FEMA 267

FEDERAL EMERGENCY MANAGEMENT AGENCY

FEMA 267b / June, 1999

Interim Guidelines Advisory No. 2

Steel Moment Frame Structures

Program to Reduce the Earthquake Hazards of

Supplement to FEMA-267

INTERIM GUIDELINES ADVISORY NO. 2 Supplement to FEMA-267 Interim Guidelines: Evaluation, Repair, Modification and Design of Welded Steel Moment Frame Structures Report No. SAC-99-01

SAC Joint Venture a partnership of: Structural Engineers Association of California (SEAOC) Applied Technology Council (ATC) California Universities for Research in Earthquake Engineering (CUREe) Prepared for SAC Joint Venture Partnership by Guidelines Development Committee Ronald O. Hamburger, Chair Thomas Sabol John D. Hooper C. Mark Saunders Robert E. Shaw Raymond H.R. Tide Lawrence D. Reaveley

Project Oversight Committee William J. Hall, Chair John N. Barsom Shirin Ader John Barsom Roger Ferch Theodore V. Galambos John Gross James R. Harris

Richard Holguin Nestor Iwankiw Roy G. Johnston Len Joseph Duane K. Miller John Theiss John H. Wiggins

SAC Project Management Committee SEAOC: William T. Holmes ATC: Christoper Rojahn CUREe: Robin Shepherd

Program Manager: Stephen A. Mahin Investigations Director: James O. Malley Product Director: Ronald O. Hamburger Federal Emergency Management Agency Project Officer: Michael Mahoney Technical Advisor: Robert D. Hanson SAC Joint Venture 555 University Avenue, Suite 126 Sacramento, California 95825 916-427-3647 June, 1999 i

THE SAC JOINT VENTURE SAC is a joint venture of the Structural Engineers Association of California (SEAOC), the Applied Technology Council (ATC), and California Universities for Research in Earthquake Engineering (CUREe,) formed specifically to address both immediate and long-term needs related to solving problems of the Welded Steel Moment Frame (WSMF) connection that became apparent as a result of the 1994 Northridge earthquake. SEAOC is a professional organization composed of more than 3,000 practicing structural engineers in California. The volunteer efforts of SEAOC’s members on various technical committees have been instrumental in the development of the earthquake design provisions contained in the Uniform Building Code as well as the National Earthquake Hazards Reduction Program (NEHRP) Provisions for Seismic Regulations for New Buildings. The Applied Technology Council is a non-profit organization founded specifically to perform problem-focused research related to structural engineering and to bridge the gap between civil engineering research and engineering practice. It has developed a number of publications of national significance including ATC 306, which serves as the basis for the NEHRP Recommended Provisions. CUREe is a nonprofit organization formed to promote and conduct research and educational activities related to earthquake hazard mitigation. CUREe’s eight institutional members are: the California Institute of Technology, Stanford University, the University of California at Berkeley, the University of California at Davis, the University of California at Irvine, the University of California at Los Angeles, the University of California at San Diego, and the University of Southern California. This collection of university earthquake research laboratory, library, computer and faculty resources is among the most extensive in the United States. The SAC Joint Venture allows these three organizations to combine their extensive and unique resources, augmented by subcontractor universities and organizations from around the nation, into an integrated team of practitioners and researchers, uniquely qualified to solve problems related to the seismic performance of WSMF structures.

