) PILE 7.9 FORUM BORED PILE 7.10 OSCILLATOR PILE 7.11 CAISSON PILES

66 66 76 82 89 93 95 103 112 117 122 127

8.0

UNDERPINNING 8.1 OLD FOUNDATION REMOVED AND NEW FOUNDATION PROVIDED 8.2 NEW FOOTING LOCATED UNDER THE EXISTING ONE 8.3 JACK PILES UNDER THE EXISTING FOUNDATION 8.4 PILES ALONGSIDE THE EXISTING FOUNDATION 8.5 NEW PILED FOUNDATION AND COLUMN 8.6 PILES THROUGH EXISTING FOUNDATION

131 133 134 136 138 139 139

9.0

PILE LOAD AND INTEGRITY TESTING 9.1 PILE LOAD TESTING 9.2 INTEGRITY TESTING OF PILES

142 142 147

10.0

FACTORS INFLUENCING THE SELECTION OF A SOIL IMPROVEMENT SYSTEM 151

11.0

CLASSIFICATION OF SOIL IMPROVEMENT SYSTEMS

43

153

12.0

SUMMARY DETAILS OF SOIL IMPROVEMENT SYSTEMS

154

13.0

TECHNICAL DETAILS OF SOIL IMPROVEMENT SYSTEMS 13.1 VIBRATORY COMPACTION 13.2 DYNAMIC COMPACTION 13.3 COMPACTION GROUTING 13.4 VIBRATORY REPLACEMENT 13.5 DYNAMIC REPLACEMENT 13.6 DRIVEN STONE COLUMNS 13.7 ACCELERATED CONSOLIDATION 13.8 JET GROUTING 13.9 LIME COLUMNS

156 156 161 166 168 171 173 175 178 180

14.0

FACTORS INFLUENCING THE SELECTION OF A LATERAL SUPPORT SYSTEM

182

15.0

CLASSIFICATION OF LATERAL SUPPORT SYSTEMS

184

16.0

SUMMARY DETAILS OF LATERAL SUPPORT SYSTEMS

188

17.0

TECHNICAL DETAILS OF LATERAL SUPPORT SYSTEMS 17.1 STEEL SHEET PILES 17.2 STEEL SOLDIERS 17.3 CONCRETE SOLDIER PILES 17.4 CONTIGUOUS AND SECANT PILE WALLS 17.5 DIAPHRAGM WALLS 17.6 PROP SUPPORTS 17.7 POST STRESSED ANCHORS 17.8 ANCHOR PILES 17.9 GEONAILS 17.10 RETICULATED MICROPILES 17.10 SOIL DOWELLING

190 190 197 200 203 206 210 212 218 220 225 227

18.0

PROBLEM SOILS AND THEIR FOUNDATION SOLUTIONS 18.1 EXPANSIVE SOILS 18.2 COLLAPSIBLE SOILS 18.3 SOFT CLAYS 18.4 DOLOMITES

228 229 234 236 238

19.0

ENVIRONMENTAL ENGINEERING 19.1 GROUNDWATER MONITORING 19.2 MONITORING OF SURFACE WATER 19.3 CONTAINMENT / REMEDIATION

241 241 242 242

20.0

DESIGN AIDS: PILING 20.1 PILE CAPACITY TO RESIST COMPRESSIVE LOAD 20.2 PILE CAPACITY TO RESIST UPLIFT LOAD 20.3 PILE CAPACITY TO RESIST LATERAL LOAD 20.4 THE DESIGN OF PILES FOR HEAVING SUBSOIL CONDITIONS 20.5 FACTORS OF SAFETY 20.6 ANALYSIS AND DESIGN OF PILE GROUPS 20.7 SETTLEMENT OF A SINGLE PILE AND PILE GROUPS

245 245 260 262 265 268 270 273

20.8 20.9

STRUCTURAL DESIGN OF PILE SHAFTS STRUCTURAL DESIGN OF PILE CAPS

280 284

21.0

DESIGN AIDS: SOIL IMPROVEMENT 21.1 SOIL COMPACTION 21.2 SOIL REPLACEMENT 21.3 ACCELERATED CONSOLIDATION

301 301 306 310

22.0

DESIGN AIDS: LATERAL SUPPORT 22.1 DESIGN PARAMETERS 22.2 EARTH PRESSURES 22.3 WATER PRESSURES AND SURCHARGE LOADS 22.4 EXTERNALLY STABILISED SYSTEMS 22.5 INTERNALLY STABILISED SYSTEMS 22.6 FACTORS OF SAFETY 22.7 MOVEMENTS ASSOCIATED WITH EXCAVATION

312 312 314 316 319 326 330 332

23.0

REFERENCE INFORMATION 23.1 NORMAL PLANT CLEARANCE REQUIREMENTS 23.2 PILING RIG DIMENSIONS 23.3 BENDING MOMENTS IN BEAMS 23.4 MENSURATION OF PLANE SURFACES 23.5 MENSURATION OF SOLIDS 23.6 PROPERTIES OF SECTIONS

334 334 336 340 341 342 343

24.0

QUALITY ASSURANCE

344

REFERENCES

347

INDEX

352

1.0

FRANKIPILE SOUTH AFRICA (PTY) LIMITED

The South African company in the worldwide Franki group was started by Mr. Wally Rowland in 1946. The initial seeds had already been sown early in 1939 but the second World War broke out in September and Wally joined up with the South African forces. At the end of the war The Franki Piling Company of South Africa, as it was initially named, was registered and the first contract was secured. This involved the installation of eight piles for a building in Paarden Eiland and a steam driven piling machine was used to install the piles which were standard Franki driven cast-in-situ piles. By 1952 Franki had branch offices in Johannesburg, Cape Town and Durban. In 1955 Wally Rowland returned to the UK to take up the position of Assistant Works Director with the British Franki company. Ian Braatvedt took over as Managing Director in 1961 and under his guidance the company grew steadily. Large contracts such as the Alusaf Bayside Smelter in Richards Bay, the Mondi Paper Mill in Durban and Iscor Steel Works in Newcastle were secured in the late sixties and early seventies and these really helped Franki to establish itself as the leading piling company in the country . In 1968 Franki started a soil investigation subsidiary which is known as Soiltech. Today it has a full complement of soil investigation and field testing equipment as well as a fully equipped soils laboratory .It has recently entered the environmental investigation field. The need to diversify into other geotechnical fields led to the formation of GeoFranki in 1987. GeoFranki's main areas of activity are lateral support, ground improvement, micropiling, grouting and cut-off walls. The South African company today has over forty major production rigs and an employee complement in excess of 600. It has offices in Johannesburg, Cape Town, Durban and Harare. It operates in Africa and the Indian Ocean Islands. FRANKI INTERNATIONAL Frankipile South Africa is a wholly owned subsidiary of SA Franki BV which is a Belgian company. SA Franki BV has a number of subsidiaries around the world and interests in many other international companies and this group is commonly referred to as Franki International. There is continual commercial and technical communication within the group as well as a common product development interest. Frankipile South Africa can thus draw on this international experience as well as obtain additional plant resources and personnel from the group if and when such a demand arises.

1

The following is a summary of the products and services which Frankipile South Africa and its subsidiaries can presently offer its clients and which are described in greater detail in this guide. FIELD INVESTIGATION

PILING

• • • • • • • • • • • • • • • • • • • •

• • • • • • • • • • •

Auger trial holes Test pits Bulk and undisturbed sampling Dynamic cone penetration test Cone penetration test Rotary core drilling Standard penetration test Vane shear test Pressuremeter test Lugeon test Piezometer installation Shelby and piston tube sampling Core orientation Rotary percussion drilling Plate load test In-situ density test Geophysical techniques Ground water monitoring Well installation Environmental investigation

LABORATORY TESTING • • • • • • • • • • • • •

Triaxial compression Unconfined compression Shear box Permeability Odeometer Grading/Sieve analysis Hydrometer Atterberg limits Bulk density Moisture density relationship California bearing ratio (CBR) PH Conductivity

Franki cast-in-situ pile Driven tube pile Precast pile Steel H-pile Timber pile Auger pile Underslurry pile Continuous Flight Auger (CFA) pile Forum Bored pile Oscillator pile Caisson pile

UNDERPINNING SOIL IMPROVEMENT • • • • • • • • •

Vibratory compaction Dynamic compaction Compaction grouting Vibratory replacement Dynamic replacement Driven stone columns Accelerated consolidation Jet grouting Lime columns

LATERAL SUPPORT • • • • • • • • •

2

Steel sheet piles Steel soldiers Concrete soldier piles Contiguous/Secant pile walls Diaphragm walls Geonails Reticulated micropiles Soil dowelling Tie back anchors

Frankipile South Africa has always adopted a policy of combining innovative design with many years of practical experience to provide the most economical solution to a geotechnical problem. The company thus maintains a strong design capability as well as its professionally run contracting activities. Whilst not directly marketing this design capability, Franki makes every use of it in negotiations with clients and in tenders where alternate designs are permitted. The fact that Franki can offer the wide selection of products and services indicated above also results in the economic optimisation of design and so it is not surprising that the company secures a large percentage of its work through negotiation and through innovative design. With this considerable expertise Franki can offer a complete package deal including investigating the site, the complete design of the foundation system and any lateral support requirements, pricing and drawing up the contract, execution of the work and final handover. It also has strategic partners it can draw on to form joint ventures where it considers the combined skills and resources of the partners will provide the client with a more comprehensive service and at a more competitive price. This is very much the case, for example, on large marine construction projects where Franki has the piling and other geotechnical skills which are often a major feature of marine work. A joint venture partner with general marine civil experience thus forms a strong combination with which to tackle the design and construction of any marine contract. This type of arrangement, however, is not limited to marine construction but can be arranged for any civil or building project with a significant geotechnical content.

3

2.0

GEOTECHNICAL INVESTIGATION

Soiltech, the division of Frankipile responsible for Geotechnical Investigations was established in 1968 and offers a complete Geotechnical Service to consulting engineers and client bodies as well as to the Company. The importance of obtaining adequate and reliable knowledge of sub-surface conditions at a sufficiently early stage cannot be over emphasised when considering: • • • •

The choice and design of an economical and technically sound foundation. Possible delays and additional expense due to inadequate soils information. Expensive foundation failures or overdesign. Potential contractor's claims based on inaccurate and/or inadequate soils information.

Soiltech is able to offer a complete geotechnical investigation service comprising: • • •

Planning of the investigation. Execution of the field work and laboratory testing. Interpretation and reporting.

The range of field work and laboratory testing that Soiltech can offer is outlined below in Table 2.0.l. Table 2.0.1- Range of Soiltech Services

FIELD INVESTIGATION AND IN-SITU TESTING • Auger trial holes. • Test pits. • Bulk and undisturbed soil sampling. • Dynamic cone penetration tests. • Cone penetration tests. • Rotary core drilling • Standard Penetration Tests. • Vane shear tests • Pressuremeter tests. • Lugeon tests • Piezometer installations • Shelby and piston tube sampling • Core orientation • Rotary percussion drilling • Vertical and horizontal plate load tests • In-situ density tests • Geophysical techniques • Monitoring well installations • Ground water monitoring and sampling

LABORATORY TESTING • • • • • • • • • • • • • •

4

Triaxial compression tests Unconfined compression tests Shear box Permeability Oedometer test Grading/sieve analysis Hydrometer Atterberg limits Moisture density relationship California bearing ratio (CBR) Specific gravity Bulk density pH Conductivity

A further important aspect of Soiltech's activities lies in the environmental engineering field. This service provides for the collection of data with respect to potentially contaminated soils, surface and ground waters. For further details refer to SECTION 19.0 ENVIRONMENTAL ENGINEERING. 2.1

GUIDE FOR PLANNING A GEOTECHNICAL INVESTIGATION

The objectives of a geotechnical investigation may embrace any combination of the following. (British Standard Code of Practice B35930: 1981): • • • • •

To assess the general suitability of the site for the proposed engineering works. To enable an adequate and economical design to be preparedTo foresee and provide against difficulties that may arise during construction owing to ground and other local conditions. To determine the causes of defects or failure in existing works and the remedial measures required. To advise on the availability and suitability of local materials for construction purposes.

Taking the above objectives into consideration, the planning of a geotechnical investigation will be influenced by the following main factors: • • • • • • •

The nature of the proposed engineering development. "If you do not know what you are looking for in a site investigation you are not likely to find much of value" (Glossop, 1968). The Geology and Geomorphology of the site. Access to and the remoteness of the site. The site topography, vegetation and drainage. The nature of adjacent developments. Knowledge of previous geotechnical investigations or foundation installations carried out in the area. In particular the opinions of persons such as local engineers, farmers and contractors. Evidence of problem soil conditions (expansive or collapsible soils, dolomites, dispersive soils, soft clays).

The cost of an adequate investigation is very low in comparison to the total cost of the project. The consequences of not providing sufficient, accurate and reliable geotechnical information, however, can have a significant effect on a project and can lead to delays and extras during construction with associated costly claims. Experienced engineers have come to realise that a thorough geotechnical investigation is invariably paid for by the client, whether it is carried out or not. If appropriate, the planning of a geotechnical investigation should be carried out in a phased approach. Phase one is an initial investigation to determine the site geology and to

5

define the problem. This is followed by phase two which is a far more extensive investigation in which the site geology is studied in greater detail and all the design parameters are determined. The phased approach will generally commence with a desk study and site reconnaissance, followed by the fieldwork and laboratory testing. Conditions vary from site to site. Consequently, a variety of techniques has been developed to enable both the geotechnical engineer and specialist contractor to select the appropriate investigation procedures. An accurate description of the soil profile forms the basis of the geotechnical investigation. In some cases this maybe all that is required. In the majority of investigations, however, it will be necessary to supplement an accurate description of the soil profile with appropriate in-situ testing and sampling, and possibly associated laboratory testing. Under appropriate conditions, particularly where the water table is at depth which is applicable to large areas within the hinterland of Africa, the drilling of large diameter trial holes and/or the forming of test pits for visual inspection by a geotechnical engineer or engineering geologist, can be carried out. The advantages of this procedure are as follows: • • • •

It allows for the soil profile to be examined in-situ in its natural stateGood quality undisturbed block samples can be cut from the auger hole or test pit sidewalls. Disturbed samples can also be taken from specific horizons identified during profiling. In-situ testing such as hand shear vane tests and horizontal plate bearing tests can be executed within the trial holes or test pits. The procedures adopted are fast and economical and provide for accurate and comprehensive evaluation of site geotechnical conditions.

Safety procedures when profiling and sampling in trial holes and test pits are extremely important. All investigation work with trial holes and test pits must be carried out in accordance with the SAICE Code of Practice for the safety of persons working in small diameter shafts and test pits for civil engineering purposes (1990). For certain projects it may be necessary to supplement the auger trial holes/test pits with additional investigation procedures. A variety of techniques are available. These could include dynamic cone penetration tests, rotary core drilling with associated sampling and insitu testing techniques (standard penetration tests, vane shear tests, lugeon tests etc.). In the coastal regions and on sites with a high water table the use of trial holes and test pits is often not feasible due to collapse of the sidewalls. In these areas the two standard methods used in a geotechnical investigation are boreholes with standard penetration tests and cone penetration tests. These are supplemented where necessary with, amongst others, rotary core drilling, vane shear tests, dynamic cone penetrometer tests and the recovery of undisturbed samples using the Shelby tube method or piston sampling. Tables 2.1.1 and 2.1.2 are provided as a guide to assist in the planning of a geotechnical investigation. These tables give typical details with regard to the field and laboratory tests

6

that could be carried out in stable soil profiles above the water table (Table 2.1.1) and in saturated variable soils (Table 2.1.2). Table 2.1.1 -Guide to planning a soils investigation in stable soil profiles above the water table (usually residual soils or cohesive transported soils) Parameter

Field Test

Description of the soil profile

Auger trial holes Test pits Boreholes with SPT Dynamic cone penetrometer (DPSH) In-situ profiling of trial holes/test pits Recover undisturbed samples from auger trial hole, test pit or borehole Vane shear test in borehole Recover undisturbed samples from auger trial hole, test pit or borehole

Consistency of the soil profile Undrained shear strength

Effective angle of friction –ǿ Effective cohesion-ć Modulus of compressibility Index property tests Permeability Collapse Heave

Level of Water Table

Laboratory Test

Cross-hole jacking test or Plate load test or Pressuremeter test Recover disturbed samples from auger trial hole, test pit or borehole Recover undisturbed samples from auger trial hole test -pit or borehole Recover undisturbed samples from auger trial hole, test -pit or borehole Recover undisturbed and/or disturbed samples from auger trial hole, test pit or borehole Drill a trial hole or a borehole, leave for a period of time for the water level to stabilise in the hole and then measure the level

7

Undrained triaxial Unconfined compression test Drained triaxial Drained shear box test Undrained triaxial with measurement of pore water pressure Oedometer test Grading analysis Atterberg limits Moisture content Falling or constant head permeability Double oedometer Collapse potential test Double oedometer Swell under load test Index property test(disturbed sample)

Table 2.1.2- Guide to planning a soils investigation in saturated, variable soils usually encountered in coastal areas or adjacent to water courses Parameter

Field Test

Description of the soil profile Consistency of the soil profile

Boreholes with SPT Dynamic cone penetrometer (DPSH) Cone penetrometer test(CPT) Boreholes with SPT Recover undisturbed samples from borehole Vane shear test in borehole Correlate with in-situ penetrometer tests Recover undisturbed samples from borehole Correlate with in-situ penetrometer tests (sandy soils only) Pressuremeter test Correlate with in-situ penetrometer tests Recover disturbed samples from borehole Recover undisturbed samples from borehole Recover undisturbed samples from borehole Recover undisturbed and/or disturbed samples from borehole

Undrained shear strength

Effective angle of friction –ǿ Effective cohesion-ć Modulus of compressibility Index property tests Permeability Collapse Heave

Level of Water Table

Laboratory Test

Drill a borehole and leave for a period of time and measure

8

Undrained triaxial Unconfined compression test

Drained triaxial Drained shear box test Undrained triaxial with measurement of pore water pressure Oedometer test Grading analysis Atterberg limits Moisture content Falling or constant head permeability Double oedometer Collapse potential test Double oedometer Swell under load test Index property test (disturbed sample)

2.2

FIELD INVESTIGATION TECHNIQUES

The field investigation techniques that Soiltech offers and which are summarised in Table 2.0.1 are discussed in more detail below. 2.2.1

AUGER TRIAL HOLES

The auger trial hole involves the drilling of a large diameter auger hole using a powerful auger machine. A qualified person is then lowered in stages down the hole by means of a small winch and is able to profile the hole by visually inspecting the sidewalls and the base. Furthermore, it is possible to cut large undisturbed samples from the sidewalls or base of the hole for later testing in the laboratory, as well as carry out cross-hole jacking tests and plate load tests as described in SECTION 2.2.7 in the trial hole excavation. Bulk sampling for the purposes of evaluating the mineral content of materials on old dumps or within soil and weathered profiles, can also readily be accomplished using this technique. Auger trial holes provide a very quick and economical method for obtaining reliable geotechnical information for a variety of engineering solutions and it is favoured by most engineers and geologists. For the successful application of this technique, however, it is essential that the sIde walls of the trial holes remain stable during drilling and profiling. It is thus not suited to areas with a high water table where collapse of the sidewalls is most likel'l. It is possible to drill a large number of trial holes in a relatively short space of time which makes this an economical form of investigation. A minimum hole diameter of 750 mm is required for in-situ profiling purposes but trial holes of up to 2000 mm in diameter are possible. Depths of up to 36 metres can be drilled in suitable materials. The technique is ideally suited to sites with deeply weathered profiles. The auger trial hole can also penetrate into soft rock and even harder fractured rock. Under suitable site conditions approximately eight 750 mm diameter auger holes to a depth of 10 metres can be drilled within a normal working day. It is also possible to profile this number of holes within the same working day. The auger rig with its crew and ancillary equipment is normally hired on a daily basis. Soiltech can arrange for the profiling of the holes by experienced qualified personnel from an independent geotechnical engineering firm should this be required. To facilitate the profiling of the trial holes, a tripod frame fitted with a winch is positioned over the trial hole, the winch being connected via a steel wire rope to a specially designed bosun's chair. All operations are carried out in accordance with the S.A.I.C.E. Code of Practice for the safety of persons working in small diameter shafts and test pits for civil engineering purposes (1990). Plastic sample bags, cling wrap, labels, tape measures and sampling tools form part of the standard equipment available on site. Under special site circumstances breathing apparatus and methanometers are made available on site.

9

AUGER RIGS FOR DRILLING TRIAL HOLES Soiltech has a variety of truck mounted auger rigs available for drilling trial holes. Details of these rigs are given in Table 2.2.1.1. Plate 2.2.1.1. shows the Williams LDH50 rig used for drilling auger trial holes. The overall dimensions of the auger rigs are given in SECTION 23.2 PILING RIG DIMENSIONS Table 2.2.1.1- Range of Auger Rigs Available Type of Auger Rig

Max. Drilling Torque (Kgm) Hotline 16MI20 3000 Soilmech RTAH 11000 Williams Digger LDH50 6818 Williams Digger LDH80 6818 WilliamsDiggerLLDHI20 13636 2.2.2

Gross Vehicle Mass (Kg) 24400 30000 28580 30000 39310

Max Depth (m) 16 32 15 24 36

Max. Hole Diameter (mm) 1000 1500 1000 2000 2000

TEST PITS

The use of test pits as an investigation technique offers the same advantages in terms of profiling and sampling as described for auger trial holes. Test pits are easily formed with a mechanical excavator or by hand, and therefore have the advantage of being relatively inexpensive. The main disadvantages are that they are limited to depths of two to three metres and cannot be used in areas of shallow water table. Test pits are therefore most appropriate in areas with a relatively deep water table where competent soils or rock are anticipated at a relatively shallow depth. They are often used to investigate areas where there is poor access for other types of equipment. In view of cost advantages, test pits are often used as a preliminary or first phase of the investigation. Where a competent soil or rock stratum occurs close to the ground surface, the profiling and sampling of test pits may provide sufficient information for design purposes and no other form of testing is required. If, on the other hand, the excavation of test pits discloses a much deeper soil profile, then it is essential to follow up the first phase with additional investigation work. This is normally carried out using techniques which can reach to greater depths, such as auger trial holes and boreholes. It is extremely important to follow the correct safety procedures when profiling and sampling in test pits, and the SAICE Code of Practice for the safety of persons working in small diameter shafts and test pits for civil engineering purposes (1990) should be strictly adhered to at all times. Experience has shown that test pits are far more prone to collapse than auger trial holes, due to the fact that a rectangular pit is less stable than a circular trial hole. Even highly experienced engineers and geologists find it difficult to assess the stability of a test pit and serious accidents have been reported.

10

Plate 2.2.1.1. A Hotline rig for drilling auger trial holes

11

2.2.3

DYNAMIC CONE PENETRATION TESTING

Dynamic Probe Light (DPL) This local standard of the Dynamic Probe Light test (ISSMFE Technical Committee on Penetration Testing, 1988) is used in many applications in South Africa. A 20 mm diameter, 60° cone is driven into the soil by an 8 kg weight dropped through 575 mm. The results are expressed as millimetres per blow. The original test (Van Vuuren, 1969) was designed for the rapid determination of the California Bearing Ratio (CBR) to depths of about one metre for investigation into road pavement performance and design. Besides the original application in the field of pavement evaluation and design, the test has also been used as a rough guide in compaction control and for estimating soil conditions for the design of shallow footings. The main advantage of this type of equipment is that it is light, portable, inexpensive to operate and provides a continuous rough record of soil consistency over the depth tested. The disadvantages are that no sample is recovered, the nature of the equipment limits its depth capability to three metres below surface and the equipment is not able to penetrate hard lenses or other obstructions (large gravel, boulders etc.). The ease and low cost with which results can be obtained, is therefore somewhat offset by the limitations of the test and the indirect approximation to soil conditions that it provides. A guide to the interpretation of the results of this test can be found in SECTION 3.0 SOIL AND ROCK CLASSIFICATION AND DESIGN PARAMETERS. Dynamic Probe Super Heavy (DPSH) In Southern Africa, considerable use is made of a local standard of the Dynamic Probe Super Heavy test (ISSMFE Technical Committee on Penetration Testing, 1988). A 60° disposable cone, 50 mm in diameter, is fitted onto the bottom of an "E" size rod and driven into the ground by a 63.5 kg hammer falling through 762 mm. The number of blows required to drive the cone through each successive 300 mm of penetration is recorded. This provides an empirical indication of consistency. Once refusal depth is reached (more than 100 blows per 300 mm), the driving rods are withdrawn by 600 mm. The disposable cone remains at the base of the hole. The rods are then re-driven with the number of blows per 300 mm being recorded. These re-drive blow counts provide an indication of the skin friction acting on the drive rods. Data collected from the DPSH test (including the re-drive figures) are presented on a report sheet. A feature of the test is that it is very economical and can be rapidly and easily performed. A major disadvantage of the test is that no soil sample is obtained. In many instances this disadvantage can be overcome by adopting a variation to the test procedure by fitting a Raymond split-spoon sampler to the "E" rods, instead of the solid cone. This technique provides a continuous disturbed representative sample of the soil profile. Any blow counts recorded during this operation cannot, however, be correlated with those of the actual DPSH test.

12

The DPSH rig is designed so that tests can be undertaken in areas that are not readily accessible, such as inside existing buildings and in narrow passage ways between buildings. Plate 2.2.3.1 shows a typical DPSH test rig. The DPSH is used under the following conditions: • • • • •

As economical supplementary data between boreholes on larger sites. On sites with erratic profiles (alluvial, colluvial or lacustrine deposits), it will locate softer areas. Probing for rock or hard strataIn conjunction with a soil profile it will provide rough consistency readings which can be plotted graphically. As the test closely approximates a driven pile, it is extensively employed for determining an estimate of skin friction and installation depths of driven cast-in-situ piles. In noncohesive materials it is very reliable, but must be used with caution in cohesive soils. The test will also indicate pile driving conditions.

Limitations of the test are: • • •

Driving refusal is frequently experienced on hard layers (such as very dense ferricretes or calcretes, boulder horizons) which may be underlain by soft soil horizons. Differences in remoulding caused by the small diameter cone on the one hand and the considerably larger piling tube on the other,can lead to erroneous prediction of pile installation depth. Similar differences may occur when excessive pore pressures are set up during the driving of a pile whereas this does not occur with the DPSH test.

A graphical presentation of this data is presented in Figure 2.2.3.1. The interpretation of the test results is generally associated with local experience. As a preliminary evaluation the blow counts can be taken as being roughly equivalent to the SPT N value (See SECTION 2.2.5). In the interpretation, however, it is essential to take into account the influence of the rod friction. 2.2.4

CONE PENETRATION TESTING (CPT)

This method was initially developed in the Netherlands in the 1930's where it was first used as a means of determining the ultimate bearing capacity of driven piles founded in sand. Over the years the test has been called the Dutch sounding test, the Dutch probe and the static cone penetration test. In terms of acceptable international standards (ISSMFE Technical Committee on Penetration Testing, 1988) it is now referred to as the cone penetration test (CPT). In the CPT test a 60° cone with a cross sectional area of 1000 mm2, usually equipped with a friction sleeve which is of the same diameter of the cone and has an surface area of 1.5 x 104 mm2, is pushed into the ground at a rate of 20 mm/sec. Separate measurements of cone

13

penetration resistance (point resistance), total penetration resistance and the side friction resistance of the friction sleeve are made continuously throughout the test. The main advantages of the CPT are that the testing procedure is relatively simple and repeatable, and the test results are more amenable to a rational analysis rather than relying entirely on empirical correlation. The CPT also gives a virtually continuous record of soil resistance values throughout the depth of penetration.

Figure 2.2.3.1 -Typical results from a DPSH test

14

The main limitations of the CPT test are as follows: • • •

Penetration depth limitations due to machine capacity. The technique is rarely effective in gravels and boulder horizons and is also not suited to weathered rock profiles. No samples are recovered.

The data obtained from the cone penetration test may be employed to: • • • •

Assist in the evaluation of the type and stratigraphy of the soils present. Interpolate ground conditions between control boreholes. Evaluate engineering parameters of soils (relative density, shear strength, compressibility characteristics, liquefaction potential). Assess driveability, bearing capacity and settlement of piled foundations.

Mechanical cone penetrometers (Begeman, 1965) have a telescopic action which requires an outer probe sleeve and an inner rod. These mechanical cone penetrometers offer the advantage of low equipment cost and simplicity of operation. They do, however, have the disadvantages of a slow incremental procedure, limited accuracy in very soft soils and labour intensive data handling and presentation. With the electrical cone penetrometer the friction sleeve and cone point advance together as a single system. The point resistance and local side shear are recorded continuously with the use of built-in load-cells. An electrical cable located inside the rods connects the load cells to recording equipment at ground surface. Electrical cones carry a high initial equipment cost and require skilled operators as well as adequate back-up for calibration and maintenance. They do, however, offer advantages over the mechanical penetrometer such as a more rapid procedure, higher accuracy and repeatability , automatic data logging, reduction and plotting. One of the important applications of the CPT test is to evaluate variations of soil type within the soil profile. With mechanical and electrical cones extensive use is made of what is known as the friction ratio as a means of soil classification (Jones, 1974, Schmertmann, 1975). The friction ratio is the ratio between sleeve friction and the point resistance and is expressed as a percentage. The most significant recent development in electric cone penetration testing is the development of the piezo-cone penetrometer test (CPTU) which incorporates a pore pressure sensor in the cone. This allows for the measurement of the pore water pressure present in the soil during penetration. Pore pressure measurements during cone penetration testing provides more details on the stratification and has added a new dimension to the interpretation of certain geotechnical parameters especially in loose or soft fine grained soil deposits. This has resulted in CPTU testing becoming a prime tool for stratification logging of soil deposits (Jones and Rust, 1982, Campanella and Robertson, 1988).

15

Further advantages of the CPTU test over the conventional CPT are as follows (Campanella and Robertson, 1988): • •

The ability to distinguish between drained, partially drained and undrained penetration. The ability to evaluate flow and consolidation characteristics. The ability to assess equilibrium groundwater conditions.

A guide to the interpretation of the results of CPT and CPTU tests can be found in SECTION 3.0 SOIL AND ROCK CLASSIFICATION AND DESIGN PARAMETERS. 2.2.5

ROTARY DRILLING, IN-SITU TESTING AND SAMPLING

The rotary drilling technique is used to drill a borehole which is normally cased through the upper soil profile. Various methods for testing and sampling the soil during the drilling of the borehole are available and described later in this section. The most common of these is the standard penetration test or SPT. Once the borehole reaches strata of rock consistency, rotary core drilling is used to recover samples. ROTARY DRILLING The borehole is typically drilled through the upper soil layers using a casing fitted with a diamond/tungsten tipped casing shoe. A drilling fluid is used to remove the cuttings and flush them to the surface where they can be sampled. This technique for advancing the borehole is called wash boring and the samples are known as wash samples. The borehole is advanced in stages with samples taken at the various depths required. Plates 2.2.5.1 and 2.2.5.2 show two types of rotary drilling rigs. When materials of rock consistency are encountered and wash boring is no longer effective, rotary core drilling is used to advance the borehole and recover core samples. The cores are drilled using a core barrel which is fitted with a diamond tipped or impregnated drill crown. The core barrel with drill crown is rotated by the drilling rig which also has the means to hydraulically crowd the drill stem. A drilling fluid is pumped through the core barrel to cool the drill bit and flush the cuttings to the surface. The conventional core barrel can recover a 1.5 metre length of core at a time. Once the core barrel is full, the drill stem with core barrel is withdrawn from the hole and the core sample is recovered and stored in a core box. Core boxes are marked with the depths drilled so that a visual inspection of the core box shows what percentage of core was recovered relative to the depth drilled. Cores are sometimes waxed to retain their natural moisture content. UCS and point load tests are often carried out on rock cores so as to determine the strength of the rock. This is an important factor when carrying out a geotechnical investigation for a contract on which piles will be required to penetrate the rock, as the piling contractors need to know the hardness of the rock to be able to assess penetration rates at the time of tender.

16

A variety of core barrels and appropriate crowns is available, allowing the driller to select the most suitable type for the particular materials being cored. Core barrel designs, such as double or triple tube, help to maximise the core recovery especially in the very soft and weathered rock strata. Heinz ( 1989) gives a detailed description of rotary core drilling techniques and equipment. Soiltech complies with the Standard Specifications for Subsurface Investigations (CSRA, 1993) in carrying out rotary drilling operations. IN-SITU TESTING Standard Penetration Test (SPT) This process was standardised in the 1920's and 1930's into what we know now as the Standard Penetration Test. In the execution of this test a standard 51 mm diameter split spoon sampler known as a Raymond Spoon is driven into the soil at the bottom of a borehole. A free-fall hammer of 63.5 kg operating off a trip mechanism and falling through a height of 762 mm provides the driving force. The number of blows required to drive the sampler each 150 mm increment of a total of 450 mm penetration is recorded. The blow count for the first 150 mm increment is discarded and the sum of the blow counts for the second and third 150 mm increments is known as the SPT "N" value. The standard penetration test has become accepted world wide as a useful test in geotechnical investigation and foundation design. SPT results in boreholes give an empirical qualitative guide to the in-situ engineering properties of cohesive and cohesionless soils and provide a sample of the soil for classification purposes. The results of the SPT can be affected by incorrect drilling and sampling procedures some of which are given below (refer also to the Canadian Foundation Engineering Manual, 1985): • • • • • • •

Inadequate cleaning of the bottom of the borehole. Driving the spoon above the bottom of the casing. Failure to maintain sufficient hydrostatic head in the borehole. Not using the standard hammer drop or correct mass. Free fall of the hammer is not obtained. The tip of the spoon is damaged. Not recording blow counts and penetration accurately.

It is thus extremely important that the drilling crew carrying out the tests is experienced in this type of work. Even then it is advisable to carry out some CPT tests close to the borehole positions to check the correlation between the two. This will give an indication as to whether the SPT values are reliable. The relationship between the SPT N value and engineering properties is empirical and some guidelines regarding the evaluation and interpretation of SPT N values are given in SECTION 3.0 SOIL AND ROCK CLASSIFICATION AND DESIGN PARAMETERS.

17

Plate 2.2.3.1 -DPSH Test Rig

Plate 2.2.5.1- Skid Mounted Rotary Core Drilling Rig

PLATE 2.2.5.2 -MOBILE B80 ROTARY CORE DRILLING RIG

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Vane Shear Tests The vane shear test is routinely used to obtain undisturbed peak and remoulded undrained shear strength. The test consists of placing a four bladed vane in the undisturbed soil and rotating it from the surface to determine the torsional force required to cause a cylindrical surface to be sheared by the vane. This force is then converted to a unit shearing resistance of the cylindrical surface as shown in Figure 2.2.5.1. A typical example of the equipment employed to apply torque to the steel rods from surface is also shown in Figure 2.2.5.1. The steel rods are housed in a sleeve in order to prevent flexing and to protect the rods. The vane which is connected to the base of the steel rods is housed within a "torpedo" attached to the base of the sleeve. The vane consists of a four bladed cruciform. For standard tests the height of the vane should be twice the diameter. The selection of the vane size is directly related to the consistency of the soil being tested, with larger vane sizes being used in the softer soils. The test procedure is to advance the vane from the bottom of the torpedo in a single thrust to the depth at which the test is to be conducted. Once the vane is in position, torque is applied in a rotational sense at a slow rate using the gear driven surface equipment. Torsional force is measured and converted to unit shearing resistance in accordance with the following assumptions: • • • • • • •

Penetration of the vane causes negligible disturbance, both in terms of changes in effective stress and shear distortion. No drainage occurs before or during shear: The soil is isotropic and homogeneous. The soil fails on a cylindrical shear surfaceThe diameter of the shear surface is equal to the width of blades. At peak and remoulded strength, there is a uniform shear stress distribution across the shear surface. There is no progressive failure, so that at maximum torque shear stress at all points on the shear surface is equal to the undrained shear strength.

The results of a vane shear test may be influenced by many factors: • • • • • •

Type of soil, especially when permeable fabric exists. Strength and anisotropy. Disturbance due to insertion of the vane. Rate of rotation or strain rate. Time lapse between insertion of the vane and the beginning of the test. Progressive/instantaneous failure of the soil around the vane.

It should be appreciated that the assumptions described above are not likely to apply at the same time. The test is therefore limited to a restricted range of material types.

19

Figure 2.2.5.1 -Vane Shear Apparatus

20

Pressuremeter Tests The pressuremeter test was originally developed by Menard (1956) and comprises a horizontal in-situ loading test carried in a borehole by means of a cylindrical expandable probe. A major difference between categories of pressuremeter tests lies in the method of installation of the device in the ground. In accordance with Mair and Wood (1987), the following two broad categories of tests can be distinguished in terms of installation method: • •

Menard type pressuremeter (MPM) test in which the device is installed in a borehole. Self-boring pressuremeter (SBP) test in which the device bores its own way into the ground usually from the bottom of a borehole.

The following parameters can be deduced from the results of the pressuremeter test. • • • •

Deformation modulus (i.e. compressibility). Undrained shear strength for clays or weak rocks. Effective angle of friction for sandsIn-situ total horizontal stress.

The degree of success in obtaining any of these parameters is essentially dependent upon the type of test and the interpretation of the data. Consideration must also be given to possible differences in the properties of soil horizons measured in a horizontal direction by the pressuremeter, and those required for many design problems which are more concerned with vertical properties. For more details with regard to pressuremeter testing and its interpretation reference should be made to Baquelin et al (1978), Windle and Wroth (1977) and Mair and Wood (1987). Lugeon Testing Lugeon testing (also known as water pressure or packer testing) is carried out to measure the permeability of the soil or rock at specific depths in a borehole. The equipment consists of two packers comprising steel tubes surrounded by inflatable rubber sleeves separated by a perforated length of steel tube. The spacing of the packers can be adjusted to the specific length of soil or rock to be tested. The minimum length of packer sleeve is 700 mm to ensure a watertight seal. The packer arrangement is connected via high pressure tubing to a suitable pump on the surface. Data collected from the system is obtained by flow metres and pressure gauges. The above arrangement is known as a double-packer system. The system can be adapted, however, for so-called single packer tests, where testing is carried out between the packer and the bottom of the hole.

21

The test consists of pumping water into the isolated zone of the borehole at three different pressures, in the following typical sequence: 1 st 2 nd 3 rd 4 th 5 th

10 min. at low pressure 10 min. at medium pressure 10 min. at high pressure 10 min. at medium pressure 10 min. at low pressure

a b c b -repeated a- repeated

The actual duration of each pressure stage is accurately timed. The pressures selected are dependant on the depth at which each test is carried out. The required pressures are maintained to an accuracy of 5% during each pressure stage. Piezometer Installations Piezometers are installed in boreholes in order to provide information regarding the at rest levels of the ground water table. In addition, ground water pressure can be measured via more specialised piezometers i.e. hydraulic, electrical and pneumatic. In general piezometers are installed into pre-cleaned holes by lowering a selected porous tip to approximately 500 mm above the bottom. The tip is surrounded by a filter of graded, washed silica sand and sealed off with a bentonite plug. The remainder of the borehole is sealed by introducing cement/bentonite grout. SAMPLING Shelby and Piston Tube Sampling This sampling technique is employed to obtain undisturbed material from soft and very soft cohesive soils. The Shelby tube used to recover the samples, consists of a thin walled stainless steel tube with an internal diameter of approximately 75 mm. The leading edge of the tube is beveled and crimped such that the entry diameter is fractional smaller than the body diameter. The tube is usually a half metre in length with the top end designed to fit into an adapter. The adapter has a one-way valve built into it to allow water to escape so as to prevent compression of the sample. The Shelby tube sampler is attached to the drill string in place of a core barrel and is lowered to the base of the borehole and pressed into the soft material using the drill rig hydraulics. The sample and tube are then raised and the sample extruded on site. The sample should be sealed and packed so as to maintain the in-situ moisture content and to resist damage during normal handling and transport. Under certain conditions, where the material to be sampled cannot be successfully obtained via the conventional Shelby tube technique, piston sampling may be employed. In these instances either a floating or rod-mounted piston is located in such a manner that the piston

22

rests on the top of the sample as it is pushed into the tube. The piston creates a vacuum which allows for retention of the sample within the tube. Core Orientation Such surveys are carried out where information is required regarding the spatial orientation of planar features, palaeontological studies, etc. The techniques employed include the following: •

•

Impression Core Orientation: this technique employs a hollow tube fixed to the base of the drill string filled with a suitable Plasticine material. The tube is lowered to the base of the pre-washed borehole and the orientator is pushed to seat onto the proud core break. The tube is withdrawn and the impression in the Plasticine matched with the bottom of the previous core run. Correct orientation is maintained during the raising and lowering of the drill string. .Integral Core Orientation: this technique involves the drilling of a pilot hole (E size or similar) to 1.5 metres below the base of the main borehole using centering bushes to centre the pilot hole in the main borehole. An orientated bar or pipe is placed into the pilot hole and cemented into position. The orientated bar is overdrilled once the cement has set. The technique can be employed in vertical or inclined holes, and is specifically used where highly fractured formations have been intersected or the impression technique cannot be employed.

2.2.6

ROTARY PERCUSSION DRILLING

There are two types of rotary drilling equipment. The one is known as a top drive rig and this consists of a drive head which remains above the surface and is connected via drill rods to a drill bit. The drive head rotates the drill string as well as imparts an impact force into the rods. The drill bit impacting on the rock chips the rock and the chips are air flushed to the surface. The other type of drilling equipment is very similar to the above but the impact force is generated by a down-the-hole hammer. This is a percussion hammer which is driven by air and which imparts a rapid series of impacts to the drill bit which is part of the hammer. The rotation drive to the drill stem is provided by a top drive head. The down-the-hole hammer is favoured for geotechnical investigation purposes because of greater versatility and sensitivity particularly when recording penetration times. The standard procedure in terms of geotechnical investigation is for percussion chips, which are flushed to the surface by compressed air, to be collected at one metre intervals. During drilling operations the operator is required to keep a record of penetration time per metre, air loss, levels of water strikes, intersection of cavities and anything else that may be of specific interest to the logger of the borehole.

23

A borehole log is compiled from the inspection of the chip samples, an evaluation of penetration times and the other relevant information supplied by the driller. The nature of the technique is such that the compilation of the borehole log can be influenced by a number of factors that can lead to inaccurate interpretation of the soil/rock conditions. Some of the more important of these factors are as follows: • • •

The highly disturbed nature of the chip samples recovered and the possibility of contamination of these samples. Total loss of sample in loose or soft layers. Incorrect interpretation of the penetration rate in relation to the hardness of the material being penetrated.

From the discussion presented above it is apparent that, in terms of geotechnical investigation, rotary percussion drilling can only be used to obtain a rough indication of the soil/rock profile and is subject to a large number of inaccuracies which include to a large extent the experience of the driller and the logger. The advantages of the rotary percussion technique are that it is relatively inexpensive when compared with rotary coring, being about one tenth of the cost of rotary coring. Drilling production is also fast when compared to rotary coring with production rates of 80 to l00 metres per day possible. It is also one of the few techniques that can be used to economically penetrate boulder horizons or layers of chert which are often encountered in dolomitic terrain. In South Africa the technique has been used successfully used as part of the overall geotechnical investigation procedures used in dolomitic terrain (Wagener, 1984). The technique is also used for the following applications: • • • •

As probe holes to determine rock head depths. As probe holes to determine the depth and extent of old mine workings. To form boreholes for the conducting of in-situ tests (pressuremeter, lugeon tests). To form boreholes for the installation of geotechnical instrumentation (piezometers, extensometers, inclinometers, etc.).

2.2.7

PLATE LOAD TESTS

Plate load tests are usually carried out to determine the compressibility and occasionally the bearing capacity of soils and rocks. The test is a convenient and direct method of obtaining these parameters and is often used in soils or rocks which cannot be sampled or where the structure (joints etc.) may control the engineering behaviour of the soil/rock mass. In its simplest form, the plate load test comprises a rigid plate placed on the surface to be tested. The load is provided by an hydraulic jack, using kentledge or an anchored beam as reaction. Figure 2.2.7.1. shows a typical test system. The plates used must be rigid and typically vary in diameter from 200mm to l000mm.

24

The following procedures are adopted for the test: • • •

The test site is carefully levelled and the plate bedded into the layer being tested using Plaster of Paris and/or bedding sandLoad is applied to the plate using a hydraulic jack in a series of predetermined steps. This application of load and the maximum load applied must be designed to conform with the type and purpose of the testing being carried out. Plate settlement is usually measured by means of dial gauges. In order to measure any tilt of the plate it is advisable to have four measuring points. The dial gauges are usually fixed to a beam supported by posts bearing on the soil some distance from the loaded area to avoid the readings being influenced by the settlement of the plate.

A variation to the standard test procedures can be implemented to allow the soil below the plate to be saturated at a specific load. The objective of this procedure is to allow the determination of any collapse properties associated with the material being tested. The widespread use of auger trial holes and test pits in Southern Africa has led to the development of light and portable horizontal plate load equipment suitable for use in trial holes and test pits. By carrying out the tests in a horizontal direction, the necessary reaction is provided by the opposing faces of the trial hole or test pit. The bearing plates on either side are of equal size and the test procedure is essentially the same as that used for vertical plate load tests. The distance between the plates is measured and the movement of each plate is taken as half the total on the assumption that the two plates have moved equally.