DISCLAIMER The purpose of this document is to serve as a supplement to the FEMA-267 publication Interim Guidelines: Evaluation, Repair, Modification and Design of Welded Steel Moment Frame Structures. This Advisory, which is intended to be used in conjunction with FEMA-267, supercedes and entirely replaces Interim Guidelines Advisory No. 1 (FEMA 267a). FEMA-267 was published to provide engineers and building officials with guidance on engineering procedures for evaluation, repair, modification and design of welded steel moment frame structures, to reduce the risks associated with earthquake-induced damage. The recommendations were developed by practicing engineers based on professional judgment and experience and a preliminary program of laboratory, field and analytical research. This preliminary research, known as the SAC Phase 1 program, commenced in November, 1994 and continued through the publication of the Interim Guidelines document. This Interim Guidelines Advisory No. 2, which updates and replaces Interim Guidelines Advisory No. 1, is based on supplementary data developed under a program of continuing research, known as the SAC Phase 2 program, as well as findings developed by other, independent researchers. Final design recommendations, superceding both FEMA-267 and this document are scheduled for publication in early 2000. Independent review and guidance in the production of both the FEMA-267, Interim Guidelines and the advisories was provided by a project oversight panel comprised of experts from industry, practice and academia. Users are cautioned that research into the behavior of these structures is continuing. Interpretation of the results of this research may invalidate or suggest the need for modification of recommendations contained herein. No warranty is offered with regard to the recommendations contained herein, either by the Federal Emergency Management Agency, the SAC Joint Venture, the individual joint venture partners, their directors, members or employees. These organizations and their employees do not assume any legal liability or responsibility for the accuracy, completeness, or usefulness of any of the information, products or processes included in this publication. The reader is cautioned to carefully review the material presented herein. Such information must be used together with sound engineering judgment when applied to specific engineering projects. This Interim Guidelines Advisory has been prepared by the SAC Joint Venture with funding provided by the Federal Emergency Management Agency, under contract number EMW-95-C-4770. The SAC Joint Venture gratefully acknowledges the support of FEMA and the leadership of Michael Mahoney and Robert Hanson, Project Officer and Technical Advisor, respectively. The SAC Joint Venture also wishes to express its gratitude to the large numbers of engineers, building officials, organizations and firms that provided substantial efforts, materials, and advice and who have contributed significantly to the progress of the Phase 2 effort.

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PREFACE Purpose The purpose of the Interim Guidelines Advisory series is to provide engineers and building officials with timely information and guidance resulting from ongoing problem-focused studies of the seismic behavior of moment-resisting steel frame structures. These advisories are intended to be supplements to FEMA-267 Interim Guidelines: Evaluation, Repair, Modification and Design of Welded Steel Moment Frame Structures first published in August 1995. The first Interim Guidelines Advisory, FEMA-267a, was published in January 1997. The specific revisions and updates to the Interim Guidelines contained in FEMA-267a were developed based on input obtained from a group of engineers and building officials actively engaged in the use of the FEMA-267 document, in the period since its initial publication in August 1995. That input was obtained during a workshop held in August 1996, in Los Angeles, California. This second Interim Guidelines Advisory has been prepared as a series of updates and revisions both to the FEMA-267, Interim Guidelines which it supplements and to the FEMA267a, Interim Guidelines Advisory publication, which it supercedes. The material contained in this Interim Guidelines Advisory No. 2 is based on the extensive analytical and laboratory research that has been conducted by the SAC Joint Venture and other researchers during the intervening period, along with recent developments in the steel construction industry. The material contained in this Advisory has been formatted to match that contained in the original Interim Guidelines, to permit the user to insert this material directly into appropriate sections of that document. This Advisory is not intended to serve as a self-contained text and should not be used as such. It does, however, completely replace the material contained in FEMA-267a. A new set of recommendations for the design, analysis, evaluation repair, retrofit and construction of moment-resisting steel frames is currently being prepared as part of the Phase 2 Program to Reduce Earthquake Hazards in Steel Moment Frame Structures. These new Seismic Design Criteria, which are anticipated to be completed early in the year 2000, will replace in their entirety the FEMA-267 Interim Guidelines and this Interim Guidelines Advisory No. 2. Background The Northridge earthquake of January 17, 1994, dramatically demonstrated that the prequalified, welded beam-to-column moment connection commonly used in the construction of welded steel moment resisting frames (WSMFs) in the period 1965-1994 was much more susceptible to damage than previously thought. The stability of moment frame structures in earthquakes is dependent on the capacity of the beam-column connection to remain intact and to resist tendencies of the beams and columns to rotate with respect to each other under the influence of lateral deflection of the structure. The prequalified connections were believed to be ductile and capable of withstanding the repeated cycles of large inelastic deformation explicitly relied upon in the building code provisions for the design of these structures. Although many affected connections were not damaged, a wide spectrum of unexpected brittle connection iii