Figure 2.2.7.1 -Example of Vertical Plate Load Arrangement

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2.2.8

IN-SITU DENSITY TESTS

In-situ density tests are mainly used for compaction control in roads and earthworks construction. In certain instances the determination of in-situ density may, however, form part of an overall geotechnical investigation fieldwork programme. Both the sand replacement method and nuclear methods are used for the determination of in-situ density. In the sand replacement method, the in-place dry density is determined by forming a hole in a layer and dividing the mass of the material removed from the hole by the volume of the hole, the latter being determined by filling the hole with a fine sand of known density .The disadvantage of this test is that the material removed from the hole needs to be dried to a constant mass, usually overnight in a suitable oven. This means that a period of at least 12 to 18 hours is required before results become available. The advantage of the test is that it gives an accurate value of in-situ dry density and in-situ moisture content. Nuclear systems for the determination of wet density and moisture content have become popular in recent years. One of the main advantages of this test procedure is that results are immediately available. The disadvantage is that there are some potential inaccuracies associated with the results produced from this test. The inaccuracies are generally associated with the measurement of moisture content and can easily be overcome by taking a sample at each test position for the laboratory determination of moisture content. To a large extent this negates the advantages of having results available immediately. On most roads and earthworks contracts the results of nuclear gauge tests are generally only accepted as a control procedure after suitable calibration with sand replacement tests has been carried out. Soiltech is able to offer both sand replacement and nuclear gauge density tests. These tests are carried out in accordance with the procedures recommended in TMH 1 ( 1986). 2.2.9

GEOPHYSICAL TECHNIQUES

Geophysical exploration is a form of field investigation in which a set of physical measurements relating to the underlying soil or rock strata is made at ground surface or in boreholes. The measurements indicate variations in space or time of certain physical properties of the soil/rock materials. Geophysics is therefore a blend of physics and geology since the physical measurements are interpreted in terms of subsurface geological conditions. The properties of soils/rock which are of significance in geophysical exploration are density, magnetic susceptibility, electrical conductivity , elasticity and thermal conductivity. Since these physical properties vary widely in soils/rocks at least one of these properties usually shows marked changes from place to place which can be measured by sufficiently sensitive instrumentation.

26

The main advantages of geophysical techniques are as follows: • •

It is possible to carry out investigations of large areas rapidly and economically. The techniques can be used to locate critical areas for further field investigation.

The disadvantage of the technique is that the results are dependent on the interpretation of physical measurements. These measurements are not in themselves geological or geotechnical parameters relative to the site subsurface conditions. It is therefore essential that geophysics is carried out and interpreted in conjunction with a carefully planned drilling programme. The main application of geophysics in geotechnical investigations is the interpolation of subsurface geological strata between carefully controlled drilling positions. The more common geophysical techniques used in geotechnical investigations are magnetics, gravity and resistivity. For more detailed information reference should be made to Darracott (1976), Bullock (1978), Griffiths and King (1965), Kleywegt and Enslin (1973) and West and Dumbleton (1975). 2.3

GEOTECHNICAL ENGINEERING LABORATORY SERVICES

Standardised and consistent soil mechanics and materials testing, often forms the basis for design and site quality control in geotechnical and materials engineering. Soiltech has a fully equipped soil mechanics and materials laboratory facility which provides a testing services to clients, consulting engineers and the Frankipile Group. A guide to testing procedures and requirements for the commonly specified soil mechanics and materials tests is presented in Table 2.3.1. All relevant road type materials testing is carried out in accordance with TMH 1 (1986). Soil mechanics testing is carried out in accordance with accepted published or International standards. In certain instances non-standard testing may be required. Through in-house expertise Soiltech can assist clients to define the testing programme and ensure that the testing is carried out to specified requirements.

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Table 2.3.1 -Guide to Laboratory Procedures and Requirements Laboratory Test Triaxial compression test Unconsolidated Undrained (UU) Consolidated Undrained with pore water pressure. measurements (CU) Consolidated drained test (CD) Shear Box Test Drained Shear Box

Consolidation Tests Consolidation test soaked at11 kPa loaded to 1600 kPa and rebounded Double oedometer test for collapse Collapse potential test. Sample loaded to 200 kPa, saturated and rebounded Double oedometer test for heave Swell under load test

Parameter Determined

Duration of Test in Days

Sample Requirements

Undrained shear strength of cohesive soils (Cu) Effective shear strength parameters c' or «I>'

3

Effective shear strength parameters c' or «I>'

7 to 10

Undisturbed: Good quality sealed block sample 300 mrn x 200 mrn x150 mm thick. Shelby tube or piston sample Disturbed or remoulded: 2 kg of representative sample

Effective shear strength parameters c' or «I>' Residual shear strength parameter «I>'

4

Compressibility characteristics

7

Compressibility and collapse characteristics over full loading spectrum Compressibility and collapse characteristics Collapse potential index Swell characteristics over full loading spectrum Swell characteristics at specified load

7

28

5 to 7

5 to 7

3 7 3

Undisturbed Good quality sealed block: sample 300 mm x 200 mm x150 mm thick. Shelby tube or piston sample Disturbed or remoulded: 2 kg of representative sample

Undisturbed Good quality sealed block sample 300 mm x 200 mm x: 150 mm thick. Shelby tube or piston sample Disturbed or remoulded: 2 kg of representative sample

Table 2.3.1 (Cont.) -Guide to Laboratory Procedures and Requirements Laboratory Test PERMEABILITY TESTS Falling head or constant head

BULK DENSITY Index properties Grading/sieve analysis Hydrometer Atterberg limits Moisture content Moisture density relationship Mod AASHTO Proctor

CALIFORNIA BEARING RATIO (CBR)

Parameter Determined

Duration of Test in Days

Coefficient of permeability

3 for sandy soils 7 to 10 for clayey soils

Sample Requirements

Undisturbed: Good quality sealed block sample 300 mm x 200 mm x150 mm thick. Shelby tube or piston sample. Disturbed or remoulded: 2 kg of representative sample Good quality sealed block sample 300 mm x 200 mm x150 mm thick

Bulk density Dry density Moisture content

3

Particle size distribution to0.075 mm Particle size distribution from0.075 mm to 0.002 mm Liquid limit, plastic limit, plasticity index Moisture content

3

2 kg sample of undisturbed or disturbed soil

Max. dry density and optimum moisture content under specified compactive effort Mod AASHTO moisture density curve. Plot of CBR vs dry density based on CBR at 3 compactive efforts (Mod AASHTO, Proctor, NRB)

2

40 kg of representative sample

6

70 kg of representative sample

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3.0 SOIL AND ROCK CLASSIFICATION AND DESIGN PARAMETERS 3.1

NOTES ON SOIL PROFILING

As indicated in SECTION 2.0 GEOTECHNICAL INVESTIGATION, an accurate description of the soil profile forms the basis of the geotechnical investigation for any engineering development. It is important that each layer is described in a consistent way to ensure accurate interpretation of the soil profile by those involved in the geotechnical design and construction process. The description of the soil in profile, based on the work of Jennings, Brink and Williams (1973), is related to the following: Designation M C C S S O

Heading Moisture Colour Consistency Structure Soil Tvpe Origin

Example Moist Reddish Brown Stiff Intact Clay Residual shale

Moisture The moisture content is assessed as: DRY, SLIGHTLY MOIST, MOIST, VERY MOIST and WET. The assessment at the moisture content is dependant on the soil type. With a moisture content of say 20%, sand will probably be described as wet, whilst clay will probably be described as slightly moist. Colour Colour is important for description and for correlation. Colour is described from the soil in profile and also from a small sample of soil made into a creamy paste with water. A profile is MOTTLED when small exposures of different colours occur. A profile is BLOTCHED when larger exposures (say 75 mm and larger) of different colour occur. Colour charts obtainable from the South African Institution of Civil Engineers illustrate the main colours as well as variations in hue and lightness of each colour. These charts illustrate the following colours.

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Blue: Green: Olive: Brown:

Dusky Pale Dusky. Pale Dark Light Dark Light Dark reddish Light reddish

Red: Grey: Orange:

Yellow:

Dusky, Dark Pale, Light. Dark Light Dark reddish Light reddish Dark yellowish Light yellowish Dark Light

Consistency Consistency is a measure of the strength or density of the soil. Observations are based on the effort required to dig into the soil or to mould it with the fingers. The consistency of cohesive soils is based on the undrained shear strength and described as VERY SOFT, SOFT, FIRM, STIFF AND VERY STIFF. Consistency vs. Undrained shear strength guidelines are set out in SECTION 3.3. Non-cohesive soil consistency is based on the angle of shearing resistance of the soil and described as VERY LOOSE, LOOSE, MEDIUM DENSE, DENSE AND VERY DENSE. Consistency vs angle of shearing resistance guidelines are given in SECTION 3.3. Structure The presence and type of discontinuities in the soil mass define the structure. Structural characteristics are generally related to cohesive soils in the following terms: INTACT FISSURED SLICKENSIDED SHATTERED MICROSHATTERED LAMINATED FOLIATED STRATIFIED

Absence of fissures and joints, though tension cracks may occur in firm samples when broken with a pick. Presence of closed joints. Highly polished fissures, usually indicative of expansive soils. Indicates fissures which have opened up and allowed entry of air, often associated with expansive soils. Shattering on a small scale with shattered fragments the size of sand grains. If well developed, the soil appears granular when cut, but the grains break down into clay and/or silt when wetted and rubbed. Indicates the presence of a highly expansive soil. Indicates that the soils show the laminated, foliated or stratified structure of the parent rock or geological process from which they were derived.

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Soil Type The soil type is described on the basis of the grain size of the individual particles. The basic grain size classes are given below. Most natural soils occur as a combination of these classes e.g. Silty clay or gravelly sand. BOULDERS GRAVEL

SAND

SILTS

CLAY

COBBLES COARSE MEDIUM FINE

COARSE MEDIUM FINE

Fragments of rock > 200 mm 60mm -200mm 20mm -60mm 6mm -20mm 2.0mm -6mm The range of size of boulders and gravel, the shape, the proportion by volume of the matrix and the description of the matrix are important. 0.6mm -2.0mm 0.2mm -0.6mm 0.06mm -O.2mm Sand particles are visible to the naked eye. 0.002mm -0.060mm Silts are barely gritty between fingers and thumb when wet, but are gritty on tongue against teeth. Silts are not easily rolled into threads when moist. Silts exhibit dilatancy when moulded with water into a pat, (i.e. it increases its volume when shearing occurs which is illustrated by the film of water on the surface being absorbed if the pat is distorted.) Silts dry moderately quickly and can be dusted off the fingers. Dry lumps possess cohesion but powder easily in the fingers. Particles less than 0.002 mm Clay particles are flaky (not powdery) when broken and will soften with the addition of water. They have a soapy or greasy feel when wetted and rubbed on the palm of the hand. Clay sticks to fingers and dries slowly. There is no dilatancy or grittiness on tongue against teeth.

32

Origin In any soil profile there are four basic categories of origin: • • • •

Rock Residual soil Pedogenic material Transported soil

In the South African context, the demarcation between residual soils and overlying transported soils is often defined by the "pebble marker". This horizon is generally characterised by a gravel layer overlying the residual soil. •

Rock

Materials described as rock comprise igneous, metamorphic or sedimentary (not pedogenic) horizons with unconfined compressive strengths of the intact or unjointed material in excess of 1000 KPa. •

Residual Soil

A residual soil is formed from in-situ decomposition of rock. Decomposition can be caused by chemical weathering or mechanical disintegration which is a function of potential evaporation (temperature, humidity, wind) and average annual precipitation. •

Pedogenic Material

Pedogenic material is residual or transported soil that has become strongly cemented or partially replaced by one of the cementing agencies. Description

Cementing Agency

Ferricrete

Iron oxide

Calcrete

Calcium carbonate

Silcrete

Silica

33

•

Transported Soil

This is soil which has been transported by a natural agency (water, wind, gravity) during relatively recent geological times (Pleistocene or Tertiary) and which has not undergone lithification into a sedimentary rock or cementation into a pedogenic material. Type Talus (scree and coarse colluvium) Hillwash (fine colluvium)

Agency Gravity

Source Rock outcrops

Run-off

Alluvium wash)

Rivers, streams and gullies

Acid crystalline Basic crystalline Arinaceous sediment Argillaceous sediment Various rocks and soils

Lacustrine Deposits

Streams terminating in lake, pan or pool

Various rocks and soils

Estuarine Deposits

Tidal rivers waters

Mixed

Littoral Deposits Aeolian Deposits

Waves Wind

(gully

and

Mixed Mixed

Resulting Soil Unsorted angular gravel and boulders Clayey sand Clay Sand Clay or silt Boulders Gravels Sands Silts Clays Sand Silt Clay Sand Silt Clay Beach sand Sand and sand

clayey

Subsurface Water Condition The water table is that level or those levels in the soil where the water in the pores of the soil occurs at atmospheric pressure, i.e. the level to which the water finds its own way in a borehole. The perched water table is a table which is only present in the soil temporarily. It will disappear and sometimes re-appear depending upon seasons or drainage conditions of the site. The permanent water table is the water table which persists throughout the seasons of the year with only minor seasonal fluctuations of level.

34

A typical soil profile and a tabulation of the various soil symbols are given in Figures 3.1.1 and 3.1.2 respectively.

Figure 3.1.1- Example of typical soil profile

35

Figure 3.1.2- Typical Soil Symbols

36

3.2

NOTES ON ROCK MASS DESCRIPTION

The accurate description of rock engineering conditions, like the requirements outlined for the soil profile, necessitates a detailed and practical method of describing samples of rock core retrieved from a rotary cored borehole, With these requirements in mind, the publication A Guide to Core Logging for Rock Engineering by the Core Logging Committee of the South African Section of The Association of Engineering Geologist (1976) has become the accepted norm for the description and interpretation of geological and rock engineering conditions in South Africa. This method of logging rock cores was based on similar principles to those for soil profiling outlined by Jennings et al (1973), but due to the complexity of rock mass behaviour subject as it is to weathering and discontinuities, the soil profiling system was modified and adapted to provide an adequate rock mass description. The core log contains descriptions of the rock mass parameters as well as discontinuity surfaces and the materials infilling these surfaces, if any. 3.2.1

DESCRIPTION OF PRIMARY ROCK MASS PARAMETERS

Six basic rock mass parameters are used in the same way as soil descriptive parameters. These are tabulated and compared to the soil descriptors below. Soil Description Moisture Colour Consistency Structure N/A Soil Type Origin

Rock Mass Description N/ A Colour Weathering Fabric and discontinuity surface spacing Hardness Rock Type Stratigraphic horizon

Colour Colour is the basic and most easily identifiable characteristic and colour variation is a primary indication of weathering. The colour of a rock mass is generally related to its mineralogy. Quartz and feldspar will give rise to light coloured rock while pyroxine and olivine give rise to dark coloured rock. Cores should be washed before logging and the colour recorded on wet, recently broken surfaces. The standard Munsell colour chart obtainable from S.A.I.C.E should be used to describe the hue and the lightness of the colour. Where variable colour exists, the dominant colour of the rock mass should be given, followed by the secondary colour which usually exhibits a pattern such as: BANDED, STREAKED, BLOTCHED, MOTTLED, SPECKLED AND STAINED. Where inclusions, such as amygdales occur they should be described together with their colour .

37

Weathering Weathering of a rock mass is a process of alteration by mechanical, chemical or biological action which significantly affects the behaviour of the rock material and the rock mass as a whole. The decomposition and the disintegration of a rock mass is given the term "weathering" and the degree of weathering is given in Table 3.2.1. Table 3.2.1 -Weathering of a rock mass Diagnostic Feature Descriptive Term Unweathered Slightly Weathered Medium Weathered

Highly Weathered Completely Weathered

Discoloration Extent None

Fracture Condition

< 20% of fracture spacing on both sides of fracture > 20% of fracture spacing on both sides of fracture

Closed or Discoloured Discoloured May contain thin filling Discoloured May contain thick filling

Throughout

-

Throughout

Original Texture

Grain Boundary Condition

Unchanged

Preserved

Tight

Partial discoloration

Preserved

Tight

Partial to complete discoloration not friable except poorly cemented rocks Friable and possibly pitted Resembles a soil

Preserved

Partial Opening

Mainly preserved Partly Preserved

Partial separation Complete separation

Surface Characteristics

For detailed definition of the five degrees of weathering reference should be made to the original publication. Fabric Fabric describes the structural and textural features of the rock material. Texture of the rock mass is governed by the size and arrangement of the individual grains. Table 3.2.2 gives recommendations on grain size terminology. Table 3.2.2 -Fabric of a rock mass Description Very fine grained

Size in mm 2.0

Recognition Individual grains cannot be seen with a hand lens Just visible as individual grains under hand lens Grains clearly visible under hand lens just visible to the naked eve Grains clearly visible to naked eye Grains measurable

38

Equivalent Soil Type Clays & Silts Fine sand Medium sand Coarse sand Gravel

Micro structural features such as foliation and banding give rise to anisotropic behaviour of the rock material for small scale features and the rock mass gives rise to this behaviour for larger scale features. A spacing of 10mm is given as the boundary between micro structure and discontinuity surface. Table 3.2.3 outlines the terms given to both Micro structure and discontinuity surface. Discontinuity Surface Spacing This describes mechanical discontinuity, weakness planes or bedding planes. Two major categories of discontinuity surface are given by features characteristic of the origin such as bedding and features as a result of movement within the rock mass such as joints. In the core log only the discontinuity surface spacing is given in the primary rock mass description. Any additional features are outlined separately. It is important to note that discontinuities caused by the drilling operation and the handling of the cores are not included in the description of the discontinuities. Table 3.2.3(a) and (b) gives descriptions of the discontinuity surfaces. Table 3.2.3(a) -Macro Features Descriptions for Structural Features: Bedding, Foliation or Flow Banding Very thickly (bedded foliated or banded) Thickly Medium Thinly Very thinly

Spacing in mm < 1 000 300- 1 000 100- 300 30- 100 10- 30

Description for Joints, Faults, or other Fractures Very widely (fractured or jointed Widely Medium . Closely Very closely

Table 3.2.3(b) -Micro Features Description for Micro-structural features: Lamination, Foliations or Cleavage Intensely laminated (foliated or cleaved) Very intensely

Spacing in mm 3- 10 80 40 20 0

SOIL STRENGTH CLASSIFICATION

Strength parameters can be assigned to soils or rocks based on: • • • •

Descriptions of the strength in the soil profile. Index property tests. Empirical relationships with penetration test results. On direct measurement in-situ or in the laboratory .

47

For the purposes of shear strength evaluation soils have been divided into two broad categories : • •

Cohesionless soils (sands, silty sands, slightly clayey sands) for which the shear strength is assumed to be represented by drained conditions in terms of the angle of shearing resistance (φ'). Cohesive soils (clays, silty clays, sandy clays etc.) for which the strength can be defined in terms of undrained (φ' = 0 and Cu equal to a finite value) and drained shear strength (φ' equal to a finite value and C' = 0 or a finite value).

In-situ Description of Soil Consistency The in-situ description of soil consistency can be used to obtain a rough estimate of φ' of a cohesionless soil and of the undrained cohesion (Cu) of cohesive soil. The undrained cohesion (Cu) is equal to half the unconfined compressive strength (UCS). It is recommended that consistency be described in accordance with Jennings et al (1973). Guidelines in this regard are given in Tables 3.3.4 and 3.3.5 for cohesive and non-cohesive soil respectively.

S.l

S.2

S.3

S.4

S.5

Table 3.3.4 -Shear strength parameters for slow draining cohesive materials Consistency Rule of thumb Field identification Unconfined Approximate Compressive SPT Strength (N) (kN/m²) V. Soft Easily moulded by fingers. Distinct < 40 12.5 > 50 40 to 50 > 19.5 75mm with 12mm bar driven with 2kg hammer. Power tools required for excavation Index Property Tests Many researches have published typical ranges and correlations between the results of index property tests (Atterberg Limits and Particle size distribution) and the effective shear strength of soils (mainly cohesive soils with plasticity index greater than 7). The following procedures are recommended: •

•

The effective angle of friction can be obtained using the relationship given by Kenney (1959). This relationship is given in Figure 3.3.4. Although this relationship is for normally consolidated soils the effective angle of friction should not be much different for over consolidated soils. If the nature of the soil is such that it is considered appropriate to use an effective cohesion greater than zero for analysis purposes then the cohesion values given in Table 3.3.6 should be used as a guide. The values given in Table 3.3.6 are for compacted soils and should be taken as the upper bound values for natural soils. It is also necessary to point out that for most soils caution must be exercised if an effective cohesion value greater than zero is to be used for design purposes. The values given in Table 3.3.6 for friction angle can also be used as a check on the values given in Figure 3.3.4.

49

Figure 3.3.4- Plasticity index vs sin φ' after Kenney (1959) Table 3.3.6- after NAVFAC DM7 (1971) Group Soil Type Symbol

Max γd

Optimum Moisture (%)

GW

19.7-21.2

11-8

Typical strength characteristics Cu C' φ' tan φ' (kPa) (kPa) (deg.) 0 0 >38 >0.78

18.1- 20.5

14- 9

0

0

>31

>0.6

17.3-19.7

16-11

50

5

34

0.67

16.5-19.7

19-11

75

10

31

0.6

15.0- 18.9

24- 12

85

12

28

0.54

15.0- 18.9

24- 12

65

10

32

0.62

11.8 -16.5

36 -19

100

12

19

0.35

GC SM SC CL ML CH

Well-graded clean gravels, gravel-sand mixtures Clayey gravels, poorly graded gravel-sand-clay Silty sands, poorly graded sand-silt mix Clayey sands poorly graded sand-clays Inorganic clays of low to medium plasticity Inorganic silts and clayey silts Inorganic clays of high plasticity

50

Shear Strength Parameters from In-situ and Penetration Tests Empirical relationships between soil shear strength (φ' for non-cohesive and Cu for cohesive) and penetration or in-situ test values have been put forward by many authors in different parts of the world. For non-cohesive soils two approaches have been adopted where relationships have been developed (i) dependent on vertical effective stress and (ii) independent of vertical effective stress. Figures 3.3.5 (a) and (b) give values of φ' independent of Po' while Figures 3.3.6 (a) and (b) show values of φ' dependant on Po' for both SPT and CPT tests.

(a)

(b)

Figure 3.3.5- φ' Independent of Vertical Stress after (a) Peck et al (1974) and (b) Kahl et al (1968)

(a)

(b)

Figure 3.3.6- φ' Dependent of Vertical Stress after (a) Mitchell et al (1978) and (b) Esopt(1974)

51

The undrained cohesion Cu for sensitive and normally consolidated clays has been studied by several authors and correlations with SPT results show a wide scatter as outlined by De Mello (1973) and Navfac DM7 (1971) and shown in Figures 3.3.7 (a) and (b) below. If accurate values are required for design, direct correlation should be obtained during the geotechnical investigation or measured on undisturbed samples in the laboratory.

(a)

(b)

Figure 3.3.7- Cu vs N correlation for soft sensitive clays after (a) De Mello (1973) and (b) Navfac DM7 (1971) Stroud (1974) gives correlations of SPT vs undrained shear strength for stiff insensitive clays which are considered to be applicable to a wide range of residual and transported clay soils in Southern Africa. These are shown in Figure 3.3.8.

Figure 3.3.8 Relationship of SPT N vs undrained shear strength after Stroud (1974)

52

Undrained shear strength correlations to CPT cone resistance values (qc) for normally and overconsolidated clays are well covered in the literature and, like the SPT test, a reasonably wide scatter is evident and is dependant on whether the clay is sensitive, normally consolidated or overconsolidated. The equation qc = Nk Cu + σvo governs the relationship of qc with undrained shear strength. For normally consolidated clay Nk ≈ 15 while for sensitive clays Nk can be as low as 5. Undrained shear strength correlations to the Dynamic Probe Light (DPL) test (outlined in SECTION 2.2.3) have been given by Brink et al (1982) and are summarised in Table 3.3.7. Table 3.3.7 -Undrained shear strength correlations with DPL test after Brink et al (1982) Sandy Materials Description SPT N (blows per 300mm) Dynamic Probe Light (DPL) (mm per blow) Very loose 75 Loose 5- 10 30- 75 Medium dense 10- 30 12.5- 30 Dense 30 –50 5 -12.5 Very dense > 50 2- 5 Clayey Materials Very soft 110 Soft 2-4 55-110 Firm 4 –8 30- 55 Stiff 8-15 15-30 Very stiff 15- 30 7- 15 3.3.3

ROCK STRENGTH CLASSIFICATION

The unconfined compressive strength (UCS) of intact rock is used as the basis for foundation design with an allowance being made for the structure of the rock mass. Table 3.3.8 and Figure 3.3.10 provide the basis for rock strength classification from profile descriptions. Point load index tests are often used to evaluate rock strength from core or block samples of intact rock. Correlation of these results with UCS values are given in Figure 3.3.9. Direct measurement of the unconfined compressive stress can also be carried out in the laboratory on core samples of intact rock.

53

Table 3.3.8 -Rock strength classification vs UCS Classification Very soft rock Soft rock Medium hard rock Hard rock Very hard rock

Field Test Can be peeled with a knife, material crumbles under firm blows with the sharp end of a geological pick Can just be scraped with a knife, indentations of 2 to 4mm with firm blows of the pick point. Cannot be scraped or peeled with a knife, hand held specimen breaks with firm blows of the pick point. Point load tests must be carried out in order to distinguish between these classifications. These results may be verified by unconfined compressive strength tests on selected samples.

Extremely hard rock

UCS (MPa) 1 to 3 3 to 10 10 to 25 25 to 70 70 to 200

> 200

Figure 3.3.9- Point Load Index correlation with UCS (after Bieniawski 1973)

54

Figure 3.3.10- Rock strength classifications vs UCS

55

3.3.4

COMPRESSIBILITY CLASSIFICATION OF SOILS

Compressibility characteristics of soils can be determined from: • • •

Direct in-situ measurement such as plate load tests. Measurements in the laboratory using for example the oedometer test. Empirical correlation with penetration tests.

Compressibility characteristics of fine grained soils are divided into immediate (elastic) settlement and long term (consolidation) settlement for drained conditions. Compressibility moduli from direct methods Soil modulus can be obtained directly from tests such as the plate load test. With such tests consideration must be given to the test procedure and its constraints. Taking these factors into consideration the compressibility modulus can then be calculated from the measured stresses and strains using Poulos and Davis (1977) for various loading geometries and soil conditions. Laboratory testing Laboratory testing to determine compressibility characteristics is well covered in the literature and reference to interpretation of tests such as the oedometer should be made. Empirical correlation with penetration tests Figures 3.3.11 (a) and (b) and Figure 3.3.12, showing Soil Modulus versus the SPT and CPT test results respectively, give empirically correlated drained modulus values for noncohesive materials. Webb (1974) has carried out extensive research on the compressibility of estuarine soils on the Natal coast and drained modulus values are given by the following equations using qc values from the CPT test as well as SPT N values : Ev' = 2.5 (qc + 3.2) MPa or Ev' = 537 (N + 15) MPa for fine to medium sands below the water table and: Ev' = 1.67 (qc + 1.6) MPa or Ev' = 358 (N + 5) MPa for clayey sands with P.I. < 15% Relationships of SPT N versus drained modulus for stiff overconsolidated clays has been presented by Stroud (1974) and are given in Figure 3.3.13. These values are particularly useful in obtaining compressibility characteristics of stiff residual soils in the Southern African region.

56

There is a wide scatter in the correlation of compressibility with penetration test values for normally consolidated and sensitive clays and values used should be regarded with caution. Drained soil modulus values for these soils show a wide scatter and the relationship of Ev' /N varies between 300 and 2000.

(a)

(b) Figure 3.3.11- Drained modulus for sands after (a) Stroud (1989) and (b) Menzenbach (1957)

57

Figure 3.3.12 -Drained modulus for non-cohesive soils based on CPT

Figure 3.3.13- Drained modulus for stiff cohesive soils after Stroud (1974)

58

3.3.5

COMPRESSIBILITY CLASSIFICATION OF ROCK

The modulus of rock material is related to the unconfined compressive stress qa with the ratio Er/qa showing a wide scatter of between 100 and 1000. A ratio of 300 should be used for design where no direct correlation for the rock type has been measured. A reduction factor is usually applied to the intact modulus to obtain the rock mass modulus since discontinuities and joint infilling can markedly affect the compressibility of the rock mass. Figure 3.3.14 shows the wide scatter of the Er/qa ratio value for various rock types and strengths.

Figure 3.3.14- Correlation of Rock Modulus with UCS After Peck (1976) and Deere (1968)

59

3.3.6

SUBGRADE MODULUS

The vertical and horizontal sub grade moduli are parameters commonly used to model the lateral restraint of piles and in pavement design. The subgrade modulus proposed by Terzaghi (1955) is defined as the deflection produced by a unit applied pressure on a 300 mm square plate and is given the unit kN/m2/m. Plate load tests are commonly carried out to determine vertical and horizontal subgrade modulus values. Table 3.3.9 gives typical values of horizontal modulus of subgrade reaction (kh) for cohesive soils. In cohesionless soils the horizontal subgrade modulus increases with depth and is given by:

kh = nh ×

Z B

where nh is known as the coefficient of modulus variation, Z is the depth in metres and B the pile breadth in metres. Values of nh are given in Table 3.3.10. Table 3.3.9 -Relationship of modulus of subgrade reaction (kh) to the undrained shear strength (Cu) of stiff overconsolidated clay Consistency Undrained Shear Strength Cu (kPa) Range of kh. (MN/m) Recommended kh (MN/m3)

Firm

Stiff

Very Stiff

50- 100 18- 36 27

100- 200 36- 72 54

> 200 > 72 100

Table 3.3.10- Factors for calculating coefficient of modulus variation (nh) for cohesionless soil Relative density 3

nh for dry or moist soil (MN/m ) (Terzaghi) nh for submerged soil (MN/m3) (Terzaghi) nh for submerged soil (MN/m3) (Reese et al)

60

Loose

Medium dense

Dense

2.5 1.4 5.3

7.5 5 16.3

20 12 34

4.0 FACTORS INFLUENCING THE SELECTION OF A PILE TYPE Before the design engineer can consider what type of pile is best suited to his project he needs to have the following basic information as a minimum. • • • •

Detailed soils information A column layout with column loads Allowable total and differential settlements Knowledge of the site and its environs

Using this information he will need to consider the following points regarding the various piling systems so that he can chose the most suitable system for the project. Structural • • • • • • • • •

Range of pile sizes to suit the column loads. Founding level to meet the pile load capacity. Founding level to meet the settlement criteria. Spacing ofpiles. The allowable rake of the piles if required. The ability to resist tension forces if required. The ability to resist horizontal forces if required. The clearance from existing buildings. The durability of the pile shafts.

Soil Profile • • • • • • • • • • •

If driven whether driving will be easy, difficult or impossible. If bored whether temporary casings will be required. If bored the difficulty in penetrating to the required depth. The presence of obstructions such as boulders. The founding level to meet pile load capacity. The founding depth to meet settlement criteria. Very soft layers which can cause problems with cast-in-situ piles. Rock sockets. Presence of ground water and at what level. The presence of aggressive ground water. The potential for pile heave during installation.

Environmental • •

The effects of noise pollution caused by piling equipment. The effects of vibration caused by pile installation.

61

Contractual • • • • • • • •

Access to and on the site for piling equipment. Headroom clearance on site for piling equipment. The cost of the piles. The cost of the pile caps and ground beams. The installation risks associated with a particular pile solution. The remoteness of the site. The availability of skills and plant to install the piling system. Adequate plant and people resources for large contracts.

Most of these points and others are covered in SECTION 6.0 SUMMARY DETAILS OF PILING SYSTEMS and in SECTION 7.0 TECHNICAL DETAILS OF PILING SYSTEMS. An initial selection can be made from SECTION 6.0 but this should be checked by reading the more detailed information given in SECTION 7.0. As one can see from the above there are a number of factors to consider and the assessment of some of these will be difficult for someone not experienced in piling. Should there be any doubt you are welcome to contact your local Franki office for advice. An incorrect choice of pile type can be an expensive and embarrassing mistake so it is advisable to make sure beforehand.

62

5.0

CLASSIFICATION OF PILING SYSTEMS

There are many various types of pile, some which are used extensively and some that are used very seldom. The suitability of the various types to the local soil conditions and the requirements of the local codes and specifications have a strong influence on what pile types are more popular in anyone country .The availability of piling plant and equipment also has a strong influence. The following is a list of the pile types that have been found to satisfy the needs of the Southern African market and which suit the local soil conditions. These have been classified into three main groups: Driven Cast-in-situ, Driven Preformed and Bored Cast- in-situ. DRIVEN CAST -IN-SITU Franki piles

(7.1)

Driven tube piles

(7.2)

DRIVEN PREFORMED Precast piles

(7.3)

Steel H-piles

(7.4)

Timber piles

(7.5)

BORED CAST -IN-SITU Auger piles

(7.6)

Underslurry piles

(7.7)

Continuous flight auger (CFA) piles

(7.8)

Forum bored piles

(7.9)

Oscillator piles

(7.10)

Caisson piles

(7.11 )

The numbers quoted on the right are the sub section numbers for a detailed description of the individual pile types and the methods of installation and which are grouped under SECTION 7.0 TECHNICAL DETAILS OF PILING SYSTEMS. With most of the main piling systems there are some variations to the installation technique which can be carried out so as to achieve a specific requirement or to overcome installation difficulties. There is a full description of these in SECTION 7.0 as well as a list of potential problems that can be experienced with each of the pile types. SECTION 6.0 SUMMARY DETAILS OF PILING SYSTEMS provides a table in which the more important details are presented in a readily referenced format. This enables easy comparison between one system and another when evaluating a suitable pile type for a specific project.

63

6.0

SUMMARY DETAILS OF PILING SYSTEMS

Ref. Pile Type

7.1

7.2

7.3 7.4 7.5

7.6 7.7 7.8 7.9

DRIVEN CAST-INSITU Franki pile Mini Light Medium Heavy Super Heavy Steel Tube pile

Nom. Typical Maximum Shaft Working Tension Diameter Load kN Load kN mm

250 355 410 520 600 150 to 600

DRIVEN PREFORMED Precast pile 250 sq300 sq Steel H-pile Different Hsections Timber pile 300 to 400 BORED CAST-INSITU Auger pile 150 – 500 600 to2000 Underslurry 900 pile to1500 CFA pile

Forum bored pile 7.10 Oscillator pile

300 to750 410 600 950 1050 1350

Max. Max Rake Depth. metres

Establishment Costs

250 500 750 1200 1600 Up to 8 MPa on shaft

75 1:4 150 250 350 450 Determined 1:4 by friction

6 15 15 15 15 15 to 50

Low Medium Medium Medium Medium Medium

1000 2000 Up to165 MPa Up to 200 kN

Determined 1:4 by friction Determined 1:4 by friction

Unlimited Medium

3 Mpa6 MPa on shaft Up to8 MPa on shaft Up to 6 MPa on shaft 600 1200 Up to 9 MPa on shaft

50

Medium

1:6

12

Medium

Determined 1:4 by friction

42

Medium

Nil

Determined Vert 42 by friction only.

High

Determined 1:10 by friction

22

Medium

200 1:6 400 1:6 Determined 1:4 by friction

12

Low

60

High

Note: Check SECTION 7.0 for more explicit information on typical working loads.

64

SUMMARY DETAILS OF PILING SYSTEMS Cost per Penetration Ability to Noise Vibration Site Area Required KN Metre Ability handle Pollution Levels if Boulders Levels not Predrilled

Normal Headroom Required Metres

Med/High Medium Medium Med/Low Med/Low High

Fair Good Good Good Good Good

Fair Good Good Good Good Good

Low Medium Medium Medium Medium Medium

Low Medium Medium High High High

Small Medium Medium Medium Medium Medium

7.2 19.2 19.2 19.2 19.2 20

Low High Low

Good Good Poor

Poor Good Poor

High High Medium

High Fair Medium

Medium Medium Medium

21 21 20

Low Low Low High High High

Good Good Fair Good Excellent Good

Good Fair Poor Fair Excellent Poor

Low Low Low Fair Low High

None None None Fair None Low

Medium Large Medium Small Large Large

15 to 30 15 to 30 30 3 30 30

65

7.0

TECHNICAL DETAILS OF PILING SYSTEMS

7.1

FRANKI DRIVEN CAST-IN-SITU PILES

The Franki pile has been used extensively throughout Southern Africa for the past 48 years and is still today one of the most popular pile types. With a wide range of pile sizes and the advantages of the enlarged base the Franki pile is suited to structures that vary from single storey residential buildings to multi-storey office blocks. There are also some interesting variations in the installation technique which have special applications. Positive features • • • • •

It is often a very economical system. There is an extensive range of pile sizes. The pile has an excellent load/deflection performance. Noise levels are relatively low. The pile has excellent tension load capacity.

Negative features • •

Vibration associated with the driving of the piling tube. Pile heave in saturated cohesive soil profiles.

The main feature of the Franki pile is the enlarged base formed at the toe of the pile. In forming the enlarged base the end bearing area is increased considerably. Furthermore, the displacement achieved when expelling the plug and forming the enlarged base compacts and pre loads the soil surrounding the base. Thus the end bearing of a Franki pile in sands develops at much lower base deflections than that of a bored pile as illustrated in Figure 7.1.1.

Figure 7.1.1 Base performance comparison between a Franki pile and a bored pile

66

PILE DETAILS Identification

Mini

Light

Medium

Heavy

Nom. diameter of pile shaft (mm) Typical working load (kN) Maximum depth (metres) Minimum pile spacing c/c(mm) Maximum rake Standard main bar reinforcing Maximum main bar reinforcing Spiral reinforcing at 150 mm pitch Nominal cover to reinforcing mm Max. tension load (kN) O.D. of piling tube (mm) I.D. of piling tube (mm) Typical hammer mass (kg)

250 or300

355

410

520

Super Heavy 600

250 or 300 500 6 15 650 900

750 15 1050

1200 15 1300

1600 15 1500

1:4 4 x10mm

1:4 4 x12mm

1:4 4 x16mm

1:4 6 x16mm

1:4 6 x20mm

6 x 16mm

8 x 16mm

8 x 20mm

8 x 25mm

12x 25mm

8 gauge

6mm

6mm

6mm

6mm

25

35

40

50

50

75 248 229 550

150 355 323 1600

250 406 366 2200

350 521 457 3300

450 610 546 4500

INSTALLATION TECHNIQUE The piling rig for installing the Franki pile has an engine, a winch, a mast, an open ended piling tube and a long cylindrical drop hammer which is located within the bore of the piling tube, the latter being held and guided by the mast. Plate 7.1.1 shows one of the rigs used to install the Franki pile. The first operation is to drive the piling tube into the ground. To be able to do this a plug of gravel or sand is formed inside the tube at its toe. This is achieved relatively easily by placing a measured quantity of gravel or sand in the tube while the tube is resting on the ground and then compacting this with short drops of the hammer. Once the plug is compacted the hammer drop is increased and the tube is driven into the ground by successive blows of the hammer falling on the plug. The plug arches in the tube thus drawing the tube into the ground while at the same time preventing the ingress of water and/or soil. The number of blows to penetrate each 250 mm is monitored so as to have a record of the driving resistance of the tube. This assists in deciding on a suitable founding level for the pile.

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FRANKI PILE INSTALLATION SEQUENCE

Figure 7.1.2 A

A plug of sand/stone is placed in the piling tube and compacted with the hammer.

B

The tube is driven by applying blows of the drop hammer to the plug which arches in the tube and draws the tube into the ground.

C

On reaching the founding level the tube is held by the extracting gear while the plug is expelled using blows of the hammer.

D

Measured quantities of relatively dry concrete are expelled from the toe of the tube thus forming an enlarged base.

E

The reinforcing cage is placed in the tube which is then filled with high slump concrete.

F

The tube is extracted by means of the extraction gear. On deeper piles the concrete level may have to be topped up during extraction.

G

The completed pile.

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The tube is normally driven to a predetermined depth at which stage the penetration rate for ten standard blows is checked. The amount of penetration is referred to as the "set". If the set is equal or less than that calculated then the tube has been driven to an adequate depth. The plug is then driven out the end of the tube by successive blows of the hammer while the extraction winch is used to hold the tube back from further penetration into the ground. The expelling of the plug is followed by the formation of the enlarged base. This is achieved by gradually feeding zero slump concrete into the tube a while at the same time continuing with measured blows of the hammer, the hammer expelling the concrete to form the enlarged base. The volume of concrete expelled and the number of blows are recorded and these are compared with the theoretically calculated energy required to provide the load capacity. The size of the enlarged base is increased until such stage as the required energy levels are met. The final operation is to use the hammer to ensure all concrete is out of the tube. The steel reinforcing cage is lowered into the tube and the tube is filled with high slump concrete. The tube is then withdrawn using the extraction winch and a system of reeved sheave blocks to gain a mechanical advantage. With deeper piles the tube has to be topped up with additional concrete during the extraction process. Once the tube is out of the ground the cut-off level of the concrete is checked and adjusted if necessary .The position of the steel reinforcing cage at the head of the pile is also checked and adjusted. The piling tube is also washed out before commencing the next pile so as to remove any latence on the inside of the tube. The above sequence of operations is illustrated in Figure 7.1.2. THE ADVANTAGE OF THE ENLARGED BASE As mentioned earlier there are numerous advantages to be gained by having the enlarged base on a Franki pile. The first of these is the increase in load bearing capacity. With the end bearing component of a pile's capacity being the product of the allowable end bearing stress and the area of the base of the pile, it is a very simple conclusion that if the end bearing area is doubled then so is the end bearing component. With a Franki pile it is possible to more than double the end bearing area by enlarging the base. This is not the only advantage related to the end bearing. By enlarging the base the surrounding soil is compacted so the ultimate end bearing stress is also increased. Another advantage is that the enlarged base has an excellent load/deflection performance at low base deflections as shown in Figure 7.1.1. This is due to the fact that the expansion of the base is a form of pre-loading of the soil surrounding the base. The main benefit that is gained from these features is that the founding level for the Franki pile is often at a considerably higher elevation than that necessary for other piling

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systems. This can and often does result in an economical piled foundation based on the use of the Franki pile while pile cap deflections remain within acceptable limits. Another major advantage of the enlarged base is the large tension loads that can be resisted by the Franki pile. The enlarged base forms an ideal positive anchorage and significant tension loads can be resisted. The Franki pile is thus often used for structures with tension pile loads such as transmission towers and chimney stacks. It is also ideally suited to founding structures in heaving clay where the enlarged base formed within a stable stratum is very effective in preventing uplift movement. To be able to resist these tension loads it is essential that the steel reinforcing is cast into the enlarged base. See Anchoring Reinforcement in base in the following section for more information on this procedure. VARIATIONS INSTALLATION TECHNIQUE Anchoring Reinforcement in Base Piles are often required to take tension loads. Because the Franki pile has an enlarged base a considerable tension capacity can be generated provided the steel reinforcement is anchored into the base. To achieve this the following variation to the installation technique takes place after the plug has been expelled. Using a 20 mm slump concrete a 1.5 to 2.0 metre length of pile shaft is formed using the rammed shaft technique described hereunder. A slightly wetter mix is used and the concrete is extruded from the tube with the hammer as the tube is slowly withdrawn. The tube is then re-plugged with normal zero slump base concrete and driven back through the shaft. As a result of this the shaft concrete is forced sideways and an enlarged base is formed. The plug is then expelled and further enlargement of the base is achieved using the standard basing technique. The reinforcing cage is then placed into the tube and the shaft concreted iRthe normal way. Franki piles made in this manner can resist considerable tension loads and tension load tests in excess of 1000 kN have been successfully carried out. Rammed Shaft The rammed shaft has an uneven surface which results in about fifty percent greater shaft friction than that of the smooth standard shaft formed using high slump concrete. The only reason for using the rammed shaft would be to increase shaft friction and as the Franki pile is generally considered an end bearing pile it is not widely used. The rammed shaft is formed by expelling successive charges of zero slump concrete out of the tube as the latter is gradually withdrawn. Each 'measured concrete charge is placed in the tube and the hammer is lowered to rest on the concrete. The tube is then withdrawn about one diameter and the concrete is given one short test blow of the hammer. The tube is then withdrawn an additional amount depending on the resistance of the ground as

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determined by the test blow. The concrete is given another three to four blows of the hammer by which stage the toe of the hammer should be level with the toe of the tube. Another charge of concrete is lowered into the tube and the process is repeated until the shaft has been cast to the correct level. Because of the stiffness of the concrete mix pile cut-off levels well below ground level are possible with the rammed shaft method. It should be noted that the reinforcing cage is placed in the tube prior to the concreting of the shaft and that the hammer operates within the reinforcing cage. The cage has to be well made so as to prevent damage from the hammer. There is also limited control on the concrete cover with this type of shaft. Problems have been experienced when using this shaft technique in very soft soil profiles. The resistance to the concrete being expelled from the tube is very low with the result that it tends to flow out in one direction more than another. This can result in abnomally shaped piles and piles with no cover to the steel in places. Similar problems can occur where the piles have been predrilled and soft saturated soil fills the annulus when the piling tube is installed. Predrilling, Jetting and Coring In the hinterland of South Africa the soils are normally not saturated and are often of dense or stiff consistency. It is thus very difficult and in many cases impossible to drive the piling tube into these soil profiles. To achieve the required penetration the pile position is first predrilled using an auger rig. The depth to which the pile is predrilled will depend on whether there is any tendency for the predrilled hole to collapse, the consistency of the soil profile and any tendency for pile heave to take place. It is standard practice to attempt to drive the tube beyond the predrilled depth so as to ensure that the pile's end bearing is not affected by the predrilling. In the coastal areas where the soil profiles are generally saturated, predrilling is often not possible. In these circumstances jetting with a water jet pipe can be used to assist the penetration of the tube through a dense stratum. A water jet is typically a 100 mm pipe with a nozzle on the toe end and connected to a high pressure high volume pump. The jet pipe is lowered down the side of the piling tube as the latter. is driven into the ground. By keeping the jetting action close to the toe of the piling tube the driving resistance is reduced considerably. To keep the piling tube plumb it is common to use two jet pipes, one either side of the tube. It is standard practice to drive the tube two to three metres beyond the level at which jetting was discontinued ceased so as to ensure full end bearing capacity. The alternative to jetting in saturated soil profiles and in particular in saturated cohesive profiles is the coring technique which is simply removing the soil from the bore of the tube while the latter is top driven into the ground using a drop hammer. The process is messy and slow but on occasions, such as, to avoid pile heave in saturated cohesive profiles, it provides the best solution.

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Conditioning the Soil In sandy soil profiles the mere act of driving the piling tube has a conditioning affect on the soil. As more piles are driven so the soil strength increases and the ground is said to "tighten up". A greater degree of conditioning, however, can be achieved in softer soil profiles by first making a sand pile and then driving back through this. This achieves a double volume displacement and thus a greater degree of conditioning. Another way a sandy soil can be conditioned is through the process of forming the enlarged base. By expelling material from the toe of the tube a sandy soil can be compacted. The material expelled can be concrete or it can be sand or gravel. If the latter, a concrete base is finally formed in the compacted zone in the normal way. Additional Depth One of the limitations with the Franki system is the limited depth to which the standard pile can be installed. This is 6 metres for the Mini and 15 metres for the Light, Medium, Heavy and Super Heavy. If the water table is not high these depths can be increased using an extension tube. This is merely an additional length of piling tube which is attached to the top of the normal piling tube and which enables additional depth to be achieved. An additional one to two metres for a Mini is common and up to 5 metres for the other sizes. Using an extension tube does, however, slow down the production rate so the cost per metre will rise. Comparison of costs with other piling systems will determine whether the use of extension tubes is an economical proposition. Because it is difficult to seal the joint between the extension tube and the main piling tube the use of this technique is not possible in areas with a high water table. Permanent Liner to Pileshaft In cases where the ground water is polluted with chemicals that are harmful to concrete it is desirable to have a permanent liner which protects the shaft concrete from attack. Liners made out of steel or plastic can be incorporated into the shaft of the pile. These liners are normally fixed to the reinforcing cage and lifted up and lowered into position with the cage. The shaft is then concreted in the normal way. Precast Concrete Shaft A Franki pile can also be formed using a precast concrete shaft with the resulting pile being referred to as a Franki Precast Composite. A precast concrete shaft is a high quality product and the Franki Precast Composite pile is ideally suited to sites with aggressive groundwater conditions. The precast shaft is also smooth and in the formation of the pile there is an annulus of loose material created around the shaft. In the case of negative friction conditions where the ground surrounding the pile is settling and causing downdrag, these two factors assist considerably in reducing the downdrag force.

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A small quantity of sand/cement grout is placed in the tube once the enlarged base has been formed. The precast shaft is lowered into the piling tube and penetrates into the grout to bear on the enlarged base. The piling tube is extracted and the gap around the precast shaft is filled with loose sand so as to provide some lateral support to the shaft. POTENTIAL PROBLEM AREAS Pile Heave This is the phenomenon in which a previously installed pile is lifted by upward movement of the soil surrounding it caused by the driving of an adjacent pile. It occurs in saturated clayey and silty soils and does not occur in clean sand. It only occurs with driven displacement type piles and not with bored piles. Where pile heave lifts the whole pile it is generally thought that bearing capacity is not affected materially. In some cases, however, pile heave has been found to be detrimental to the bearing capacity of the pile. This is believed to be due to the separation of the shaft from the base due to insufficient tension transfer mainly caused by low bond figures in the green concrete. By test loading a pile that has heaved one can establish whether the heave has been detrimental to the pile's bearing capacity. If this proves negative then the problem can be ignored although any simple measures to reduce the heave are nevertheless advisable. Predrilling or coring a pile as discussed in the above VARIATION TO INSTALLATION TECHNIQUE section can largely reduce the amount of heave because displacement takes place over a reduced depth. If predrilling or coring is carried out for the full pile depth then the risk of shaft-base separation is largely eliminated. Heave may still take place but this will lift the base of the pile and, from experience, is not usually detrimental to the pile's bearing capacity. Another simple measure that can reduce or eliminate pile heave is to leave the pile for three to four days before driving the pile immediately alongside. This time delay enables the bond between the reinforcing bars and the concrete to build up. Welding a shear key to the steel to improve the bond is also a possibility. When considering the use of a driven displacement pile in saturated cohesive soils the number and spacing of the piles in the group is an important consideration. A single tank base with a large number of piles in one group is a real potential problem and more suited to a bored pile solution. An open warehouse type structure with groups of two and three piles should not present a major problem. The spacing of the piles in the pile cap should be increased if pile heave is a potential problem.

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Vibration During Driving and Forming of the Enlarged Base The act of driving a pile causes a certain amount of vibration. In general the smaller the pile the smaller the energy applied and the less the vibration. Exceptions to this rule have occurred and are thought to occur when the frequency of driving is resonant with the natural frequency of the ground. Normally, however, the vibrations experienced with a Light pile would be much lower than those experienced with a Heavy. The vibration levels generated by the Franki system are generally not that severe. They can and have resulted in minor cracks forming in buildings and the extension of existing cracks. The discomfort of feeling the vibration is not normally a problem but the longer the contract the more sensitive people become. Measures such as predrilling, jetting and coring can be used to reduce the levels of vibration. The vibration experienced when forming the base is however always there but generally of a lower level than when driving because the energy levels are lower. Contracts close to residential buildings should be avoided unless the piles are predrilled, jetted or cored. The more sensitive parts of city centres should also be avoided. Noise Pollution Noise levels are not much above that of the main engine noise. There is the thump from the hammer impact but this takes place inside the tube so is fairly muffled. The odd clang of a wire rope hitting the side of the mast does not seem to worry people. Noise is generally not a problem with the Franki system. Artesian Conditions The risk of artesian conditions is very low but they are known to occur. When a cast-in-situ pile is formed through an artesian layer the ground water which is under pressure tends to travel up the side of the pile and in so doing washes out the cement in the concrete. The amount of defective shaft resulting from this action will depend on how strong the water source is. In the one recorded case of artesian conditions the effect on the pile shafts was serious. Artesian conditions should be identified and reported by the drilling operator during the geotechnical investigation. This condition should then be fully investigated. A Franki pile with a precast shaft or a permanent casing or one of the preformed piling systems could provide a better choice of pile type if artesian conditions are present.

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Plate 7.1:1. Franki Crawler Rig

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7.2

DRIVEN TUBE PILES

This system is not used extensively mainly due to the high cost of the steel tubes. There are, however, situations where the positive features of the system outweigh the costs. Small pile sizes are commonly used for underpinning houses and light buildings with limited headroom and poor access. Medium pile sizes are commonly used for piling new column foundations within existing buildings or in difficult access areas. The larger sizes are used mostly for river bridge foundations and in marine construction. Positive features • • • • • • •

There is an extensive range of pile sizes. The system can achieve considerable depths ( >60 metres in a suitable profile ). The pile is permanently cased and thus ideal for river and marine work. It can be installed in limited headroom. It can be installed in areas with very difficult access. The shaft is cast in the dry so quality control is good. Noise levels are not high.

Negative Features • •

It is a relatively expensive system. There is vibration associated with the driving of the tube.

Steel piling tubes can be installed either open ended or closed ended. With the closed ended technique the toe end of the piling tube is sealed off with a steel plate so that there is full displacement during driving. It is this more common technique which is covered in this section. The use of open ended steel tube piles is associated with temporary staging structures as well as marine work and is not covered in this text due to its specialised nature. PILE DETAILS The working load shaft stress is generally in the range 8 to 10 MPa. Shaft stresses of up to 16 MPa have been used on deep piles where there is a significant friction component and the piles are driven onto a competent founding stratum. The strength of the shaft concrete has to be raised in line with the higher shaft stress. Pile Diameter Typical working load (kN) Typical max. depth (m) Pile spacing c/c (mm) Maximum rake Typical main reinforcing

150 125 12 500 1:4 4x10mm

250 400 20 750 1:4 4x12mm

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400 1000 40 1200 1:4 5x16mm

500 1750 50 1500 1:4 6x20mm

600 2500 50 1750 1:4 8x20mm

STEEL PILING TUBES Either barrel type tubes or spirally manufactured tubes can be used provided the manufacturing process, and in particular the welding, has been carried out to high standards with stringent quality control. The tubes should be manufactured according to SABS 719 Grade B which is the equivalent of the American API 5L Grade B specification which is used worldwide. Before installation can commence the sections of piling tube have to be prepared and welded up. The lead section of tube has a steel plate or rock shoe welded to the toe end and a splicing band welded to the top end. Follower tubes need only the splicing band to be welded to one end. The wall thickness of the tube will be in the 5 to 10 mm range. The smaller diameters will most likely have a 5 or 6mm wall thickness throughout whereas the larger diameters will have a 10 mm wall at the toe reducing to 6mm at the top. The splicing bands should be about 250 mm wide and should be made out of plate of the same thickness as the wall of the tube. If two tubes of different thickness are being welded together then the plate thickness should match the thicker of the two. Ideally there should not be more than a 2 mm difference in the wall thicknesses of two sections of tube to be welded together. The diameter of the end plate should exceed that of the tube by the OD plus 12 mm. Alternatively, the plate can be welded on the inside of the tube in which case the friction component of the pile's bearing capacity will be greater. The end plate can have a rock shoe type arrangement welded to it. Rock shoes should be used if a boulder layer has to be penetrated or if the pile is to be driven onto a sloping rock face. An effective rock shoe can be manufactured using a central pin and four heavy gussets welded to the end plate as shown in Figure 7.2.1.

Figure 7.2.1 Driven tube pile details

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The tubes can be either top or bottom driven during the installation process. If bottom driving is used there is obviously more stress placed on the welds, both circumferential as well as longitudinal. Bottom driving is more efficient, however, so is often resorted to for the final drive even if the initial penetration is achieved with a diesel or hydraulic hammer driving on a helmet on the top of the tube. INSTALLATION TECHNIQUE Bottom Driven The lead section of tube is lifted up and positioned in the leader of the piling rig. A leader guiding mechanism must be clamped to the tube and the leader adjusted for verticality or rake. With the toe of the tube resting on the ground on the pile position a measured quantity of semi dry concrete is discharged into the tube. The initial quantity should fill the tube about 3 to 4 diameters. The hammer is a cylindrical drop hammer which operates within the bore of the piling tube. The plug of concrete is compacted using a few short drops of the hammer. The drop is then gradually increased but is normally not greater than 2.5 metres. The plug material becomes pulverised during driving and this lowers the efficiency of the blow of the hammer. For this reason and to prevent the tube from splitting the plug has to be continually refreshed by adding additional plug material throughout the driving operation. The next section of tube is welded on when the top of leader tube is at a convenient height for welding. Further sections are welded on as and when required. When the toe of the tube is approaching the founding stratum measurements of the penetration rate are made and the set checked occasionally. When the required set has been achieved this signifies the completion of the driving and the hammer is removed from the tube. It is possible to inspect the internal bore of the tube using a light or a mirror and reflected sunlight if necessary . The reinforcing cage is then lowered into position and the shaft of the pile concreted using a high slump self-compacting mix. It is common practice to reinforce only the upper 12 metres of the pile shaft because of the permanent thick wall casing. Top Driven If top driving is used the tube is lifted into the leader of the piling rig as previously, the guides are fixed and the leader is adjusted for verticality or rake. The helmet and hammer are then lowered onto the head of the tube and alignment of the hammer and tube is checked. Driving commences with the operation of the hammer and is continued save for welding on follower tubes until the required founding stratum has been reached and an adequate set achieved.

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Top driving of long closed ended steel piling tubes with a thin wall is not as efficient as bottom driving in hard driving conditions as a large proportion of the hammer's energy is absorbed by the tube itself. If the driving is particularly hard or the piles are heavily loaded then it may be necessary to resort to bottom driving to achieve the required depth or set. The efficiency of bottom driving is estimated to be between 15 and 25 percent. Once the tube is driven it is a simple matter to place the reinforcing cage and concrete the shaft as described above. A typical pile detail is shown in Figure 7.2.1. Plate 7.2.1 shows driven tube piles being installed on a river bridge foundation. VARIATIONS IN INSTALLATION TECHNIQUE Predrilling and Jetting If hard driving is experienced this can be relieved to some extent by predrilling or jetting the pile through the dense layers. This relief can only be achieved over the top 15 metres if jetting and 36 metres if predrilling. POTENTIAL PROBLEM AREAS Split Pile Tubes There is no doubt that this is the main risk when installing this pile type. The splitting of the odd casing is to be expected and there are ways to remedy this. Ifa number of casings split, however, then the problem can be very serious. The risk of splitting a casing should not be high provided the casings are manufactured to high standards of quality control, the appropriate casing thicknesses are used and the correct driving procedures are adhered to. As the welds in the casing are potential weak areas it is sensible to keep the number of welds to a minimum. Tubes made up of long barrels are thus preferred over those made using the spiral technique. The standard of on- site welding must be high and only skilled welders should be used. During driving it is most important to refreshen the plug regularly. The impact of the hammer tends to enlarge "the casing so if the plug is not refreshed the casing will continue to expand until it fractures. By refreshing the plug the point of impact of the hammer is raised and another section of tube comes under the bursting stresses caused by driving. A sufficient length of heavier wall casing must be used at the toe of the pile so that the impact of the hammer stays within this section. As stated earlier the risk of casing fracture can be reduced if top driving is used. The efficiency of top driving is, however, much lower and top driving is very noisy.

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Pile Heave Being a displacement type pile, pile heave may be experienced in saturated cohesive soil profiles. Heave checks should be carried out and if heave is occurring then the concreting of the piles should be delayed until all the piles in a group have been driven. They should then all be re-driven so as to eliminate the heave and any negative effects it may have on the pile's bearing capacity. Vibration During Driving As large hammers are used on the larger pile sizes a considerable amount of vibration can occur with this pile type. The severity of the problem can be reduced by using short hammer drops, predrilling the pile positions or jetting alongside and below the tube during the driving operation. Noise Pollution With bottom driving the noise levels are not high and similar to that experienced with a Franki pile. Top driving on the other hand is noisy and should be avoided in built up residential and commercial areas.

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Plate 7.2.1 Driven tube piles being installed on a river bridge foundation

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7.3

PRECAST PILES

The Precast pile was one of the first piling systems to be used in South Africa there being a record of the piling to the old Putt Bridge in Port Alfred (now demolished) where the precast piles were installed using a steam operated piling machine back in 1908. Modern technology has introduced the jointing of precast piles which has overcome the original depth limitations and increased the use for the pile type. Precast piles have a wide use from bridges to commercial and industrial buildings. Because of the high noise levels associated with the driving of precast piles they are seldom used in heavily populated residential areas or in downtown city centres. Positive features • • •

It can provide an economical solution especially in deeper soil profilesA precast pile shaft is a higher quality product than an in-situ shaftInstallation is quick and control on site is good.

Negative features • • • • •

There is considerable noise associated with the driving of the pile. }~ There is vibration associated with the driving of the pile. Damage or fracture of pile shafts can occur during driving. Pile lengths have to be predetermined well in advanceIncorrect assessment of shaft lengths can result in high waste.

PILE DETAILS Pile Size Typical working load (kN) Maximum depth (metres) Minimum-pile spacing (mm) Maximum rake Standard main bar reinforcing Maximum main bar reinforcing Cover to main bars (mm)

250 mm square 1000 Unlimited 1200 1:4 4x20mm

350 mm Square 2000 Unlimited 1500 1:4 8xl6mm

4x25mm

8x20mm

30

40

The above two sizes are the most common in jointed precast piles. If single length piles are to be used then the size can be made to suit the contract. The cost of the shutters will, however, be considerable so this is only economical on a large contract. Common sizes for single length precast piles are 350 and 400 square and 350 by 400 rectangular. Single pile lengths of up to 16 metres in ordinary reinforced concrete are common. Longer pile lengths of up to 30 metres are possible using prestressing but this technology is not in common use in Southern Africa.

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PRECAST SHAFT MANUFACTURE The shafts of precast piles are normally cast in a factory and transported to the site as this is the most economical arrangement. On larger contracts especially in remote areas it may well be more economical to set up a casting yard on site. Before deciding on this the local aggregates should be tested so as to ensure that concrete of a sufficiently high quality can be produced. The moulds in which the shafts are cast are made of steel and have a slight taper on them so the shaft can be extracted without having to strip the mould. The moulds are sometimes designed to allow the passage of steam through them for accelerated curing. It is extremely important that the mechanical joints are very accurately aligned so that when sections are coupled together on site the shaft of the combined pile is straight. The moulds are fitted with devices for locating and aligning the mechanical joints and rock shoes. The concrete used in the manufacture of precast concrete pile shafts has to be of the highest quality and thus the aggregates used have to be likewise. The stone in particular has to be of good quality and clean even if this means washing the stone. As the design 28 day strength is 60 MPa the cement content is of the order of 450 kg per cubic metre. Additives are generally used to assist in obtaining these high strengths. The main bar reinforcing steel is high tensile with a mild steel spiral which is generally at 150 pitch except near the ends of the shaft unit where the spacing is reduced to cope with the higher bursting stresses. To ensure that the full moment resistance of the shaft is transmitted through the mechanical joint, splice bars which are threaded into the joint plates extend for the full bond length back from the joint plate. For handling the pile once it has been cast lifting lugs are cast into the pile at fifth points from each end. Piles are often steam cured by passing steam through the shutters while the pile bed is covered with a heat retaining blanket. The timing of the steaming process and the control of the temperatures must be carefully controlled in accordance with recommended steam curing procedure. Piles that are removed from the casting beds are stacked one upon the other in a holding yard while the final curing takes place. When stacking precast pile shafts it is important to support the piles on bearers at fifth points from the ends. All the bearers for a stack of precast shafts must be exactly in line with those of the bottom row or else moments are induced into the shafts and cracking can take place. The overall stability of a stack of precast shafts is also important. When transporting the pile shafts and when stacking on site it is also important to stack the piles correctly. When lifting the pile shaft for driving a single lift point one quarter from the end is normally used. Plate 7.3.1 shows a precast pile manufacturing operation.

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MECHANICAL PILE JOINTS There are a number of internationally patented mechanical pile joints available and there are also a few locally developed joints. It is not the subject of this text to examine these in detail but the structural essentials of a good joint are: • • • •

an ability to transfer the full shear force and bending moment of the concrete section without overstressing any component of the joint and its splice bars. an ability to transfer the full axial compression and tension forces both static and dynamic without overstressing any component of the joint and its splice barsan ability to transfer the full axial compression force both static and dynamic without overstressing in the pileshaft concrete or causing spalling. the faces of the joint must be at right angles to the longitudinal axis of the pileshaft so as to ensure that the jointed shafts have one common longitudinal axis. For this to be achieved the joint must have a device for accurate location in the shutter when casting the pile.

Plate 7.3.2 shows a mechanical precast pile joint. An alternative form of mechanical joint is the welded joint. In this form the joint consists of two 25 mm thick mild steel end plates one of which is chamfered to take the weld. The normal group of splice bars are threaded into the end plates so as to transfer the moment. When installing the pile a single run of an automatic feed welding machine is normally sufficient to ensure an adequately strong joint. ROCK SHOES Precast piles that are required to penetrate cobble or boulder layers or required to found on hard rock or a sloping rock face are normally fitted with rock shoes. Rock shoes can vary from a simple end plate which is merely there to protect the concrete to a sophisticated shoe with a hardened steel point which is design to key into a sloping rock face. Plate 7.3.3 shows a typical rock shoe. INSTALLATION TECHNIQUE The piles are delivered to site and stacked close to the pile positions. The lead pile section is picked up by the rig and located in the piling helmet which is fitted to the leader of the piling rig. The pile section together with the helmet is then lowered so the toe of the pile is on the pile position. The leader is then adjusted for verticality or rake and the driving hammer is lowered onto the piling helmet. The pile can be driven with a drop, diesel or hydraulic hammer. Once the pile is a metre in the ground the verticality or rake is checked again and final adjustments are made. For the remainder of the driving operation the technique is to follow the pile as any attempt to try and correct the position and rake tolerances can result in cracking of the pile shaft through induced moments.

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When the first section of pile shaft is in the ground the second section is lifted up and located in the helmet as was the first. The shaft with the helmet is lowered onto the first pile shaft assisted by a locating pin. The wedges that lock the two joint plates together are then driven home. The alignment of the leader is checked and driving recommenced. If additional lengths are required the process is repeated for the additional shaft lengths. The precast pile is traditionally driven to a set as are most driven piles. The set for long slender precast piles should be calculated using the Wave Equation method. The traditional Hiley formula for calculating pile sets can be used for short precast piles. The final set and temporary compression diagram should be recorded as part as the pile record. It is recommended that the set on some of the piles on a contract are checked 24 hours later so as to ensure that there is no increase in the set which can occur in certain clayey soil profiles. If this is found to be the case the pile must be redriven. In sandy soil profiles the set taken 24 hours later is normally considerably less than that taken at the time of driving due to the fact that the static friction is much greater than the dynamic. Plate 7.3.4 shows precast pile installation in progress. VARIATIONS IN INSTALLATION TECHNIQUE Predrilling and Jetting To obtain relief from hard driving and to assist in reducing the noise levels predrilling and jetting may be resorted to. Slipcoating of Pile Shafts Shell Slipcote is a form of bitumen which is used to coat the sides of preformed piles with the object of reducing the friction on the sides of the pile. It is used in negative friction situations where settlement of ground around a pile results in additional load being transferred to the pile by the fact that the friction is acting in reverse. Hence the term negative friction. The slip coating of precast pile shafts is quite an involved process. The shafts have first to be coated with a priming coat which can be brushed or sprayed on. The shafts are then placed in steel moulds which allow a 10 mm gap all round the shaft. Spacers are used under the shaft and on the sides so as to ensure the 10 mm gap. The slipcoat bitumen is heated and poured into the moulds so that it fills the 10 mm gap. The mould when full also allows for a 10 mm thickness on the upper surface. Once the bitumen has cooled the pile shaft is lifted out of the mould and stored. When cool the slipcoat layer it is quite stiff and is not wiped off during the driving of the pile. The product's ability to reduce friction is very impressive and negative drawdown can be virtually eliminated.

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POTENTIAL PROBLEM AREAS Pile Heave Precast piles; being displacement type piles, can experience pile heave. If heave has been measured then a selected number of piles that have heaved should be redriven to check whether the set has increased from that of the initial drive. If it has then all the piles should be redriven. Noise Pollution The driving of precast piles is a noisy process and due consideration must be given to the potential for objections from the public and possible legal intervention. This should only present a problem in residential and downtown city areas. Vibration The vibration caused by driving can be problem but it is very difficult to forecast beforehand how severe the vibration will be. The general approach from a contracting point of view is to take the risk which is relatively low and sort the problem out in the unlikely event of one arising. Due cognizance should be taken, however, of the condition of surrounding buildings and whether these buildings are occupied by businesses whose operations are susceptible to vibration such as photographic studios and lens grinding. Shaft Breakage A precast shaft has to withstand fairly severe driving stresses during the installation process. As a result it is not uncommon to have the occasional shaft fracture during driving. With well made pile shafts and the correct driving technique the fracture rate should not be greater than one percent. Extremely hard driving over a considerable depth will increase the risk of shaft breakage. The affected pile is normally replaced with one or two additional piles using a revised pile layout. Sonic testing of precast piles is a simple and effective way for checking whether the shafts have been damaged or fractured during driving. See SECTION 9.0 PILE LOAD AND INTEGRITY TESTING for details of this form of testing.

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Plate 7.3.1 Precast pile manufacture

Plate 7.3.2 Precast pile joint

Plate 7.3.1 Precast pile rock shoe

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Plate 7.3.4 Precast pile installation

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7.4

STEEL H-PILE

Steel H-Pile sections are rolled by steel mills in South Africa and are thus available for installation as piles. The cost of steel is relatively high, however, so it is not a widely used pile type due to the fact it is not normally an economical solution. It has advantages in that the steel sections can be driven through soil profiles which have minor obstructions in the form of cobbles and small boulders as well as fill materials containing builders rubble. Positive features • • •

The pile has good penetrating ability. The pile can be installed to considerable depthH-sections are ideal for use as soldier piles.

Negative features • • •

Steel H-sections are relatively expensive. Noise levels are high if the pile is driven. The end bearing capacity is limited unless driven onto rock.

PILE DETAILS The H-section is rolled in two forms known as Universal Columns and Universal Bearing Piles. Details of these sections are given in Figure 7.4.1. The working load shaft stress can be up to 125 MPa for Grade 43 steel and up to 165 MPa for Grade 50B steel. Using the higher stress the range of working loads varies between 500 to 3300 kN for the sections available. INSTALLATION TECHNIQUE The first section of shaft is lifted up by the piling rig and located into the piling helmet which is then released from the hammer. The first shaft section together with the helmet is then lowered down onto the pile position and the hammer is lowered so as to rest on the helmet. The leader of the piling rig is adjusted for verticality or rake. The pile can be driven with either a vibratory, drop, diesel or hydraulic hammer. Once the first shaft section has been driven the second one, which has been prepared for welding, is lifted up and then lowered so as to line up with the first section. Guiding lugs tack welded to the first section assist in this operation. A butt weld is then carried out. Once the initial weld has been made the piling rig can move off and carry on driving another pile while the full weld is completed. Driving is then continued and the cycle is repeated if further lengths are required.

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Figure 7.4.1 Details of Universal column and bearing pile sections

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The pile is driven to a set which has been calculated to provide adequate bearing capacity. The set should be checked after 24 hours so as to determine whether there has been any relaxation. If there has the pile should be driven until the required set has been achieved and again checked after 24 hours. The head of the pile is finally trimmed to the correct level. Shear transfer cleats or lugs may need to be welded to the head of the pile so as to ensure transfer of load from the pile cap into the pile. VARIATIONS IN INSTALLATION TECHNIQUE Predrilling and placing In the case of soldier piles the alternative to driving is to auger a hole and drop the steel HPile sections in. This is common practice in all soil profiles which are cohesive and not saturated thus allowing the augering of the hole to take place without collapse. One of the main advantages of placing the H-Pile sections in predrilled holes is the fact that position and verticality tolerances attainable can be very stringent whereas with driving this is not the case. Noise reduction is also an important factor. Having placed and fixed the H-Pile section in position the annulus around it is filled with concrete or grout. Below the excavation level this concrete is full strength but above the excavation level a weak concrete or grout is used. The reason for this change is to facilitate the removal of the weak material from in front of the H-pile section on the excavated face thereby achieving a plane face to the excavation. Predrilling and Jetting To obtain relief from hard driving and to assist in reducing noise levels predrilling or jetting can be resorted to. POTENTIAL PROBLEM AREAS Noise Pollution The driving impact of steel on steel is unfortunately one of the noisiest operations in piling. Even with a cushioning helmet on top of the H-Pile section there is still a high noise level. Certain measures have been used in Europe to reduce the noise levels and these involve enclosing the leader of the piling rig with light sheet metal lined with sound absorbing material. This has been reported as being fairly successful. The use of an hydraulic or vibratory hammers would also assist with regard to noise levels. The safest solution is to use the Steel H-Piles only on sites where noise pollution is not a problem.

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Pile Heave Because there is limited displacement when driving steel H-Piles there is a very low risk of pile heave. If heave is measured then the piles should be re-driven. Vibration The vibration caused by driving can be problem but it is very difficult to forecast beforehand how severe the vibration will be. Because of the limited displacement associated with steel H-Piles the risk is even lower than with a precast pile. The general approach from a contracting point of view is to take the risk which is relatively low and sort the problem out in the unlikely event of one arising. Due cognizance should be taken, however, of the condition of surrounding buildings and whether these buildings are occupied by businesses whose operations are susceptible to vibration such as photographic studios and lens grinding. Bent Pile Shafts Because of the elastic properties of steel it has been observed that the shafts of steel H-Piles can be deflected and bent during driving particularly if there are obstructions in the ground. Unfortunately there is no easy way in which to monitor whether bending has taken place and if so to what degree other than if the toe of the pile emerges from the ground surface! A pile load test programme will indicate whether the performance of the piles is acceptable but this is an expensive exercise. A study of the founding levels and driving records may highlight discrepancies and suggest a limited number of piles for load testing.

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7.5

TIMBER PILES

This pile type is not often used in South Africa but it has been included in this text for the sake of completeness. Permanent timber piles have been used in harbour construction in Southern Africa but because of severe attack by borers their life was limited even with very hard woods. The use of timber piles in permanent works is presently limited to very light structures such as marina jetties. They are also used for temporary staging over water and stitching earth embankments. The timber used is generally one of the gums such as saligna. Positive features • •

Timber piles are relatively cheap. They are relatively light to transport and handle.

Negative features • • • •

The available sizes and lengths are limited. Load capacity is very limited. Timber is su.bject to attack by various insects and organisms. There is risk of the timber pile splitting during driving.

PILE DETAILS The sizes of readily available timber piles is dictated by what one can purchase from the Department of Forestry or from one of the commercial suppliers. Smaller sizes are readily available but the larger sizes have to be ordered and delays can be experienced. Timber piles used in temporary staging are typically 12 metres in length with a tip diameter of 250 mm. Larger sizes are available locally but are more difficult to source. Special hard woods can be procured from some West African countries but these are expensive. Timber piles driven as a temporary staging are normally friction piles with a capacity normally limited to 200 kN. With the larger sizes of timber pile and dense soil conditions this could be increased. PREPARATION OF TIMBER PILES The timber pile should be de-barked prior to driving. The supplier has the equipment to do this so the timber piles should be ordered in a de-barked condition. If they are to be used in permanent works they should be treated. Each timber pile should be fitted with tightly fitting steel band at both ends to prevent the it splitting during driving. The head of the timber pile should be squared off so that the hammer impacts on a plane surface which is at right angles to the axis of the pile. The toe end is sometimes fitted with a form of rock shoe to provide additional protection to the tip end during driving.

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INSTALLATION TECHNIQUE The timber pile is lifted up using the winch of the piling rig and located in the helmet which is then detached from the hammer. The pole and helmet are then lowered onto the pile position. The mast is adjusted for verticality or rake and driving is commenced. After about one metre of penetration the alignment of pile and leader is checked and the latter adjusted if necessary .The pile is then driven to the full depth. With temporary staging work over water the piles can either be driven by a rig mounted on a barge or by a special cantilever piling rig which is supported on the previously driven piles. A crane type rig may also be used. Drop hammers are generally used for driving timber piles and the mass of the hammer is in the 1500 to 2000 kg range. Piles can be raked but the rake is normally limited to 1:6. It is often a requirement of the contract that all temporary works have to be removed. Timber piles can be extracted but the cost of establishing equipment to do so and the operation itself can be expensive. The alternative is to cut them off at riverbed or seabed level and this can be achieved using a diver and a pneumatic saw. A small explosive charge can also be used to cut a timber pile. VARIATIONS IN INSTALLATION TECHNIQUE Installation using Jetting In soft sands and silty sands as are found in lagoons and river estuaries timber piles can be installed using jetting only. A jetting system with both water and air feed is used to loosen the soil around and below the timber pile which sinks into the ground under its own weight. This is a particularly simple and inexpensive way of installing timber piles for light structures such as marinas and similar pleasure craft jetties. POTENTIAL PROBLEM AREAS Splitting of the Timber The constraining of the ends of the pile with tight fitting steel bands or 16 gauge steel wire tightly bound around the section is essential if the risk of splitting is to be reduced. Even then the timber may still split as there may be a zone of weakness away from the end of the pile. Once splitting starts to occur the driving should be terminated and the timber pile extracted. Inadequate Pile Length In sometimes happens that the timber pile is of insufficient length in a softer area of the site. In such circumstances the pile can be de-rated and additional piles installed. It is possible to splice on an additional section of pile but this is expensive. A change of pile type to an open ended steel cased pile is the other possibility.

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7.6

AUGER PILES

This pile type is very common in South Africa as it is ideally suited to the partially saturated cohesive and residual soils found in large areas of this part of the continent. A number of the large industrial projects in South Africa such as the Sasol oil from coal plants, Eskom power stations and Iscor steel works are founded on this pile type. There is a wide range of pile sizes and thus the pile type is suited to both large and small structures as well as large and small contracts. Positive features • • • • • • •

In conditions suited to the use of the auger pile it provides an economical solution. The system also provides an economical solution to heaving soil profilesNoise levels are low and limited to engine noise from the piling rig. There is no vibration associated with auger piles. There is a considerable range of pile sizes from 100 to 2000 mm diameter. Auger piles can be installed to depths of up to 42 metres. The system can handle boulder layers.

Negative features • •

There is a risk that the sidewalls of the pile excavations may collapse resulting in the use of costly temporary casings. This risk can be eliminated in most instances by carrying out an adequate geotechnical investigation using an auger rig to drill trial holes. Below the water table there is a risk of excessive water ingress into the pile excavation which can seriously inhibit progress and can lead to a change in pile type.

Note that both the above negative features can and should be checked as part of the geotechnical investigation. This is normally done and these risks are thus largely eliminated. PILE DETAILS The range of auger piles sizes as noted above is more extensive than with any other pile type. Auger piles can be as small as 100 mm and as large as 2000 mm in diameter. The following sizes are the more common: 200, 250, 300, 350, 400, 450, 500, 600, 750, 900, 1050, 1200, 1350, 1500, 1650, 1800 and 2000 mm in diameter. Any other size in this range is possible but a special flight may have to be manufactured for any non-standard size. The safe working loads of auger piles founded on competent material can be calculated using a shaft stress of 3 MPa for pile diameters less than 600 mm and 6 MPa for pile diameters of 600 mm and greater. The reason for the difference in the shaft stress is the fact that with piles of 600 mm diameter and greater the base of the pile can be cleaned out carefully by hand and thus one can be sure that there is sound end bearing. Smaller sizes

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are normally cleaned out using a cleaning bucket but this is not entirely successful as the cleaning bucket will always leave a certain amount of loose material on the base of the pile. Because of this the end bearing is generally ignored in the calculation of the pile's capacity for diameters smaller than 600 mm. Hence the recommended lower design figure for the shaft stress for the smaller diameter piles. Pile Diameter mm 200 250 300 350 400 450 500 600 750 900 1050 1200 1350 1500 1650 1800 2000

Working Load kN 95 145 210 285 375 475 585 1650 2650 3800 4700 7350 8550 10600 14400 15250 18850

Minimum Pile Spacing mm 500 625 750 875 1000 1125 1250 1500 1875 2250 2500 3125 .3375 3750 4375 4500 5000

Typical Reinforcement Main Bars 4 x 8 mm 4 x 10 mm 4 x 12 mm 5 x 12 mm 4 x 16 mm 5 x 16 mm 4 x 20 mm 5 x 20 mm 5 x 25 mm 7 x 25 mm 8 x 25 mm 13x25mm 15 x 25 mm 18 x 25 mm 15 x 32 mm 16 x 32 mm 20 x 32 mm

Piles that are socketed into rock can be designed for higher loads due to the increased capacity of the socket friction as well as end bearing. The shaft stress for socketed piles can be increased to a maximum of 8 MPa provided the sockets are cleaned out and inspected to ensure the competency of the rock at founding level. INSTALLATION TECHNIQUE The drilling section of the auger rig consists of an engine powering a winch and an auger drive head. The latter drives the drill stem which is known as a kelly and the kelly is in turn attached to the drill bit or auger flight. The rig has both a drilling speed and a spin-off speed, the latter being used to spin the spoil off the flight. The installation of the pile thus commences with the drilling of the hole. The auger rig is positioned over the pile position and the mast is checked for verticality or rake. Drilling is started by rotating the auger flight and allowing the flight to penetrate into the soil. Once

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the auger flight is loaded with soil, rotation is stopped and the winch is used to withdraw the flight from the hole. Once clear of the hole the flight is spun and the spoil gets flung off clear of the hole. The flight is then lowered down the hole and the process is repeated. The excavated spoil is continually removed from around the hole during the drilling operation. Different types of drilling tools can be fitted to the kelly. These tools have been developed for penetrating the various types of soil strata that can be encountered. Tools for the penetration of rock are also available. The pile hole is drilled to the required depth which is often onto rock. With pile diameters 600 mm and greater the auger rig is then moved off the pile position and a small tripod rig with a safety winch is erected over the hole. A man is then lowered down the hole by means of the safety winch and he sets about cleaning the remaining loose soil from the bottom of the hole. This material is removed using the winch and a small spoil bucket. This operation is carried out in accordance with the SAICE code of practice for the safety of persons working in small diameter shafts and test pits for civil engineering purposes (1990). With pile sizes less than 600 mm in diameter the holes cannot be cleaned out by hand. In these instances a cleaning bucket is used to remove as much of the loose material as possible after which anything remaining is compacted by a plate fixed to the end of the kelly bar. Once either of the above cleaning operations has been completed the hole is checked by a supervisor. The steel reinforcing cage fitted with cover spacers is then lowered down the hole. The shaft of the pile is concreted using a self compacting concrete mix which is discharged into the hole via a short funnel. The purpose of the funnel is to ensure that the concrete flows down the centre of the reinforcing cage and not onto the cage as this leads to segregation as well as poor compaction. After the shaft has been concreted a poker vibrator is used to assist the compaction of the top three metres where there is insufficient head for self compaction. The smaller pile sizes such as 100, 150 and 200 mm diameter can be drilled by hand in suitable soil profiles using a hand auger purchased from a hardware store. One can also hire mechanised handheld augers for drilling these sizes. Tractor mounted augers are available for drilling sizes from 200 up to 400 mm diameter. Franki South Africa owns a number of larger auger rigs. These are mainly truck mounted rigs which can be driven to the sit-e. The following table gives details of these machines.

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Rig Type Hotline16M120 Soilmec RTAH Williams LDH80 Williams LLDH120

Max. Diam. mm 1200 1500 2200

Max. Depth metres 16 32 24

Torque kN metres 3000 11900 6818

Rig Mass kg 24400 30000 28580

2200

36

13636

39310

A selection of photographs of auger rigs and auger pile contracts is shown in Plates 7.6.1 to 7.6.6. VARIATIONS TO THE INSTALLATION TECHNIQUE Underreams An underream is an enlargement of the base of the pile. In certain circumstances it may be considered desirable to increase the pile's end bearing area by forming an underream. With the auger system there are mechanical tools that can be attached to the kelly which will excavate the underream. This can only be achieved in materials that are not too dense or stiff especially on the larger diameters. In the harder materials the underreams can be formed manually with the assistance of air tools. Underreams must be cleaned out by hand prior to concreting. By implication, therefore, underreams are only used for piles with a diameter of 600 mm or greater. The alternative to an underream is a rock socket and as this can be drilled more easily and quicker the tendency is to use sockets as opposed to underreams. Piles to Resist Ground Heave There are large areas of South Africa where the soils are described as active. The volume of active soils increases with increase in moisture content and decreases with decrease in moisture content. Increase in soil volume results in the ground surface rising and this is referred to as ground heave. When soil heaves in an area which has been piled there is a friction transfer between the soil and the pile which imparts an uplift force on the pile shaft. Where this force exceeds the axial compression load on the pile the pile shaft will be in tension. If the tension force is not high this can be resisted by the shaft reinforcement and the anchorage provided by the length of pile below the active zone. In circumstances where potentially very large tension forces can occur, a method to reduce the friction transfer to a level that can be resisted by the shaft reinforcement has been developed. This method involves the creation of an annulus around the shaft of the pile into which a low strength material is placed. The theory is that the low strength material is only capable of transferring low friction forces and thus the total friction transfer will be similarly reduced. The annulus is created by drilling a pile excavation which is say 200 mm larger in diameter than the required shaft diameter. A thin walled metal liner of the required

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shaft diameter is then placed in the excavation thereby creating an annulus. This is then filled with a suitable low strength material such as vermiculite or polystyrene aggregate. The annulus is only created over the depth which is considered to be active. The remainder of the pile below the active zone is constructed in the normal way as this is designed to provide the anchorage against any tension force in the pile shaft which may still develop. See SECTION 18.0 PROBLEM SOILS AND THEIR FOUNDATION SOLUTIONS for more information on piled foundations in heaving soils. POTENTIAL PROBLEMS Collapse of Sidewalls If there is an area of the site where sidewall collapse occurs then the pile holes need to be temporary cased through the collapse zone. To be able to do this and still end up with the correct size pile, drilling down to the top of the collapse zone has to be undertaken using a flight about 100 mm larger in diameter than the pile size. A steel casing which has a slightly smaller diameter is then lowered into the hole. A vibratory hammer is clamped to the top of the casing and it is driven down through the collapse zone. Using the size flight for the pile diameter the spoil inside the casing is drilled out and the remainder of the pile hole drilled. The pile is cleaned out and concreted in the manner described above, the concrete filling the temporary casing to a level above the cut-off level. The casing is finally extracted using the vibratory hammer and the concrete flows out of the casing so as to fill the larger diameter. As a number of additional plant items are required for handling the temporary casings their use increases the cost of the operation quite significantly and to an extent that may render the solution uneconomical when compared to other possible pile types. If only a small section of the site has the problem then the overall economics will still favour the Auger pile. This may not be the case if the whole site has a collapse problem. Water Ingress The occurrence of ingress of ground water into a pile excavation is fairly common particularly in the soft rock at the toe of the pile. This water seeps into the pile excavation and collects in the bottom of the hole and a decision has to be made as to whether the pile can be concreted successfully in these conditions. The answer depends on the rate of inflow. If the flow is limited then a pump lowered to the bottom of the hole during the cleaning operation will remove the water allowing cleaning to proceed. When this has been achieved a large quantity of concrete is discharged rapidly into the hole via a concreting shute as soon as the pump is removed. The steel reinforcing cage is then lowered into the hole and bedded into the concrete. The remainder of the shaft is then concreted.

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During the short time it takes to remove the pump and discharge the first concrete a quantity of water will enter the hole. This will not affect the performance of the pile provided that the concrete that is discharged into the pile hole is concentrated in the centre and the concrete flow rate is high. This will limit the amount of water that is mixed with the concrete at the base of the pile. Some water will of course get mixed with the concrete during the concreting operation but this can be compensated for by using additional cement in the mix. If the rate of inflow is too fast then it is better to cast the pile under water. For this to be achieved it will be necessary to temporary case the hole for the full depth prior to concreting to prevent material that may collapse off the sidewalls due to the presence of the water from collecting on the bottom of the hole. The concreting operation is then carried out using a tremie pipe which is standard procedure in piling and the casing is extracted. Another alternative in this situation would be to drill the piles under a head of bentonite as covered under SECTION 7.7 UNDERSLURRY PILES. This would eliminate the problem but would increase the cost. Boulders The problems presented by boulders depend on factors such as the size, hardness and concentration of the boulders, the type of matrix and whether there is any water present in the boulder layer. If the size of the boulders is less than one third the pile diameter, the concentration is plus/minus two or three boulders per metre of depth and the matrix is soft or loose the auger rig will drill through the layer albeit with a slower penetration rate. If the concentration increases to tightly packed the penetration rate will decrease even further. If the matrix also changes to very stiff or very dense then penetration will be very slow if at all possible. It will not be possible to drill out boulders larger than one half the pile diameter. In these cases personnel are lowered down the pile excavation and a sling is attached to the boulder by drilling a hole through it. A crane is then used to lift the boulder out of the excavation. If a large boulder is encountered then it will be necessary to split the boulder into smaller pieces before they can be removed. Piling in these conditions can be very slow and thus expensive but the problem can be solved. Auger piling in boulders below the water table can still be feasible provided the ingress of water can be handled by pumps to allow access to the boulder by personnel. If the ingress is too fast for this to happen then there is no alternative but to change to the Oscillator Piling System which is capable of handling this type of problem. For details of the Oscillator system see SECTION 7.10.

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Plate 7.6.1 Williams Digger drilling

Plate 7.6.2 Typical auger flight

Plate 7.6.3 Auger piling in progress

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Plate 7.6.4 Installing temporary casing

Plate7.6.5 Hole cleaning operation

Plate 7.6.6 Exposed auger piles

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7.7

UNDERSLURRY PILES

This system is commonly used for heavily loaded structures where the soil profile is saturated over all or part of the profile resulting in unstable sidewall conditions for conventional auger piles. The term Underslurry indicates that the pile is excavated under a head of bentonite slurry which prevents collapse of the pile excavation. Both auger and grab excavation methods can be used resulting in circular and rectangular cross-section respectively. These two methods are referred to as Auger Underslurry and Barettes. Positive features • • • • • •

Economical pile solution for heavy structures on saturated soil profiles. Noise levels are low and limited to engine noise from the equipment. There is no vibration associated with Underslurry piles. A good range of pile sizes from 900 mm to 1500 mm diameter. Underslurry piles can be installed to depths of up to 42 metres. Barette type piles can be shaped.

Negative features • • • •

High establishment costs. Large site area required. Suited to heavily loaded structures only. Only vertical piles can be constructed.

PILE DETAILS Auger Underslurry Pile Diameter mm 900 1000 1100 1250 1350 1500

Working Load kN 4000 4700 5700 7350 8500 10600

Minimum Pile Spacing mm 2250 2500 2750 3125 3375 3750

Typical Reinforcement Main bars 7 x 25 mm 8 x 25 mm 10 x 25 mm 13x 25mm 15 x 25 mm 18 x 25 mm

The above working loads are calculated using a shaft stress of 6.0 MPa. If the piles are socketed into bedrock or there is considerable shaft friction then the loads on the piles can be increased but should not exceed 8 MPa on the shaft. The pile's tension capacity is dependent on friction and the amount of reinforcement. The maximum depth to which Auger Underslurry piles can be installed is 42 metres and the piles should be vertical as it is not possible to control the drilling on the rake so as to ensure a straight pile axis.