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fractures did occur, ranging from isolated fractures through or adjacent to the welds of beam flanges to columns, to large fractures extending across the full depth of the columns. At the time this damage was discovered, the structural steel industry and engineering profession had little understanding of the specific causes of this damage, the implications of this damage for building safety, or even if reliable methods existed to repair the damage which had been discovered. Although the connection failures did not result in any casualties or collapses, and many WSMF buildings were not damaged, the incidence of damage was sufficiently pervasive in regions of strong ground motion to cause wide-spread concern by structural engineers and building officials with regard to the safety of these structures in future earthquakes. In response to these concerns, the Federal Emergency Management Agency (FEMA) entered into a cooperative agreement with the SAC Joint Venture to perform problem-focused study of the seismic performance of welded steel moment connections and to develop interim recommendations for professional practice. Specifically, these recommendations were intended to address the inspection of earthquake affected buildings to determine if they had sustained significant damage; the repair of damaged buildings; the upgrade of existing buildings to improve their probable future performance; and the design of new structures to provide more reliable seismic performance. Within weeks of receipt of notification of FEMA’s intent to enter into this agreement, the SAC Joint Venture published a series of two design advisories (SAC, 1994a; SAC, 1994b). These design advisories presented a series of papers, prepared by engineers and researchers engaged in the investigation of the damaged structures and presenting individual opinions as to the causes of the damage, potential methods of repair, and possible designs for more reliable connections in the future. In February 1995, Design Advisory No. 3 (SAC, 1995a) was published. This third advisory presented a synthesis of the data presented in the earlier publications, together with the preliminary recommendations developed in an industry workshop, attended by more than 50 practicing engineers, industry representatives and researchers, on methods of inspecting, repairing and designing WSMF structures. At the time this third advisory was published, significant disagreement remained within the industry and the profession as to the specific causes of the damage observed and appropriate methods of repair given that the damage had occurred. Consequently, the preliminary recommendations were presented as a series of issue statements, followed by the consensus opinions of the workshop attendees, where consensus existed, and by majority and dissenting opinions where such consensus could not be formed. During the first half of 1995, an intensive program of research was conducted to more definitively explore the pertinent issues. This research included literature surveys, data collection on affected structures, statistical evaluation of the collected data, analytical studies of damaged and undamaged buildings and laboratory testing of a series of full-scale beam-column assemblies representing typical pre-Northridge design and construction practice as well as various repair, upgrade and alternative design details. The findings of this research (SAC 1995c, SAC 1995d, SAC 1995e, SAC 1995f, SAC 1995g, SAC 1996) formed the basis for the development of FEMA 267 - Interim Guidelines: Evaluation, Repair, Modification, and Design of Welded Steel Moment Frame Structures (SAC, 1995b), which was published in August, 1995. FEMA 267 provided the first definite, albeit interim, recommendations for practice, following the discovery of connection damage in the Northridge earthquake.

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As a result of these and supplemental studies conducted by the SAC Joint Venture, as well as independent research conducted by others, it is now known that a large number of factors contributed to the damage sustained by steel frame buildings in the Northridge earthquake. These included: • design practice that favored the use of relatively few frame bays to resist lateral seismic demands, resulting in much larger member and connection geometries than had previously been tested; • standard detailing practice which resulted in the development of large inelastic demands at the beam to column connections; • detailing practice that often resulted in large stress concentrations in the beam-column connection, as well as inherent stress risers and notches in zones of high stress; • the common use of welding procedures that resulted in deposition of low toughness weld metal in the critical beam flange to column flange joints; • relatively poor levels of quality control and assurance in the construction process, resulting in welded joints that did not conform to the applicable quality standards; • excessively weak and flexible column panel zones that resulted in large secondary stresses in the beam flange to column flange joints; • large variations in the strengths of rolled shape members relative to specified values; • an inherent inability of material to yield under conditions of high tri-axial restraint such as exist at the center of the beam flange to column flange joints. With the identification of these factors it was possible for FEMA 267 to present a recommended methodology for the design and construction of moment-resisting steel frames to provide connections capable of more reliable seismic performance. This methodology included the following recommendations: • proportion the beam-column connection such that inelastic behavior occurs at a distance remote from the column face, minimizing demands on the highly restrained column material and the welded joints; • specify weld filler metals with rated toughness values for critical welded joints; • detail connections to incorporate beam flange continuity plates, to minimize stress concentrations; • remove backing bars and weld tabs from critical joints to minimize the potential for stress risers and notch effects and also to improve the reliability with which flaws at the weld root can be observed and repaired;