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PILE DETAILS (cont.) Barettes Pile Width mm 600 800 1000 1200

Pile Breadth mm 2200 2200 2200 2200

Shaft Area m2

Working Load kN

1.24 1.68 2.12 2.56

7500 10000 12500 15000

Min. Pile Spacing mm 2250 2500 2750 3000

Typical Reinforcing Main Bars 14 x 25 mm 16 x 25 mm 22 x 25 mm 26 x 25 mm

The above working loads are calculated using a shaft stress of about 4.5 MPa. Because of the grab's limited ability to penetrate materials of very soft rock consistency it is recommended that this be a maximum unless a socket is formed. The pile's tension capacity is dependent on friction and the amount of shaft reinforcement. The maximum depth to which Barettes can be installed is really unlimited but a practical limit of 60 metres is suggested. Barettes cannot be installed on a rake. BENTONITE SLURRY Bentonite slurry is a mixture of bentonite and water. Bentonite is a form of Montmorillonite clay. It is mined in various parts of the world including South Africa and is processed into a powder which is shipped in bulk or in 40kg sacks. Bentonite slurry is a mixture of approximately 5 percent of bentonite powder by mass with water. It has to be mixed in a specially designed mixer as the powder tends to float on water like talcum powder. The mixed slurry has an Specific Gravity of about 1.04. A bentonite slurry has various unusual qualities the main one being that it is a thixotropic liquid. This means that the clay particles do not settle out of suspension with time. Sand and other heavy particles mixed with the bentonite slurry during the excavation process are also held in suspension and do not settle to the bottom of the excavation. The bentonite slurry when maintained at a positive head of at least 1.5 metres relative to the water table builds up a layer of clay particles on the walls of the pile excavation. This layer which is normally 2 to 3 mm thick is known as the cake and is relatively impermeable. The cake coupled with the positive head is sufficient to prevent collapse of the sidewalls of the pile excavation which is the main advantage of using bentonite slurry . Bentonite slurry is mixed on site in a mixing plant and stored in large tanks. It is fed to the pile excavation during the drilling operation and is pumped back to the storage tanks during the concreting operation. Sand becomes mixed with the slurry during the drilling and the sand content can be as high as 30 percent by weight. Prior to concreting this should be reduced to less than 3 percent. De-sanding is achieved by means of cyclones and vibrating sieves.

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During the pile installation process the viscosity and the pH of the bentonite slurry may be altered by the drilling, the cement in the concrete and the ground water. These properties are routinely checked during the piling operations and various chemicals are .added to maintain the bentonite slurry within the specification. If the pH of the slurry rises above 12.5 the slurry will flocculate and lose its properties. This contaminated slurry has to be removed to waste. INSTALLATION TECHNIQUE Auger Underslurry The auger rig is set up on the pile position and the mast is checked for verticality. A shallow excavation is formed after which a short length of steel casing is screwed into the ground by the auger machine. This casing is known as a starter casing and it is installed to protect the top of the pile excavation from collapse. The starter casing is filled with bentonite slurry after which the drilling of the pile excavation continues once the mast has been checked for position and verticality. During the drilling operation t4e excavation is kept full of bentonite slurry at all times. The drilling bucket and kelly displace bentonite slurry when they are lowered into the excavation and to overcome this a surge tank is created around the head of the pile so as to contain the rise in level. On reaching the founding level the excavation is cleaned out as best as possible with a cleaning bucket attached to the kelly. Further cleaning of the base of the excavation is achieved using an airlift pump the suction of which is moved across the full area of the base. This airlifting operation is also used to change the bentonite slurry over, with the contaminated slurry being pumped to the storage tanks for processing while clean bentonite is fed into the pile excavation. It is preferable to have clean bentonite in the excavation during concreting. On completion of the cleaning operation the steel reinforcing cage is lowered into the pile excavation. Wide concrete cover spacers are used to prevent the spacers penetrating into the soft sidewalls of the excavation. The pile shaft is then concreted using standard tremie techniques. Measurements of the level of the concrete in the excavation are taken regularly to check that the tremie is immersed in the concrete at all times. The pile should be concreted a minimum of one metre above the cut-off level to achieve some degree of compaction at the head of the pile. The installation sequence is shown in diagrammatic form in Figure 7.7.1. Some aspects of Auger Underslurry piling operations are shown in Plates 7.7.1 to 7.7.5.

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AUGER UNDERSLURRY PILE INSTALLATION SEQUENCE

Figure 7.7.1 A

A short length of starter casing is drilled into the ground on the pile position. The starter casing is filled with bentonite slurry and drilling commenced.

B

The drilling continues until the founding level is reached, the level of the slurry in the excavation being maintained at all times.

C

An airlift pump is lowered into the excavation and is used to clean the base of the pile and to pump the contaminated slurry to storage tanks for treatment.

D

The reinforcing cage is placed in position and the pile shaft concreted using a tremie pipe. Displaced bentonite slurry is pumped to storage tanks.

E

The concreting of the pile shaft is carried out in one continuous operation.

F

The completed pile after the starter casing has been extracted.

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Barettes A Barette is constructed in a very similar manner to that described above except that the excavation is carried out using a grab and not an auger rig. The grab is normally of the cable or hydraulic type and excavates a rectangular hole. The width of the hole can be varied by changing the jaws on the grab. The depth of excavation is virtually unlimited. To achieve penetration into material of rock consistency, chisels can be used to break up the rock for removal by the grab. This is a costly exercise, however, and normally the piles are founded on the rock and not socketed into rock. Excavations using a grab can be carried out in multiple passes in the same manner as that for a diaphragm wall. In this way the shape of the excavation can be extended and altered. These larger units are referred to as load bearing panels and can be arranged to form part of a diaphragm wall as well as performing a load bearing function. The grab for excavating Barettes is shown in Plates 7.7.6. VARIATIONS IN INSTALLATION TECHNIQUE Integrity Testing As the whole pile installation is carried out under bentonite slurry some form of integrity testing is often considered desirable. Rotary core drilling can be used to check the contact between the concrete and the rock. The integrity of the pile shaft itself can be checked with either nuclear or sonic methods as described in SECTION 9.0 PILE LOAD AND INTEGRITY TESTING. To enable these tests to be carried out three or four small diameter steel tubes are cast into the pile. These tubes are normally fixed to the reinforcing cage and lowered with the cage. Composite Steel Pile Shafts In Europe a novel building technique is sometimes used to accelerate the construction of multi-storey buildings with basements as well as to reduce movements associated with basement excavation. In essence this technique calls for the construction of the basement of the building from the ground floor downwards while at the same time constructing the building itself from the ground floor upwards. The saving in time with this form of construction is considerable. If the building is founded on piles then a variation of the Underslurry system is used to make this revolutionary building system possible. This variation involves the use of large steel column sections as pile shafts over the depth of the basement. The advantage gained by this variation is that brackets, cleats etc. can be welded to the columns to allow the floor beams to be joined to the columns. Each floor level is completed as excavation proceeds downwards to form the bracing between the surrounding basement walls.

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These steel columns are cast into the concrete pile shaft with suitable shear connectors. Special adjustable guiding devices have been developed for the accurate positioning of the steel column sections in the excavation. This is necessary so as to ensure that the prefabricated steelwork fits at each floor level. The system has not been used in South Africa to date but the expertise is available within the Franki Group should there be a suitable project where the savings in construction time are meaningful to the client. POTENTIAL PROBLEM AREAS Collapse of the Pile Shaft Excavation There is a risk of collapse of the sidewalls of the excavation but this can be reduced to a minimum by observing correct procedures and ensuring that the bentonite slurry itself is within the specification and that the differential head is not less than 1.5 metres at any stage. A minor localised collapse can occur occasionally but this should not be detrimental to the pile. A major collapse of a pile shaft excavation is serious and the quick fix is to backfill the excavation as quickly as possible. The cause of the collapse should be determined before any further piling work is carried out. Additional care should be exercised with long panel excavations, excavations in extremely soft clays and silts and excavations carried out in close proximity to existing loaded foundations.. Delays Before Concreting Ideally a pile should be concreted the same day as its excavation is completed. If there are delays before the concrete is placed the bentonite cake tends to increase in thickness and this has been found to be detrimental to the pile's friction capacity. If a delay is unavoidable then the pile excavation should be reamed out again using the excavation tools prior to concreting the pile. Tremie Concreting This is an operation which needs a large amount of experience on the part of the contractor for it to run smoothly and produce a sound pile shaft. Unfortunately even the best trained crews can run into problems especially if the concrete itself is not one hundred percent. The most likely problem is a blockage in the tremie. This requires the tremie to be removed and cleaned out. If only a small amount of concrete has been discharged into the pile the best plan of action is to remove the steel reinforcing cage and permanent liner and drill out the wet concrete. Once cleaned out the steel cage can be installed and the pile concreted.

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If a large amount of concrete has already been placed when the blockage occurs the tremie should be removed and cleaned out. A water tight end cap should be placed on the toe of the tremie and latter lowered until it penetrates the wet concrete by at least two metres. In this operation, as with all tremie operations, it is vital that the joints in the tremie pipe are absolutely watertight. This can be checked by shining reflected sunlight down the tremie pipe using a mirror. The tremie is then filled with concrete and raised slightly at which time the end cap comes off due to the weight of the concrete. Thereafter the concreting operation can proceed as normal. Re-starting a tremie operation is a tricky process. For this reason the end result should be checked using one of the integrity testing methods. The alternative is to abandon the pile and install a replacement pile. Concrete bleeding For the concrete to flow through the tremie it has to have a slump of about 200 mm. To obtain this slump without having excess water in the mix is extremely difficult if not impossible with the available aggregates and admixtures. On deep piles the excess water in the concrete bleeds out and makes its way up through the concrete to the surface. The channel or run it forms in so doing is normally found near the centre of the pile where the tremie tends to form a zone of excess mortar. The concrete in this zone often shows signs of excess water and may have had some of the cement washed out of it. This is a problem which cannot be completely eliminated with deep piles cast using a tremie but certain measures can be taken to ensure it is reduced to a minimum. By adopting suitable techniques the percentage area of the pile shaft affected in this manner can be kept low so that the overall load bearing capacity of the pile is not affected. Before commencing the contract the gradings of the available aggregates should be determined and trial mixes designed and tested to arrive at the optimum mix design. During the contract there should be adequate site control to ensure the mix is stringently adhered to and in particular that no additional water is added to the mixer trucks while they are on site.

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Plate 7.7.1 Drilling under bentonite slurry

Plate 7.7.2 Bentonite slurry mixing, storage and de-sanding equipment

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Plate 7.7.3 Testing of bentonite slurry

Plate 7.7.4 Tremie concreting operation

Plate 7.7.5 Typical Underslurry Contract

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7.8

CONTINUOUS FLIGHT AUGER (CFA) PILE

This piling system is a fast and economical one which has no vibration and limited noise levels associated with it. Unfortunately it also has some strong negative features and it is problems associated with these that have reduced the popularity of the system. Positive features • • • •

High productions levels are attainable in suitable soil conditions. The system is economical in suitable soil profilesNoise levels are low and limited to the engine noise of the piling rigs. There is no vibration associated with CFA pile.

Negative features • • • • • •

The manufacture of the pile is largely in the hands of the operator and there needs to be adequate monitoring of concrete/grout flow and pressure as well as the extraction rate of the flight so as to be completely satisfied that the piles are made correctly. In sandy soils below the water table there is a reduction in soil strength in the immediate vicinity of the pile due to the drilling operation and as a result of this the load/deflection performance of a CF A pile is inferior to that of a driven pile. The steel reinforcing cage is inserted into the wet concrete/grout after the pile is cast. Soil falling off the flight can contaminate the concrete/grout of the pile shaftCFA piles have to be cast to ground level. This results in waste of concrete and excess trimming if there are deep cut-off levels. There is very limited indication of soil strength during the drilling operation.

PILE DETAILS Pile Diameter mm 300 350 400 450 500 600 750

Working Load kN

Min. Pile Spacing mm

350 450 600 800 1000 1400 2200

750 875 1000 1125 1250 1500 1875

Typical Reinforcement Main bars 5 x 12 mm. 4 x 16 mm. 4 x 20 mm. 5 x 20 mm. 4 x 25 mm. 5 x 25 mm. 8 x 25 mm.

The maximum depth of CF A piles is 22 metres at present and the recommended maximum rake 1: 10. The above pile working loads are calculated using a shaft stress of 5 MPa. Where piles are founded on rock consideration can be given to increasing the shaft stress up to 6 MPa. The pile's tension capacity must be determined from a calculation of friction.

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INSTALLATION TECHNIQUE The piling rig used to install CFA piles has the means to rotate, lift and lower a hollow stemmed continuous flight auger which is the main mechanical feature of the system. The hollow stem of the auger is blocked off at the toe by means of a suitable plug prior to the flight being lowered onto the pile position. The leader of the piling rig is adjusted for verticality or rake. The flight is rotated and at the same time allowed to penetrate into the soil. The penetration rate has to be controlled by the rig operator so as not to stall the auger head. On the other hand too slow a penetration rate will result in excess soil being removed with resulting de-compression of the soil profile which is undesirable. During the drilling operation the only indication of the strength of the soil profile is gained from a measure of the torque. This is not a sensitive measurement and only gives an overall indication which is insufficient for pile depth determination. CF A piles are thus normally installed to a predetermined depth. When this depth has been attained the flight is lifted slightly and the pumping of the concrete/grout commenced. The pressure of the concrete/grout blows the plug out the end of the flight and concrete/grout flows out to form the shaft. The flight is thereafter extracted at a controlled rate to match that of the flow of the concrete/grout, If extraction is too fast a necked pile shaft can be formed. If it is too slow then the flight tends to become stuck in the ground. Rotation of the flight during the concreting/grouting operation is normally kept to a minimum. It may be necessary to rotate the flight during the initial extraction stages but thereafter rotation is ceased. CFA piles are always cast to ground level so as to ensure an adequate head for compaction and for maintaining the correct pile diameter. Once the concreting/grouting operation has been completed the rig moves away from the pile position. The head of the pile is cleaned up and any latence and soil are removed to expose sound concrete/grout. The reinforcing cage is then lifted up and placed into the concrete/grout. Vibrators attached to the cage are sometimes used to assist in penetrating the cage into concrete. Figure 7.8.1 shows the CFA piling method in diagrammatic form. In some instances it may be desirable to use a sand/cement grout instead of concrete and in fact some contractors use grout in preference to concrete as it is easier to pump. Provided the mix attains the desired strength both are equally acceptable. When grout is used the water in the grout slowly migrates into the surrounding soil if it is sandy and this results in a drop in level of the grout at the head of the pile. Additional grout has to be added so as to maintain the cast level of the pile. This occurs to a much lesser degree with concrete. Plate 7.8.1 shows CFA piling operations in progress.

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CONTINUOUS FLIGHT AUGER (CFA) PILE INSTALLATION SEQUENCE

Figure 7.8.1 A

The hollow-stemmed continuous flight auger is drilled into the ground by means of the drive head. Penetration of the auger into the ground is slower than the combination of pitch and rotation so a certain amount of soil is ejected at the surface. This decompression is necessary so as to allow the auger to penetrate.

B

The auger is drilled down to the founding level after which the concrete/grout .pump is connected to the hollow-stemmed flight by means of high pressure hoses.

C

The concrete/grout is pumped down the hollow-stemmed flight as the latter is gradually withdrawn. A high level of control is necessary so as to ensure that the rate of extraction matches the rate of flow of the concrete/grout. If the extraction rate is too fast a necked pile shaft will be formed. If it is too slow there will be a pressure build up which forces grout up the sides of the auger which can lead to the flight becoming stuck.

D

The steel reinforcing cage is lowered into the wet concrete/grout in the pile shaft until it is at the correct level. This completes the installation sequence.

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VARIATIONS IN INSTALLATION TECHNIQUE Intermediate Bentonite Stage After drilling is complete the auger is withdrawn as bentonite slurry is pumped into the excavation through the hollow stemmed auger. The slurry prevents the hole from collapsing. A full length reinforcing cage can then be placed in position and the shaft concreted or grouted using a tremie. The advantage gained by using this variation is the fact that a heavy reinforcing cage can be placed right to the toe of the pile which may not be possible with the conventional technique. POTENTIAL PROBLEM AREAS Necked pile shafts Most of the problems experienced with this pile type in the past are those associated with the incorrect rate of extraction of the flight which results in necked pile shafts. It is important that adequate monitoring of the flow rate of the concrete/grout and the matching extraction rate of the flight is carried out to ensure a sound pile shaft. Reduction in pile capacity As mentioned above the drilling of the auger into the ground does result in de-compression of the soil. Overdrilling of the pile must be avoided or limited to a minimum as this causes further decompression. The power of the auger drive head also has a bearing on the degree of decompression with the more powerful heads giving better results as they have the power to penetrate the auger into the ground with limited decompression. Because of the decompression the ultimate unit friction transfer for CF A piles is normally about 60 to 75 percent of that for driven displacement piles. The end bearing component of CFA piles founded in sandy soils below the water table is significantly affected by the drilling operation. For this reason the end bearing is often ignored in these soil conditions and the pile is designed as a friction pile. Load tests have shown that in fact there is an end bearing component but it requires a pile toe movement of about five percent of the pile diameter for it to start to perform so in effect the pile acts as if it is a friction pile in the range of acceptable head deflections. CFA piles founded in or on materials which do not decompress such as stiff cohesive soils and rock have the full end bearing component and may be designed as such. Obstructions Such as Boulders The auger should penetrate boulders which are less than one third of the pile diameter and are not tightly packed. Larger boulders present a problem and the only solution is to move the pile position. The CFA system is regarded as being sensitive to obstructions and this should be borne in mind when considering its use.

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Plate 7.8.1 CFA rig drilling

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7.9

FORUM BORED PILE

The main feature of this system is the relatively small and light piling equipment used to install the pile. This makes the system suited to low headroom conditions and sites with limited or difficult access. Site establishment costs are low so the system is also suited to small contracts. Positive features • • • • •

Site establishment costs are low. The piling rig can operate in limited headroom and difficult access conditions. The pile has an enlarged base similar to the Franki system. The pile has good tension capacity. A rock socket can be formed as an alternative to the enlarged base.

Negative features • • • • • •

High cost per metre of pile installedLow production rates. System not suited to cohesionless soils below the water table. Only two sizes of pile available. Depth of piles limited to about 12 metres depending on soil conditions. Noise and vibration levels, whilst not high, can be a problem.

PILE DETAILS Pile Size Nominal diameter of pile shaft (mm) Typical working load (kN) Maximum depth (meters) Minimum pile spacing (mm) Maximum rake Standard main bar reinforcing Maximum main bar reinforcing Spiral reinforcing at 150 mm pitch Nominal cover to reinforcing steel (mm) Typical tension load (kN) OD of piling tube (mm) ID of piling tube (mm) Typical hammer mass (kg)

410 Diam. 410 600 12 1000 1:6 4x16mm 8x20mm 6mm 40 200 406 366 660

* Depends on the soil conditions. Greater depths are possible.

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600 Diam. 600 1200 12 1500 1:6 6x16mm 12x25mm 6mm 50 400 610 546 1600

INSTALLATION TECHNIQUE The plant used for the installation of the Forum bored pile consists of a tripod type mast with a winch and engine mounted on a frame fixed to one of the legs of the tripod. The system makes use of a temporary piling tube which is made out of one metre long screw- coupled sections. Various excavation tools are used for removing the spoil from the bore of the tube as well as for driving the tube. An internal drop hammer is provided for forming the enlarged base and a separate hydraulic jacking unit is used for extracting the tubes. Once the rig has been set up correctly the first section of casing is driven. The spoil is removed from this first section after which another section of tube is added by means of the screw coupling. The tube is again driven and the spoil removed. The cycle is repeated until the founding level is reached. After cleaning out the hole a small charge of zero slump concrete is discharged into the tube. The tube is raised slightly by means of the jacking unit and the concrete is expelled from the tube by blows of the hammer. Further charges of concrete are expelled in a similar manner thus forming an enlarged base of the correct size. Having ensured that all the concrete has been driven out of the tube the steel reinforcing cage is lowered into the tube which is then filled with high slump concrete. Using the jacking unit the tube is extracted with individual sections being removed as this operation continues. The tubes are washed clean ready for use on the next pile. With the tube fully extracted the cut-off level is checked and adjusted. The installation sequence is shown diagrammatically in Figure 7.9.1. Plates 7.9.1 and 7.9.2 show a typical Forum Bored Pile operation in progress. Plate 7.9.3 shows the rig operating in difficult access conditions. VARIATIONS IN INSTALLATION TECHNIQUE Rock Socket The system has the facility to form a rock socket as an alternative to the enlarged base. Various chisels are used to penetrate the rock which should have an unconfined compression strength of not greater than 10 MPa. Conventional tools are used for removing the spoil and cleaning out the socket. The remaining concreting operation is carried out as detailed above. Rock sockets are mainly used as anchorage against uplift. A typical example is an expansive clay profile overlying rock where a rock socket provides the resistance to uplift.

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FORUM BORED PILE INSTALLATION SEQUENCE

Figure 7.9.1 A

Driving the temporary casing which is made up out of one metre long screw coupled sections.

B

Coring out the soil from the bore of the casing using the coring tool.

C

Forming the enlarged base by driving out semi-dry concrete using the hammer. Hydraulic jacks prevent the casing from penetrating further into the ground.

D

The steel reinforcing cage is placed in position and the tube filled with high slump concrete.

E

The temporary casing is jacked out with each casing segment being removed.

F

The completed pile.

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Penetrating Obstructions The same chiseling technique used to form a rock socket can be used to remove obstructions. Because of the small tube size and limited weight of the chisel there is naturally a limit to the size and hardness of obstructions that can be penetrated. As a guide obstructions up to one tube diameter and a hardness of up to medium hard rock can be penetrated without undue difficulty. Permanently Sleeved Pile Shafts The installation technique can be altered to incorporate a permanent liner for the pile shaft. This is sometimes desirable in aggressive ground water conditions and the liner can be made out of chemically resistant materials. The liner is placed in position after the formation of the enlarged base. It is sometimes fixed to the reinforcing steel cage and both are inserted as a unit. The liner is filled with high slump concrete after which the tube is then extracted. The annulus between the liner and the piling tube is filled with a sand cement grout. . POTENTIAL PROBLEMS Boiling Sand and water are said to "boil" into a pile excavation when they flow in past the toe of the piling tube due to the differential head between the water table on the outside of the tube and the water level inside the tube. From this it is clear that boiling only takes place when a water table is present. Any suction effect caused by the excavating tools will encourage boiling to take place. If boiling takes place during excavation of the pile there is a risk of ground settlement around the pile position. Such settlement can undermine an existing foundation and needs to be avoided. To achieve this the temporary casing must be driven well ahead of the level of excavation inside the casing. Boiling can also seriously hamper the cleaning out of the pile prior to concreting. In sandy soil profiles below the water table it is important that there is a cohesive layer immediately above the founding level into which the tube can be sealed. The pile can then be cleaned out successfully and the base formed. If such a clay layer does not exist then a driven tube pile may be the better solution. .

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Plate 7.9.1 Boring in progress

Plate 7.9.2 Concreting operation

Plate 7.9.3 Forum bored piles in difficult access conditions

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7.10

OSCILLATOR PILE

The main feature of the Oscillator pile is it's ability to penetrate through rock and boulder layers and to socket into bedrock. These attributes make it an ideal pile for large river bridges and for marine construction. The pile diameters are relatively large and the range of pile sizes limited. The system is thus only economical when used under heavily loaded structures. A permanent metal liner can be incorporated into the pile and this feature is often specified on river bridges so as to protect the wet concrete of the pile shaft. Positive features • • • • • • •

Large pile sizes for large pile loads. The ability to penetrate substantial boulder layers and other obstructions. The ability to form rock sockets. The use of thin walled permanent liners. Noise levels are low and limited to the noise from the equipment. There is no vibration associated with Oscillator piles. Depths of up to 60 metres are possible.

Negative features • • •

It is an expensive pile type. A large working platform area is required. There is a limited range of pile sizes.

PILE DETAILS Thefo11owing table applies to unlined Oscillator piles. Pile Details Nom. Shaft diam. Unlined (mm) Typical working load Maximum depth (metres) Minimum pile spacing Maximum rake Typical main bar reinforcing Typical spiral reinforcing (mm) Nom. cover to reinforcing (mm) O. D. of piling tube (mm) I. D. of piling tube (mm)

1080 mm Diam 1080 6500 25 2700 1:4 15x25mm 8 75 1080 980

l200mm Diam 1200 8000 25 3000 1:4 15x32mm 8 75 1200 1100

1500mm Diam 1500 13000 25 3750 1:4 18x32mm 10 75 1500 1400

The above working loads are based on a shaft stress of7.5 MPa. The pile's tension capacity depends on the calculated value of friction. The maximum depth of 25 metres stated in the table is limited due to concreting problems with greater depths.

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The following table applies to lined Oscillator piles. Pile Details Nom. Shaft diam. Unlined (mm) Typical working load Maximum depth (metres) Minimum pile spacing Maximum rake Typical main bar reinforcing Typical spiral reinforcing (mm) Nom. cover to reinforcing (mm) O. D. of piling tube (mm) I. D. of piling tube (mm)

950 mm Diam 950 6500 60 2700 1:4 15x25mm 8 75 1080 980

1100 mm Diam 1100 8000 60 3000 1:4 15x32mm 8 75 1200 1100

1350 mm Diam 1350 13000 60 3750 1:4 18x32mm 10 75 1500 1400

The above working loads have been calculated using a shaft stress of 9 MPa. The pile's tension capacity depends on the calculated value of friction. INSTALLATION TECHNIQUE The equipment consists of an oscillator which has the ability to clamp the piling casing, to move it rotationally through about 15 to 20 degrees and to lower and raise it, all movements being achieved by the use of hydraulic rams. A crane equipped with various excavation tools is used for removing the spoil from the bore of the piling tube and for forming the rock socket. Once the oscillator machine has been set up on the pile position the initial casing which is fitted with a cutting edge is inserted into the oscillator and clamped. By oscillating the casing backwards and forwards and by raising the vertical rams the tube penetrates the soil under its own weight. Excavation of the spoil from the bore of the tube using a grab proceeds concurrently with casing penetration. The casings are fitted with mechanical joints which allow them to be joined together. The boring process continues with additional casings added as an when required. Penetration of boulder layers is achieved by means of large chisels in combination with the cutting edge of the piling casing. Once the bedrock is reached the casing is allowed to penetrate into the rock to form a seal against the possible ingress of running sand. The socket is formed by chiselling the rock and removing the chips by means of a suction baler. This same unit is used to finally clean the bottom of the socket prior to concreting.

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The reinforcing cage fitted with roller spacers is lowered into position. If the pile has a permanent liner this is normally attached to the reinforcing cage and the two are lowered as a unit. Splicing of rebar cages and permanent liners is necessary on the deeper piles. If the pile excavation is dry then the concrete is deposited into the pile by means of a chute which prevents the concrete from striking the reinforcing cage as it descends to the toe of the pile or the surface of the concrete. A minimum slump of 150 mm should be used and the mix should be designed to be self compacting. The head of the pile is normally vibrated to assist compaction. On river bridges the pile excavation is invariably full of water. The concrete has thus to be placed under water using a tremie pipe which has a diameter of about 200mm. The tremie is made up of different lengths joined together with a watertight screw thread coupling. A high slump concrete is used for casting the pile shafts as it has to flow down the tremie and then self compact. Some of the temporary pile casing can be removed during the concreting operation. Once the concreting of the pile shaft has been completed the remainder of the pile casing is withdrawn. Both the tremie pipe and the piling casings are washed thoroughly after each pile is cast. If the pile shaft does not have a permanent liner the depth that is concreted inside the temporary casing should be limited. The reason for this is that the concrete will tend to lose its workability with time due to the pressure of the head of concrete and drainage paths at the toe of the casing and at the casing joints which allow the water in the concrete to drain away. Concrete which has lost its workability will tend to arch in the pile casing during extraction which could cause defects in the pile shaft. Experience has shown that with a depth of about 25 metres the pile can be successfully concreted without these defects. Retarders and plasticizers have limited benefit in this situation but their use is still recommended. Aspects of the installation process are shown in Plates 7.10.1 to 7.10.3. VARIATIONS IN INSTALLATION TECHNIQUE Integrity Testing It has become common practice to carry out various integrity tests on these piles as they are heavily loaded and conventional load testing is very expensive. The contact between the base of the pile and the rock is normally checked with rotary core drilling. The integrity of the shaft itself is checked using either nuclear or sonic methods as described in SECTION 9.0 PILE LOAD AND INTEGRITY TESTING. To enable these tests to be carried out three or four small diameter steel tubes are cast into the pile. These tubes are spaced equidistant around the perimeter of the pile and are normally fixed to the reinforcing cage and lowered with the cage.

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POTENTIAL PROBLEM AREAS Stuck Casing There is a slight risk of this occurring on deep piles especially if the diameter is 1500mm. To reduce the risk the casing should be kept moving so as to lubricate the outside and reduce friction. Stuck Chisels Chisels used in this type of work are large and heavy. They are dropped a couple of metres so as to break into the rock. They sometimes get wedged in the casing or in the socket itself. It may take a few days to dislodge the chisel but occasionally this may prove impossible. If this occurs the pile will have to be abandoned, the casing filled with sand and withdrawn. A revised pile layout will have to be designed and the pile installed on an alternate position. Tremie Concreting This is an operation which needs a large amount of experience on the part of the contractor for it to run smoothly and produce a sound pile shaft. Unfortunately even the best trained crews can run into problems especially if the concrete itself is not one hundred percent. The most likely problem is a blockage in the tremie. This requires the tremie to be removed and cleaned out. If only a small amount of concrete has been discharged into the pile the best plan of action is to remove the steel reinforcing cage and permanent liner and to dig out the wet concrete using an excavating grab and chisel. Once cleaned out a new steel cage and liner (the old one will most likely be damaged) can be installed and the pile concreted. If a large amount of concrete has already been placed when the blockage occurs the tremie should be removed and cleaned out. A water tight end cap should be placed on the toe of the tremie and latter lowered until it penetrates the wet concrete by at least a metre. In this operation, as with all tremie operations, it is vital that the joints in the tremie pipe are absolutely watertight. This can be checked by shining reflected sunlight down the tremie pipe using a mirror. The tremie is then filled with concrete and raised slightly at which time the end cap comes off due to the weight of the concrete. Thereafter the concreting operation can proceed as normal. Re-starting a tremie operation is a tricky process. For this reason the end result should be checked using one of the integrity testing methods. The alternative is to abandon the pile and install a replacement pile.

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Plate 7.10.1 Grab excavating a pile Plate 7.10.2 Concreting in progress

Plate 7.10.3 Typical Oscillator contract

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7.11

CAISSON PILES

This is no longer a widely used piling system because of the high cost of the permanent casing. It has been included in this text as it has been popular in the past and it can still provide a solution to a unique problem. . Positive features • • •

Considerable depths can be achieved (> 65 metres). Risk during concreting is lower. Larger range of pile sizes.

Negative: features • • •

Expensive system due to high cost of casing. System only suited to heavy loads due to large pile diameters. Risk of casing damage in boulder layers and on bedrock.

PILE DETAILS Daim. of pile shaft (mm) Typical working load (kN) Maximum depth (metres) Minimum pile spacing (mm) Typical main bar reinforcing Typical spiral reinforcing (mm) Cover to reinforcing (mm) Casing wall thickness (mm)

1000 4700 65 2000 12x25mm 8 75 10

1200 6800 65 2400 18x25mm 10 75 12

1350 8600 65 2700 12x32mm 10 75 12

1500 10600 65 3000 18x32mm 10 75 16

The above nominal working loads are based on a shaft stress of 6 MPa. This can be increased to as high as 10 MPa provided the piles are socketed into rock and the concrete strength is adequate. The quoted pile diameters are typical and may vary from the above as the casings can be made to any diameter. The range of sizes should, however, be within that given. The maximum depth of installation will depend on the soil conditions. Depths greater than 65 metres are possible. Heavier wall thicknesses are required for the deeper piles. INSTALLATION TECHNIQUE There are two methods for installing the casing: driving it or installing it with an oscillator machine. As the latter is covered under Oscillator Piles the driving technique will be covered here.

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The pile driving rig is set up on the pile position. The casing is lifted into the leader of the piling rig and a tube guidance mechanism is attached to the tube. A driving helmet is lowered onto the top of the casing and the piling hammer in turn rests on the helmet. The hammer is normally a diesel hammer. When the position and rake have been checked and adjusted, driving commences and continues until the top of the casing is about a metre above the ground. Plate 7.11.1 shows the casing being driven on a river bridge site. The spoil from this first section of casing is then removed using a grab after which a second length of casing is welded onto the first section. The cycle of drive, excavate and weld is repeated until the founding rock stratum is reached. A chisel is then used to ensure that the full pile cross-section is on rock by penetrating the rock a small amount. The casing is often re-driven to ensure that there is a seal between the rock and the casing. If a full rock socket is required this is normally achieved by chiselling with a heavy chisel while the rock chips are removed using a suction baler. The base of the pile is cleaned using a suction baler or an air lift. The steel reinforcing cage is then lowered into position followed by the tremie pipe. The concreting of the shaft by means of the tremie is carried out in the normal way. There is an alternate method for casting these piles. This involves casting a concrete plug in the bottom of the pile casing. When this has set the pile is dewatered and any latence on the top of the concrete plug cleaned off. The steel reinforcing cage is then inserted and the pile is concreted in the dry using trunking to guide the concrete to the bottom without segregating on the steel cage. In some instances the pile may be able to be dewatered without the need for a plug as the casing has been sealed completely on or into the rock. This is often the case where there is a layer of decomposed rock overlying the founding stratum. VARIATIONS IN INSTALLATION TECHNIQUE Sockets in Extremely Hard Rock Penetrating extremely hard massive rock to form a socket can be very expensive. Attempts to use explosives have not been very successful mainly due to the fact that this causes shattering of the rock which weakens it and forms fissures which can allow the ingress of sand. The alternative is to eliminate the socket and install some large steel dowels in place of the socket. In this way tension transfer can be achieved without shattering and weakening the rock and at lower cost. The dowels can also be installed after the shaft of the pile has been cast thereby allowing the main piling plant to move on to the next pile.

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POTENTIAL PROBLEM AREAS Damaged Casings If the casings are driven there is a risk that the casing wall may buckle. The risk of this happening in a soil profile is very low unless the driving is very hard. When the leading edge of casing, however, intersects an obstruction such as a boulder there is a great risk of damage. The same applies when the leading edge comes into contact with the uneven surface of the founding stratum. For this reason the leading edge of the pile casing is normally strengthened to reduce the risk of damage. Even then the piling foreman has to monitor the driving of the casing extremely carefully to avoid serious damage to the casing. There is much less risk of casing damage if the casings are installed using an oscillator and this technique should be used if there are obstructions present in the profile. The wall thickness of the casing, however, must be adequate to withstand the torque of the machine. Casings can also be damaged due to collapse induced by a large differential water pressure if the casings are dewatered. If dewatering is envisaged then the casing wall thickness should be calculated accordingly. Sand Ingress In river bridges the rock of the founding stratum has often been exposed to scour and has a clean hard surface. Furthermore there is often clean river sand overlying the rock. A problem can arise with sealing the casing against the rock to stop the inflow of sand. This problem can be exacerbated by the rake angle of the pile and a steeply sloping rock face. Excavation under a head of bentonite slurry may help solve the problem. Forming the cutting edge of the casing to approximate that of the rock face has also been resorted to in the past for driven casings. A sophisticated cutting edge using tungsten inserts and the oscillator for rotation is probably the best solution for ensuring adequate penetration for creating the seal. The chiselling of the base rock can also cause fracturing of the rock beyond the casing. Fissures thus formed can provide a path via which sand enters the pile socket. The risk of undetected loose material at the pile toe must be avoided and inspection by a diver should be undertaken before concreting the pile shaft if there is any doubt.

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Plate 7.11.1 Caisson pile installation on a river bridge

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8.0

UNDERPINNING

The causes of cracking or other forms of distress in a building structure are usually due to either differential movement of the foundation or differential shrinkage or temperature movements in the structure. Differential movement of the foundations can either be caused by uneven settlement of the foundation or by differential ground heave under the foundation. Before remedial measures are prescribed to stabilise foundation movement on a damaged structure it is essential that an adequate investigation is carried out to ascertain the nature of the foundation movement and in particular whether the movement is due to settlement or heave. If the differential movement is due to varying degrees of ground heave then the underpinning systems described in this section will not provide a complete solution to the problem. In fact it is debatable whether an underpinning solution is at all applicable in these circumstances. If the differential movement is due to uneven settlement of the foundation then an underpinning system is a feasible way in which to arrest further movement and stabilise the structure. Before deciding which underpinning system is best suited to solving the problem, the reasons for the settlement should be determined as well as details of the soil profile over the site. The history of the site coupled with knowledge of the local geological conditions can also assist in evaluating the cause of the settlement. When the investigation is complete, consideration must be given to which underpinning methods could be used and which is likely to be the most economical. There are six basic underpinning methods that should be considered as set out below. 8.1

Providing temporary support to the structure, removing the existing footing and replacing it with a new footing with greater bearing capacity followed by the removal of the temporary support.

8.2

Excavation under sections of the existing foundation, constructing a new footing with better bearing capacity and ensuring load transfer to the new footing, with or without preloading.

8.3

Excavation under the existing foundation, jacking in piles and ensuring load transfer to the new piles, with or without preloading.

8.4

Installing piles alongside the foundation and casting a new section of footing which is keyed into the existing foundation.

8.5

Constructing a new foundation with or without piles and transferring load to this foundation by means of an additional column or similar structural member.

8.6

Installing piles through the existing foundation and keying the heads of the piles to the existing foundation.

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FACTORS INFLUENCING THE CHOICE OF UNDERPINNING SYSTEM Which of the above methods is best suited to the problem on hand depends on a number of factors. The following is a list of the more important factors together with comments as to how these influence the choice of underpinning method. Soil profile It is necessary to know the soil profile to understand why the foundation movement is occurring in the first instance. Once this understanding has been obtained and a decision has been made that underpinning is the correct solution, the soils information is then vital in deciding at what level any remedial foundation should be founded so as to achieve the objective of preventing further settlement. If the use of piles is considered then detailed soils information will be required to assess the size and founding level of the piles, which in turn can influence the choice of pile type. Load on the foundation For basic design calculations it is essential to have a good estimate of the load on the foundation. Structural details of the existing foundation It is important to have this information so that the effect of any structural changes to the foundation can be checked. The use of piles along the perimeter of a footing could, for example, increase the bending moment in the footing and the reinforcement needs to be checked for this increased moment. Structural details of the existing superstructure Where additional columns and/or structural elements are envisaged it will be necessary to have these details. Level of the water table If excavation below the foundation is planned then it is important to determine the level of the water table. A shallow water table might well preclude excavation beneath the existing foundation due to risk of collapse and the difficulty and high cost of shoring in these conditions. Access If any underpinning work is planned it is important to check that the workmen with their equipment can gain access to and around the foundation so that the construction procedure can be adhered to. The presence of services should also be checked and taken into account in the planning.

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Available headroom This is important, particularly if a piled solution is being considered. Inconvenience Inconvenience in the form of noise, vibration, security, site pollution and limited use of space and may affect the public, the landlord and the tenant. Due consideration must be applied to these aspects when choosing a method and pile type for underpinning. How critical further settlement would be There is a difference in performance between an underpinning method that employs preloading and one that does not. If pre-loading is not carried out then additional settlement can be expected during load transfer to the new foundation. It will be necessary to assess what additional settlement will take place and whether this additional movement will be detrimental to the structure. Whether it is necessary to raise the footing that has settled On some occasions it may be considered desirable to raise the level of the footing which is being underpinned. The normal requirement is to raise the foundation back to its original level or even slightly above it. It is a high risk operation, however, and one that should only be specified in exceptional circumstances. Careful planning and a high level of site supervision are needed for it to be successful. Cost of the underpinning system As with all forms of construction the cost is very important. Detailed cost estimates need to be carried out to determine the most economical system. Programme In some cases it may be that time is more important than cost. This can be especially true if an owner is loosing rental because the tenants cannot occupy leased premises or their trade is badly affected by the underpinning operations. 8.1

OLD FOUNDATION REMOVED AND NEW FOUNDATION PROVIDED

To be able to do this it will be necessary to provide temporary support to the load bearing member which is being underpinned. This can be achieved by fixing some brackets to the sides of the member and spreading the load through beams to stub columns supported on temporary footings or piles. This is illustrated in Figure 8.1.1.

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Once the load is temporarily supported the member can be cut through below the support point and the existing foundation removed. The new foundation is then constructed with a stub column connecting to the existing one. The temporary support is then removed. This form of underpinning is not common. 11 is also expensive because of the necessity to provide temporary support and remove it afterwards. 11 is thus only used in exceptional circumstances. 8.2

NEW FOOTING LOCATED UNDER THE EXISTING ONE

This is a more common form of underpinning and one which is considerably less expensive than that described in SECTION 8.1. 11 is used where the water table is low enough to allow excavation under the existing foundation and where adequate bearing capacity can be obtained at a depth of one to two metres below the soffit of the existing foundation. The method involves an initial excavation in a narrow pit alongside the existing foundation. This pit must be excavated down to the anticipated new founding level and must be just large enough to allow reasonable working space. A narrow trench is then excavated from the pit extending under the foundation and down to the founding stratum. The base of this excavation is cleaned up and a reinforced concrete footing cast. A stub column of concrete or load bearing brickwork is built on top of the footing, extending up to just under the existing foundation. The gap between the top of the column and the soffit of the existing foundation is filled with rammed sand/cement grout. The pit is finally backfilled. The method is illustrated in Figure 8.2.1. If pre-loading of the base is considered desirable, a gap large enough for an hydraulic jack is left between the top of the column and the soffit of the existing foundation. A pair of hydraulic jacks is positioned, one on either side of the column. The jacks are extended and a load of up to 50 percent in excess of the design load is applied to the column via the jacks. The level of the existing foundation is monitored to ensure that it is not lifting excessively and if so the load is adjusted accordingly. Once the .load has been held on for say 30 minutes, the load is reduced to between 100 and 125 percent of the working load and a fabricated steel stub column is placed in the gap between the jacks. Steel shims are used to finally ensure that the steel stub column is fixed firmly between the main column and the soffit of the existing foundation. The jacks are released and steel stub column is encased in concrete to protect it. The final operation is the backfilling of the pit. There are normally a number of underpinning points located at various positions under the structure. These are normally spaced at between two and four metre centres, the spacing depending mainly on the structural integrity of the wall and its footing to span between points and the design load of the new footing relative to the applied load from the structure.

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Figure 8.1.1 Old foundation removed and new foundation provided

Figure 8.2.1 New footing located under existing one.

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Before carrying out the underpinning it is necessary to plan the sequence of work. The more heavily loaded points should be installed first and adjacent points should not be tackled simultaneously. Accurate bench marks should be established at various points on the structure and these should be checked regularly during the underpinning operation. 8.3

JACK PILES UNDER THE EXISTING FOUNDATION

In cases where an adequate founding level is greater than about 1.5 metres below the soffit of the existing foundation consideration must be given to installing piles instead of constructing a new footing at great depth. A common way to install piles directly under the existing foundation is to jack them in short sections using an hydraulic jack and the load of the structure as reaction. A steel tube pile is commonly used for this purpose as the steel tube can be readily cut into short sections or pile elements and then welded together again as the pile is jacked into the ground. Precast elements have also been used as pile elements for underpinning but this is no longer common practice. The first step of this method involves the excavation of a narrow pit alongside the existing foundation. The size of the pit should be just large enough to allow working space. The depth of this pit should be about 1200 mm below the existing foundation. This allows for a 400 mm long jack and a 500 mm long pile element with the welded joint about 300 mm above the ground. The tube pile elements are normally jacked in open ended in which case the soil will tend to push up inside the tube during excavation. This soil is not normally removed as it is difficult to do so. In such instances the steel tube forms the structural body of the pile. The process is illustrated in Figure 8.3.1. In softer soils consideration should be given to welding a plate on the end of the first pile element. This will increase the pile's bearing capacity as well as leave the bore of the tube empty so that it can be filled with concrete or grout. The pile is normally jacked to refusal under a load in excess of the desired working load. This proof load can be up to 50 percent more than the required working load but this must be adjusted during the underpinning operation depending on the monitoring of the movements of the existing foundation. The proof load should be held on the pile for at least 30 minutes or longer if movement of the pile is detected. When the pile has been installed the tube is trimmed to the correct elevation and a steel plate is welded on to form a top bearing plate. A short section of steel RSJ is placed on top of the bearing plate and the jack in turn positioned on top of the RSJ. This is illustrated in Figure 8.3.2. The jack is extended and a load of 125 percent of the working load applied to the pile.

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Figure 8.3.1 A pile being jacked under an existing foundation

Figure 8.3.2 Final proof loading of a jacked pile

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Two short steel column sections are then placed one either side of the jack between the RSJ and the soffit of the existing foundation, shims being used to ensure a tight fit. The load on the jack is released and the load of the structure is transferred via the steel columns and RSJ to the pile. The steel columns and RSJ are normally encased in concrete to protect them. The final operation is the backfilling of the pit. There are often a number of underpinning points located at various positions under the structure. These are normally spaced at between two and four metre centres, the spacing dependent on the structural ability of the wall and its footing to span between points and the capacity of the underpinning pile. Before carrying out the underpinning it is necessary to plan the sequence of work. The more heavily loaded points should be installed first and adjacent points should not be tackled simultaneously. Accurate bench marks should be established at various points on the structure and checked regularly during the underpinning operation. 8.4

PILES ALONGSIDE THE EXISTING FOUNDATION

This can be a more economical way in which to provide a piled underpinning solution than the jack pile system described above in SECTION 8.3. It is also a quicker method than the jacking system and piles of much greater capacity can be installed and to greater depths. Most of the pile types listed in SECTION 5.0 can be used with this method and the choice of pile type will be influenced by the factors enumerated in SECTION 4.0. In most cases, however, it is the limited access, working area and headroom constraints which will dictate which pile type can be installed. These restraints can be very severe and as a result special piling rigs and systems have been developed for the purpose. An example of this is shown on Plate 8.4.1. This little rig, which is mounted on the chassis of a wheelbarrow, installs a driven micropile in the form of a steel tube 100 mm in diameter. It can manoeuvre through doorways in a house and into very narrow areas such as between the toilet and the bath. Another small rig used for this type of work is the conventional diamond drilling rig which can be used to install a drilled micropile of 100 mm diameter. There are many ways in which the load can be transferred onto the piles. One of the simplest methods is shown in Figure 8.4.1(a) and involves a short cantilever, tied and keyed into the existing footing. Another common method involves constructing a pile capping beam under the existing foundation as shown in Figure 8.4.1 (b). The construction of a completely new pile cap over the top of the existing footing as shown in Figure 8.4.1 (c) is yet another method. The combination of the correct choice of size and type of pile together with one of the above methods for transferring the load to the piles can be used for a very wide variety of underpinning problems. The method is suited to the underpinning of a house foundation at the lower end of the scale and to underpinning a large bridge at the upper end.