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• qualify connection configurations through a program of full-scale inelastic testing of representative beam-column assemblies, fabricated in the same manner as is proposed for use in the structure; • increased participation of the design professional in the specification and surveillance of welding procedures and the quality assurance process for welded joints. In the time since the publication of FEMA-267, SAC has continued, under funding provided by FEMA, to perform problem-focused study of the performance of moment resisting connections of various configurations. This work, which is generally referred to as the SAC Phase II program, includes detailed analytical evaluations of buildings and connections, parametric studies into the effects on connection performance of connection configuration, base and weld metal strength, toughness and ductility, as well as additional large scale testing of connection assemblies. The intent of this study is to support development of final guidelines that will present more reliable and economical performance-based methods for: • identification of damaged structures following an earthquake and determination of the extent, severity and consequences of such damage; • design of effective repairs for damaged structures; • identification of existing structures that are vulnerable to unacceptable levels of damage in future earthquakes; • design of structural upgrades for existing vulnerable structures; • design of new structures that are suitably resistant to earthquake induced damage; • procedures for construction quality assurance that are consistent with the levels of reliability intended by the design criteria. This Phase II program of research, which is being conducted by the SAC Joint Venture in parallel and coordination with work by other researchers, is anticipated to be complete in late 1999. It is the intent of FEMA and the SAC Joint Venture to ensure that pertinent information and findings from this program are made available to the user community in a timely manner through the publication of this series of design advisory documents. This Interim Guidelines Advisory No. 2 is the second such publication. Format This Advisory has been prepared as a series of updates and revisions to the FEMA-267, Interim Guidelines publication. It has been formatted in a manner intended to facilitate the identification of changes to the original FEMA-267 text. Only those sections of FEMA-267 that are being revised at this time are included. Other sections of FEMA-267 remain in effect as the current best recommendations of the SAC Joint Venture. This Advisory replaces the earlier Interim Guidelines Advisory, FEMA-267a, in its entirety.

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To facilitate coordination of this Advisory with FEMA-267, the existing system of chapter and section numbering has been retained. The Table of Contents lists all sections of the chapters being revised, including those sections for which no revisions are included. Within the body of this document, a section heading is provided for each section of the chapter; however, if no revision to the section is currently being made, this is indicated immediately beneath the section heading. To facilitate reading of this document, where a revision is made to a section in FEMA 267, the entire text of that section is included herein. Where existing text from FEMA-267 is reproduced in this document, without edit, it is shown in normal face type for guidelines, and in italicized type for commentary. Where existing text is being deleted, this is shown in strike through format. A single strikethrough indicates text deleted in the first advisory, FEMA-267a. A double strikethrough indicates text deleted in this current advisory. New text is shown in underline format. A single underline identifies text added in the first advisory, FEMA-267a. A double underline identifies text added in this current advisory. When a modification has been made to a portion of text, relative to FEMA-267, this will also be noted by the presence of a vertical line at the outside margin of the page. The following two paragraphs illustrate these conventions for guideline and commentary text, respectively. This sentence is representative of typical guideline text, that has been reprinted from FEMA-267 without change.This sentence, is representative of the way in which text being deleted from FEMA-267 in this Interim Guidelines Advisory is identified. This sentence illustrates the way in which text deleted from FEMA-267 in the previous Interim Guidelines Advisory is identified. This sentence illustrates the way in which text being added to FEMA-267 in this Interim Guidelines Advisory is identified.This sentence illustrates the way in which text added to FEMA-267 in the previous Interim Guidelines Advisory is identified. Commentary: This sentence is representative of typical commentary text, that has been reprinted from FEMA-267 without change. This sentence is representative of the way in which commentary text being deleted from FEMA-267 in this Interim Guidelines Advisory is identified. However, this sentence, is representative of the way in which text being deleted from FEMA-267 commentary in the previous Advisory is identified. This sentence indicates the way in which text added to the FEMA-267 commentary in this Advisory is shown.This final sentence illustrates the way in which text added in previous advisory, FEMA-267a, is identified. Intent This Interim Guidelines Advisory, together with the Interim Guidelines they modify, are primarily intended for two different groups of potential users: a) Engineers engaged in evaluation, repair, and upgrade of existing WSMF buildings and in the design of new WSMF buildings incorporating either Special Moment-Resisting Frames or Ordinary Moment-Resisting Frames utilizing welded beam-column connections. The