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8.5

NEW PILED FOUNDATION AND COLUMN

There are occasions when the need to underpin is due to the fact that additional load is to be applied to the foundation and not because settlement has occurred. A typical example is the addition of another floor or two to an existing building. Any of the underpinning techniques covered in this section could be used but sometimes the structure itself also has to be strengthened. In these circumstances a complete new foundation with or without piles as well as a new column are constructed. This can be difficult work as low headroom and limited access restraints can place serious limitations on the size of piling rig that can be used and this also limits the choice of pile type. Here again unconventional solutions may have to be resorted to. An example is driving steel tube piles through the roof of a single storey building with a large crane. Another is the driving of 3 metre long sections of mechanically jointed precast pile with a crane especially adapted to operate in a 6 metre headroom. Solutions to problems of this nature are best discussed with one of Franki ' s Engineers and solutions developed for the particular application. 8.6

PILES THROUGH EXISTING FOUNDATION

If a pile can be installed through an existing foundation and subsequently keyed into it then there is the possibility that no additional structural concrete work will be required. The hole through the foundation can be cut with a diamond coring tool. The pile itself can be either driven or bored. Because of the task of drilling through the existing foundation and the fact that its structural integrity could be impaired, the size of hole and thus the size of the pile is normally limited. Piles with a diameter of up to 200 mm are used although micropiles with a diameter of about 100 mm are more common. The micropile rig shown in Plate 8.4.1 is often used for installing these piles. The structural transfer of load from the foundation to the pile is achieved by grouting up the hole in the foundation using an expanding grout once the pile has been installed and trimmed to the correct level. With shallow footings it is also possible to form an enlargement of the head of the pile so that it bears on the soffit of the footing. This is illustrated in Figure 8.6.1. This technique is fast and economical but is generally suited to light structures only. PILES TYPES USED FOR UNDERPINNING Any of the eleven pile types listed in SECTION 5.0 could be used for underpinning. Factors such as those listed in SECTION 4.0 will assist in determining the choice of pile type. The steel tube pile, however, is a frequent choice for underpinning as its features meet most of the requirements for underpinning. Special pile types developed for underpinning such as the drilled micropile and the jacked Mega pile can also be considered. Underpinning is an area suited to innovative solutions and a discussion with your local Franki office could be well worthwhile.

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Figures 8.4.1 (a), (b) and (c) Piles alongside the existing foundation

Figure 8.6.1 Piles through the existing foundation

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Plate 8.4.1 Miniature piling rig for driving micropiles

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9.0

PILE LOAD AND INTEGRITY TESTING

9.1

PILE LOAD TESTING

The load testing of piles is a well established practice and is often specified on medium and large piling contracts. The most common form of load test is the static compression test in which a load is gradually applied to the head of the pile while the deflection of the pile head is monitored. Static test loading can also be carried out in tension as well as laterally. Piles can also be tested using dynamic or semi-dynamic load testing procedures but these are not in present use in Southern Africa. The main objectives in carrying out a load test are as follows: • To verify that the pile's load/deflection performance meets the contract specification. • To verify the assumed pile design parameters. • To establish the pile's load/deflection characteristics. • To assess the pile's ultimate capacity, if possible. • To study the pile -soil interaction. • To check the structural integrity of the pile shaft. STATIC LOAD TESTING Static load testing of piles can be carried out on working piles or on trial piles specially installed for the purpose. Most pile load testing is carried out on working piles as the cost of installing additional piles for testing purposes is prohibitive on all but the very large contracts. Working Piles When load testing a working pile the load has to be limited so as not to damage the pile. For this reason the maximum test load is normally limited to one and halftimes the design working load. In most cases this can be increased to twice the design working load without risk of damage to the pile provided the load test is set up and executed in the correct manner. Tension and lateral load tests are normally limited to one and a half times the design working load. The principle reason for carrying out a load test on a working pile is to check the pile's performance compared to that specified in the contract documents. For this reason it is common practice for load tests on working piles to be part of the piling contract. The secondary benefits to be gained from such a test are the confirmation of the pile design parameters and a check on the pile's structural integrity. More economical means of testing pile integrity have been developed, however, and these are described in SECTION 9.2. The load test can also highlight whether installation problems, such as pile heave, have had any effect on the pile's performance.

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Trial Piles Trial piles are specially constructed for the purpose of carrying out load tests and thus they can be loaded to failure. Testing a pile to its ultimate load capacity provides more accurate and meaningful design data which can result in achieving further economies in the foundation design. For structures sensitive to settlement the results from an ultimate test can provide the basis for a more accurate specification of working load settlement acceptance criteria. On large contracts the substantial cost of trial piles and an extensive load testing programme can often be recovered many times over through achieving economies in design. A trial pile programme also offers the opportunity to install more than one pile type and to compare their performance. This will assist in deciding which pile type will provide the most economical solution. Trial piles can also be installed to different depths so that the optimum founding level can be confirmed. Special instrumentation in the form of strain gauges cast into the pile shaft at various levels can also increase the amount of data that can be obtained form these tests. STATIC LOAD TESTING PROCEDURES Two compressive load test procedures are outlined in detail in SABS 1200F. These are termed the British and Danish procedures and both describe a series of test load cycles in which the pile is loaded gradually in increments and then unloaded in a similar way. Most specifications call for the use of the British method or a variation thereof as the Danish procedure is very time consuming. A common procedure which is satisfactory for most soil profiles is a variation of the British method. This involves a first cycle in which the load is increased in 25 percent increments up to the design working load, held for 12 hours and then unloaded in 25 percent increments back to zero. The intermediate load increments are maintained until two successive readings 30 minutes apart show that the head deflection has not changed by more than 0.1 mm. The load is kept at zero for a period of one hour after which the deflection is checked and the second load cycle is begun. The second cycle is similar to the first but the maximum load is 1.5 times the design working load. After unloading from the second cycle, the residual deflection is monitored over a 12 hour period. A typical pile performance specification will state that the pile head should not deflect more than 8 mm under the design working load and not more than 15 mm under 1.5 times the design working load. The residual deflection after the second cycle should not exceed 6 mm. If the piles are very long and/or slender these figures may have to be adjusted to allow for the elastic shortening of the pile shaft. Allowance should also be made in the calculation of the allowable residual deflection for friction preventing the recovery of the pile. A typical result from a three cycle load test is shown in Figure 9.1.1. Procedures for tension and lateral load tests are not given in SABS 1200F. A two cycle load test procedure similar to that described above could be programmed for these tests.

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STATIC LOAD TEST CONFIGURATION For piles loaded in compression the load should be applied concentrically to the pile head using hydraulic jacks. A loading beam is used to transfer the load from the jacks to the source of reaction which can be kentledge, anchor piles or anchors. The installation of anchors and anchor piles can take a considerable time whereas kentledge can often be provided at short notice. If a number of tests are required then kentledge will be the most economical form of reaction. Load tests of up to 500 tons are possible with kentledge. For greater loads, anchors or anchor piles must be used. The movement of the pile head is monitored using a minimum of two deflectometers reading to 1/100th of a millimetre resting on glass plates cemented to the pile head. The deflectometers are supported on a reference beam which in turn is mounted on posts driven into the ground at least two metres from the test pile. As the reference beam itself may move due to temperature effects, its deflection is sometimes monitored with dial gauges as well. It is advisable to have a backup system for monitoring pile head movement and this is normally provided by a precise dumpy level and scale rules fixed to the pile head. A suitable reference bench mark is located well away from the test area. In most instances it is sufficiently accurate to obtain the load on the pile from the product of the hydraulic pressure in the jack(s) and the area of the bore of the jack(s). It is common practice to have the jacks calibrated and the calibration certificate submitted with the test load results. A calibrated load cell should be used if a more accurate measure of the load is required. A typical test arrangement using anchor piles is shown in Figure 9.1.1.

Figure 9.1.1 -Typical compression test arrangement

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One of the load test configurations that can be used for a tension load test is shown in Figure 9.1.2. A central rod which is cast into the pile or fixed to the reinforcement, passes through a centre hole jack with a nut providing the seat for the jack ram. The jack rests on a beam which transfers the load onto two spread footings, one either side of the pile. The measurement of load and deflection is achieved in a similar manner to that described for the compression test load above.

Figure 9.1.2. -Tension load test configuration A lateral load test is most easily carried out by jacking two test piles apart. Measurement of lateral movement should be recorded by at least two deflectometers mounted on a reference beam with supports are well clear of the zone of influence. It is important for the accuracy of the test that the lateral load is applied through the centre line of the piles thus eliminating any torsional effects. Such an arrangement is shown in Figure 9.1.3.

Figure 9.1.3- Lateral load test configuration

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LOAD TEST RESULTS A typical result from a compression load test is shown in Figure 9.1.4. The pile was loaded in three cycles to 1.0, 1.5 and 2.0 times the design working load with pile head deflections of about 6, 10 and 15 mm and residual deflections of 2.5, 3.7 and 6.5 mm respectively. These results indicate a successful load test. Similar curves can be plotted for tension and lateral load tests.

Figure 9.1.4- Typical compression load test results

INTERPRETATION OF LOAD TEST RESULTS There are a number of methods outlined in the literature that can be used for analysing a load test result. These methods are generally aimed at predicting the ultimate pile capacity and splitting the piles capacity into friction and end bearing components. Two such methods are described in Chin and Vail (1973) and Van Weele (1957). More recently methods for modelling the performance of piles on a computer have been developed and one of these methods is described in Everett (1991). With a computer model the various parameters can be altered until the predicted load/deflection curve matches the actual. Once this has been achieved the computer model is an accurate model of the pile and every aspect of the piles performance such as ultimate load, load distribution between friction and end bearing as well as end bearing performance can be obtained. The model can then be used to predict the performance of similar piles of varying sizes and depths in the same soil profile.

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9.2

INTEGRITY TESTING OF PILES

Due to the limitations, both from a logistic and economic point of view, of testing a representative sample of working piles using the static load test methods outlined above, nondestructive integrity test methods have been developed to aid the detection of defects in the structural integrity of pile shafts. It must be realised, however, that the results of these tests are not definitive and are subject to interpretation but they do provide another tool with which to make an initial assessment. Any defect indicated by integrity testing should be investigated further using more positive testing methods. Integrity testing is at present carried out using either sonic or nuclear technology. The sonic methods involve either the collection of data from reflections of a sonic wave generated by tapping the head of the pile or by the collection of data by a collector of sonic waves generated by an emitter, both the emitter and collector being lowered down the pile in separate small diameter tubes cast in for this purpose. These two methods are referred to as Sonic Impact and Sonic Logging. Nuclear testing is very similar to Sonic Logging except that a nuclear isotope is used to generate the signal. INTEGRITY TESTING USING SONIC IMPACT With this test a sonic wave is propagated down the longitudinal axis of the pile while a transducer is held against the surface of the head of the pile. The sonic wave will be reflected off the toe of the pile or off any intermediate defects and these reflected waves are picked up by the transducer. A computerised signal processing unit records the reflected waves and prints out this record which is known as a reflectogram. In this way a major defect in the shaft of a cast-in-situ pile or a crack in the shaft of a precast pile can be detected. The time taken for the wave to travel down the pile shaft and for the reflected wave to travel back up is recorded by the equipment and from this the length of the sound pile shaft can be ascertained. The system can thus be used to check the depths of existing piles as well as the integrity of the shafts. The reflectogram shown in Figure 9.2.1 was taken from a sound pile. The blip in the curve at depth zero is a record of the initial blow to the head of the pile to generate the wave. The second major blip is at a depth of 10.5 metres as set out on the horizontal axis and this shows the reflected wave from the toe of the pile. The reflectogram in Figure 9.2.2 was taken from a defective pile and the difference in the two curves is very noticeable. The defect in this pile is indicated to be at a depth of about 2.5 metres. Interpretation of these reflectograms has to be carried out by an experienced technician with a considerable amount of experience. This method requires exposure of the pile head which will have to be brushed clean or even trimmed so as to expose sound concrete. The tests can be carried out by one person and forty to fifty piles can be tested in a day provided the heads have been prepared. The test is thus a very quick and economical one.

147

All pile types can be tested but there are depth limitations. If the depth to diameter ratio exceeds about 40 to 50 then the reflected wave may be damped out by the friction on the pile shaft. The test is thus not successful on long slender piles where the ratio might be as high as 100. Joints in precast piles can also present a problem as the wave tends to reflect off the joint. Not all pile joints, however, have this problem. There are a few variations in the type of equipment for carrying out this type of test which the various suppliers claim have certain advantages. In essence, however, the ability to detect a defect suffers from the same limitations.

Figure 9.2.1 -Reflectogram of a sound pile

Figure 9.2.2 -Reflectogram of a defective pile

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INTEGRITY TESTING USING SONIC CORING The sonic coring method involves the lowering of electronic equipment down tubes cast in the shaft of the pile. These tubes are about 75 mm in diameter and can be made of steel or plastic. Normally three or four such tubes are cast in the pile at even spacing around the perimeter. The tubes have to filled with water before a test. The electronic equipment consists of a sonic wave transmitter, a sonic wave receiver and a computer for storing the data and plotting out the results. To carry out a test the transmitter is lowered down one of the tubes while the receiver is lowered down one of the other tubes. Both the transmitter and the receiver are lowered at the same time by the same winch which has a means of automatically recording the depth. The sonic waves travel through the concrete between the transmitter and the receiver. If there is a defect, the pattern of the trace diagram is deformed as is shown in Figure 9.2.3. The test only covers a narrow band between the two tubes. If there are four tubes cast into the pile then a total of six different records can be taken. If.there is a major defect this should be detected by at least one of these. On smaller piles three tubes are used and on larger piles six or more tubes can be used. A typical arrangement of tubes and the resultant coverage is shown in Figure 9.2.4. If a defect is detected with the equipment further investigation in the form of visual inspection, rotary core drilling or a load test should be carried out to determine the exact nature of the defect and whether the performance of the pile is affected. A single sonic coring test result should not be used as the final arbiter of the integrity of a pile. The disadvantages of this form of test are the cost of the tubes and the fact that the piles to be tested have to be selected prior to casting. For large diameter piles which are normally heavily loaded, these disadvantages are often outweighed by the comfort of knowing that the integrity of the piles has been checked. INTEGRITY TESTING USING A NUCLEAR ISOTOPE The nuclear method is similar to the sonic coring in that the device is lowered down a tube cast in the pile. The equipment available in Southern Africa consists of a dual gamma ray emitter and detector. This obtains a measure of the density of the concrete by recording the amount of radiation reflected back. The device is lowered down the tube by means of a winch which automatically records the depth. The signal is recorded and processed in a similar manner to that in the Sonic Coring method. This type of test suffers from the same disadvantages as that of the Sonic Coring method. It has one additional disadvantage in that a nuclear isotope has to be transported from the laboratory to the site with the strict controls that are placed on the transporting of such materials.

149

Figure 9.2.3- Typical arrangement of the Sonic Coring method

Figure 9.2.4 -Typical arrangement of monitoring tubes

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10.0 FACTORS INFLUENCING THE SELECTION OF A SOIL IMPROVEMENT SYSTEM As with piling, the design engineer needs to have certain minimum information before he can assess which, if any, soil improvement solution is best suited to a particular project. The following is a summary list of this information. • • • • •

Detailed soils information including gradings and Atterberg limits. All loads and loading conditions. Allowable total and differential settlements. Knowledge of the site and its environs. Details of the various soil improvement systems.

Once the information is available consideration has to be given to the following points so that the most suitable system can be selected for the project. Structural • • • •

Allowable bearing pressures on the improved soil. The layout and size of the foundation footings. The number and spacing of treatment points. The effect of the process on the surrounding buildings, if any.

Soil Profile • • • • • • • •

Sections of the soil profile through the site. Grading of the soils in the profile. The presence and level of the water table. The depth of treatment to meet bearing and settlement requirements. The degree of improvement necessary to meet bearing and settlement requirements. The presence of very soft layers which cannot be compacted mechanically. The ease or difficulty in achieving the required depthThe presence of obstructions such as boulders, rubble etc.

Environmental • • •

The sensitivity of the environs to vibration. The sensitivity of the environs to noise. Problems associated with large volumes of water used in the treatment process.

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Contractual • • • • • • • •

Access to and from the site for equipment. Headroom clearance on site for equipment. The cost of soil improvement. The cost of the footings. The installation risks associated with each system. The remoteness of the site. The availability of the skills and plant to carry out the work. The availability of suitable material as replacement fill, if applicable.

Most of these points are covered in SECTION 12.0 SUMMARY DETAILS OF SOIL IMPROVEMENT SYSTEMS and in SECTION 13.0 DETAILS OF SOIL IMPROVEMENT SYSTEMS. An initial selection can be made from SECTION 12.0 but this should be checked by reading the more detailed information given in SECTION 13.0. There are a number of factors to be considered and the assessment of these will be difficult for someone not experienced with soil improvement techniques. Should there be any doubt you are welcome to contact your nearest Franki office for advice.

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11.0 CLASSIFICATION OF SOIL IMPROVEMENT SYSTEMS The use of soil improvement techniques to solve geotechnical problems is on the increase and a number of various methods have been developed for this purpose. The following is a classification and listing of the soil improvement systems that are on Franki's product list. The section number for detailed information is quoted on the right. Soil Compaction • • •

Vibratory compaction Dynamic compaction Compaction grouting

(13.1) (13.2) (13.3)

Soil Replacement • • •

Vibratory replacement Dynamic replacement Driven stone columns

(13.4) (13.5) (13.6)

Consolidation •

Accelerated consolidation

(13.7)

In-situ columns • •

Jet grouting Lime columns

(13.8) (13.9)

There are other methods of improving the soil such as soil mixing, chemical grouting and soil reinforcement. The latter involves techniques such as soil nailing, reticulated micropiles and other forms of soil reinforcement. Soil nailing and reticulated micropiles are also lateral support systems and as such are covered under SECTIONS 17.7 AND 17.8. The other methods are not covered in this text. A summary of the details of these various methods is given in a readily referenced format in SECTION 12.0 SUMMARY DETAILS OF SOIL IMPROVEMENT SYSTEMS. For a full description of each of the systems and its method of application refer to SECTION 13.0 TECHNICAL DETAILS OF SOIL IMPROVEMENT SYSTEMS.

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12.0 SUMMARY DETAILS OF SOIL IMPROVEMENT SYSTEMS Sect. No.

Soil Improvement system

SOIL COMPACTION 3.1 Vibratory compaction 3.2 Dynamic compaction 13.3 Compaction grouting SOIL REPLACEMENT 13.4 Vibratory replacement 13.5 Dynamic replacement 13.6 Driven stone columns CONSOLIDATION 13.7 Accelerated Consolidation IN-SITU COLUMN 13.8 Jet grouting 13.9

Lime Columns

Approx. Column Diameter (mm)

Spacing of Compaction Points (mm)

Typical Bearing Capacity

N/A

1 500 to 2500

100 to 200 kPa

N/A

3000 to 7000

1 00 to 200 kPa

150 to 200

1000 to 2000

100 to 200 kPa

900 to 1200

1500 to 2500

1500 to 2500

2500 to 5000

400 to 600

1200 to 2000

300 to 500 kN per point 300 to 600 kPa on column area 300 to 750 kPa on column area

N/A

1500 to 3000

N/A

500 to 2000

Varies with application Varies with application

Depends on cube strength of grout See Broms (1993)

500 to 600

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SUMMARY DETAILS OF SOIL IMPROVEMENT SYSTEMS Normal Maximum Depth (metres) 20 12 10 15 8 15 25 20 15

Unit Cost Pollution

Noise Pollution Level

Vibration Required Level

Site Area Headroom

Normal Requirements (metres)

Low Low High High Medium High Low High Medium

Low Medium Low Low Medium Medium Low Low Low

Medium High Low Medium High Medium Low Low Low

Medium Medium Small Medium Medium Medium Medium Medium Medium

25-30 25-30 6-12 20-25 25-30 20 25-30 15-25 15-25

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13.0 TECHNICAL DETAILS OF SOIL IMPROVEMENT SYSTEMS 13.1

VIBRATORY COMPACTION

Vibratory compaction is achieved using a vibratory immersion probe of one form or another. Compaction to considerable depths is possible. The degree of soil improvement is largely dependent on the grading of soil, the natural resonant frequency of. the soil and the level of energy available. Where the grading of the soil is suitable this system achieves very effective compaction at a competitive rate. Positive features • • • • •

Fast and economical system. The degree of compaction can be readily checked. Noise and vibration levels are low. Compaction to depths of 20 metres is possible. Suited to soil profiles with a high water table.

Negative features • •

The degree of compaction is sensitive to grading. Not suited to materials with high silt and or clay content.

SUITABLE SOIL PROFILES Not all soils are suited to compaction by deep vibrators. It is important to carry out a grading analysis and to study the grading of the soil so as to assess its suitability before deciding \\hether to use this technique. There are two methods that can be used for this purpose. One of these was proposed by Brown (1977) and involves the calculation of a suitability number β . The formula for calculating this number is as follows:

β = 1.7

3

(d 50 )

2

+

1

(d 20 )

2

+

1

(d10 )2

Where d50, d20 and d10 are the particle sizes in millimetres at 50, 20 and 10 percent passing by mass. A number less than 10 indicates a soil that is highly suitable for compaction by vibratory means whereas a number in excess of 30 indicates an unsuitable soil. The second method was proposed by Mitchell and Katti (1981) and is illustrated in Figure 13.1.1. This consists of a grading envelope which shows the limits for the most desirable grading. The degree of compaction is very sensitive to the amount of silt and clay in the soil which should not exceed 15 percent. The clay fraction should not exceed 3 percent.

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Figure 13.1.1 Soil grading suitable for vibratory compaction. Mitchell and Katti (1981) When the fines exceed the percentages stated above the increase in pore water pressure resulting from re-arrangement of the grains cannot dissipate quickly and there is a tendency for liquefaction. Under these conditions it is not possible to achieve any significant compaction with what is a relatively quick compaction method. Unfortunately the soils in Southern Africa generally have a significant silt and clay content so compaction of this nature is not widely used. The more common soil improvement process for these conditions is Vibrocompaction which is a replacement process and is covered in SECTION 13.4 VIBRATORY REPLACEMENT. COMPACTION DETAILS Compaction points are normally spaced at between 1.5 and 2.5 metres centre to centre. The actual spacing is best decided upon by carrying out test compaction patterns and monitoring the results by carrying out pre and post compaction soil strength measurements using the CPT and SPT methods as well as carrying out plate load tests on the compacted soil. If adequate compaction is achieved then a spread footing founded on the improved soil can be designed using a bearing pressure of up to 200 kPa. The zone of soil compacted should extend beyond the edges of the footing by about 10 percent of the depth treated. Normally the upper 1000 mm will not be compacted using this technique and this must either be removed or compacted in-situ using an impact roller or dynamic compaction.

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INSTALLATION TECHNIQUE There are two basic types of vibrator used for deep compaction: one which has horizontal amplitude and one with vertical amplitude. The former type has a built in motor with eccentric weights which rotate about a vertical axis thus providing horizontal amplitude. The latter type consists of a separate vibrator unit with eccentric weights which rotate about the horizontal axis thus imparting a vertical amplitude. The vibrator clamps to a long slender metal section which is called a probe. Horizontal Amplitude The vibrator used with this system is a large immersion type with the motor and eccentric weights located at the lower end of the unit. The eccentric weights rotate about a vertical axis so the amplitude of the vibration is in the horizontal plane. The vibrating section is coupled to a follower section by means of a flexible coupling. A suspension point for handling the vibrator is located at the upper end of the follower section. The whole vibrator unit and part of the follower section becomes immersed in the ground during the compaction process. The vibrator section has two or more water jets. Two high volume jets are located at the tip of the vibrator and are used for aiding penetration of the vibrator into the ground. On some vibrators there are another two low volume jets located above the vibrator and these are used to feed water into the cavity to keep the sidewalls stable. The hydraulic fluid for the motor and the water for the jets are fed down from the head of the vibrator through pipes located in the hollow core of the unit. Large quantities of water are used in this process and this can present a site control problem. A crane is normally used to lower and raise the vibrator. With the vibrator set up on the compaction position the motor is started and the high volume water jets are activated. The crane lowers the vibrator into the ground. When the vibrator has penetrated to the full depth the water flow through the high volume jets is shut off leaving the upper low volume jets to feed water into the cavity. The vibrator is raised slowly and then lowered again into the soil which flows into the cavity under the tip of the vibrator. The raising and lowering of the vibrator continues in a repetitive cycle as it is gradually withdrawn. The water flow must be controlled and at some stage it must be shut off completely. The effectiveness of the compaction can be monitored at all times using an oil pressure gauge mounted in the cab of the crane. The compaction process must continue up to ground level or a minimum of one metre above the footing soffit level. Vertical Amplitude The probe is a long steel section to which the vibrator is clamped at the head. The cross sectional shape of the probe can vary considerably and there are some patented types on the market. Not all probes have a constant cross section and some are fitted with wings which protrude from a central column.

158

The vibrator, which is either electrically or hydraulically powered, is clamped onto the head of the probe using a hydraulically activated mechanism. The whole unit is suspended from a crane which lowers and raises the unit as required. With this system only the probe enters the ground. The probe with the vibrator attached is set up over a compaction position. The vibrator is activated and the probe is lowered slowly into the ground. There are no water jets or any other means for assisting penetration so the vibrator has to be powerful enough to drive the probe to the full depth. Once full penetration is achieved, the probe is lifted about a metre and then lowered again. This lifting and lowering is continued in a cyclic manner as the probe is gradually withdrawn. An indication of the degree of compaction can be obtained by monitoring the electric current in the case of electrically powered vibrators and the hydraulic pressure in the case of hydraulically powered vibrators. The probe is not capable of compacting the upper one metre of soil due to limited containment so this must be compacted using an impact roller or dynamic compaction or alternatively, the level of the footing soffit must be below this depth. Plate 13.1.1 shows a typical horizontal amplitude vibrator and Plate 13.1.2 a typical vertical amplitude unit with a y -probe. VARIATIONS IN INSTALLATION TECHNIQUE Variable Frequency Vibration All soil profiles have a natural resonant frequency which varies depending on the profile. It has been found that the best compaction results are achieved when the vibrator is operating at this resonant frequency. In Europe variable frequency vibrators have been developed and some contracts have been completed. This is the latest in deep compaction technology but it has not been introduced to Southern Africa as yet. POTENTIAL PROBLEM AREAS Variable soil profile Soft layers of silt, peat and clay cannot be compacted using the probe or for that matter any form of mechanical compaction. Even silty and clayey sands can prove difficult to compact. Soil profiles in Southern Africa often have strata of these materials present with the result that only a certain percentage of the profile can be compacted. This situation is not normally acceptable and the vibratory compaction solution has to be rejected for this reason. When selecting a soil improvement system for a site it is important to determine whether any of these soft layers are present. A common solution to the variable soil profile problem is the use of the vibratory replacement form of compaction which is covered in the SECTION 13.4 and which is more commonly known as Vibrocompaction.

159

Plate 13.1.1 A typical horizontal amplitude vibrator

Plate 13.1.2 A typical vertical amplitude vibrator

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13.2

DYNAMIC COMPACTION

Dynamic compaction is achieved by dropping a large weight known as a pounder from a considerable height onto the soil to be compacted. It is a compaction system that is very effective in the right conditions. It is fast and economical and suited to the lighter loaded structures such as shopping centres, industrial buildings and low-rise residential buildings as well as earthfills for road and dams where it will increase the bearing capacity of the in- situ soil and reduce settlement potential. Another major use is to reduce liquefaction potential. A wide range of soils can be compacted including fills which contain rubble. It is often used to solve collapsible soil problems under large loaded areas. Positive features • • • •

It is a fast and very economical compaction system. Most soil profiles can be compacted. Compaction can be achieved both above and below the water table. Fills contaminated with rubble, boulders and rocks can be compacted.

Negative features • •

The impact shock wave can cause damage to surrounding buildings. The depth of improvement with locally available equipment is limited.

SUITABLE SOIL PROFILES All soil types with the exception of soft silts, clays and peats can be compacted using the system. Materials both above and below the water table can be compacted. The Dynamic Replacement system is often used where there are soft silts, clays and peat as is covered in SECTION 13.5 DYNAMIC REPLACEMENT. DETAILS OF COMPACTION POINTS The depth of compaction is a function of the weight of the pounder and the height of the drop. For the normal energy levels this approximates to the following relationship. D = k x √Wh Where D

is the depth of compaction in metres.

k

is an influence factor which varies between 0.375 and 0.7.

W

is the weight of the pounder in tonnes.

h

is the drop height in metres.

Experience has shown that the depth of influence and the degree of improvement are also influenced by the shape of the pounder. A wide range of pounders has been developed for varying site conditions with different pounders often being used on different phases of the same contract.

161

The compaction is usually carried out in three different phases known as the primary , secondary and ironing phases, in this order. Compaction of the deepest layer is achieved with the primary phase. The secondary phase achieves compaction mainly in the intermediate layers. The ironing phase ensures overlapping of the initial phases by compacting the shallow layers between the initial prints. Figure 13.2.1 illustrates how the various phases compact the different levels in a soil profile. The initial choice of spacing for the primary compaction points is based on experience but one expects these to be between six and ten metres apart. Various field tests are carried out during the early stages of the contract so as to check on the level of compaction being achieved and the spacing of the primary points while the energy input is also determined at this stage according to the results of the tests. Once the primary phase is complete work proceeds on the secondary phase points. These are positioned midway between the primary phase points. Here again the energy input of the secondary phase points is determined by the results of field tests. The final ironing phase is aimed at compacting the upper two to four metres. In this phase the drop height of the pounder is limited (say 4 to 8 metres) and the whole area is compacted. INSTALLATION TECHNIQUE The equipment consists of a heavy weight which is referred to as a pounder and a means of lifting and dropping this weight, which is usually a crane or a special frame fitted with a linear winch. Earth moving equipment is used to backfill the craters formed by the dynamic compaction and to re-establish site levels. The compaction process involves the repeated lifting and dropping of the pounder on a compaction point. The number of times the pounder is dropped on one point is determined through tests on site. The sequence of points follows the primary, secondary and ironing phases as set out above. Plate 13.2.1 shows a typical pounder, Plate 13.2.2 shows a crater formed by the impact of the pounder and Plate 13.2.3 shows a dynamic compaction contract in progress. VARIATIONS IN INSTALLATION TECHNIQUE Collapsible Sands A typical collapsible sand has a relatively open grain structure with individual grains connected by a clay bridge. When the soil moisture content rises the clay bridges soften with a resultant loss in shear strength. The overall shear strength of the soil is thus high when it is dry and low when it is wet. This change in the shear strength due to wetting can lead to the collapse of the grain structure of a soil under load, resulting in foundation failure.

162

Figure 13.2.1 Compaction patterns for primary, secondary and ironing phases

163

If one attempts to compact a collapsible sand when it is dry a high level of energy is required to do so. By soaking the area to be compacted with water before the compaction is to be carried out, the energy levels required for compaction are reduced dramatically and compaction using the dynamic compaction method can be readily achieved. POTENTIAL PROBLEM AREAS Excess Pore Water Condition Under saturated or near saturated conditions, the pore water pressure will increase with each blow of the pounder. If it becomes excessive the pounder will have little compacting effect as the blow is being cushioned by the pore water. In these circumstances further compaction in these areas may have to be delayed until the pore water pressures have dissipated. In coarse grained materials the dissipation of excess pore water pressures takes place virtually immediately. With saturated clays on the other hand, dissipation could take weeks and such delays often make the dynamic compaction option impractical. Damage to Surrounding Buildings Due to its nature the generation of vibration by dynamic compaction is inevitable. While in open areas this is of little significance, problems can be experienced in developed areas unless correct precautions are taken. The magnitude of the vibrations and the transmission thereof is greatly dependent on the nature of the materials being compacted, the depth of the compaction, the nature of the underlying materials, the presence of the water table and the energy input procedure. As a general rule a saturated layer underlain by a hard rock will give the highest energy/vibration transmission. Techniques that have been developed to control and isolate vibration include the excavation of isolation trenches, compacting from reduced levels, reducing the energy input per blow and more recently the development of low vibration pounders. As a result of this dynamic compaction has been carried out successfully immediately adjacent to existing structures.

164

Plate 13.2.1 A typical pounder Plate

13.2.2 Crater formed by pounder

Plate 13.2.3 A dynamic compaction contract in progress

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13.3

COMPACTION GROUTING

Compaction Grouting is a soil compaction technique in which the density of the soil is improved by introducing a thick grout under pressure into the soil. The thick grout forms an enlarged bulb or series of bulbs in the soil and in so doing, it displaces the soil immediately surrounding the bulb, thereby increasing its density. It is a relatively expensive technique but one ideally suited to remedial work associated with soils of low density such as poorly compacted fills. Positive features • • • • • •

Small rigs can get into difficult access and low headroom conditions. No vibration. Noise levels limited to the engine noise only. Negative features Relatively expensive techniqueLow production rate so suited more to small contracts.

SUITABLE SOIL PROFILES The ideal soils for compaction grouting are loose sandy soils and gravels, above or below the water table. Silty and clayey sands as well as partially saturated clays and silts can also be treated using compaction grouting, provided the soil mass has good drainage characteristics. The process cannot compact saturated clays. SOIL IMPROVEMENT DETAILS The diameter of the grout pipe is normally in the 50 to 100 mm range. The centres at which the points are arranged are in the 1.0 to 4.0 metre range but 1.5 to 2.0 is more common. If compaction near the surface is required, the points have to be positioned at the closer spacing. The full depth of a stratum can be compacted as a series of enlarged grout bulbs can be formed to cover the full depth. The overall maximum depth of treatment is normally limited to about ten metres with conventional equipment. INSTALLATION TECHNIQUE The grout pipe can be installed using either driving or drilling techniques. The sequence of grouting is generally planned as a series of primary and secondary compaction points. All the primary points are drilled and grouted first followed by the secondary points some days later. The secondary points are positioned midway between the primary points. The presence of the primary compaction points act as a containment when grouting the secondary points. A tertiary stage could be used as well if found necessary.

166

The grouting of each compaction point can be carried out from the bottom up, which is referred to as upstage grouting, or from the top down, which is known as downstage grouting. It is possible, and sometimes desirable, to use a combination of the two. With upstage grouting, the expanded bulbs are formed as the grouting tube is gradually withdrawn. With downstage grouting, the uppermost bulb is formed first and after the initial set has taken place, the grout pipe is drilled through the bulb to a lower depth where the next bulb is expanded. It is sometimes beneficial to form the top bulbs first in a downstage operation, followed by upstage grouting of the balance of the stratum. This technique is advantageous when there is limited overburden, as the upper enlarged bulbs act as a containing mechanism. The pressure at which the grout is injected has to be carefully monitored as excessive pressure will cause fracturing of the soil and resulting ground heave. The pumps generally have a pressure capability of 40 bar but a limiting pressure of 20 bar at the head of the grout pipe is a typical figure for deep compaction. One of the objectives when compaction grouting a soil mass is to attempt to even out the volume of grout injected over the whole area. The percentage replacement should be decided on and the volumes of grout controlled according to this figure. Ground heave must be monitored and the volumes reduced if heave is taking place. A sand/cement grout with a slump of between 25 and 75 mm is used for compaction grouting. It does not have to meet any strength requirements as the objective is not to form a structural element in the ground but to compact the ground itself. Cement contents can vary from zero to 500 kg per cubic metre but 300 kg per cubic metre is more typical. Flyash is often used as a substitute for up to 50 percent of the cement as flyash extends the working life of the grout and improves workability. Retarders are also used for the same purpose. The grading of the sand is important to ensure the workability of the mix, even under high pressure. Often two or more sands are blended to produce the ideal grading. If a well graded sand is not available, a bentonite slurry can be blended with the sand and the cement partially substituted by fly ash to aid the workability. Pumping rates should also be carefully monitored and controlled. The pumping rate should be in the range between 15 and l00 litres per minute. The rate should be lower in soils with poor drainage characteristics and when the compaction process is carried out close to the ground surface. Higher rates can be used in free draining soils with significant cover. VARIATIONS IN THE INSTALLATION PROCEDURE Raising Footings that have settled The fact that compaction grouting causes ground heave can be used to an advantage in that footings that have settled can be raised ~gain using the compaction grouting technique. While the level of the footing is closely monitored further pumping of grout is undertaken until the footing has risen to the original level.

167

13.4

VIBRATORY REPLACEMENT

This method is commonly referred to as Vibrocompaction. It is a replacement method of soil improvement and crushed stone is the most common form of replacement material. The system has been used in the coastal areas of Southern Africa for the past thirty years and has proven to be a reliable product. It is ideally suited to structures with a large area of uniform distributed loading such as tank bases, the ground floors of warehouses and industrial buildings as well as road embankments. The solution has also been used on bridges, multistorey buildings and silos. Positive features • • • • •

Well-proven soil improvement system. Low noise levels limited to engine noise only. Low vibration levels except close to vibrator. Can provide an economical form of foundation. Fast installation rate.

Negative features •

Uses large quantities of water which needs good site management.

SUITABLE SOIL PROFILES As described in more detail later, the Vibrocompaction method forms a compacted stone column in the ground which behaves under load in a similar manner to that of a pile. In forming the stone column the vibrator will also compact the in-situ soil providing it has a suitable grading. See SECTION 13.1 for methods to assess this suitability. The lack of improvement in individual soil layers is, however, not a problem because the stone column can transfer the load through these layers. If the layers are very soft or are thicker than one third of the column diameter, the stability of the stone column should be checked. A method for carrying out this check is given in Section 21.0 DESIGN AIDS SOIL IMPROVEMENT. The process is commonly used in sand and silty sand soil profiles in the coastal areas of Southern Africa. SOIL IMPROVEMENT DETAILS The vibrocompaction columns are generally about 1000 to 1100 mm in diameter. They are designed to carry loads of between 300 and 500 kN.. The spacing varies between 1500 and 2500 mm with 1500 and 1750 being the more common spacings. The spacing is often the result of field tests carried out on site to determine the effectiveness of the soil improvement.

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INSTALLATION TECHNIQUE The vibrator is of the immersion type with the motor and eccentric weights built into the unit. The vibrator section is coupled to the follower section by means of a flexible coupling. There are water jets located at the tip of the vibrator so as to assist initial penetration and to keep the annulus around the vibrator clear during the compaction stage. Fins welded to the vibrator assist in preventing rotation of the vibrator in the ground, which is a natural reaction to the spin of the motor. The vibrator is set up over the compaction point and the motor and water jets are activated. The vibrator is then lowered slowly so that it penetrates the ground. A slight tension is maintained in the crane cable so as to keep the vibrator in the vertical plane. When the vibrator has reached the treatment depth, it is surged up and down a few times so as to enlarge the annular gap around it. The water flow to the jets is then reduced, the vibrator is lifted about 1.5 metres and a quantity of crushed stone is fed into the crater formed around the vibrator. The vibrator is lowered so that it penetrates into the stone which is now lying at the bottom of the hole. The vibrator compacts the stone to a high degree as monitored during the process by means of the current drawn by electrical vibrators or alternatively the hydraulic pressure on hydraulic vibrators. The cycle of lifting the vibrator, feeding in the stone and lowering the vibrator again, is repeated until the complete stone column has been formed. The vibrator is shown in Plate 13.1.1 in SECTION 13.1. A Vibrocompaction contract in progress is shown in Plate 13.4.1. POTENTIAL PROBLEM AREAS Very Soft Soils The presence of very soft layers is generally one of the reasons for choosing the vibratory replacement method in the first place but these same layers can still present a problem if they are excessively thick. The stone column manufactured by the process needs the lateral support of the ground to be able to carry axial load. Soft layers do not provide strong lateral support and thus the vertical carrying capacity of the stone column can be limited. If the soft layers are less than one third of the column diameter then there is limited concern provided there are only one or two such layers. If the soft layers are much thicker then the stability of the stone column should be checked. See SECTION 21.0. Silty or Clayey Soil Profile For the compaction of saturated soils to take place the pore water has to dissipate. If the soil profile is saturated and has a large silt or clay content, then excessive pore water pressures will develop and the whole soil mass will become liquefied. Under these conditions the formation of stone columns will not be possible.

169

Plate 13.4.1 A vibratory replacement (Vibrocompaction) contract in progress

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13.5

DYNAMIC REPLACEMENT

This technique is used in very soft cohesive soil profiles where the compaction of the in- situ material is not possible. The soft soil is replaced by columns of stone, rubble or other suitable materials which are driven using a special pounder designed for this purpose. These large stone columns can form the foundation for many types of structures including low-rise buildings, earth dam walls, road embankments etc. Positive features • • •

Economical solution in difficult soil conditions. Noise levels limited to the engine noise of the plant. Large diameter of stone column with significant load carrying capacity.

Negative features • • •

Limited depth of installation using conventional equipment (6 to 8 metres). Shock wave from pounder impact. Ground heave must be monitored.

TYPICAL SOIL PROFILE The ideal soil profile has an upper sandy stratum which is one to two metres thick underlain by two to six metres of soft silt or peat underlain by competent soil or rock. As this is a replacement process the grading of the in-situ sandy soils is not of significance although the process will compact the in-situ materials provided they have a suitable grading. The stability of the stone column may need to be checked if the soft stratum is very soft and/or its depth is greater than half a column diameter. See SECTION 21.0. DETAILS OF DYNAMIC REPLACEMENT POINTS The diameter of the stone columns is between 1.5 and 2.5 metres. Larger diameters are possible with purpose made equipment. The minimum spacing between the edges of the columns should be 1000 mm. The load capacity of each individual column can be calculated using a shaft stress of between 300 kPa for very soft profiles to as high as 600 kPa for more competent profiles. The cut-off level for the capping footing should be a minimum of 1000 mm below the natural ground surface. INSTALLATION TECHNIQUE The equipment used for carrying out dynamic replacement consists of a pounder and device for lifting and dropping the pounder. The latter is normally a crawler crane but special lifting frames fitted with linear winches are sometimes employed for the larger contracts and heavier pounders.

171

The crane is set up on position and the pounder is dropped from a reduced height so as to form a initial crater. This crater is then filled with rock or rubble after which the pounder is dropped the full height. The energy drives the stone into the ground displacing the in-situ material. Additional charges of stone are added to the crater each time it re-forms and thereby the stone column is driven deeper and deeper. Records are kept of the quantity of stone used and checks are made regarding the depth of the stone column. The column has to be driven so that it penetrates through the soft stratum and into the denser founding stratum. Once this has been achieved the crane moves on to the next position. If the stone columns are closely spaced, excess displaced material will be forced to the surface and allowance should be made for the removal of this material from site. POTENTIAL PROBLEM AREAS Inadequate Penetration In a dynamic replacement solution it is essential that the stone be driven down through the full depth of the soft layer. Should the problem of lack of penetration occur the solution is either to increase the mass of the pounder or to reduce the density of the soil by pre-drilling prior to commencing the dynamic replacement operation.

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13.6

DRIVEN STONE COLUMNS

This method involves the driving of a steel piling tube using the bottom driving technique used in installing Franki type piles (See SECTION 7.1 ). Once the tube has been driven the stone column is formed by expelling measured quantities of stone out the tube using the internal drop hammer. The displacement caused during the driving of the tube and the forming of the stone column results in compaction of the surrounding soil. The stone column acts as a structural member in much the same manner as a pile shaft. Stone columns have been used for the foundations of light to medium structures including buildings and bridges. Positive features • • •

Depths of up to 15 metres can be treated It is a clean system. Noise levels are low.

Negative features • • • •

Production rates are low. The relative cost of the method is high. A medium level of vibration is associated with the system. Ground heave is a potential problem.

SUITABLE SOIL PROFILES The method can be used in any soil profile that will not heave during the driving of the tube. It should thus be avoided in saturated cohesive soil profiles as these soils are the most problematic when it comes to ground heave caused by displacement. DETAILS OF STONE COLUMNS The stone columns can be made with either a 410, 520 or a 600 mm diameter piling tube. The minimum spacing should not be less than 2.5 times the tube diameter. The allowable working loads can be based on a shaft stress of between 300 and 750 kPa depending on the strength of the surrounding soil. INSTALLATION TECHNIQUE The equipment consists of a piling rig, piling tube and internal drop hammer. The piling tube is located in the mast of the piling rig. The piling rig is set up on position and the tube is lowered onto the ground. A plug is formed at the toe of the tube by placing a charge of crushed stone into the tube and compacting this with a few blows of the hammer. The drop height of the hammer is then increased and the tube is driven into the ground.

173

On reaching the required depth the tube is held by the piling rig while the hammer is used to drive the plug out of the tube. A charge of crushed stone aggregate is then placed in the tube, the tube is extracted a little using the extraction winch and the stone is expelled using blows of the hammer. This cycle is repeated as the tube is gradually withdrawn forming a continuous stone column over the full depth. VARIATIONS IN INSTALLATION TECHNIQUE Compacting a pile founding stratum A situation can arise where there is a suitable founding stratum for a piled foundation but considerable benefit can be derived from additional compaction of that stratum prior to the installation of the piles. In order to achieve this compaction, stone columns are formed in the stratum material for the full depth of the stratum. The degree of compaction can be controlled by varying the spacing of the stone columns as well as the quantity of stone expelled to form the column. POTENTIAL PROBLEM AREAS Ground Heave The driving of the piling tube causes displacement. In saturated silty and clayey soils this displacement can often cause ground heave where the soil being displaced moves outwards and up. This upward movement of the soil imparts a tension into the previously installed stone columns. As they are not capable of resisting tensile forces, the columns part and a gap is formed. This is obviously detrimental to the load bearing performance of the columns and thus the columns so affected have to be rejected. In most cases the only solution to the problem of displacement heave is a change to some other soil improvement or piling system.

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13.7

ACCELERATED CONSOLIDATION

When a load is applied to a saturated soil there is an instantaneous increase in the pore water pressure. The rate of dissipation of the pore water and the consolidation associated with it depends on the permeability of the soil and the length of the drainage path. When the soil is relatively impermeable and the drainage path is long, it will take a considerable time for the pore water pressures to normalise and for full consolidation to take place. Accelerated consolidation is a technique which involves the introduction of drains into the soil to reduce the length of the drainage path and thus decrease the time taken for consolidation to take place. It is used in situations where a large loaded area, such as a road embankment or material stockpile, is underlain by a considerable depth of very soft silt or clay. Positive features • • • •

A well established method to improve considerable depths of soft cohesive soils. Minimal vibration associated with installation of the soil drainsNoise levels are low and limited to engine noise during installation of the drains. A relatively low cost method.

Negative features • •

A time consuming method. The cost of importing and removing any surcharge material.

DETAILS OF SOIL DRAINS FOR ACCELERATED CONSOLIDATION There are three main types of drain used for accelerating consolidation: The sand drain, the sandwick drain and the band drain. The band drain has become the most popular and it is also the most economical due to its fast rate of installation. The soil profile suitable for band drain installation, however, needs to be fairly soft or loose as the mandrel installing the drain is driven into the ground. Drilling methods are mainly used for sand and sandwick drain installation and thus they can be installed in stiffer and denser soil profiles. The Sand Drain This was one of the first systems used for consolidation drains and is simply a column of highly permeable sand formed in the ground. A hole is initially formed by either drilling or driving and this is filled with sand of a suitable grading. If the drilling technique is used then the drains can be installed in a fairly stiff or dense soil profile. A temporary casing is often used to keep the hole open. The casing is extracted during the placing of the sand to reduce the risk of arching in the tube. The diameter of these drains is normally in the 150 to 250 mm range but larger sizes can be used if necessary although the cost will be considerable greater .

175

Various problems have been experienced with this form of drain which have seriously reduced the efficiency of the drains. In some cases disturbance of the soil during the installation procedure has reduced the permeability of the soil immediately around the drain thus reducing the effectiveness of the system. In other cases the column of sand has become necked due to incorrect installation procedure or due to high lateral soil pressures and this has also reduced the effectiveness of the drains. These negative case histories have tended to reduce the use of sand drains to situations where there is confidence that these problems will not occur. The Sandwick Drain The sandwick drain is very similar to the sand drain with the exception that the sand is contained within a sock made out of a geofabric. The sand filled sock is referred to as the sandwick. The hole is formed in the ground in a similar manner to that used for the sand drain and then the sandwick is lowered into the hole. The diameter of the wick is between 50 and 75 mm so the installation can be readily achieved by conventional drilling techniques. The presence of the sock assists the contractor with the installation of the drain and reduces the risk of discontinuities and necking. Problems can still be experienced with the installation procedure affecting the permeability of the soil immediately around the drain. The Band Drain The band drain has taken over from sand type drains to a large degree. The main reasons for this are the fact that band drains are very economical, they are strong and able to resist necking, squeezing and buckling and the small mandrel used to install them causes a minimum of disturbance to the surrounding soil. In ideal conditions the speed of installation is very fast with it taking only two to three minutes to install each drain. The soil profile does, however, need to be soft or loose as the mandrel used to install the drain is driven into the ground by means of a vibrator. The band drain itself is about l00 mm wide and between 2 and 7 mm thick. It consists of a strip of flexible cardboard or plastic which has longitudinal drainage channels formed in it. In some cases this strip is fitted with a surrounding filter sleeve. The band drain is supplied in large rolls. The equipment for installing band drains consists of a crane with a leader, a vibrator and a hollow mandrel. The mandrel is rectangular in shape with a length exceeding the depth to which the drains need to be installed. Connected to it at the head and at various intervals over its length are guides located in the leader. The vibrator is clamped to the head of the mandrel. The band drain is fed in at the top of the mandrel and emerges at the bottom, allowing a plastic anchoring shoe to be attached to it. The band drain is supplied in a coil which is mounted on a spool on the leader.

176

The vibrator drives the mandrel with the anchor shoe and band drain into the ground. The coil of band drain unwinds as the mandrel penetrates. When the desired depth has been reached the mandrel is extracted leaving the band drain behind in the ground. Once the mandrel is clear of the ground a cut is made through the band drain just above ground level. Another anchor shoe is attached to the piece protruding from the tip of the mandrel and the crane moves to the next position. The installation of the mandrel does cause some disturbance to the surrounding soil but experience has shown that this is limited to 2d where d is the diameter with the same circumference as that of the band drain. Other forms of Drains Whilst the three types of drain mention above are the most common used there are other forms which can be considered. Stone columns have a low permeability and can function as drains as well as act as structural columns (See SECTION 13.6). Lime columns can also act in a similar manner (See SECTION 13.9). Drainage Blanket The water which flows up through the drains under a preloaded area has to be led off to the sides of such an area. This is achieved by placing a layer of sand with a high permeability over the tops of the drains prior to placing the remainder of the fill or surcharge material. This layer, which is normally about 500 mm thick, is referred to as a drainage blanket. In certain instances there may be a natural drainage layer at the surface in which case there is no need for any additional measures to be taken. Design The design of drainage systems is covered in SECTION 21.0.

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13.8

JET GROUTING

Jet grouting involves the mixing and partial replacement of the in-situ soil with cement slurry as opposed to the conventional grouting which involves the injection of cement slurry into the voids in the soil. In its simplest form the process involves the ejection from a rotating grout tube of cement slurry under very high pressure. The jet cuts a path outwards from the grout tube in a radial manner, the cement mixing with the coarse particles in the soil while replacing the fines. The combination of rotation and gradual withdrawal enables a large diameter grout column to be formed in the ground. Positive features • • • •

A wide range of soil types can be treated. It is a vibrationless system. Noise levels are low and limited to engine noise only. It has features which provide unique solutions to difficult geotechnical problems.

Negative features • It is a relatively expensive technique. SUITABLE SOIL PROFILES The ideal soil type for jet grouting is a clean loose medium to coarse sand. The sand particles are readily eroded away by the grout jet and thus the jet is able to penetrate up to half a metre with a single jet and up to a metre with air and water assisted grouting. Gravels are also amenable to treatment using jet grouting especially the finer gravels. Larger particles such as cobbles will tend to shield the jet and limit the size of the grouted column and make the cross-section very irregular. Cohesion in the soil tends to reduce the ease with which the particles are eroded. The diameter of the grout column will thus reduce as the silt and/or clay content increases. In silty sands the reduction in diameter can be of the order of 15 to 30 percent. With purely cohesive soils such as silts and clays the diameter is even further reduced to roughly half that in clean sand. The stiffness of the cohesive soil is also important and only soils with a very soft and soft consistency (SPT value of up to 6 ) should be regarded as being suitable. Jet grouting is unaffected by the presence of a water table. INSTALLATION TECHNIQUE The equipment consists of a crawler mounted drill rig, grout mixing plant, the high pressure grout pump and the grouting tube fitted with high pressure jets. The grout tube which is usually 50 to 75 mm in diameter is either drilled in by the machine itself or can placed in a predrilled hole to speed up the process. A temporary casing or a bentonite slurry is used to keep the holes open. The holes can be vertical or inclined at a rake.

178

The jetting operation involves the pumping of a cement slurry under high pressure so that it emerges from the jet at the base of the grout tube at a high velocity. If a cylindrical column is required the jet tube is rotated and gradually raised at a constant rate. Other shapes, such as flat panels, can be formed by not rotating the grout tube. The simplest form of jet grouting involves a single jet of cement slurry which is pumped in at pressures of up to 600 bar. This has been developed further with the introduction of a jet of air which acts as a shroud around the cement slurry jet and is referred to as the double jet system. An air pressure of between 2 and 15 bar is used. This system has been further improved with the addition of a high pressure water jet for eroding away the soil, known as the triple jet system, which can achieve twice the radial penetration of the single jet system. The water pressure is high ( up to 500 bar) but the grout pressure is reduced to between 5 and 30 bar. Any excess suspension flows up the annular gap between the grout tube and the soil to the surface from where it is channeled into settling ponds for subsequent removal from site. The addition of an air jet assists this excess material to rise to the surface and keep the annulus clear. Blockage of the annulus can cause a build up of pressure which can result in soil fracturing and resultant ground heave and should be avoided. The diameter of the grout column is a function of the speed of withdrawal, the soil type, the system of jets and the pressures used. Different pressures and/or rates of withdrawal have to be used in the different soil strata to produce a grout column of reasonably constant diameter. Field tests are necessary to determine what these parameters should be for the various soil strata. APPLICATIONS OF JET GROUTING Jet grouting has a wide range of applications which include: • • • • • • •

Forming of grout columns to support structural loads. Forming a contiguous wall to a caisson or cofferdam. Forming a contiguous wall for lateral support. Forming cut-off walls for groundwater controlForming a base seal to an excavation below the water table. Underpinning of foundations. Sealing between piles on a contiguous pile wall construction.

VARIATIONS IN INSTALLATION TECHNIQUE Horizontal Jet Grouting It is possible to install jet grouted columns in the horizontal plane. This technique has been used in various parts of the world to form a protective portal arch over a soft ground tunnel excavation.

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13.9

LIME COLUMNS

Lime columns are formed by physically mixing unslaked lime (CaO) with the soil. This mixing is achieved in-situ by means of a specially designed mixing tool fitted on the end of a hollow kelly driven by an auger machine. The system is mainly suited to the improvement of soft clay or silt profiles. Lime columns can be used in these soil types to improve bearing capacity and reduce settlement. They have also been used in a secant pile form for lateral support as well as for embankment stabilisation. The addition of lime also increases the permeability of the clay and lime columns can be used as drains for the acceleration of consolidation. Positive features • • • •

Versatile and flexible system Fast and economical in certain soil profiles It is a vibrationless system. Noise levels are limited to engine noise only.

Negative features • •

Bulk supplies of unslaked lime must be economically available. Equipment not readily available in Southern Africa.

THE STABILISATION EFFECT OF LIME The shear strength of the soil decreases during the mixing operation. Thereafter a short term increase in the shear strength of stabilised clays and silts is caused by flocculation of the clay and by a reduction in the water content. This is followed by a long term increase in shear strength resulting from various pozzolanic reactions when the lime reacts with silicates and aluminates in the clay. Observations have shown that approximately one third of the final shear strength is achieved after one month and seventy five percent after three months. The undrained shear strength of lime stabilised clay will increase from 10 to 50 times the initial shear strength. In favourable conditions undrained shear strengths of up to one MPa have been achieved. INSTALLATION TECHNIQUE Lime columns normally have a diameter of 500 to 600 mm. The spacing of the columns depends on the application. For lateral support they can be installed in a contiguous layout. Under foundations a spacing of between 1200 and 2000 mm is common. Larger spacings are sometimes used for other applications. The equipment used is normally a powerful top drive drill rig or an auger rig. The machine is fitted with a hollow kelly which allows the powdered lime to be blown down it to the mixer head at the bottom. The mixer head is the same diameter as the required lime

180

column and consists of a pair of blades either side of the kelly located on a diameter. The blades are shaped so as to obtain good mixing of the lime with the soil. The installation commences with the drilling of the mixer blade into the ground down to the full depth of treatment. With the blade rotating the lime is blown in as the kelly is gradually withdrawn. The withdrawal rate has to be controlled so that the lime content is relatively constant throughout the length of the column. The quantity of lime is normally in the range 6 to 8 percent relative to the dry weight of the soil. Other materials such as gypsum, cement and fly-ash can be blended with the lime and have given improved results in organic clays with high water content. LIME COLUMN APPLICATIONS Lime columns have been used extensively in countries such as Sweden, Norway and Finland where there are considerable depths of soft clay. The main applications are: Foundations for Structures Lime columns with a diameter of 500 to 600 mm are installed in groups under the foundation footings in a similar layout to piles. The columns are not, however, designed to carry the full foundation load but act in combination with the surrounding soil to provide the required bearing. The solution is only suited to lightly loaded structures as the strength of the lime columns is not high. The lime columns act more as settlement reducers and to reduce the differential settlement. Lateral Support Lime columns can be installed to form a secant pile wall for lateral support. A typical application is in the form of a circular cofferdam which acts as the pit for a pipe jacking operation. The pipe can be jacked through the lime columns which is not the case with concrete piles. Lime columns can also be installed behind a sheet pile or diaphragm wall to improve the stability of the wall as well as in embankments. Drains Lime columns installed in a deep soft clay profile under a road embankment will not only increase the settlement rate but will also act to control the settlement. Columns placed outside the loaded area will contribute to increasing the overall shear resistance on any potential failure surface. For more information on lime columns refer to Broms (1993).

181

14.0 FACTORS INFLUENCING THE SELECTION OF A LATERAL SUPPORT SYSTEM In this context a lateral support system is considered to be a system which provides stability to any surface excavation or constructed slope (basement excavations, cut slopes, fill slopes etc. ). With such a broad definition it is apparent that there is a wide range of lateral support systems that could be used for slope stabilisation. Certain authors (Tomlinson, 1970) consider the selection and design of lateral support systems to be an art rather than a science. Whether one agrees with this or not it must be accepted that the selection, design and installation of a lateral support system requires the application of considerable skill and experience and sound engineering judgement. The following information can be considered as the minimum required to evaluate lateral support requirements for a project: • • •

A detailed layout of the proposed surface excavation or constructed slope. Detailed geotechnical information including shear strength parameters. Knowledge of the site and its environs.

Using this information the designer will need to consider the following aspects in evaluating the most suitable lateral support system: Site Conditions • • •

Topography before and after construction. Possible variations in site geology or soil profile close to the site. Ground water conditions.

Proposed Development • • •

Geometry and depth of the excavation or slope. Proximity of excavation to site boundaries. Possibility of incorporating the lateral support system into the permanent works.

Adjacent Developments • • • • •

Presence of buried services in close proximity to the site. Surcharge loading (dead, live or transient). Feasibility of installing support systems into adjacent property. The extent of ground movement acceptable during and after construction. The sensitivity of adjacent developments to ground movements.

182

Contractual • • • •

Access to and from site for plant. Availability of skills and plant to install the lateral support system. Availability of materials for the lateral support system. The remoteness of the site.

With the knowledge of the above facts an initial evaluation of a suitable lateral support system can be made from the details given in SECTION 16.0. SUMMARY DETAILS OF LATERAL SUPPORT SYSTEMS but this should be checked using the more detailed information given in SECTION 17.0. TECHNICAL DETAILS OF LATERAL SUPPORT SYSTEMS. As can be seen from the above there are numerous factors that need to be taken into consideration in deciding on a suitable lateral support system and the assessment of these factors will be difficult for someone not experienced in lateral support. An incorrect choice can lead to delays with resulting extra costs to a project. If there is any doubt contact the nearest Franki office for expert advice and assistance.

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15.0 CLASSIFICATION OF LATERAL SUPPORT SYSTEMS The different lateral support systems which can currently be installed by Frankipile South Africa have been classified in accordance with the system proposed by O'Rourke and Jones (1990). EXTERNALLY STABILISED SYSTEMS An externally stabi1ised system uses an external structural wall against which stabilising forces are mobilised. The structural wall forms an integral part of the overall system and is either designed as a cantilever to provide the primary method of support or to act as a support system against which braced or tie back forces are mobilised. Typical examples of externally stabilised systems are shown in Figure 15.1. A cantilever wall is probably the simplest method of providing support. The wall relies on the resistance generated by penetration below the excavation depth to provide the required support. The wall system should be installed before excavation commences and should remain in contact with the soil at all times. The disadvantages are that the cantilever system is generally only suitable for limited depths of excavation (3 to 5m), and besides the wall stiffness itself, there is no other positive method of controlling adjacent ground movements due to excavation. Space restrictions can also limit the maximum width of the wall, which in turn limits the height and stiffness of the cantilever . For excavations deeper than 3 to 5m additional stability can be provided by braced or tied back support systems. Bracing is usually in the form of horizontal or raking props. In recent years post-stressed anchors have become the most popular tied back support system. This type of system allows a positive force to be mobilised onto the wall element, which in turn transfers the force onto the excavation face. A major advantage is that the post stressing procedure can be used to limit adjacent ground movements. Anchored walls are suitable for stabilising a wide variety of soil and rock types. Walls with multiple anchors have been used to support very deep excavations (25m or greater) and excavations with deep seated failure surfaces. In most multiple anchor wall applications, the anchors generally act as the primary support system to provide overall stability, with the wall system acting as a structural element against which the anchor forces are mobilised. The various elements associated with externally stabilised systems have been classified as follows: WALL SYSTEMS Steel sheet piles Steel soldiers Concrete soldier piles Contiguous arid secant pile walls Diaphragm walls

(17.1) (17.2) (17.3) (17.4) (17.5)

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BRACED AND TIED BACK SYSTEMS Prop supports Post stressed anchors Anchor piles

(17.6) (17.7) (17.8)

The number given on the right refers to the specific sub-sections in SECTION 17.0. TECHNICAL DETAILS OF LATERAL SUPPORT SYSTEMS in which detailed descriptions are given for the individual support systems. The more important factors relating to each of the lateral support systems are summarised in tabular form in SECTION 16.0. SUMMARY DETAILS OF LATERAL SUPPORT SYSTEMS. This allows for a quick and easy comparison between the various systems when evaluating a suitable system for a specific project.

Figure 15.1. Examples of Externally Stabilised Systems

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INTERNALLY STABILISED SYSTEMS An internally stabilised system involves the installation of reinforcing elements within a soil/rock mass which extend beyond any potential failure surface. This systems relies on shear transfer along potential failure surfaces to mobilise stabilising forces (in tension, shear or bending) within the reinforcing elements. The fact that shear transfer has to take place along a potential failure surface, implicitly implies that some movement is required to mobilise stabilising forces within the reinforcing elements. Figure 15.2. shows typical applications for the following three internally stabilised lateral support systems that can be provided by Frankipile South Africa. GeoNails. Reticulated micropiles. Soil doweling

(17.9) (17.10) (17.11)

The number on the right refers to the specific sub-sections in SECTION 17.0. TECHNICAL DETAILS OF LATERAL SUPPORT SYSTEMS in which detailed descriptions are given for the individual support systems. The more important factors relating to each of the lateral support systems are summarised in tabular form in SECTION 16.0. SUMMARY DETAILS OF LATERAL SUPPORT SYSTEMS. This allows for a quick and easy comparison between the various systems when evaluating a suitable system for a specific project. Although there are some fundamental differences in the mechanical action of the three internally stabilised systems, the illustrations given in Figure 15.2 show that there are circumstances where more than one technique may be used as lateral support. Research work (Jewe111980 and 1991) has, however, shown that internally stabilised systems work most efficiently when the reinforcements are angled across a potential failure surface so that they are loaded mainly in tension. The following general conclusions may therefore be arrived at in this regard: • •

•

Where a steep slope is to be excavated (detail (a) in Figure 15.2) it is most efficient to use GeoNails installed close to the horizontal. To stabilise such a slope reticulated micropil.es will require a much higher density of reinforcement. In marginally stable relatively flat slopes (detail (b) in Figure 15.2) where overall stability must be improved, then either GeoNails or reticulated micropiles could be used. Other factors (access for plant, geological conditions) may be more significant in deciding on a final reinforcement system. In flat slopes of soft clay soils for example, where stability is governed by a well defined failure surface (detail (c) in Figure 15.2) then large diameter soil dowels would be most appropriate.

186

Figure 15.2 -Examples and applications of internally stabilised Lateral Support systems

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16.0 SUMMARY DETAILS OF LATERAL SUPPORT SYSTEMS Ref.

Type

Nominal Size (mm)

Nominal Spacing (m)

Additional Secondary Support

Normal max. Depth that can be supported (m)

Continuous

None

1 to 2.5 1 to 2.5

Timber lagging or gunite Gunite

Contiguous

None

Continuous

None

Cantilever- 3 Braced 10 Anchored- 15+ Cantilever- 3 Braced 10 Anchored -20+ Cantilever- 4 to 5 Braced 10 Anchored- 25+ Cantilever- 4 to 5 Braced 10 Anchored- 25+ Cantilever -4 to 6 Braced 10 Anchored- 25+

1.0 to l.5 vertical and horizontal 0.15 to l.0

Gunite

15

None or gunite None or gunite

8

EXTERNALLY STABILISED 17.1. Steel sheet piles Per supplier's detail 17.2

Steel soldiers

17.3

Concrete soldier piles

Standard H section, RSJ or channel profiles 300 to 1200 diameter

17.4

Contiguous and secant pile walls

300 to 1200 diameter

17.5

Diaphragm walls

Width 400, 600, 800, 1000, 1200 INTERNALLY STABILISED 17.9 GeoNails 80tol00 diameter 17;10 17.11

Reticulated micro- piles Soil doweling

80 to 250 diameter 450 to 1200 diameter

1.0 to 3.0

Note: For specific details on bracing and anchoring refer to SECTIONS 17.6 TO 17.8.

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8

SUMMARY DETAILS OF LATERAL SUPPORT SYSTEMS Vertical load bearing capacity Poor Fair Good Good Good Poor Good Good

Establishment costs

Cost per Noise m2 pollution

Site area Flexibility in the event of required Obstructions Water and collapse

Medium Medium

High Medium

Medium Medium

Poor Poor to Fair

Good Poor

Medium Medium Large Small

Poor Poor Fair to Good Good

Poor Fair to Good Good Poor

Small Medium

Good Poor

Fair Fair

Medium Medium High Low Medium Medium Medium

High High if driven, else low Medium Low High Low High Low to Low to Low Medium Medium Low Medium Low to high

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17.0 TECHNICAL DETAILS OF LATERAL SUPPORT SYSTEMS 17.1

STEEL SHEET PILES

One of the easiest and quickest ways in which to form a watertight retaining wall in soft or loose saturated soil profiles is to use steel sheet piles. These are steel sections which have the facility to interlock one with another and which can be driven into the ground to form a watertight wall. The sheet pile sections can be extracted once they have performed the function for which they were installed and this can reduce the cost considerably. The steel sheet piles can either be used as cantilever walls or as braced/tied back walls. Positive features • • •

Fast method for forming a wall or cofferdam in soft or loose saturated soil profiles Sheet piles can be extracted and used many times thus reducing costs. The extraction of temporary works is often a requirement satisfied by steel sheet piles.

Negative features • • • •

Sheet piles are imported and are expensive. Delays in procurement can delay the start of a contract. Installation needs to be carried out by persons skilled in the operation. High noise levels associated with installation.

Typical uses for which sheet piles are installed are : • • • • • •

Temporary cofferdam for the construction of the pilecap for a pier in the river. Temporary cofferdam for the construction of a pump house below ground level. Temporary cofferdam for the construction of a basement to a building. Permanent wall as part of the construction of a harbour quay. Permanent cut-off wall to restrict the flow of ground water. Temporary or permanent retaining walls.

The cost of steel sheet piles is high so their use is generally only justified economically for temporary support of excavations below the water table in soft saturated soil profiles. Despite the high costs, sheet piles are often used for certain permanent works or in situations where the speed of installation is a distinct advantage. STEEL SHEET PILE SECTIONS Steel sheet piles are hot rolled steel sections with a shape typical of that shown in Figure 17.1.1. Each sheet pile has a pair of clutches formed in a way that allows the male clutch of one sheet to interlock with the female clutch of another. When a number of steel sheet piles are interlocked they form a wall commonly referred to as a steel sheet pile wall.

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Each manufacturer has a range of sizes to meet various requirements. Figure 17.1.1 shows such a range from one of the European manufacturers. The sheet pile sections can be ordered in whatever length is required. There is, however, a maximum recommended length which is dictated mainly by the handling and driving. These lengths are given for each section profile in Table 17.1.1. Moment of Inertia of Section 4 cm per metre width 4110 13513 23885 39831 49262 92298

Maximum recommended Length metres 6 14 18 23 24 26

Table 17.1.1 -Maximum recommended lengths of steel sheet piles Ninety degree corners, forty five degree corners and junction sheet piles are also manufactured so as to enable the installation of sheet pile walls to a desired plan configuration such as a rectangular cofferdam with cross walls. Other shapes can be fabricated locally from standard sheet pile sections. Steel sheet pile sections are not manufactured on the African continent and are normally imported from Europe. They can be manufactured from various grades of steel as set out in Table 17.1.2 in which the recommended working stresses are also quoted. A small amount (0.25 to 0.50 percent) of copper can also be included in the steel for the purpose of increasing the steel's resistance to corrosion. The length of a sheet pile can be successfully extended by welding on an additional piece. Grade of Steel

Ultimate Stress MPa

Grade 43A Medium Tensile Grade 50/B High Yield

430/510 510/590 490/620

Working Stress Permanent Works MPa 125 160 170

Working Stress Temporary Works MPa 155 195 215

Table 17.1.2 -Details of Grades of Steel and Working Stresses When shipped from the manufacturer the sheet pile sections are plain and uncoated. Should there be a requirement to coat the sheet piles with a protective coating, this is normally carried out locally or on site. The piles are sandblasted prior to applying the coating. A tar epoxy is one of the most effective coatings.

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Figure 17.1.1. Belval steel sheet pile sections

INSTALLATION TECHNIQUES Steel sheet piles are normally installed by driving them into the ground. The driving can be achieved using a vibratory type hammer, a drop hammer or any diesel, hydraulic or air hammer. Prior to driving, the sheets have to be lifted up one at a time and threaded one end into the other. A crane is used for this purpose and a scaffold platform is often constructed to

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provide support for the assembled sheets until such time as they are driven. For smaller cofferdams it is advisable to assemble the whole cofferdam prior to driving. Driving has to be carried out in a controlled manner as the sheet piles have a tendency to fan out, resulting in non-verticality in the direction of the line of the wall. A slight nonverticality is not critical on a straight wall but it becomes a problem when the wall has to make a 90 degree corner. These problems can be overcome with planned pre-assembly and controlled driving where all the sheets in a section are driven down one or two metres at a time. External lateral force can be provided to the head of the sheet pile by means of a steel cable and a jack as an additional measure to control verticality. If verticality is not maintained in driving a section of sheet pile, then this can be corrected by driving a special tapered pile. This should be considered when approaching a comer or when closing a cofferdam. Tapered piles can be fabricated from standard sheet pile sections. When using a conventional hammer a helmet is provided to protect the head of a sheet pile during driving. The helmet should be close fitting and rigid if it is to distribute the blow of the hammer evenly over the area of the section. The buckling of the heads of steel sheet piles during driving can occur in very hard driving. With vibratory hammers the hammer clamps the steel sheet pile hydraulically and thus there is no need for a helmet. Vibratory hammers are limited, however, in their driving ability in certain soil profiles. Conventional hammers should thus be used in hard driving conditions or where the vibratory hammer fails to drive the sheet pile to the required depth. The toe of a sheet pile can be damaged when driving through a cobble or boulder layer or when driving onto an uneven hard bedrock surface. Measures to limit the risk of damage include the use of sheet piles made from high tensile steel, reinforcement of the toe of the sheet pile by welding on additional strengthening plates and reducing the energy per blow of the hammer. Most steel sheet pile sections have a small amount of play in the clutches so a steel sheet pile wall can be installed on a radius. Special straight web sheet piles can be used to form a circular cofferdam. The interlocks between the sheet piles, if installed correctly, are watertight. Leakage on a clutch can occur if the correct driving procedures are not followed. A split clutch in which the male clutch disengages from the female during driving can also occur with poor driving technique. Steel sheet piles can be extracted from the ground once they have performed the function for which they were installed. Various types of equipment are available for extracting steel sheet piles the most common being a vibratory extractor. Impact extractors powered by air

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or diesel are also available and are more suited when the extraction proves to be very difficult. A large crane, in combination with an extractor, is used to provide the extracting force. Plates 17.1.1 and 17.1.2 show steel sheet piling operations in progress. LIFE OF STEEL SHEET PILES The life of a permanent steel sheet pile wall depends very much on the corrosiveness of the surrounding environment and what protective measures are used. Unfortunately the coast of Southern Africa has a highly corrosive environment and thus steel sheet piles are not often used for permanent marine structures with a long life expectancy. Where they are used for marine work, the section of sheet pile above the seabed level should be coated with a good protective layer. In the zone above low water mark the protection coating should be of the highest quality. In fresh unpolluted water the life of unprotected mild steel sheet piles is reported to be in excess of 80 years. This can be increased by using high yield steel and/or copper content. Above the minimum water level the protection given to the sheet pile should be matched to the corrosiveness of the environment. An examination of steel sheet piles extracted a considerable period after installation has shown that very little, if any, corrosion takes place below the ground level even with marine structures. VARIATIONS IN INSTALLATION TECHNIQUE Composite sheet pile walls The use of conventional sheet pile sections is limited by the size and moment of inertia of the sections. By reinforcing the sheet pile wall by means of additional steel sections which are welded to the sheet piles, the depths that can be retained can be increased. Another variation is the use of circular steel pipe piles in combination with sheet pile sections to achieve greater depths or retainment. Figure 17.1.2 shows how these two alternative variations look in plan. They are often used on marine construction. POTENTIAL PROBLEM AREAS Split clutches A split clutch, in which the male clutch disengages from the female during driving, can occur with poor driving technique. This can be a serious problem which may necessitate the injection of chemical grouts behind the sheet pile wall to cut off the water flow through the opening formed by the split clutch.

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Sealing on uneven bedrock surface The sealing of a steel sheet pile wall on an uneven bedrock surface is another potential problem area where loose sand overlies the bedrock. A positive solution is to chemically grout the sand overlying the rock and behind the wall before excavation takes place. Draw down of surrounding soil The driving of steel sheet piles tends to compact the loose soil either side of the wall causing settlement which is referred to as draw down. When driving steel sheet piles immediately alongside a structure with shallow foundations, this draw down can cause foundation settlement resulting in cracking of the structure. The draw down settlement can also result in overturning of adjacent boundary walls. Vibratory hammers are likely to cause more draw down that conventional hammers. Noise pollution Driving steel sheet piles with a conventional hammer is an extremely noisy operation and should not be attempted in environmentally sensitive areas. Driving with a vibratory hammer is less noisy than with a conventional hammer. Encasing the leader of the piling rig with sound proofing materials has been successful in reducing the noise associated from the driving operation to an acceptable level.

Figure 17.1.2 Composite Sheet Pile Wall Sections

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Plate 17.1.1. Driving steel sheet piles

Plate 17.1.2. Partially completed steel sheet pile wall

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17.2

STEEL SOLDIERS

Over the years steel soldiers comprising single or twin H-sections, joists or channels have been commonly used for cantilever walls, braced or tied back walls or as a structural element to transfer anchor forces onto the face of an excavation. Positive Features • • • • • •

Driven steel soldiers can be used in soft unstable soil profilesHand over hand techniques can be used where driving or predrilling is not possible. Driven or predrilled soldiers can be incorporated into the permanent works. Driven or predrilled soldiers can be designed to carry vertical loads. Use of twin sections facilitates installation of anchor systems. The profile of steel soldiers facilitates the installation of secondary support between soldiers.

Negative Features • • •

Steel sections are relatively expensive. Noise levels are high if sections are driven. Limited stiffness for cantilever walls.

INSTALLATION TECHNIQUE Steel soldiers which are used either as a cantilever support system or as an anchored structural element are generally installed prior to the commencement of excavation. The steel sections can either be driven into the ground or they can be placed in an auger drilled hole. With the latter technique the steel section is embedded in concrete below the excavation level and a weak grout above the excavation level. The weak grout facilitates subsequent trimming during the main bulk excavation. The type of installation procedure that is adopted will generally depend on the nature of the ground conditions. In soft or loose unstable soil profiles, driving is the preferred procedure. When deciding on which steel section to use, due consideration must be given to the possibility of the section buckling during driving if it is too slender. Other negative factors are noise problems associated with driving, the limited bearing area for passive resistance below excavation level and possible problems with minimum depth requirements. The latter two factors are particularly significant where the steel soldiers are to be used as cantilever support systems, or where the soldiers are required to carry vertical loads. Installation by predrilling will be mainly suited to soil profiles that remain stable during the predrilling operation. For more specific details regarding installation techniques reference should be made to SECTION 7.4 STEEL H-PILES.

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Steel soldiers are generally installed at 1.0 to 2.5 metre spacing. This implies that the soil being supported has some inherent stability and will arch between the soldiers. For certain soil profiles (stiff clays or dense residual soils) arching between the soldiers may provide sufficient secondary support for short and medium term stability. Soft clays and even some non-cohesive soils may remain stable for sufficient length of time to enable secondary support to be provided between the soldiers. Timber lagging or a lightly reinforced 50 to 100 mm thick gunite skin is generally used for secondary support. Where the steel soldiers are to be used with anchors it is common to use twin sections welded together with a suitable gap between the sections to allow for anchor installation. The anchor head is then stressed directly onto the twin soldier section and in this way the anchor forces are transferred to the excavation face without the use of expensive waler systems. Under certain circumstances it may not be practical to drive the soldiers or to predrill a hole. Problems in this regard are usually associated with access difficulties or ground formations where driving or predrilling is not possible. Under these circumstances pre- stressed anchors are usually used as a tie back system to provide overall stability, with steel soldiers being used as a structural element to transfer the anchor forces onto the excavated face. Soldiers and anchors are then installed as the excavation proceeds, adopting a construction procedure commonly referred to as the "hand over hand" method. This requires that the excavation be formed in shallow stable benches with the anchors and soldiers for each bench being installed as soon as possible after excavation. After stressing of the anchors the next bench is excavated and the procedure repeated with the soldiers being welded together to form a continuous vertical structural element. Depending on the nature of the ground conditions it may be necessary to provide temporary stability by forming berms at a suitable slope angle with the soldiers and anchors being installed in slots formed through the berms. Once all the anchors are stressed, the berm can then be completely removed and the procedure repeated at the next level. This procedure is obviously only suitable for anchored walls and cannot be used for cantilever walls. Experience with soldier installation procedures for anchored walls has shown that, in instances where adjacent ground movements are to be kept to a minimum, driving or predrilling of the soldiers is preferable to using the hand over hand construction technique. A further advantage for deep excavations supported with multiple anchor systems, is that installation by driving or predrilling can also speed up overall excavation and lateral support construction procedures. In particular, the cantilever support provided by the driven or predrilled soldiers, allows excavation to proceed rapidly to a depth of2.0 and 3.0 metres before installation of the first row of anchors. The stiffness provided by steel soldiers also allows unrestricted excavation to be carried out between rows of anchors, thus avoiding time delays associated with berm and slot type installation procedures. Plate 17.2.1 shows a typical steel soldier pile installation with anchor tie backs for the support of an excavation.

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STRUCTURAL DESIGN CONSIDERATIONS In certain instances the steel soldiers may be incorporated as part of the final structure. This will generally require some form of corrosion protection for the soldiers. Secondary support between the soldiers is then usually provided by a suitably designed structural gunite skin spanning between the soldiers. Where the soldiers are driven or predrilled, they can also be designed to carry vertical loads. This will, however, require suitable penetration below the excavation level. In the design of the steel soldiers the maximum axial, bending and shear stresses during all phases of construction should be considered and the design should then be carried out in accordance with the appropriate Code of Practice. Plastic design methods are used under certain temporary conditions. Where the steel soldiers are to be incorporated as part of the final structure, the final structural loading and support conditions may be the controlling factors in the design.

Plate 17.2.1 Steel soldiers with anchor tie backs and timber lagging

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17.3

CONCRETE SOLDIER PILES

As for driven or predrilled steel soldiers concrete soldier piles can be installed to provide support either as cantilever walls, braced or tied back walls or to act as structural elements against which anchor forces are mobilised. Positive Features • • • •

Good flexibility in relation to available pile types and diameters. Large diameter piles can provide increased stiffness for certain applications. Easily incorporated into permanent works. Can be designed to carry vertical loads.

Negative Features • •

Sleeves within the pile or a waler system is required for anchor installation. Large diameter piles will encroach into available space.

INSTALLATION TECHNIQUE Concrete perimeter piles are usually installed prior to commencement of excavation at between 1.0 and 2.5 metre spacing. Usually a 0.3 to 1.5 metre gap is left between piles. This implies that this system is most suited to relatively stable soil profiles which allows the soil to arch between the piles for at least a sufficient time period for secondary support to be installed. Taking this into consideration, the following pile types are usually the most suitable for use as concrete soldier piles: • • • •

Auger piles CFA piles Forum bored piles Franki piles

Reference should be made to SECTION 7.0. TECHNICAL DETAILS OF PILING SYSTEMS to obtain specific details for these pile types, the soil conditions for which they are most suited and installation methods. Where the concrete soldier piles are to be used as a cantilever support system or where the piles are required to carry vertical loads, it is necessary to ensure that sufficient embedment of the pile shaft is obtained below the bottom of the excavation. Whether this can be achieved needs to be carefully evaluated in relation to the soil profile and the type of pile being used. For anchored walls the anchors can be installed between piles, with loads being transferred to the piles with a suitable waler system. Alternatively steel sleeves can be attached to the pile reinforcing to allow for the installation of anchors through the piles. This is generally a less expensive option since walers are then not required. This procedure does, however,

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require that precision work to be carried out during pile installation to ensure the sleeves are at the correct level and orientation. A further alternative is to use twin channels, joists or H-beam sections as pile reinforcement. These sections are welded together with a suitable gap and, after excavation to a selected level, the front concrete of the pile is broken away to expose the steel sections. The anchors are then installed through the sections. This procedure is generally only suitable where auger piles are used in a stable soil profile although special techniques can also be adopted to use this procedure with Franki, Forum bored and CFA piles. As for driven or predrilled steel soldier piles experience has shown that, for deep excavations with multiple anchor systems, the use of concrete soldier piles can significantly reduce adjacent ground movements and speed up overall excavation and lateral support construction procedures. The greater stiffness associated with concrete soldier piles is more effective than steel soldier piles in this regard. As indicated above the use of concrete soldier piles requires that the soil is stable enough to arch between the piles for at least a short time period. For certain soil profiles (stiff clays or dense residual soils), arching between the soldiers may provide sufficient secondary support for short and medium stability. In less stable soil profiles (soft clays or loose soils etc. ), secondary support in the form of a lightly reinforced 50 to 100 mm gunite skin is provided between the piles. Plate 17.3.1 shows a typical concrete soldier piles with anchor tie backs. STRUCTURAL DESIGN CONSIDERATIONS In most cases the soldier piles are incorporated as part of the final structure. Secondary support between the piles is then usually provided by a suitably designed structural gunite skin spanning between the piles. Where vertical loads are to be carried, it is necessary to ensure that the piles are installed to a sufficient depth below final excavation level. In the design of the piles the maximum axial, bending and shear stresses during all phases of construction should be considered and the design should then be carried out in accordance with the appropriate Code of Practice. Where the piles are to be incorporated as part of the final structure, the final structural loading and support conditions may be the controlling factors in the design.

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Plate 17.3.1 -Concrete soldier piles with anchor tie backs

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17.4

CONTIGUOUS AND SECANT PILE WALLS

A contiguous pile wall is simply a row of concrete soldier piles installed so that each pile is in contact or near contact with piles on either side of it. The positive and negative features given in SECTIONS 17.3. for concrete soldier piles, are therefore also applicable to this system. There is, however, a relatively large increase in cost for this system in comparison with concrete soldier piles. The technique is therefore generally only used in unstable soil profiles (soft saturated clays or sands) which do not have an ability to arch between adjacent piles. In certain instances the increased overall stiffness of a contiguous pile wall as compared to steel soldiers or concrete soldier piles, may be a significant factor in reducing the risk of adjacent ground movements. Under these circumstances, a contiguous pile wall may be preferred even when relatively stable soil conditions occur. This wall is not water tight unless some specific steps are taken to achieve this, so leaching and/or piping of non cohesive soils through gaps between the piles can be a problem below the water table. A special case of a contiguous pile wall is called a secant wall. Here the piling is carried out in a sequence in which subsequent piles are cut into the previously installed piles thereby affecting a seal between the units. INSTALLATION TECHNIQUE A contiguous pile wall is generally selected in preference to steel soldiers or concrete soldier piles due to the unstable nature of the soil profile or the presence of a high water table. The pile type selected for this system should therefore be capable of dealing with these conditions. In general terms, bored pile systems such as augured underslurry, CF A, Forum bored or temporary cased auger piles are preferred. Franki piles can be used where there is a particular reason for requiring a driven pile type. Reference should be made to SECTION 7.0. TECHNICAL DETAILS OF PILING SYSTEM to obtain specific details for these pile types. Whichever pile type is used, the installation sequence is to first install odd numbered piles and, when these have set, the even numbered piles. The spacing between the odd numbered piles has to be carefully chosen so as to allow unobstructed installation of the even numbered piles, whilst at the same time limiting the gap between the piles. Special measures must be taken to improve the plan and verticality tolerances of pile installation. A plan tolerance of 1:25 mm and a verticality tolerance of 1: 100 should be possible in soil profiles free of obstructions. The following techniques can be used to improve water tightness of a contiguous wall. • •

The gap between the piles can be drilled and grouted in situ. Odd numbered piles can be installed slightly back of the centreline and cast in a cement bentonite mix. Even numbered piles are then drilled so that they cut into the

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•

bentonite cement columns thus forming a seal. This solution can only be used with an auger/bored pile system (Auger Underslurry or CFA piles). The contiguous piles can be formed in a secant wall manner where even numbered piles cut a secant into odd numbered piles thus forming a seal. This solution can only be used with an auger/bored pile system (auger underslurry or CFA piles) or an oscillator pile system.

Typical details used for contiguous and secant pile walls are shown schematically in Figure 17.4.1. A contiguous auger pile with anchor tie backs is shown in Plate 17.4.1. A contiguous cantilever wall formed with CF A piles is shown in Plate 17.4.2. STRUCTURAL DESIGN CONSIDERATIONS Due to the expense associated with the installation of a concrete contiguous pile wall the system invariably forms part of the permanent works. In the design of the piles the maximum axial, shear and other stresses during all phases of construction as well as those due to final structural loading should be taken into consideration. Design should then be carried out in accordance with the appropriate Code of Practice.

Figure 17.4.1- Typical contiguous and secant pile wall details

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Plate 17.4.1 -Contiguous auger piles with anchor tie backs

Plate 17.4.2 -Contiguous cantilever wall formed with CF A piles

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17.5

DIAPHRAGM WALLS

A diaphragm wall is a reinforced concrete wall constructed in the ground using underslurry techniques. Walls with widths of between 300 and 1200 mm can be formed in this way to depths in excess of 30 metres. Positive Features • • • • • • • • •

Walls can be installed to considerable depths. Walls with substantial thickness can be formed. The system is flexible in plan layout. The wall can easily be incorporated into the permanent works. The wall or certain sections can be designed to carry vertical load. Basement construction time can be reduced. Economical, positive solution for large deep basements in saturated and unstable soil profiles. Noise levels limited to engine noise only. No vibration during installation.

Negative Features. • •

Not normally economical for small, shallow basements. The system needs a relatively large site area.

Under certain conditions diaphragm walls may be used as cantilever, braced or tied back walls in preference to the systems discussed in the previous sections. The conditions that usually dictate that diaphragm walls should be used are as follows: • • •

In very unstable soil profiles below the water table where continuous support and watertight conditions are required to prevent mud flows, piping and erosion of the soils. Where construction time is important and the use of a diaphragm wall can shorten the programme. In conditions where deeper than normal cantilever support may be required. These conditions could occur where the wall is to act only as a cantilever, or where a very deep initial excavation is required before the first braced or tie back supports can be installed.

INSTALLATION TECHNIQUES A diaphragm wall is constructed in a series of separate adjoining panels with each panel keyed into the adjacent panels. The reason for constructing the wall in panels is the limitation on the length of excavation that will remain stable under a head of bentonite. The minimum panel length is dictated by the size of the excavation grab. The stability of the trench sidewalls, the plan shape of the panel and problems associated with the flow of concrete during concreting control the maximum length of panels.

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For anchored walls, steel sleeves are attached to the reinforcing cage to allow for the anchors to be installed through the wall panels. Typical details for diaphragm wall construction are given in Figure 17.5.1. A completed diaphragm wall with anchor tie backs is shown in Plate 17.5.1. STRUCTURAL DESIGN CONSIDERATIONS A diaphragm wall is usually designed as part of the permanent structure. In the design of the wall, the maximum axial, bending and shear stresses during all phases of construction as well as those due to final structural loading, should be taken into consideration. Design should then be carried out in accordance with the appropriate Code of Practice.

Figure 17.5.1 -Typical details for diaphragm wall construction

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The correct layout of panels is an essential part of a diaphragm wall design. The layout must allow effective excavation and concreting of the panels and must take into consideration factors such as the size of the grab, the plan shape of the wall, the dimensions of steel cages and site logistics. Conventional diaphragm wall construction commences with the construction of a pair of guide walls, one on either side of the main wall. These guide walls are usually formed in concrete and are about 1.2 metres deep. They provide guidance to the excavating grab, support to the sidewalls at the surface as well as a convenient platform for controlling the concreting operation. The internal guide wall is removed during the main excavation. The construction of the diaphragm wall may be started in more than one place with the initial panels known as starter panels. A panel constructed adjacent to a another panel is termed an intermediate panel, and one constructed between two existing panels, a closure panel. When concreting a panel, the ends of the panel have to be formed so that the one keys into the other. To achieve this a shaped steel form is placed in position at one or both ends of the panel prior to concreting. These forms are known as "stop ends". Once the concrete has taken its initial set, the stop ends are gradually withdrawn leaving the end face of the concrete panel with the desired key. Starter panels need to have two stop ends, intermediate panels only one and closure panels have no stop ends. Excavation is carried out under a head of bentonite slurry by means of a grab suspended from a crane in a similar manner to that described for underslurry barettes under SECTION 7.7 UNDERSLURRY PILES. The breadth of the grab is generally between 2.0 and 3.0 metres and the widths between 300 and 1200 mm. Panels are normally 2 to 5 metres in length and thus require more than one pass of the grab. The trench is kept topped up with bentonite slurry during both the excavating and concreting operations. Once excavation of the panel is complete, the bentonite in the trench is processed to reduce the density and adjust the pH. The required stop ends are placed in position followed by the steel reinforcing cage and the tremie pipe. A pump is located in the trench for pumping the slurry back to storage tanks during concreting. The concrete operation is carried out using normal tremie concrete techniques and a concrete mix with a 200 mm slump. The level of the concrete should be cast at least 750 mm above the required cut-off level. Once the concrete has taken its initial set any stop ends are gradually extracted leaving the end face of the concrete panel with the desired key. The panels are excavated and concreted according to the planned sequence until the full diaphragm wall is complete. Where the diaphragm wall is to be used as a cantilever support system or where the wall is required to carry vertical loads, it is necessary to ensure that there is sufficient embedment below the bottom of the excavation. In many instances the wall is also designed to act as a groundwater barrier and a minimum embedment depth is then usually required for this purpose.