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recommendations for new construction are applicable to all WSMF construction expected to resist earthquake demands through plastic behavior. b) Regulators and building departments responsible for control of the evaluation, repair, and occupancy of WSMF buildings that have been subjected to strong ground motion and for regulation of the design, construction, and inspection of new WSMF buildings. The fundamental goal of the information presented in the Interim Guidelines as modified by this Advisory is to help identify and reduce the risks associated with earthquake-induced fractures in WSMF buildings through provision of timely information on how to inspect existing buildings for damage, repair damage if found, upgrade existing buildings and design new buildings. The information presented here primarily addresses the issue of beam-to-column connection integrity under the severe inelastic demands that can be produced by building response to strong ground motion. Users are referred to the applicable provisions of the locally prevailing building code for information with regard to other aspects of building construction and earthquake damage control. Limitations The information presented in this Interim Guidelines Advisory, together with that contained in the Interim Guidelines it modifies, is based on limited research conducted since the Northridge Earthquake, review of past research and the considerable experience and judgment of the professionals engaged by SAC to prepare and review this document. Additional research on such topics as the effect of floor slabs on frame behavior, the effect of weld metal and base metal toughness, the efficacy of various beam-column connection details and the validity of current standard testing protocols for prediction of earthquake performance of structures is continuing as part of the Phase 2 program and is expected to provide important information not available at the time this Advisory was formulated. Therefore, many of the recommendations cited herein may change as a result of forthcoming research results. The recommendations presented herein represent the group consensus of the committee of Guideline Writers retained by SAC following independent review by the Project Oversight Committee. They may not reflect the individual opinions of any single participant. They do not necessarily represent the opinions of the SAC Joint Venture, the Joint Venture partners, or the sponsoring agencies. Users are cautioned that available information on the nature of the WSMF problem is in a rapid stage of development and any information presented herein must be used with caution and sound engineering judgment.

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TABLE OF CONTENTS

1

3

4

THE SAC JOINT VENTURE DISCLAIMER PREFACE Purpose Background Format Intent Limitations

ii ii iii iii iii vi vii viii

INTRODUCTION 1.1 Purpose 1.2 Scope 1.3 Background 1.4 The SAC Joint Venture 1.5 Sponsors 1.6 Summary of Phase I Research 1.7 Intent 1.8 Limitations 1.9 Use of the Guidelines

1-1 1-1 1-1 1-8 1-8 1-8 1-8 1-9 1-9

CLASSIFICATIONS AND IMPLICATIONS OF DAMAGE 3.1 Summary of Earthquake Damage 3.2 Damage Types 3.2.1 Girder Damage 3.2.2 Column Flange Damage 3.2.3 Weld Damage, Defects and Discontinuities 3.2.4 Shear Tab Damage 3.2.5 Panel Zone Damage 3.2.6 Other Damage 3.3 Safety Implications 3.4 Economic Implications

3-1 3-1 3-1 3-1 3-1 3-4 3-4 3-4 3-5 3-7

POST-EARTHQUAKE EVALUATION 4.1 Scope 4.2 Preliminary Evaluation 4.2.1 Evaluation Process 4.2.1.1 Ground Motion 4.2.1.2 Additional Indicators 4.2.2 Evaluation Schedule 4.2.3 Connection Inspections 4.2.3.1 Analytical Evaluation 4.2.3.2 Buildings with Enhanced Connections 4.2.4 Previous Evaluations and Inspections

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4.4 4.5

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Detailed Evaluation Procedure 4.3.1 Eight Step Inspection and Evaluation Procedure 4.3.2 Step 1 - Categorize Connections By Group 4.3.3 Step 2 - Select Samples of Connections for Inspection 4.3.3.1 Method A - Random Selection 4.3.3.2 Method B - Deterministic Selection 4.3.3.3 Method C - Analytical Selection 4.3.4 Step 3- Inspect the Selected Samples of Connections 4.3.4.1 Damage Characterization 4.3.5 Step 4 - Inspect Connections Adjacent to Damaged Connections 4.3.6 Step 5 - Determine Average Damage Index for the Group 4.3.7 Step 6 - Determine the Probability that the Connections in a Group at a Floor Level Sustained Excessive Damage 4.3.7.1 Some Connections In Group Not Inspected 4.3.7.2 All Connections in Group Inspected 4.3.8 Step 7 - Determine Recommended Recovery Strategies for the Building 4.3.9 Step 8 - Evaluation Report Alternative Group Selection for Torsional Response Qualified Independent Engineering Review 4.5.1 Timing of Independent Review 4.5.2 Qualifications and Terms of Employment 4.5.3 Scope of Review 4.5.4 Reports 4.5.5 Responses and Corrective Actions 4.5.6 Distribution of Reports 4.5.7 Engineer of Record 4.5.8 Resolution of Differences