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Plate 17.5.1 Diaphragm wall with anchor tie backs

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17.6

PROP SUPPORTS

A bracing system comprising prop supports can be used to support the various wall systems described in SECTION 17.1 TO 17.5. A system of horizontal struts to provide cross bracing is common in trench excavations and other excavations of limited width. Inclined raking struts are used to support walls where the distance is too great for horizontal struts. These systems are illustrated in Figure 17.6.1. The struts can be made of steel, concrete or timber depending on the loads to be carried. Positive Features • • •

No encroachment into adjacent property. No specialist expertise is required for installation of the system. Simple and quick construction procedure for smaller excavations.

Negative Features • •

Restricts access and construction working space. Specialist procedures such as pre-stressing of struts may be required to limit adjacent ground movements on larger excavations.

INSTALLATION TECHNIQUES The first phase of construction comprises the installation of one of the wall systems described in SECTION 17.1 to 17 .5. With horizontal struts excavation usually proceeds until the first level of support is reached. If necessary a horizontal waler is attached to the wall and the strut is tightly attached to the waler or directly onto the wall. The excavation then proceeds to the next level. Horizontal and vertical spacing of the struts is a function of the support forces required and the type of wall system that has been used. To limit deflections, attention must be paid to construction and design details regarding stiffness of the struts and the connection between the struts and the walers or wall. Struts can be pre-stressed in instances where it is necessary to specifically restrict adjacent ground movement. Pre-stressing is usually carried out to about fifty percent of the anticipated working load. Inclined raking struts are used to support walls where the distance is too great for horizontal strut support. The usual installation procedure is to phase the excavation so that a berm is left behind to support the wall element. The foundation system for the raking strut is then installed and, if necessary , a horizontal waler is attached to the wall. The wall is then braced by installing the raking strut with suitable connections to the foundation and the wall. This is illustrated in Figure 17.6.1. If required, the berm is lowered and the next level of raking struts installed. This will be dependent on the type and height of wall to be supported. Horizontal spacing of the struts will be a function of the support forces required and the type of wall being supported. To control deflections, attention must be paid to the construction and design details regarding stiffness of the struts and the

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connection between the struts and the wall. Particular attention must be paid to the design of the foundation system for the raking strut. Under certain circumstances it may be necessary to transfer large horizontal forces into the foundation system. In poor soil conditions a piled foundation system may be required to support the raking struts. VARIATIONS IN INSTALLATION TECHNIQUE Raking piles can be installed to act as props as illustrated in Figure 17.6.1. This procedure can speed up basement excavation since it is not necessary to leave berms and wait for the raking struts to be installed before removal of the berms.

Figure 17.6.1 Typical Prop Support Systems

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17.7

POST STRESSED ANCHORS

Post stressed anchors are frequently used as tie backs for lateral support of deep excavations. Anchors can be used as a tie back for one of the wall systems described in SECTIONS 17.1 to 17.5. An anchor system can also provide primary support with the wall providing secondary support as well as acting as a structural element which transfers the anchor forces onto the excavated face. Positive Features • • • • •

Post stressing will assist in limiting adjacent ground movements. Suitable for most soil and rock types. Usually the only procedure that can stabilise deep-seated failures. High stabilising forces can be mobilised with multiple anchor systems. Provides an unobstructed area for basement construction.

Negative Features • • • • •

Anchors will encroach into neighbouring property. Considerable expertise required for drilled anchor installation. A structural element is required to transfer anchor forces onto the excavated face. The corrosion protection of permanent anchors is expensive. De-stressing of temporary anchors must be allowed for.

ANCHOR TYPES The various components used for post stressed anchors are shown in Figure 17.7.1. The following components and construction procedures are used by Frankipile South Africa for post stressed anchors: • •

•

Tendons comprising threaded bars or strandThe procedures used to form the fixed anchorages are shown in Figure 17.7.2. Dead man fixed anchorages (Type A) are used under specific conditions generally when the anchorage position is close to ground surface. A straight shafted gravity grouted anchor (Type B) is used in very stiff/very dense soils or in rock. A re-injectable anchor using a high pressure grouting procedure (Type C) is the most common system adopted since this procedure is suitable for most soil and rock types. The fixed anchorage is formed by hydro fracturing and/or compaction of the surrounding soil or rock. The high pressure grouting is generally carried out using a tube-a-manchette system. The free anchorage is normally formed by suitably sheathing the tendon.

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•

The anchor head consists of a device which can post stress the tendon against a suitable bearing plate and lock the tendon at the required load. Various anchor wedge systems are available for strand anchors. A nut is used for threaded bars.

Anchors can be used for temporary or permanent applications. In general terms a temporary anchor will require little or no corrosion protection whereas specific attention will have to be paid to corrosion protection for permanent anchors. The terms temporary and permanent are dependant on a number of factors (service life, corrosion environment, consequences of failure etc. ) and it is therefore not possible to provide an exact definition. The classification system given in the SAICE Geotechnical Division Code of Practice (1989) Lateral Support in Surface Excavations is recommended as a classification system in this regard. This classification system is reproduced as Table 17.7.1. The procedures that can be used for corrosion protection are many and varied and require high quality construction procedures. Some guidelines for corrosion protection procedures are given in Table 17.7.1. Where permanent anchors are required it is recommended that discussions be held with GeoFranki to decide on a suitable system for a specific project.

Figure 17.7.1. Components for post stressed anchors MANUFACTURE AND TRANSPORT Where possible anchor capacity, fixed lengths and free lengths for a project should be predetermined prior to commencing site operations. This enables the anchors to be preassembled off site with obvious quality control benefits, especially in the case of permanent anchors where attention to detail is essential to ensure reliable corrosion protection measures. Transport and on site storage are equally important to ensure the anchor integrity prior to installation.

213

Figure 17.7.2. Procedures to form fixed anchorages

214

SERVICE LIFE

TEMP. Up to months

six

Six months to two years PERM. Over two years

RESULT FAILURE No 0

Not serious

CONDITIONS PERTAINING TO OF CORROSION MATERIAL ENVIRONMENT AROUND FREE LENGTH No. No. No 0 Non0 Rock A corrosive

1

Serious

1

Corrosive

1

Sand

B

2

Catastrophic

2

Very corrosive

2

Clay

C

LATER ACCESS

No. X

Needed / Restressable anchor Not needed / nonstressable anchor

Y

TOTAL THE SCORES IN THE COLUMNS AND REFER TO THE PROTECTION ALTERNATIVE TYPE CODE BELOW PROTECTION ALTERNATIVES FIXED ANCHOR LENGTH Type Code Treatment 0-1, ABC, XY Tendon grouted bare 0-6. AY 2-3, ABC, XY Tendon epoxy coated 4-5, ABC, XY

5-6, ABC, XY

Tendon grouted into corrugated sheath while grouting anchor

Tendon pregrouted in corrugated sheath

FREE LENGTH Type Code Treatment 0, AXY Bare tendon empty hole 1-6, AY Bare tendon grouted in hole 0, BC, XY 1, Sheathed tendon A, XY in empty hole

HEAD AND UPPER TENDON Type Code Treatment 0-1, ABC XY Bare head

1, BC, XY 2, ABC, XY 3-6, ABC, XY 5-6, ABC, XY

2, ABC, XY

Painted head

3-4, ABC, X

Epoxy head

Sheathed tendon in grouted hole

2-6, ABC, X

Greased and sheathed tendon in empty hole Greased and sheathed tendon in grouted hole Greased and sheathed tendon in common sheath in grouted hole

5-6, ABC, Y

Head covered with cap filled with grease Head concreted or grouted into box out

coated

FOR EXAMPLE: Temp. Anchors up to four months in very corrosive conditions in sand with serious consequences of failure and tested at one month intervals. Score: O + 1 + 2 = 3BX FIXED ANCHOR LENGTH: Tendon epoxy coated before grouting into holes. FREE LENGTH: Tendon greased and sheathed before grouting into hole. HEAD AND UPPER TENDON: Epoxy coated. Table 17.7.1 Corrosion protection guide. After SAICE Geotechnical Division Code of Practice (1989) Lateral Support in Surface Excavations

215

INSTALLATION TECHNIQUES Pre-assembled anchors are installed into predrilled holes drilled by rotary or percussion techniques. It is important that holes are thoroughly cleaned to ensure efficient bond between the anchor grout and in-situ rock or soil. For this reason it is normal to drill holes 0.3 to 0.5 metres deeper than the actual total anchor lengths to provide a sump for drilling debris. Temporary casings can be used during the drilling operation in instances where stability of the drilled hole is of concern. Plate 17.7.1 shows a crawler rig installing temporary anchors. Grouting procedures are adapted to suit site conditions and the procedures used must ensure full grout cover between the anchor and the surrounding soil or rock without the risk of grout contamination. Where necessary temporary or permanent casings are used to ensure that this is achieved. Normal practice is to home the anchor system into a fully grouted hole. High pressure grouting of the fixed anchorage using a tube-a-manchette system has become a common procedure for most anchor installations in soils and soft or fractured rocks. The tube-a-manchette forms part of the overall anchor assembly. The stressing and testing of anchors is one of the most important phases of any anchor installation. During this phase each anchor is tested to a specified percentage above its design load and the performance of the anchor recorded on a stress/strain graph which is compared to predicted performance. In addition, creep relaxation is recorded which allows a prediction of the long term behaviour of each individual anchor. The correct stressing and testing of anchors is an extremely important aspect in the overall installation procedure. By adopting the correct procedure each anchor on a project is tested to a specified percentage above the required working loads. This effectively confirms the suitability of the entire anchor assembly with the exception of the corrosion protection aspects. The standard stressing procedures for different anchor types and applications are well documented in the SAICE Geotechnical Division Code of Practice (1989) lateral support in surface excavations and reference should be made to this document in this regard. ANCHOR CAPACITY The procedures used for anchor installations are such that anchor capacity is usually controlled by the permissible tendon stress which should be determined in accordance with the recommendations given in the SAICE Geotechnical Division Code of Practice (1989) for lateral support of surface excavations. This requires that the stress of the tendon when locked off should not exceed 70% of the characteristic strength and that during stressing and testing the stress should not exceed 80% of the characteristic strength. This later criteria is usually the controlling factor for both temporary and permanent anchors which are usually stressed to 125% and 150% of working load. Typical anchor working loads using this criteria is presented in Table 17.7.2.

216

Typical Working Load (kN)

Tendon System

Temporary

Permanent

20 mm

105

85

25 mm

165

135

1 strand

170

140

2 strands

340

280

3 strands

510

420

4 strands

680

560

5 strands

850

700

Threaded Bars Characteristic Strength 525 MPa

Strand 15.7 mm 12 mm

No movement Design parameter

100% 7 - 15%

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18.2

COLLAPSIBLE SOILS

Nature of the Problem Behaviour of soil with a collapsing fabric was first studied by Knight (1961) and the basic concept of collapse settlement is illustrated in Figure 18.2.1.

Figure 18.2.1 -Mechanism of collapse settlement Several conditions must be satisfied for collapse settlement to occur: • The soil must have a collapsible fabric. Soils of low in-situ dry density which are silty or sandy commonly exhibit collapsible fabric• An initial condition of partial saturation must be present. This condition is applicable to the upper horizons of the soil profile in most areas of Southern Africa. • An increase in moisture content must occur so that a loss of shear strength of bridging colloidal materials can be effected• The imposed pressure exerted on the soil fabric by the structure must exceed the overburden pressure. Problems associated with construction on collapsible soils are not only confined to buildings with shallow foundation structures but to roads, airfields and railways as well as earth dams and reservoirs. Distribution of Collapsible Soils As with expansive soils, collapsible soils occur in both transported and residual soils. Since the problem of collapse can occur in all types of transported soil, foundation problems due to collapse can occur anywhere in Southern Africa but more commonly and with greater severity in areas where aeolian sands have been deposited. Deep deposits of Kalahari silty sands in the arid western regions produce collapsible soil up to 20 metres in thickness. Residual soils with a collapsible grain structure are mainly confined to Granites

234

of the Basement Complex. These collapsing Granites are confined to old erosion surfaces and areas with an annual water surplus. Figure 18.1.1 in shows areas where collapsing soils are common. Evaluation and Prediction The identification and quantification of collapse settlement of the soil fabric requires detailed field identification of the collapsible horizons as well as laboratory or in-situ testing to quantify the magnitude of collapse settlement. The recording of the soil profile discussed in SECTION 3.1 is the first step in the correct identification of the problem. Careful inspection with a hand lens as well as identification of the origin of the soil will substantially aid the engineer in identifying the collapse phenomenon. Inspection of the damage to surrounding structures will also be invaluable in alerting the investigator to the problem of collapse. Oedometer testing in the laboratory is the most common method used to predict collapse settlements. The collapse potential test is the simplest of the types of oedometer test used and provides the engineer with a measure of the collapse potential but not a parameter for estimating collapse settlement. Table 18.2.1 provides a guide for the severity of collapse. Table 18.2.1 -Collapse Potential Collapse Potential Severity of Problem 0% -1% No problem 1% -5% Moderate problem 5% -10% Problem 10% -20% Severe problem > 20% Very severe problem Single Oedometer testing is used with the sample at natural moisture content loaded to the anticipated applied stress from the structure and then soaked. The measured consolidation due to collapse, provides a conservative estimate of the likely collapse settlements. The double oedometer test developed by Jennings and Knight (1958) is a fairly complicated procedure requiring corrections of the consolidation curves and careful interpretation. Plate load testing methods in which a 300 mm diameter plate is loaded to the anticipated bearing pressure and the surrounding soils then soaked, can be successfully used to predict soil modulus values at natural moisture and soaked conditions. The collapse settlements can be calculated for the specific structure and applied loads from the measured soaked modulus. Engineering Solutions Soil treatment methods for collapsing soils generally revolve around methods of compaction either by removal and recompaction or by compaction in-situ. For shallow

235

collapsible soil profiles, compaction using vibratory or impact rollers can be considered and has been used successfully for collapsible soil depths up to 1,5 m. The use of this method has not always been successful and trials should be carried out to ensure the desired compaction can be achieved with the equipment proposed. Dynamic compaction described in SECTION 13.2 is a technique which has been greatly advanced recently and can be successfully used for most collapsing profiles up to 10 m deep. Other methods of compaction such as Vibroflotation have been used successfully and can be considered. For collapsing profiles > 2 m deep dynamic compaction will provide the most cost effective solution for medium and large areas requiring treatment. Like expansive soils, engineered solutions for collapsible soils comprise either piled foundations or stiffened rafts. •

•

18.3

Most methods of piling can be used for collapsible soil profiles but because the profile is generally partially saturated and slightly cohesive, auger piles or driven cast in-situ piles are generally the most economical pile type. The driven cast in situ pile has the advantage that compaction of the surrounding collapsing soil is achieved during pile installation. Design methods for raft foundations on collapsing soils have been developed by Lytton (1972). Tromp (1979) has used the method of assuming a "soft spot" of chosen diameter. A soft spot is an assumed area of collapsible soil softened by a localised increase in moisture content. Raft types such as the Boucell is ideally suited to collapsing soils since the raft is equally stiff in both directions of bending and is therefore not sensitive to the position of the soft spot. SOFT CLAYS

Nature of the Problem As a problem soil type soft clays have no specific connotation to the Southern African region and are more widely distributed in other areas of the world such as Scandinavia. These soils exhibit very low shear strength, high compressibility and lead to severe time related settlement problems. In Southern Africa these clays are often partially saturated and overconsolidated. Some typical undrained shear strengths and compressibility values associated with these soils which are mainly confined to the eastern seaboard of Southern Africa, are given in Table 18.3.1. The problems associated with these soils are stability and settlement related. Instability and large settlements for heavy loadings such as road embankments, present engineering problems to infrastructural developments. Most building structures located on these soils demand a piled foundation solution.

236

Area Deposit South Coast Durban Richards Bay Area South Coast Durban Richards Bay

Table 18.3.1 Typical Properties of Soft Clays Typical shear strength values Material Undrained CU Description (kPa) c'(kPa) Black silt clay 10 - 25 5 Black silt clay 10 - 25 0 Sandy silty clay 15 - 35 10-20 Black silty clay 10 – 20 5 – 15 Typical compressibility values Material Description Compressibility mv (m2/MN) Light grey sandy clay 0.3 - 1.0 Black silt clay 0.5 -1.5 Black silt clay 0.5 - 2.0 Light grey sandy clay 0.2 - 1.0 Black silt clay 0.5 - 2.0

Drained φ' 25 25 25-30 20 - 25 Consolidation cv(m2/vear) 5 - 10 1-S 0.5 - 2.0 5 - 10 1-5

Evaluation and Prediction Testing and evaluation of these soils is done using normal site investigation techniques and soil mechanics principles. Testing using small diameter rotary cored boreholes with in-situ SPT and shear vane tests in conjunction with CPT tests are the most cost-effective methods of investigation. There are numerous empirical and rigorous methods for establishing geotechnical design parameters for these soils from the in-situ or laboratory test results. Engineering Solutions There are few methods available to increase the strength or stiffness of soft clays but dynamic compaction and/or dynamic replacement methods have been used successfully in estuarine sediments which are generally slightly cohesive with lenticular deposits of silts and clays. Lime columns have been used in other parts of the world to cement soft clay horizons and enhance their properties. A piled foundation is generally used for medium and heavy structures in areas underlain by soft clays, the piles being founded on bedrock or dense sand horizons underlying the soft clays. The choice of pile types suited to soft clay conditions can be assessed using the parameters set out in SECTION 4.0. For piles of low and medium capacity , driven jointed precast piles have been used extensively in the Durban and Richard's Bay areas. Large diameter bored piles founded on or socketed into bedrock have been used for heavily loaded structures. In certain areas where a sand stratum overlies the soft clays it is sometimes possible to found light structures on Franki piles with an enlarged base or on footings founded on soil improved by Vibrocompaction.

237

Negative skin friction due to settlement of soft horizons under surcharge loading can occur in these soft clay profiles and the additional load imposed on the pile must be taken into account when carrying out the pile design. The alternative is to eliminate the effects of negative friction by isolating the pile shaft using Shell pileslip which is a bitumen product developed specifically for this purpose. This has been done in the Richard's Bay area. For embankments and deep fills where piling is not an economical option, several methods of construction have been successfully used. The installation of vertical drains to shorten drainage paths and rapidly dissipate pore pressures is economical but requires phasing of the construction process and careful monitoring of movements and pore pressures. The use of preloading is both economical and effective if the necessary pre-planning can be implemented. Slope flattening or berms are often used to decrease the shear stress where shear strength is a problem during the construction phase. 18.4

DOLOMITES

Nature of the Problem Solution cavities within water soluble Dolomitic rock masses result in the formation of sinkholes or subsidences being formed above the cavities due to changes in the ground water regime. Lowering of the water table with the resultant removal of water from voided areas of the Dolomitic bedrock causes subsurface erosion into the voids by infiltration of surface water. Cavities within the dolomitic residuum move upwards during this erosion process until the cavity reaches the surface resulting in the formation of a sinkhole. A compaction subsidence as opposed to a sinkhole occurs where the cavities are filled with highly compressible Dolomitic residuum called WAD. The phenomenon is graphically illustrated in Figure 18.4.1.

Figure 18.4.1 -Typical dolomitic condition

238

Distribution of Dolomitic Rocks The occurance of Dolomitic conditions is limited to areas underlain by this rock type. Dolomitic rocks of the Chuniespoort, Campbell and Witwatersrand groups are limited to the Transvaal and Northern Cape and are shown in Figure 18.4.2.

Figure 18.4.2 -Distribution of Dolomitic rocks in Southern Africa Evaluation and Prediction The investigation of dolomitic areas and the evaluation of the stability of these areas requires personnel with expertise and experience in this field of geotechnical engineering. The investigation of dolomitic areas incorporates both geophysical and direct methods of investigation. Gravity surveys are the most common and reliable geophysical method available and are carried out in conjunction with direct methods such as percussion drilling, auger trial holes and backactor test pits. The use of borehole cameras lowered into small diameter percussion drilled holes facilitates the inspection of areas where cavities are likely. Classification and evaluation of dolomitic areas is generally done on an empirical basis. Classification of sites is done according to Wagner's (1982) method where the depth of overburden provides the basis for the site being designated as a Class A, B or C site. The average thickness of overburden C from ground level to top of pinnacles gives classifications which reflect the risk of instability as outlined in Table 18.4.1.

239

Table 18.4.1 -Dolomite Classification Class A : Class B : Class C :

Pinnacle and boulder dolomite at or near the surface. C 15 m

Engineering Solutions No economical methods of soil treatment have been developed for Dolomitic areas since the detection and backfilling of voided areas is extremely difficult. Control of surface water to prevent erosion into cavities as well as strict control of groundwater movement are the most effective methods of prevention. Engineered solutions comprise either the formation of a raft or the use of piled foundations. The use of soil rafts or mattresses developed by Wagner (1963) is commonly used and is suitable for light and medium structures. Piles are generally used for heavy and movement sensitive structures. The three pile types generally suited to the difficult conditions encountered are oscillator piles of large diameter and capacity , predrilled precast concrete piles and predrilled steel H-piles. The predrilling is carried out under a head of drilling foam using large diameter down-the-hole percussion hammers. Driven cast in-situ piles founded in a dense horizon of sufficient thickness can also be cost effective as a shallow piled solution for light and medium structures.

240

19.0 ENVIRONMENTAL ENGINEERING Frankipile South Africa, through its specialist divisions Soiltech and GeoFranki, offer the following range of services related to the engineering of the environment: • • • • •

Core drilling for the interpretation of fracture patterns and aquifers. Permeability testing. Ground water monitoring and sampling. Monitoring and sampling of surface water bodies. Containment / remediation of contaminated areas.

Core drilling and permeability testing are envirotechnical services that also form part of procedures used in general geotechnical investigations. The techniques adopted for these two procedures are described in detail in SECTION 2.0 GEOTECHNICAL INVESTIGATION. The other services are specific to envirotechnical operations and are discussed in detail in this section. 19.1

GROUNDWATER MONITORING

The frequent reporting in the media of major pollution scares, the general tightening of legislation regarding environmental issues and the greater public awareness of pollution, has highlighted the need to monitor groundwater for potential contamination. For far too long industry has not been aware of the problem of groundwater contamination or conveniently ignored it. This situation is now changing and industry has to get its house in order. To firstly ascertain whether the groundwater is contaminated, samples of water have to be obtained. Tests on these samples will show the degree of contamination and the area over which contamination has occurred. The groundwater can also be monitored over a period of time to determine the possible source of the pollutants and other factors such as flow patterns and the effectiveness of remedial measures. This involves the installation of monitoring wells and the use of specialised electronic equipment. Soiltech provides a complete service with respect to sampling and monitoring of groundwater. Installation of Wells The wells are installed using a variety of techniques depending on sub-surface conditions. Rotary percussion or auger drilling are the most common methods of well installation. The work must be carried out to stringent standards to prevent cross contamination of the samples taken from the various wells. For this reason Soiltech operates a quality assurance programme which is implemented in accordance with ISO 9000 requirements. Work instructions and quality control procedures for individual contracts are prepared to suit client requirements and engineering specifications, thereby assuring the end-user of the highest quality results.

241

Monitoring Soiltech collects its water monitoring data via the computerised GrantlYSI 3800 water quality logger unit. This equipment records conductivity, pH, dissolved oxygen, salinity, Eh (ORP) and temperature. The electrode readings are compensated automatically for temperature and pressure. Readings are collected via a sonde, using pump purging techniques within the monitoring wells. Pump purging is carried out along guidelines supplied by the Water Research Commission and Department of Water Affairs and Forestry. The pumping equipment consists of a Grundfos MP 1 environmental submersible pump. Field data can be either printed out in tabular form directly from the logger unit, or down loaded into a P .C. with the ability then to graph and analyse data further. The ongoing monitoring of water and industrial ,effluents can assist in determining the source of pollution incidents so that remedial action can be taken timeously. The need for expensive and time-consuming laboratory sample testing is reduced, since field staff can determine on site whether certain basic parameters lie outside pre-defined limits, allowing straightforward interpretation and trend analysis. When water samples are required for additional laboratory testing, these are collected and preserved according to the latest specifications for transit to approved analytical laboratories. If necessary , the monitoring can be supplemented by the collection and analysis of contaminated soils using Soiltech's geotechnical equipment. The various sampling procedures are discussed in SECTION 2.0 GEOTECHNICAL INVESTIGATION. 19.2

MONITORING OF SURFACE WATER

The same techniques as described for groundwater monitoring are used for the monitoring of open surface water bodies, except that samples are taken directly from the surface water body. 19.3

CONTAINMENT / REMEDIATION

With the growth in awareness of the consequences of pollution to groundwater systems, the need to control groundwater flow by the construction of hydraulic barriers has received renewed attention. In groundwater pollution control, the barrier often performs two functions. The first is to isolate the polluted groundwater by controlling the seepage into downstream areas. The second is the containment of the contaminated water allowing in-situ treatment by dewatering or recirculation in the case of stabilisation applications.

242

In the last decade, Franki have constructed various forms of hydraulic barriers to control groundwater flow. Slurry walls, a diaphragm walls and steel sheet pile walls have all been used for this purpose. 19.3.1

SLURRY WALLS

Slurry walls are generally employed to control ground water movement in soil horizons for pollution control measures as listed above, or as hydraulic cut-off for an earthfill dam or similar structure. They can be constructed under dry stable conditions as well as in unstable soil profiles below the water table. Installation Techniques Where the depth of the slurry wall is less than six metres, it can be excavated as a continuous trench using standard trench excavating equipment. For deeper walls, the excavation is carried out in vertical panels utilising grabbing techniques, as described in SECTION 17.5 UNDERSLURRY WALLS. During excavation the level of the slurry in the trench must be kept a minimum of 1.5 metres above the level of the surrounding water table. For the shallow trenches, either a cement/bentonite or a sand/bentonite slurry can be used. With a sand/bentonite slurry , the trench is excavated initially using a conventional bentonite slurry .On completion of the excavation, the sand/bentonite slurry is discharged into the trench using a tremie, and the bentonite slurry pumped to storage. If a cement/bentonite slurry is chosen, the slurry is left in the trench to set after completion of the excavation, eliminating the need to change the slurry . For deeper walls excavated by grab, preference is given to a cement/bentonite slurry due to its self hardening nature. This makes it possible to excavate into a previously constructed panel without endangering the stability of the previously constructed work. As mentioned above, the cement/bentonite slurry is left in the trench to set after completion of the excavation. The trench can vary in width between 400 and 1200 mm with 600 mm being the most common. Depths in excess of 20m are achievable using grabbing techniques. With excavations deeper than 12m, special precautions are often necessary to ensure the verticality and effective overlapping of adjacent panels. Mix Design The prime consideration of a slurry mix is a low penneability which remains stable under the geotechnical and groundwater conditions anticipated on site. For this reason it is essential to carry out a geotechnical investigation of the site which includes the sampling of the ground water. Permeabilities of the order of 10-6 -10-8 cm/sec can be achieved, with a sand/bentonite mix generally having a lower permeability than the cement/bentonite alternative. In designing a sand/bentonite mix, it is imperative that a well graded "dirty" sand is used to prevent the

243

risk of internal piping of the bentonite particles between the coarser grained sand particles. A silt content of at least 20% should be aimed at. With a cement/bentonite slurry, the quantity of cement will depend on the required strength and flexibility of the cut-off wall. A higher cement content will result in a higher strength slurry with increased brittleness and reduced flexibility. The use of slagment in combination with cement, partially offsets these negative factors with the added benefit of a reduction in permeability. The slurry mixing procedure is important and will greatly control the performance of the product. The following basic procedure must be adhered to:• • •

Bentonite must be fully hydrated with an uncontaminated water prior to use. The cementitious addition should be colloidally mixed with uncontaminated water prior to blending with the bentonite. A homogeneous mix must be obtained by thorough blending and mixing prior to introduction to the trench.

Chemical Resistance In the case of polluted groundwater, the resistance of the bentonite slurry to the contamination in the groundwater must be checked, preferably by testing to ensure the long term performance of the cut off wall under site conditions. Variations in Installation Technique While a slurry trench can effectively control the horizontal migration of water into or out of the area to be isolated, water flow can occur through the base of the contained area or into the contained area from surface. A permeable rock formation below the slurry wall may require grouting to extend the cut-off to competent rock. A clay or similar capping may be necessary to control the inflow of water into the isolated zone as well as outflow. If a capping for the containment area is envisaged, or the barrier is to be extended above surface by constructing a concrete or earth wall, attention should be paid to the detail of joining the two barriers systems. In the case of the cement/bentonite slurry, allowance should also be made to accommodate the inevitable cracking that occurs in the upper 300 to 500 mm of the cement/bentonite slurry. 19.3.2

OTHER BARRIER SYSTEMS

In addition to the above, steel sheet piles (See SECTION 17.1 ), diaphragm walls (See SECTION 17.7), grouting and dewatering provide possible alternatives for controlling groundwater flow, be it for contamination control, water preservation or flood control. It is recommended that discussions be held with Frankipile to optimise the barrier best suited to the circumstances of a given situation.

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20.0 DESIGN AIDS: PILING The capacity of a piled foundation to support load is governed both by the structural strength of the pile itself and by the strength of the soil surrounding the pile shaft and the base of the pile. The structural strength of the pile is controlled using materials with known properties and the design is carried out using structural design principals assuming the pile to be a column subjected to vertical and lateral loading with the soil providing restraint. The capacity of the soil to carry the loads transferred to it by the pile is influenced by several factors, the most important of which are : • • •

The soil type and its stress history. The strength and stiffness of the soil. The method of installation of the pile which results in changes to the stress regime and strength of the soil surrounding the pile.

20.1

PILE CAPACITY TO RESIST COMPRESSIVE LOAD

20.1.1

P1LE BEHAVIOUR UNDER LOAD

The behaviour of a pile under load is a complex soil-pile interaction problem and rigorous methods have not, as yet, been developed to model pile behaviour. Structural loads are generally imposed at the head of the pile. This load is transmitted along the pile shaft and transferred into the surrounding soil. At small loads, the transfer of load occurs almost entirely along the pile shaft in friction. With increased load, friction transfer approaches the ultimate while an increasing share of the load goes onto the base of the pile. Full pile shaft capacity is mobilised at relatively small deflections (less than l5mm) while full pile base capacity is mobilised at relatively large deflections (approximately 10% of the pile base diameter). The proportioning of load transfer into the pile shaft and pile base will depend on several factors, the most important of which are the pile type and geometry , method of installation and soil profile. Figure 20.1.1 shows the idealised behaviour of a pile under increasing load. From start to point A the load is resisted almost entirely by friction on the pile shaft. Between A and B the friction still increases slightly but reaches a maximum, whereas the end bearing resistance starts to build up. If the load is removed at this stage the pile head will recover to virtually its original position indicating an elastic behaviour. At point C, where the pile toe deflection is approximately ten percent of the pile base diameter, the end bearing resistance has reached its ultimate and the pile is on the point of failure. Small increase in load from C to D results in a large increase in the pile head deflection as well as pile toe deflection. On removal of the load the pile will not recover to its original position due mainly to the plastic deformation at the base of the pile.

245

Figure 20.1.1 -Idealisation of pile behaviour The static calculation of a pile's ultimate capacity considers the contribution of pile shaft and pile base separately and the basic equation for the ultimate pile resistance is given as. Qp = Qs + Qb

(20.la)

The static method of calculation can be used for both soil displacement and replacement type piles while dynamic methods of calculation using pile driving formulae are limited to driven displacement type piles. See SECTION 20.14. Equation 20.1 a ignores the mass of the pile itself in contributing to the applied load which is a valid assumption for normal applied loading and pile geometry.

246

20.1.2

STATIC CALCULATION OF PILE CAPACITY

Equation 20.l a outlines the procedure adopted in determining the ultimate pile capacity. The implicit assumption that the pile's shaft friction capacity (Qs) and base end bearing capacity (Qb) are not interdependent, is valid for normal pile geometry. The evaluation of the ultimate shaft capacity Qs is carried out by integrating the pile-soil shear strength along the shaft using the equation: τ = ca + σn tan φs (20.lb) where

τ

=

pile-soil shear strength.

ca

=

pile shaft adhesion

σn

=

normal stress between pile and soil.

φs

=

angle of friction between pile and soil.

The ultimate base capacity is calculated using bearing capacity theory with the equation: Qb = Ab (cuNc + σv Nq +0.5 γ dNγ) Where

(20.lc)

Ab

=

area of base.

cu

=

cohesion of soil.

σv

=

vertical stress in soil at pile base.

γ

=

unit weight of soil.

d

=

pile diameter

Nc,Nq & Nγ

=

bearing capacity factors dependant on soil properties and pile geometry.

Calculation of pile capacity is divided into three main soil categories : • • •

Cohesive soils Non-cohesive soils Rock

A further classification based on the method of installation of the pile is generally used in the static method of calculation. Two broad methods of pile installation are considered : Soil Displacement piles and Soil Replacement piles. In this section, soil displacement piles will be referred to as driven piles, and soil replacement piles as bored piles. Values of strength and compressibility parameters for the various types of soil and rock are given in SECTION 3.3. PILES IN COHESIVE SOILS For piles in clay, the undrained capacity is generally taken to be the critical value. The undrained shear strength Cu is used for the calculation of both the ultimate shaft and base capacities. In stiff over consolidated clays the drained rather than the undrained load capacity may be the critical value and effective stress parameters can be used.

247

Base Capacity The base end bearing capacity of both driven and bored piles is given by the equation : Qb = Nc Cu Ab Where

(20.1d) Cu

=

undrained cohesion at the pile toe

Ab

=

area of pile base

Nc

=

bearing capacity factor genera1ly = 9 for penetration of at least five pile diameter into the bearing stratum. Nc values up to 20 have been measured for driven piles with an expanded base.

Figure 20.1.2 shows the variation of Nc with the depth of penetration after Skempton (1951).

Figure 20.1.2 -Bearing Capacity factors after Skempton (1951)

248

Shaft Capacity Equation 20.le gives the ultimate pile shaft capacity in cohesive soils: Qs = Ca As Where

(20.le) Ca

=

average pile-soil adhesion over pile shaft length

As

=

surface area of pile shaft.

The calculation of the ultimate pile shaft capacity in clay is influenced by the nature of the cohesive soil as well as the method of installation and type of pile. Driven Piles The semi-empirical relationship of the undrained pile-soil adhesion (Ca) and the undrained shear strength (Cu) has been studied by several authors for driven displacement piles and is generally equal or greater than unity for soft clays and decreases markedly with an increase in the undrained shear strength. The relationship giving the shaft adhesion factor α defined in equation 20.lf for varying shear strengths of clay is given in Figure 20.1.3. α=

Ca Cu

(20.1f)

Figure 20.1.3 -Pile Adhesion factors after Tomlinson (1970) for driven piles in clay

249

Bored Piles The skin friction of bored piles is calculated using Equation 20.le. The shaft adhesion factor α relating the pile soil adhesion Ca and undrained shear strength Cu has been extensively studied locally and abroad for both residual and transported clay soils. Values of α once again vary considerably but α is generally between 0.2 and 0.8 with a trend of α increasing in value with a decrease in undrained shear strength. A tabulation of typical values is given in Table 20.1.1. If accurate values of a are required for the determination of pile capacity , pile testing will be required to determine the value for the particular site or measured values for similar founding conditions used. Table 20.1.1 -Typical values of factor a and pile adhesion Ca Undrained Shear Strength Cu (kpa) 0.4

3

m > 0.4

 H  0.28 m 2 n 0.64 Q L σh   = and Ph = 2 2 2 m2 +1  QL  m + n

(

)

)

(

)

 H 2  1.77 m 2 n 2 σh   = m2 + n 2  Q p 

(

)

Section A-A σ΄h = σh cos2 ( 1.1θ ) Figure 22.3.1 Horizontal pressures due to point and line load surcharges

318

22.4

EXTERNALLY STABILISED SYSTEMS

22.4.1

CANTILEVER RETAINING WALLS

A Rankine earth pressure distribution is recommended for cantilever retaining walls. A typical pressure distribution diagram is shown in Figure 22.4.1. It is assumed that the wall will fail by rotating about a point just above the toe of the wall, and that the active pressure is balanced by the passive pressure. Design guidelines are given in Figure 22.4.1. Only pressures induced by active and passive states within the soil are illustrated in Figure 22.4.1. Where water occurs or surcharge loads are present, the relevant water pressure and imposed surcharge pressures must be included (see SECTION 22.3). In the design procedure for the overall stability of a cantilever wall, it is recommended that appropriate safety factors be applied to the soil strength parameters rather than to the passive forces only. A more detailed discussion on factors of safety is given in SECTION 22.6. Computer software, which allows the wall to be modelled as beam elements and the soil as a system of springs (subgrade reaction model), is readily available. This procedure would be the most suitable to determine bending moments and shear forces that need to be resisted by the cantilever wall structural element. For a simple cantilever wall, a first estimate of the maximum bending moment can be obtained by the method given in Figure 22.4.1. For final designs, the values obtained from this procedure need to be checked by a more rigorous analysis. In terms of the structural design, it is recommended that a cantilever wall system be analysed with a factor of safety of 1.0 and that the bending moments and shears derived from this calculation are adopted as "working" moments and shears. These bending moments and shears should then be multiplied by an appropriate load factor for ultimate limit state design to structural codes. 22.4.2

BRACED WALLS

Semi-empirical pressure distributions such as those proposed by Peck (1969), are used to determine the magnitude and distribution of lateral support pressures for braced excavations. These distributions are given in Figure 22.4.2. Strut loads for specific spacings can be estimated directly from the pressure distributions given in Figure 22.4.3. It is important to emphasise, however, that the distributions given are for end of construction conditions and support forces at each excavation level should be considered in sequence. The wall elements are usually designed as continuous members supported at strut levels.

319

(a) - Deflected shape with rotation about point C. (b) - Theoretical pressure distribution. (c) - Simplified pressure distribution. Pp is the required passive resistance for stability and is obtained by taking moments about point C. The available passive resistance for the assumed depth d is then checked and compared with the required Pp If necessary d is adjusted and the calculation repeated. ds is the required penetration depth and is equal to 1.2d to 1.5d. For the simplified pressure distribution shown and ignoring wall friction:

d=

H KP

0.67

−1

L is the span of the wall for the calculation of the maximum bending moment. L= K x H where K is obtained as follows: φ'

20°

30°

35°

40°

K

2.0

1.5

1.4

1.3

Figure 22.4.1 Pressure distribution for a cantilever retaining wall

320

Figure 22.4.2 Pressure distributions for braced excavations after Peck (1969) 22.4.3

TIED BACK W ALLS

Design procedures for tied back walls vary from the relatively simple procedures that are used for walls with a single tie back to complicated soil/structure interaction analyses for multi-tied walls. Either the free-earth or fixed-earth support methods can be used for the design of single tied back walls. A Rankine pressure distribution is generally used in the analysis. Typical pressure distributions and design details are given in Figures 22.4.3 and 22.4.4. The design of multi-tied walls is a complex soil/structure interaction problem in which the earth pressure distribution which governs the design depends on the wall stiffness, the method of construction, the tie back spacing and pre-stress load. Sophisticated programmes such as FREW ( 1987) and WALLAP ( 1989) can be used to carry out rigorous numerical analyses of multi-tied walls. The procedures given below are recommended in the absence of sophisticated software. These procedures should be carried out for all stages as the excavation and tie back installation proceeds. •

Evaluate the magnitude of support force required using limit equilibrium techniques involving the analysis of single or multiple wedges or circular arc failures. A typical single wedge analysis is illustrated in Figure 22.4.4. The single wedge analysis can be used with confidence for routine designs with uniform geometry and simple surcharge loads. Software is available for the analyses of complex geometric problems. In carrying out this analysis suitable safety factors should be applied to the soil shear strength parameters (see also SECTION 22.6).

321

Free-earth support assumes that the passive resistance in front of the wall is sufficient to resist forward movement at the toe but not sufficient to prevent rotation. Free-earth support is recommended for loose cohesionless soils, silts and clays. (a)

deflected shape.

(b)

Assumed pressure distribution.

(c)

Bending moment distribution

T is the required tie back force. Pp is the theoretical passive resistance required for stability of the wall and can be determined by taking moments about E. The available passive resistance for the assumed depth d is then checked and compared with the required Pp. If necessary d is adjusted and the calculation repeated. For equilibrium Σ horizontal forces = 0. This allows T to be determined. For the pressure distribution shown: T=PA-Pp The maximum bending moment is calculated using the assumed pressure distribution and assuming that the wall is simply supported at E and D. Figure 22.4.3 Pressure distribution for free-earth support

322

Fixed-earth support assumes that the passive pressure in front of the wall is sufficient to prevent both forward movement and rotation at the toe. Fixed-earth support is recommended for dense sands and gravels. (a)

Deflected shape.

(b)

Assumed pressure distribution.

(c)

Bending moment distribution.

For analysis purposes the wall is considered to be two equivalent beams EF and FC connected by a pin joint at F. The following relationship between x/H and φ' is used to determine the position of F. φ'(degrees)

20

25

30

35

40

x/H

0.25

0.15

0.08

0.033

–0.01

The tie back force, T, is calculated by equating moments about E to calculate the reaction at F and then determining T for horizontal equilibrium of beam EF. The required depth, d, is calculated by equating moments about C and solving for (d -x). An increase of 20 to 50% is made to the calculated value of d to allow for the length CD. The maximum bending moment is determined by considering the wall as a beam simply supported at E and F. Figure 22.4.4. Pressure distribution for fixed-earth support

323

•

•

•

Check the magnitude of the support forces using a suitable earth pressure distribution. The earth pressure coefficient used in this calculation will vary between K. and Ko depending on anticipated and allowable wall movements and surcharge loading. These coefficients should be determined using an appropriate factor of safety applied to the soil shear strength parameters (see SECTION 22.6). In general terms a triangular distribution of earth pressure will be the most appropriate. Once the procedures given above have been used to determine the magnitude of support force required then the distribution of the anchor forces needs to be decided upon. In less complex cases this can often be decided upon using judgement decisions based on geometric considerations and taking due cognisance of the soil/rock profile and surcharge loads. A simple evaluation of anchor forces can also be obtained by considering the contributory area of the earth pressure diagram applicable to each anchor. More rigorous analyses such as that given by Littlejohn, Jack and Sliwinski (1971) can also be used under appropriate conditions. Consideration also needs to be given to the overall anchor arrangement in terms of other factors such as minimum free lengths and minimum depth below ground surface. For guidelines in this regard reference should be made to the SAICE Code of Practice for Lateral Support in Surface Excavations ( 1989). The support system as a whole should be checked for overall stability. This will comprise a series of analyses in which the system containing the wall and anchors is checked for stability along selected circular and non-circular potential failure surfaces.

The wall elements for both single and multi-tied walls should be designed to resist the induced bending moments and shear forces from the overall anchor and earth pressure system during all stages of excavation and construction. Rowe (1952) has shown that the flexibility of the wall is an important consideration for single tied back walls in which the support forces mobilised are a combination of passive resistance from the earth below excavation level and the force provided by the brace or tie back system. Depending on the flexibility of the wall there is a reduction in bending moment due to redistribution of soil pressures associated with arching of the soil being supported. Typical bending moment reduction factors are given in Table 22.4.1. For multi-tied walls the wall elements can be designed as continuous members supported at tie back positions with a load distribution in accordance with the assumed earth pressure distribution. The most suitable procedure for multi-tied walls is to use a subgrade reaction model where the wall is modelled as beam elements and the soil as a system of springs. In the structural design of the wall elements the overall lateral support system should be analysed with a factor of safety of 1.0 applied to the lateral pressures and the bending moments and shear forces derived should be adopted as "working" moments and shears. These moments and shears should then be multiplied by an appropriate load factor for ultimate limit state design to structural codes.

324

Table 22.4.1. Bending moments reduction factors for single tied back walls after Reynolds (1981) Ratio of wall thickness to span 0.02 0.10 0.20 0.30 0.40

Bending moment reduction factor φ΄=20o φ΄=30o φ΄=35o 0.70 0.56 0.48 0.80 0.69 0.62 0.86 0.78 0.72 0.90 0.83 0.78 0.91 0.87 0.83

Known

Unknown

P is the surcharge W is the weight of sliding wedge = ½ γ H2 cot B γ is the density of wedge H is the depth of excavation S is the shear resistance of retained material along plane of rupture. S = ( cH/sinB) x N tan φ' A is the angle of inclination of the anchor φ' is the effective angle of friction c' is the effective cohesion

T=

B is the angle of inclination of potential plane of rupture (B should be varied and plotted against values of T) T is the anchor force N is the normal force on the wedge N = (P+ γ H/2)HcosB.cotB + T sin (A + B)

(P+ γH/ 2) HcosB(F− cotBtanφ′) − (c′H/sinB) sin (A + B) tanφ + Fcos(A + B)

where F is the factor of safety required. Figure 22.4.5 Forces acting in single wedge mechanism of failure after BS 8081 (1989)

325

22.5

INTERNALLY STABILISED SYSTEMS

The design procedures for internally stabilised systems are no different to any other lateral support system. To guarantee a sufficient margin of safety, equilibrium should be checked for all possible failure mechanisms using safe assumptions for the properties of the soil and reinforcement and taking due consideration of water pressures and surcharge loadings. 22.5.1

GEONAILS

A GeoNail system should be checked for both internal and overall stability in accordance with the procedures given below. These procedures should be applied to all stages of construction. Internal Stability This is the most important design check for a GeoNail reinforced soil mass. Limit equilibrium procedures are generally adopted in which the equilibrium of all possible failure surfaces through the GeoNail reinforced zone are checked. It is essential to demonstrate that equilibrium can be maintained on all possible failure surfaces, and it is not sufficient to focus on a single "critical surface". It is questionable whether a stability analysis using a general method of slices is appropriate to check internal stability of a GeoNail reinforced soil mass. This is due to the unknown influence of the GeoNails on the forces between individual slices. Limit equilibrium analyses using rigid body mechanisms would appear to be the most suitable method for checking equilibrium since there is no need for assumptions on internal slices. The simplest method in this regard is a single wedge analysis. This type of analysis is illustrated in Figure 22.5.1, and can be used with confidence for designs with uniform geometry and soil profiles and simple surcharge loads. A two-part wedge analysis may be appropriate with complex geometric conditions and high surcharge loads, particularly if these loads occur some distance beyond the crest of the excavation. For more specific details with regard to design procedures, reference should be made to Gassler and Gudehaus (1981), Shen et al (1982), Gassler (1988) and Long et al (1990). It is necessary to take into consideration that the internal stability of a GeoNail structure depends on the geometry, surcharge loadings, shear resistance of the soil and the tensile strength and pullout resistance of the GeoNails. Taking these factors into consideration, it is apparent that the concept of a single factor of safety, as applied to conventional slope stability analyses, is not satisfactory for a GeoNail reinforced system. The most appropriate design procedure is to carry out the analysis using the appropriate geometry, water pressures and surcharge loads and to select safe values for the soil shear strength and tensile and pullout resistance of the GeoNails. These safe values are obtained by applying the appropriate factors of safety to the relevant parameters (see SECTION 22.6.). The design process should then demonstrate satisfactory equilibrium by determining the required forces to maintain equilibrium and then designing a GeoNail layout so that the available forces provided by the GeoNail system exceeds the required forces on all the potential failure mechanisms.

326

Q is the surcharge on the wedge. W is the weight of the sliding wedge

FNact =

R is the reaction force on the failure plane.

Tm Sh

n

∑l i =l

i

θ is the angle of inclination of the failure plane. T is the mean force per metre length of GeoNail

Sh is the horizontal spacing of the GeoNail

φ is the angle of shearing resistance of soil.

FNact is the force provided by the GeoNails

FNreq is the required force for stability

li is the length of nails beyond the failure plane.

Figure 22.5.1 Single wedge stability analysis of a GeoNail structure. After SAICE Code of Practice (1989) Lateral Support in Surface Excavations The value of pullout resistance of the GeoNails is probably the most significant factor in determining available forces. The assumption that ultimate bond (τult) is related to the effective cohesion and the normal effective stress multiplied by the effective angle of friction (τult = c' + kσv' tan φ') appears to give excessively conservative values in most instances. Work carried out by Heymann et al (1992) has shown that for insensitive cohesive soils and soft rocks the ultimate bond for GeoNails can be obtained using the same procedures as those used to predict the ultimate skin friction for piles. For recommendations in this regard reference should be made to SECTION 20.0 DESIGN AIDS: PILING. Heymann et al (1992) has also shown that for sandy residual soils of low to moderate plasticity such as residual granites and sandstones, a lower bound value for the ultimate bond tu1t can be obtained from: τult in kPa = 4 < φ'

327

For many soil types there does however not appear to be a satisfactory method of accurately predicting ultimate bond values. For most projects the most satisfactory procedure is to make the best possible design assumptions, carry out suitable in-situ pull out tests as soon as possible after the commencement of the project, and then revise the design. Overall Stability This comprises a check on the stability of the overall GeoNail structure. This is firstly carried out using the analogy that the GeoNail structure acts as a homogenous and resistant unit to support the soil behind in a manner similar to a gravity retaining wall. A second design check should then be carried out comprising a series of analyses in which the overall GeoNail structure is checked for stability along selected external circular and non circular failure surfaces. These concepts are illustrated in Figure 22.5.3

Figure 22.5.3 Design check for overall stability of a GeoNail structure

328

22.5.2

RETICULATED MICROPILES

Design procedures for reticulated micropiles are described by Lizzi (1983 and 1989). In these publications the author stresses that the design of reticulated micropiles is not amenable to rigorous theoretical analysis and that the design approach is empirical and based on engineering judgement obtained from back-analysis of real cases. To quote from Lizzi (1989): "In spite of a very large quantity of studies and calculations, some of them with the help of sophisticated computers, we cannot state that the matter is theoretically under control". The following general design guidelines, which are based mainly on Lizzi (1983) and 1989), should be applied to a reticulated micropile support system: •

•

•

The reticulated micropile structure should be considered to act as a gravity retaining structure. The lateral thrust on the structure due to earth pressure, water pressure and surcharge loads can be calculated from the recommendations given previously in SECTIONS 22.1,22.2 and 22.3. The earth pressure coefficient used in the evaluation of lateral thrust will fall between Ka and Ko but probably closer to Ko. In the determination of the earth pressure coefficient a suitable factor of safety should be applied to the soil shear strength parameters (see SECTION 22.6). The next step in the design process is to choose a suitable geometric layout. The layout usually consists of closely spaced vertical and raked piles (see SECTION 17.10). Lizzi (1983) recommends that sliding stability of the reticulated micropile structure should be resisted by the soil only, through friction and cohesion, and that shear along the piles be ignored. This would seem to be a conservative approach but can be used as a basis in deciding on the overall geometry of the pile layout system at any critical sliding surface. An evaluation can be made of the compressive, tensile and shear forces within individual piles. Lizzi (1983) recommends a procedure, which he considers to be analogous to reinforced concrete, whereby the compressive loads due to the lateral earth pressures are carried jointly by the piles and the surrounding soil in accordance with an amplification factor. This amplification factor is defined as the ratio of the elastic modulus of the pile to that of the soil. The tensile loads induced by the lateral earth pressures are carried only by the piles. These tensile and compressive loads are to be carried only by the length of pile below any critical failure surface. Reference should be made to SECTION 20.0. DESIGN AIDS: PILING for procedures to evaluate compressive and tensile capacity of the piles. Lizzi (1989) indicates that the pile loads calculated are approximate but that this is not particularly significant since the loads are usually small due to the high density of piles. If a pile does become overstressed there will also be load transfer to the remaining piles. A similar design approach to evaluate individual pile loads is described by Dash and Jovino (1980), except that the concept of an amplification factor is not used.

329

•

It is necessary to ensure that there is sufficient stability against sliding on any critical surface. As indicated previously, Lizzi (1983) recommends that sliding be resisted by the soil only through friction and cohesion. This approach is probably conservative. Dash and Jovino (1980) take the opposite viewpoint and assume that sliding is resisted entirely by shear on the piles. It would seem to be sensible to include shear resistance from the piles in analysing sliding stability. This is particularly the case if one considers that the piles are very much stiffer than the surrounding soil and will tend to attract load even if only small movements takes place along potential sliding surfaces.

22.5.3

SOIL DOWELLING

With soil dowelling heavily reinforced large diameter piles are installed into marginally stable slopes to provide stability through shear and bending resistance. The forces that need to be resisted by a soil dowelling system are usually of large magnitude. The failure surface in this instance is usually well defined and the magnitude of the forces to be resisted can be evaluated using conventional slope stability analysis. The recommendations given in SECTIONS 22.1 and 22.6 should be used in carrying out the slope stability analysis. Once the forces to be supported have been determined the design of the soil dowels should be carried using conventional procedures for laterally loaded piled foundations. Reference should be made to SECTION 20.0 DESIGN AIDS: PILING in this regard. 22.6

FACTORS OF SAFETY

In deciding on a minimum factor of safety for the design of a lateral support system the design engineer needs to take the following into consideration: • • • • • •

The reliability of the measured or assumed values of the relevant soil or rock parameters involved in the analysis. The reliability and accuracy of the mathematical model used in the analysis. The magnitude of surcharge loads and the confidence level in the prediction of these loads. Previous experience in similar geotechnical conditions. Previous experience with the lateral support system to be utilised. The consequences of failure. It is convenient to consider failure in this context in terms of limit state theory .The extreme event in this case would be the ultimate limit state where complete collapse of the lateral support system occurs. For most lateral support systems the serviceability limit state will however be the most important criteria. With the serviceability limit state failure can be considered to occur when movement of the ground being supported exceeds an allowable value. This allowable value will obviously be much higher for the support of virgin ground or a street face compared to the support of a heavily loaded, settlement sensitive structure.

330

The above factors have been taken into consideration in compiling a set of recommended minimum factors of safety for various lateral support applications. These recommended minimum values are given in Table 22.6.1. It is important to emphasise that these minimum values should be used as guideline values by the design engineer and each project should be treated on its merits. Table 22.6.1. Guidelines for minimum factors of safety

Type of support system

Recommended minimum factor of safety Temporary Support Permanent Support Low surcharge and/or movement not critical

High surcharge and/or movement critical

Low surcharge and/or movement not critical

High surcharge and/or movement critical

1.25

1.5

1.5

1.5-2.0

1.25

1.25

1.5

1.5 - 2.0

Externally stabilised Soil and rock effective shear strength parameters Tie back capacity Internally stabilised Soil and rock effective shear strength parameters

1.25

GeoNail bond

2.0

Tensile or compressive capacity of micropiles

2.0

1.5

Only applicable under certain circumstances

2.0 - 2.5 2.0-2.5

331

Generally not applicable

22.7

MOVEMENTS ASSOCIATED WITH EXCAVATION

An essential point in considering movements associated with excavation is that no excavation, however well supported, can be made without causing some ground movement. The ground movements are generally due to lateral yield of the soil/rock towards the excavation with an associated component of vertical movement. The magnitude of movement that occurs is related to the height of the excavation to be retained, the nature of the soil/rock strata being retained and the magnitude of surcharge loads. It is possible to exercise some control over the magnitude of the movement by the selection of the correct lateral support system and construction techniques. For example, if a support system is required for an excavation adjacent to a settlement sensitive structure with relatively high surcharge loads then the designer would not choose a passive support system such as GeoNails but rather and active system with post-stressed anchors. The potential for movement can even be further reduced by using closely spaced large diameter concrete soldier piles as a wall element in conjunction with the post-stressed anchors and by adopting carefully phased excavation procedures. Prediction of movements cannot be made with the classical theories used in lateral support design. Experience in recent years indicates that under certain conditions it may be possible to arrive at reasonably close estimates of movements using finite element techniques. The use of finite elements has however not been well proven in practice and at present the technique is mainly used for research purposes. At present the conventional approach in the prediction of movements is to use empirical methods based on observational data of previous excavations. The chart presented as Figure 22.7.1 has been used in certain applications. This chart was derived by Peck (1969) from data obtained from a number of instrumented excavations in a variety of soil conditions. Movements much smaller than those that would be predicted from Figure 22.7.1 have been observed in the monitoring of numerous basement excavations in Southern Africa. This may be due to the fact that the majority of the basement excavations that have been monitored are in residual soils above the water table. A further reason may be due to the advances in lateral support design and construction procedures that have occurred subsequent to the work carried out by Peck (1969). Empirical values, based on local experience, for the prediction of horizontal movement at the crest of excavations are given in Table 22.7.1. Local experience has also shown that vertical settlement at the crest of an excavation is usually smaller than the horizontal movement. It is important to emphasise that the values given in Table 22.7.1 should be used by the designer as a guide to obtain an indication of the magnitude of movement that can be anticipated. In instances where it is necessary to exercise some control over the magnitude of movement, the designer should consult his nearest Franki office to decide on the most suitable lateral support system and construction techniques.

332

Zone I:

Sand and soft to hard clay, average workmanship

Zone II:

Very soft to soft clay

Zone III:

Very soft to soft clay to a significant depth below bottom of excavation

Figure 22.7.1 Settlement adjacent to open cuts in various soil conditions after Peck (1969) Type of Support System

Horizontal movement as a percentage of excavation height

Externally Stabilised Cantilever retaining walls Walls with prop supports Tied back walls' Internally Stabilised GeoNail Systems

0.5% 0.2 to 0.5% 0.05 to 0.15% 0.1 to 0.3%

333

23.0 REFERENCE INFORMATION 23.1

NORMAL PLANT CLEARANCE REQUIREMENTS

NOTES: 1.

Clearances given are the absolute minimum for vertical piles. For raking piles additional clearance may be required and this should be discussed with your local Franki office.

2.

Headroom dimensions refer to standard systems and may be reduced under certain circumstances and using special or modified equipment.

3.

Working in close proximity to other structures will reduce the production rate and thus increase the cost. Avoid working to the minimum clearances if at all possible.

4.

Special attention must be given in the case of the working area being in close proximity to an overhead powerline. The requirements of the appropriate authority must be obtained and adhered to.

334

23.1

NORMAL PLANT CLEARANCE REQUIREMENTS

PILE TYPE Franki Driven

RIG TYPE

A mm

B mm

C mm

D mm

Mini

400

1000

3300

7200

Roller “F” Type

550

1650

5850

17000

SA81

800

2400

7800

19200

SA83

550

2050

7000

18500

Ull2LCrane

600

950

11000

19000

22RB Crane

600

950

9700

13500

LDH8O

700

850

12300

19300

LLDH12O

700

850

14200

25500

Soilmec CA/NCK 605 2B

700

900

10000

30500

Hotline 16M/120

550

1250

9950

14700

Forum Bored

Tripod Rig

300

1200

4400

4450

Precast

KH15O

700

750

8000

20900

CFA

KH15O

700

750

8000

20900

Sanwa/NCK6052B

500

750

11000

26500

1300 Oscil./U112L

1600

1450

11100

25000

1500 Oscil./U112L

1700

1550

11600

25000

Auger

Oscillator

NOTE:

Dimension A includes a clearance of l00mm Dimension B includes a clearance of l50mm Dimension C includes a clearance of 200mm

335

23.2

PILING RIG DIMENSIONS

336

23.2

PILING RIG DIMENSIONS

337

23.2

PILING RIG DIMENSIONS

338

23.2

PILING RIG DIMENSIONS

339

23.3

BENDING MOMENTS IN BEAMS

Type

Total Load W

Cantilever

Concentrated at end

Maximum Bending Moment WL

Uniformly distributed Freely Supported

Concentrated at Centre Uniformly Distributed

One end fixed Other end freely supported

Varying uniformly from zero at one end to maximum at the other end Concentrated at the centre Uniformly Distributed

Both ends fixed

Concentrated at the centre Uniformly Distributed

340

WL 2 WL 4 WL 8 0.128WL

3WL 16 WL 8 WL 8 WL 12

Maximum Deflection WL3 3EI WL3 8 EI WL3 48 EI 5WL3 384 EI WL3 0.013 EI WL3 0.00932 EI WL3 0.0054 EI 3 WL 192 EI WL3 384 EI

23.4

MENSURATION OF PLANE SURFACES

341

23.5

MENSURATION OF SOLIDS

342

23.6

PROPERTIES OF SECTIONS

343

24.0 QUALITY ASSURANCE To ensure that the Frankipile South Africa slogan "Quality is our Foundation" is more than a mere statement, a rigorous quality assurance programme has been developed by the company for use in the design and installation of its products as well as the services that it offers. This programme is also flexible and can accommodate any quality assurance requirements that a client might impose on a project. The programme governs all the company's activities by means of standardised procedures in accordance with three levels of control. The three levels are referred to as Level I, Level 2 and Level 3. The first level is a basic level quality assurance and is applied to all the Company's activities other than where one of the other levels is applied. Level 2 is used when there are special circumstances such as difficult ground conditions or a particularly difficult construction method and where a higher level of control is desirable. If there are Quality Assurance requirements in the contract document then Level 3, the highest level, is applied. The following is a more detailed account of these three levels. LEVEL l This is the basic level and covers the company's everyday operations which are anticipated to have no undue complications. It is the minimum level of quality assurance that the company applies to all its operations. It covers the following: (a)

Contract Review Prior to the submission of a tender for a contract the specified conditions are reviewed to ensure they can be met. This review is recorded formally.

(b)

Document Control At this level the documentation of the foundation is formally controlled, ensuring that all staff are working with the latest drawings and specifications. All official documentation such as correspondence with the Engineer/ Architect/Employer any transmittal of contract drawings, material and product test results and the like are formally recorded and sent under cover of a receipted transmittal note. This ensures that work is executed to the latest instructions and recorded properly for traceability.

(c)

Concrete Testing In addition all concrete and grouts are tested regularly to ensure that, the strength, slump and any other requirement is met. All testing of concrete and grout will be

344

executed as per specifications and recorded and transmitted formally. Results will be recorded as per instructions form the Engineer . (d)

Training All personnel will have received the requisite training in the correct installation of the company's products according to the company's product manuals. This training is recorded on the personal files of the staff concerned.

LEVEL 2 This level is used when special systems are to be installed or where complications might be expected in the execution due to uncertain ground conditions or where any other aspect of the contract suggests a higher level of control is desirable. It introduces the concept of executing the work in accordance with written procedures. This means that all operations are reduced to written standard procedures with the recording of different steps in the process to ensure compliance with the specifications and/or design criteria. These procedures are all set out on a standard format. LEVEL 3 This is the highest level of quality assurance and is applied to contracts where such a level is a requirement of the contract documents. At present this is the case with large complicated contracts such as the construction of powerstations, refineries and metalurgical plants. On these contracts everything related to a construction activity is formally documented. A particular feature would be the traceability of materials back to their source and obliges suppliers and subcontractors to fulfill all the requirements of the specifications and the quality assurance programme. It is the responsibility of the main contractor to audit suppliers and subcontractors in this regard. The following are some of the quality assurance activities covered by this level: (a)

Purchasing The purchase of built-in materials are controlled to ensure compliance with the Specifications.

(b)

Process Control This requires that all construction and manufacturing operations be covered by detailed written procedures. The process control requirements are also applied at Level 2.

(c)

Inspection, Measuring and Test Equipment This ensures that such equipment is identified, registered, inspected, calibrated, and maintained at regular intervals.

345

(d)

Non-conformance This deals with the identification, documentation, evaluation and reworking of nonconforming materials and system elements.

(e)

Internal Quality Control It is also a requirement of the quality assurance programme that all operations are audited on a regular basis to ensure compliance by all concerned.

It should be noted in particular that the quality assurance system, at whatever level, requires that any non-conformity or anomaly is to be fully recorded. Such recording is done on standardised forms for transmittal to the relevant parties for analysis and disposition. This disposition needs to be approved by the designer. By using its Quality Assurance programme Frankipile is able to ensure that the quality of its products and services is constantly monitored. In addition the standardised recording of installation procedures allows the company to monitor the efficacy of the work procedures and to modify them accordingly. When errors or incorrect procedures are identified it is then able to ensure that corrective action is taken to avoid further non-conformance. Corrective action may involve the adaptation of the procedure so as to eliminate the nonconforming condition. It may also involve changing to another foundation system if it is found that the non-conformance cannot be eliminated through adaptation. Re-training of staff may also form part of the corrective action. Because most of their work is carried out below ground level Frankipile South Africa considers the use of its Quality Assurance programme essential if it is to ensure the quality of the products and services described in this book. By its use the company is able to guarantee its products with confidence.

346

REFERENCES Baguelin, F., Jezequel, J.F. and Shields, D.H. (1978), The Pressuremeter and Foundation Engineering Trans. Tech Pub, Clausthal. Barron, R.A. (1948), Consolidation of Fine-grained Soils by Drain Wells, Transactions ASCE (113), pp718-754. Barton, N. and Choubey, V. (1977), The Shear Strength of Rock Joints in Theory and Practice, Rock Mechanics, Springer -Verlag. Berezantzev, V.G., Khristoforov, V. and Golubkov, V. (1961), Load Bearing Capacity and Deformation of Piled Foundations, Proc. 5th Int. Conf. ISSMFE, 2, 11-15. Bieniawski, Z.T. (1973), Engineering Classification of Jointed Rock Masses, Trans. South African Institution of Civil Engineers, No. 15, pp 335-343. Blight, G.E. (1984), Uplift Forces Measured on Piles in Expansive Clay, Int. Conf. on Expansive Soils, Adelaide, South Australia. Bowles, J.E. (1977), Foundation Analysis and Design, McGraw-Hill., New York. Brackley, I.J.A. (1975), The Interrelationship of the Factors Affecting Heave of an Expansive Unsaturated Soil, PhD thesis, University of Natal, Durban, South Africa. Brackley, I.J.A. (1980), Prediction of Soil Heave from Suction Measurements, Proc. 7th ISSMFE Regional Conference for Africa, Accra, Ghana, Vol1, pp 159-166. Brinch Hansen, J. (1961), The Ultimate Resistance of Rigid Piles Against Transversal Forces, Geoteknisk Institut, Bulletin No. 12, Copenhagen. Brink, A.B.A., Partridge, T.C. and Williams, A.A.B. (1982), Soil Survey for Engineering, Clarendon Press, Oxford. Broms, B.B. (1964)a, Lateral Resistance of Piles in non-Cohesive Soils, Journal of the Soil Mechanics and Foundation Division, ASCE, Vol. 90, SM2, pp 27-63. Broms, B.B. (1964)b, Lateral Resistance of Piles in Cohesive Soils, Journal of the Soil Mechanics and Foundation Division, ASCE, Vol. 90, SM3, pp 123-156. Broms, B.B. (1993), Chapter 4- Lime Stabilisation, Ground Improvement, Edited by M.P. Moseley, CRC Press, Boca Raton. Brown, R.E. (1977), Vibrocompaction of Cohesionless Soils, Journal of the Geotechnical Engineering Division, ASCE, No. GT-12, pp 1437-1451. BS 5930 (1981), Code of Practice for Site Investigations, British Standards Institution, London. BS 8081 (1989), Ground Anchorages, British Standard Institution, London. Bullock, S.J. (1978), The Case for Using Multi-Channel Seismic Equipment and Techniques for Site Investigation, Bulletin SA Ass. Eng. Geologists, XV, pp 19-35. Burland, J.B. and Burbridge, M.C. (1985), Settlement of Foundations on Sand and Gravel, Proc. I.C.E., Part 1, No.78, pp 1325-1371. Burland, J.B., Broms, B.B. and De Mello, V.F.B. (1977), Behaviour of Foundations and Structures, State of the Art Review, Proc. 9th Int. Conf., ISSMFE, Tokyo, 3, 395-546. Campanella, R.G. and Robertson, P.K., (1988), Current Status of the Piezocone Test, Proc. ISOPT I, Vol1, Orlando, Florida, pp 93-116. Canadian Founding Engineering Manual (1985), Canadian Geotechnical Society, 2nd Edition. Chin, F.K. and Vail, A.J. (1973), Behaviour of Piles in Alluvium, Proc. 8th Int. Conf. ISSMFE, Moscow, Vo12. Collins, L.E. (1953), Preliminary Theory for the Design of Underrearned Piles, Trans. South Arican Institution ofCivil Engineers, Vo13, No.11. Core Logging Committee (1978), A Guide to Core Logging for Rock Engineering, Bulletin SA Ass. ofEng. Geologists, No.15, pp 295-328. CSRA (1993), Standard Specifications for Subsurface Investigations, Department of Transport, Pretoria, South Africa.

347

REFERENCES D' Appolonia, E. (1953), Loose Sands -Their Compaction by Vibroflotation, Spec. Tech. Pub. 156, American Soc. for Testing Materials, pp 138-154. Darracott, B.W. (1976), Seismic Surveys and Civil Engineering, Trans. South African Institution of Civil Engineers, 18(2), pp 35-40. Dash, H. and Jovino P.L. (1980). Construction of a Root-pile Wall at Monesson, Pennsylvania, Transportation Research Record 749, Transportation Research Board, National Academy of Sciences, Washington D.C. De Mello, V.F.B. (1971), The Standard Penetration Test State-of-the-Art Report, 4th ISSMFE PanAmerican Conference, Puerto Rico, 1, 1-86. Deere, D.U. (1968), Geological Considerations, Chapter I in K.G. Stagg and O.C. Zienkiewicz, Rock Mechanics in Engineering Practice, Wiley, New York, pp 1-20. Donaldson, G.W. (1973), The Prediction of Differential Movement on Expansive Soils, Proc. Third Int. Conf. on Expansive Soils, Haifa, Israel, Voll, pp 289-293. Esopt I (1974), Various Authors, Proceedings of the Conference on the Settlement of Structures, Pentech Press, Cambridge. Everett, J.P. (1991), Load Transfer Functions and Pile Performance Modelling, Proc. ISSMFE loth Regional Conference for Africa, Maseru, Voll, pp 229-234. Fox, E.N. (1948), The Mean Elastic Settlement of a Uniformly Loaded Area at a Depth Below the Ground Surface, Proc. 2nd Int. Conf., ISSMFE, Rotterdam, 1,29-32. GassIer, G. and Gudehus, G (1981), Soil Nailing -Some Aspects ofaNew Technique, Proc. lOth Int. Conf. ISSMFE, Stockholm, Vo13. Glassop, R. ( 1968), The Rise of Geotechnology and its Influence on Engineering Practice, 8th Rankine Lecture, Geotechnique 18, pp 105-50. Greenwood, D.A. (1970), Mechanical Improvement of Soils Below Ground Surface, Proc. Conference on Ground Engineering, Inst. of Civil Engineers, London., pp 11-22. Griffiths, D.H. and King, R.F. (1965), Applied Geophysics for Engineers and Geologists, Pergamon Press, Oxford. Gudehus, G. (1983), Design Concepts for Pile Dowels in Clay Slopes, Discussion Speciality Session 5, Proc. 8th ISSMFE European Conf., Helsinki, Vo13. Heinz, W.F. (1985), Diamond Drilling Handbook, Sigma Press, Johannesburg. Heymann, G.H., Rohde, A.W., Schwartz, K. and Friedlaender, E. (1992), Soil Nail PullOut Resistance in Residual Soils, Proc. Int. Symposium on Earth Reinforcement Practice, Fukuoka, Japan. Hoek, E. and Bray, J.W. (1977), Rock Slope Engineering, Inst. Min. and Met. London. Hughes, J.M.O. and Withers, N.J. (1974), Reinforcing Soft Cohesive Soil and Stone Columns, Ground Engineering, May, pp 42-49. ISSMFE Technical Committee on Penetration Testing (1988), Proceedings ISOPT I, Voll, Orlando, Florida. Jennings, J.E. and Kerrich, J.E. (1962), The Heaving of Buildings and the Associated Economic Consequences with Particular Reference to the Orange Free State Gold Fields, The Civil Engineer in South Africa, November, Vo14, No II. Jennings, J.E. and Knight, K. (1956), Recent Experiences with the Consolidation Tests as a Means of Identifying Conditions of Heaving or Collapse of Foundations on Partially Saturated Soils, Trans. South African Inst. Civil Engineers. Vo16, No 8, pp 255-256. Jennings, J.E. and Knight, K. (1975), A Guide to Construction on or with Materials Exhibiting Additional Settlement due to Collapse of Grain Structure, Proceedings ISSMFE 6th Regional Conference for Africa, Durban. Jennings, J.E., Brink, A.B.A. and Williams, A.A.B. (1973), Revised Guide to Soil Profiling for Civil Engineering Purposes in Southern Africa, Trans. South African. Inst. Civil Eng., 15,3-12.

348

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349

REFERENCES O'Rourke, T.D. and Jones C.J.F.P. (1990), An Overview of Earth Retention Systems: 19701990, ASCE Speciality Conference on Earth Retaining Structures, Cornell. Peck, R.B. (1969), Deep Excavations and Tunnelling in Soft Ground, State of the art Report. 7th Int. Conf. ISSMFE, Mexico. Peck, R.B., Hansen, W.E., and Thorburn, T.H. (1974), Foundation Engineering, 2nd Edition, Wiley, New York. . Perloff, W.H. and Baron, W. (1976), Soil Mechanics Principles and Applications, The Ronald Press Company, New York. Pidgeon, J.T. (1983), The Design of Stiffened Raft Foundations on Expansive Soils, Ground Profile No 33, Journal of the Geotechnical Engineering Division of SAICE. Poulos, H.G. and Davis, E.H. (1980), Pile Foundations Analysis and Design, John Wiley and Sons, New York. Poulos, H.G. and Davis, E.H. (1974), Elastic Solutions for Soil and Rock Mechanics, Wiley, New York Randolph, M.F. (1977), A Theoretical Study of the Performance of Piles, Ph.D. Thesis, Cambridge University. Reese, L.C. (1977), Laterally Loaded Piles: Program Documentation, Journal of the Geotechnical Engineering Division., ASCE, Vol. 103, No. GT4: 287-305. Reynolds, C.E. and Steedman, J.C. (1981), Reinforced Concrete Designers Handbook, Ninth Edition, Viewpoint Publications. Rowe, P.W. (1952), Anchored Sheet Piled Walls, Proc. Inst. ofCivil Eng., London, Part 1, Vol. 1. Rowe, R.K. and Arimitage, H.H. (1987), Theoretical Solutions for Axial Deformation of Drilled Shafts in Rock, Vol24 ,pp 114-142. Rust, E. and Jones, G.A. (1990), Prediction of Performance of Embankments on Soft Alluvial Deposits using the Piezometer Probe (CPTU),. RDAC Research Project No89/14, South African Road Board. SAICE Code of Practice for the Safety of Persons Working in Small Diameter Shafts and Test Pits for Civil Engineering Purposes ( 1990), Geotechnical Division SAICE. Schmertmann, J. (1975), Measurement of In-situ Shear Strength, Geotechnical Engineering Division, ASCE, Proceedings of the Speciality Conference on In-situ Measurement of Soil Properties, Raleigh, North Carolina, Vol. 2, pp 57-138. Schmertmann, J .H. ( 1969), Dutch Friction-cone Penetration Exploration of Research Area at Field 5 Eglin Air Force Base Florida, US Army Waterways Experimental Station, Vicksburg (Mississippi), Contract Report S-69-4, 1969. Schmertmann, J.H. (1970), Static Cone to Compute Settlement Over Sand, Journal of the Geotechnical and Foundation Engineering Division, ASCE, 96 (SM3), 1011-1043. Seed, H.B., Woodward, J.R. and Lundgren, R. (1978), Clay Mineralogical Aspects of the Atterberg Limits, Journal of the Soil Mechanics and Foundation Division, ASCE, Vo190, No. 5m4. Shen, C.K., Herrmann, L.R., Romstad, K.M., Bang, S., Kim, Y.S. and Denatale, J.S. (1981), In-situ Earth Reinforcement Lateral Support System, Report 81-03, Dept. of Civil Engineering, University of California, Davis. Skempton, A.W. (1951), The Bearing Capacity of Clays, Building Research Congress, London, Institiute Civil Engineers, div. I: 180. Skempton, A.W. (1953), Discussion: Piles and Pile Foundations, Settlement of Pile Foundations, Proc. 3rd Int. Conf. ISSMFE, Vol. 3: 172. Smith, E.A.K. (1960), Pile Driving Analysis by the Wave Equation, Journal of the Geotechnical and Foundation Engineering Division, ASCE, Vol. 86, SM4: 35-61. Stroud, M.A. (1974), The Standard Penetration Test in Insensitive Clays and Soft Rocks, Proc. European Symposium on Penetration Testing (ESOPT 1),367-375.

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REFERENCES Stroud, M.A. (1989), The Standard Penetration Test -Its Applications and Interpretation. Proc. I.C.E. Conf. on Penetration Testing in the U.K., Birmingham. Terzaghi, K. & Peck, R.B. (1967), Soil Mechanics in Engineering Practice, Wiley, New York. Terzaghi, K. (1955), Evaluation of Coefficients of Sub-grade Reaction, Geotechnique, Vol 5: 297. Thorburn, S. (1975), Building Structures Supported by Stabilised Ground, Geotechnique, (25), 1. TMHI (1986), Standard Methods of Resting Road Construction Materials, NITRR of the CSIR, Pretoria, South Africa. Tomlinson, M.J. (1970), Lateral Support of Deep Excavations, Ground Engineering, The Institution of Civil Engineers, London. Tomlinson, M.J. (1970), Some Effects of Pile Driving on Skin Friction, Conf. on the Behaviour of Piles, Inst. Civ. Engineers., London: 59-66. Tomlinson, M.J. (1977), Pile Design and Construction Practice, Viewpoint Publications, Hertfordshire. Touma, F.T. & Reese, L.C. (1974), Behaviour of Bored Piles in Sand, Journal of the Geotechnical and Foundation Engineering Div., ASCE, Vol. 100, No. GT7: 749-761. Tromp, B.E. (1985), Design of Stiffened Raft Foundations for Houses on Collapsing Hillwash in Southern Suburbs of Johannesburg, Personal Communication. Tschebotarioff, G.P. (1973), Foundations, Retaining and Earth Structures, 2nd Ed., McGraw-Hill Book Company, New York. Underground Construction in Soft Ground (1994), Int. Symposium on Underground Construction in soft ground, ISSMFE, Technical Committee TC-28, New Delhi, India. Van der Merwe, D.H. (1964), The prediction of Heave from the Plasticity Index and Percentage Clay Fraction of Soils, Trans. SA Inst. Civ. Eng., Vo16, No 6, pp 103-107. Van Vuuren, D.J. (1969), Rapid Determination of CBR with the Portable Dynamic Cone Penetrometer, The Rhodesian Engineer, September, Paper No.105. Van Weele, A.F. (1957), A Method of Seperating the Bearing Capacity of a Test Pile into Skin Friction and Point Resistance, Proc. 4th Int. Conf. ISSMFE, Vo12: 76. Vesic, A.S. (1972), Expansion of Cavities in Infinite Soil Mass, Journal of the Geotechnical and Foundation Engineering Division, ASCE, Vol. 98, No. SM3. Wagener F. von M. (1984), Engineering Construction on Dolomite, DSc. Thesis, Department of Civil Engineering, University of Natal. Wagener, F von M. (1981), Engineering Evaluation of Dolomitic Areas, Seminar on the Engineering Geology of Dolomite Areas, Department of Geology, University ofPretoria. Wagener, F von M. (1982), Engineering Construction on Dolomite, PhD thesis, University of Natal, Published by Geotechnical Division, SAICE, Johannesburg. Webb, D.L. (1970), Settlement of Structures on Deep Alluvial Sand Sediments in Durban South Africa; Proc. BGS Conf. on In-situ Investigations in Soil and Rock., London. West, G., and Dumbleton, M.J. (1975), An Assessment of Geophysics in Site Investigations for Roads in Britain, Transportation and Road Research Laboratory Report KR680. Weston, D.J. (1979), Expansive Roadbed Treatment for Southern Africa, Proc. 4th Int. Conf. on Expansive Soils, Denver, Colorado 1980. pp 339-360. Whitlow, R. (1990), Basic Soil Mechanics, Longman Group, U.K. Williams, A.A.B. and Donaldson, G. (1973), Dev. Relating to Building on Expansive Soils in SA 1973-1980, Proc. 4th Int. Conf. on Expansive Soils, Denver, pp 834-844. Williams, A.F., Johnston, I.W., and Donald, I.B. (1980), The Design of Socketed Piles in Weak Rock, In: Structural Foundations on Rock, Edited by P.J.N. Pells, A.A. Balkema, Rotterdam. Windle, D. and Wroth, C.P. (1977), The Use of a Self Boring Pressuremeter to Determine the Undrained Properties of Clays, Ground Engineering. Winterkorn, H.F. and Fang, H. (Ed) (1975), Foundation Engineering Handbook, Van Nostrand Reinhold, New York.

351

INDEX Cone penetration test (CPT), 6,8,1316,17,43,46,49,51,53,56 Cone penetrometer, 15 Cone penetrometer, electric, 15 Consistency,7,8,13,30,31,37,48,49,60 Consolidation, 153 Consolidation tests, 28 Contiguous pile walls, 184, 188, 189,203205 Continuous flight auger (CFA) piles, 64, 65,112-116 Core barrel, 16, 17 Core orientation, 23 Core recovery, 41 Corrosion protection: Ground anchors, 215 Steel sheet piles, 194 Cut-offwal\s,243-244 D Deadman anchors, 214 Deformation modulus, 56,57,58,59 Density tests, 26,29 Design : Braced walls, 319-325 Dynamic compaction, 303-305 GeoNails, 326-328 Lateral support, 312-333 Piles, 245-269 Pile caps, 284-294 Pile groups, 270-279 Pile shafts, 280-284,295-300 Soil improvement, 301-311 Soil replacement, 306-311 Vibratory compaction, 301-302

A Accelerated consolidation, 153,154,155, 175-177,310,311 Active earth pressure, 315 Allowable bearing pressure, 151,154 Anchored walls, 185,321-325 Anchors: Corrosion protection, 215 Deadman, 214 Soft ground, 212-217 Anchor piles, 185,218-219 Angle of friction φ,7,8,21,28,49,50,51 Artesian conditions, 74 At-rest pressure, 314 Atterberg limits, 7,8,29,49 Auger piles, 63,64,95-102 Auger trial holes, 6,7,9,25 B Band drains, 176-177,310 Barette, 103,104,107 Bending moments for beams, 340 Bentonite, 104 Bentonite slurry, 104,115,208 Bored cast-in-situ piles, 63,64,65 Boreholes,7,8,13,15,23,24 Boulders,65,100,115 Braced walls, 319-320 Bulk density, 29 C Caisson piles, 63,64,127-130 California bearing ratio (CBR), 29 Cantilever walls, 184,185,197,200,206 219 Cavities,238 Clay: Dispersive, 228 Expansive, 229-233 Cofferdams, 190 Cohesion, 7,8,28 Cohesionless soils, 48 Cohesive soils, 48 Collapse potential index, 28 Collapse tests, 7,8,28 Collapsible soils, 162,234-236 Compaction, 161 Compaction grouting, 153,154,155,166167,306 Concrete soldier piles, 184,188,189,200202 352

Compaction,153,154,155,166-167 Jet,153,154,155,178-179 Gunite, 183,198,201 Gypsum, 181 H Hand augers, 97 Heave: Piles, 73,80,86,92 Soil,47,98,229-233 Heave tests, 7,8 Hydrometer,29 I Inclinometers, 24 Index property tests, 7,8,29,47,49 In-situ density tests, 26 Integrity testing ofpiles, 107,124,147-150 J Jet grouting, 153,154,155,178-179 Jetting,71,85,91,94 Joints, in rock, 40 " K Kelly, 96 L Laboratory testing, 2,4,27-28,56 Lateral load test, 145

Diaphragm walls, 184,188,189,206-209 Dolomites,238-240 Downdrag, 195 Drainage blanket, 177 Drilled micropile, 225-226 Driven cast-in-situ piles, 63,64,65 Driven stone columns, 153,154,155,173174,306-309 Driven tube piles, 64,65, 76-80 Driven preformed piles, 63,64,65 Dry density, 49,50 Dynamic compaction, 153,154,155,161165,303-305 Dynamic cone penetration test (DPSH), 6,7,8,12-13 Dynamic probe light (DPL), 12,53 Dynamic replacement, 153,154,155,171172 E Earth pressures: 314-318 Active, 315 At rest, 314 Passive, 316 End bearing capacity, 248,251,252,254, 255,256,257 ,258 Environmental investigations, 241-242 Expansive soils, 29,44,47,229-233 Extensometers, 24 F Factor of safety, 268,330,331 Field investigation, 4,9-27 Fixed earth support, 323 Forum bored piles, 64,65,117-121 Fox correction factors, 275 Franki piles, 64,65,66- 75 Franki precast composite pile, 72 Free earth support, 322 Friction angle, 7,8,21,28,49,50,51 Friction capacity, 249,250,252-258 Friction ratio, 15 G GeoNails, 186,188, 189,220-224,326-328 Geotechnical investigation, 2,4-27 Geophysical techniques, 26-27 Grading analysis, 7,29 Ground anchors, 212-217 Groundwater monitoring, 241-242 Groundwater table, 7,8,34 Group index, 43 Grouting: Anchor,216

353

Lateral, 262-265 Tension, 260-262 Classification, 63 Design aids, 245-269 Driveability, 15 Driving formulae, 259-260 End bearing, 248,251,252,254,255, 256,257,258 Factors of safety, 268 Friction,249,250,252,253,254, 255,256,257,258 Heave, 73,80,86,92 Integrity testing, 107,124,147-150 Load testing, 142-146 Load transfer functions, 273-275 Penetrating ability, 65 Selection, 61-62 Settlement, 273-275 Shaft capacity, 249,250,252,253,254, 255,256,257,258 Technical details, 66-130 Tension,66,70,117 Pile shaft: Bending moment curves, 295-300 Design, 280-284 Design curves, 295-300 Reinforcement, 294 Pile types and classification, 63,64,65 Piling equipment: Normal clearance requirements, 334, 335 Overall dimensions, 336-339

Lateral support: Classification, 184-187 Design aids, 312-333 Selection, 182-183 Technical details, 190-227 Lime columns, 153,154,155,180-181 Liquefaction, 15 Liquid limit, 29 Load testing of piles: 142-146 Compression, 142-146 Tension, 145 Lateral, 145 Lugeon test, 21-22 M Mensuration of plane surfaces, 341 Mensuration of solids, 342 Mod AASHTO test, 29 Modulus of: Compressibility , 7,8,21,56,57,58,59 Subgrade reaction, 60 Moisture content, 7,8,29 Moisture density relationship, 2,29 Movement associated with excavation, 332- 333 N Negative friction ,85,238 Noise pollution, 65,74,80,86,91,155,189 195 O Oedometer test, 7,8,28 Oscillator piles, 64,65,122-126 P Particle size, 44 Passive earth pressure, 316 Pedogenic material, 33 Permeability, 7,8,29 Piezo-cone penetration test (CPTU), 15, 43,46 Piezometers, 22,24 Pile bearing capacity, 15 Pile caps: Design and detailing, 284-288,294 Typical details, 289-293 Pile: Base capacity, 248,251,252,254,255, 256,257,258 Bearing capacity, 245-258 Behaviour,245-246,273-275 Capacity: Compression,247-258 In heaving sub-soil, 265-268 354

Shear box test, 7,8,28 Shear strength, 15,48,49 Sheet piling, 184,188,189,190-196 Shelby tube samples, 6,22,28 Site investigation, 2,4-27 Slip coating of piles, 85 Slurry waIls, 243-244 Soft clays, 236-238 Soil: Classification: 30,43,46 From cone penetrometer, 43,46 Standard, 30,43 Cohesionless, 48, Cohesive, 48 Colour, 30 Compaction, 7,8,153 Compressibility , 15,28,56,59 Consistency,7,8,30,31,48 Consolidation,153,154,155,175-177 Densification, 7,8 Description, 7,47 Expansive, 29,44,47,229-233 Grading, 7,29 Liquefaction, 15 Moisture, 7,9,30 Origin, 30,33 Pedogenic, 33 Pressures,314-318 Profiling, 30 Profile, typical, 35 Profiling, symbols, 36 Replacement,153

Piston samples, 6,22,28 Plastic limit, 29 Plasticity index, 29,50 Plate load test, 7,24-25,56 Point load index test, 16,53,54 Precast piles, 64,82-88 Predrilling,71,73,74,85,91 Pressuremeter test, 7,8,21 Problem soils, 228-240 Procter test, 29 Prop supports, 185,210-211 Poststressed anchors, 185,212-217 Q Quality assurance, 344-346 R Raymond spoon, 12,17 Reinforcement areas, 289 Relative density, 15 Residual soils, 33 Reticulated micropiles, 186,188,189, 225226 Rock, 13,33 Rock: Classification, 30 Colour,37 Cores, 16,17 Dicontinuities, 40-41 Fabric, 37,38 Hardness, 37,39 Stratigraphy,37,39 Strength classification, 53-55 Type,37,39 Weathering, 37,38 Rock mass description, 37-42, Rock quality designation {RQD), 41 Rock shoes, 77,84 Rock socket design, 254-255 Rotary core drilling, 6 Rotary drilling, 16 Rotary percussion drilling, 23-24 S Samples: Block,9,28,29 Disturbed, 7,8,28,29 Piston, 6,22,28 Raymond spoon, 12,17 Shelby tube, 6,22,28 Undisturbed, 6,7,8,28,29 Sand drains, 175 Sand wick drains, 176 Secant pile wall, 184,188,189,203-205 355

Trial piles, 143 Triaxial compression test, 7,8,28 Tube-a-manchette,212

Soil (Cont.): Residual, 33 Sampling,22 Strength classification, 47 Structure, 30,31 Transported, 33 Type, 15,30,32 Soil dowelling, 186,188,189,227,330 Soil improvement; 151-181 Classification, 153 Design aids, 301-311 Selection,151-152 Technical details, 156-181 Soil nails, 220-224,326-328 Soiltech,4,9,26,27 Soldier piles, 184,188,189,197-199,200202 Steel H-piles, 64,65,89-92 Steel sheet piles, 184,188,189,190-196 Standard penetration test: 6,7, 13,17 SPT "N" number, 17,48,49,51,52,53, 56 Stone columns, 153,154,155,173-174, 306309 Subgrade modulus, 60 Surcharge loads, 317 Swelling potential, 44 Swell under load test, 7,8,28 T Tension test load, 145 Test pits, 7,10,25 Tiebacks, 185,212-217 Tied back walls, 321-325 Timber piles, 64,65,93-94 Transported soil, 33 Tremie concrete, 108,125

U Unconfined compression strength {UCS), 8,16,48,53,54,55,59 Underpinning,131-141 Underreams, 98 Underslurry piles, 64,65,103-111 Undisturbed samples, 6,7,8,28,29 Undrained cohesion, 48,50,52 Undrained shear strength, 7,8,21,28,52, 53,60 Unified soil classification system, 43,45 V Vane shear test, 6,7,8, 19,20 Vibration, 65,74,80,86,92,155 Vibratory compaction, 153,154,155,156160,301,302 Vibratory replacement, 153,154,155, 168170,306-309 Vibrocompaction,153,154,155,168170,306-309 W Walers,198,200,210,211 Washboring, 16 Waterpressure,316 Water table, 7,8,34 Wave equation, 260 Wedge analysis, 326 Wells,241 Wick drains, 176 Y Y-probe compaction, 159,160

356