POST-EARTHQUAKE INSPECTION 5.1 Connection Types Requiring Inspection 5.1.1 Welded Steel Moment Frame (WSMF) Connections 5.1.2 Gravity Connections 5.1.3 Other Connection Types 5.2 Preparation 5.2.1 Preliminary Document Review and Evaluation 5.2.1.1 Document Collection and Review 5.2.1.2 Preliminary Building Walk-Through 5.2.1.3 Structural Analysis 5.2.1.4 Vertical Plumbness Check 5.2.2 Connection Exposure 5.3 Inspection Program 5.3.1 Visual Inspection (VI) 5.3.1.1 Top Flange 5.3.1.2 Bottom Flange

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5.3.2 5.3.3 5.3.4 5.3.5 6

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5.3.1.3 Column and Continuity Plates 5.3.1.4 Beam Web Shear Connection Nondestructive Testing (NDT) Inspector Qualification Post-Earthquake Field Inspection Report Written Report

5-6 5-7 5-7 5-9 5-9 5-9

POST-EARTHQUAKE REPAIR AND MODIFICATION 6.1 Scope 6.2 Shoring 6.3 Repair Details 6.4 Preparation 6.5 Execution 6.6 Structural Modification 6.6.1 Definition of Modification 6.6.2 Damaged vs. Undamaged Connections 6.6.3 Criteria 6.6.4 Strength and Stiffness 6.6.4.1 Strength 6.6.4.2 Stiffness 6.6.5 Plastic Rotation Capacity 6.6.6 Connection Qualification and Design 6.6.6.1 Qualification Test Protocol 6.6.6.2 Acceptance Criteria 6.6.6.3 Calculations 6.6.6.3.1 Material Strength Properties 6.6.6.3.2 Determine Plastic Hinge Location 6.6.6.3.3 Determine Probable Plastic Moment at Hinges 6.6.6.3.4 Determine Beam Shear 6.6.6.3.5 Determine Strength Demands on Connection 6.6.6.3.6 Check Strong Column - Weak Beam Conditions 6.6.6.3.7 Check Column Panel Zone 6.6.7 Modification Details 6.6.7.1 Haunch at Bottom Flange 6.6.7.2 Top and Bottom Haunch 6.6.7.3 Cover Plate Sections 6.6.7.4 Upstanding Ribs 6.6.7.5 Side-Plate Connections 6.6.7.6 Bolted Brackets

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NEW CONSTRUCTION 7.1 Scope 7.2 General - Welded Steel Frame Design Criteria 7.2.1 Criteria 7.2.2 Strength and Stiffness

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7.3

7.4

7.5

7.6 7.7 7.8

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7.2.2.1 Strength 7.2.2.2 Stiffness 7.2.3 Configuration 7.2.4 Plastic Rotation Capacity 7.2.5 Redundancy 7.2.6 System Performance 7.2.7 Special Systems Connection Design and Qualification Procedures - General 7.3.1 Connection Performance Intent 7.3.2 Qualification by Testing 7.3.3 Design by Calculation Guidelines for Connection Qualification by Testing 7.4.1 Testing Protocol 7.4.2 Acceptance Criteria Guidelines for Connection Design by Calculation 7.5.1 Material Strength Properties 7.5.2 Design Procedure - Strengthened Connections 7.5.2.1 Determine Plastic Hinge Locations 7.5.2.2 Determine Probable Plastic Moment at Hinge 7.5.2.3 Determine Shear at Plastic Hinge 7.5.2.4 Determine Strength Demands at Critical Sections 7.5.2.5 Check for Strong Column - Weak Beam Condition 7.5.2.6 Check Column Panel Zone 7.5.3 Design Procedure - Reduced Beam Section Connections 7.5.3.1 Determine Reduced Section and Plastic Hinge Locations 7.5.3.2 Determine Strength and Probable Plastic Moment in RBS 7.5.3.3 Strong Column - Weak Beam Condition 7.5.3.4 Column Panel Zone 7.5.3.5 Lateral Bracing 7.5.3.6 Welded Attachments Metallurgy & Welding Quality Control / Quality Assurance Guidelines on Other Connection Design Issues 7.8.1 Design of Panel Zones 7.8.2 Design of Web Connections to Column Flanges 7.8.3 Design of Continuity Plates 7.8.4 Design of Weak Column and Weak Way Connections Moment Frame Connections for Consideration in New Construction 7.9.1 Cover Plate Connections 7.9.2 Flange Rib Connections 7.9.3 Bottom Haunch Connections 7.9.4 Top and Bottom Haunch Connections 7.9.5 Side-Plate Connections 7.9.6 Reduced Beam Section Connections 7.9.7 Slip-Friction Energy Dissipating Connections

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7.9.8 Column Tree Connections 7.9.9 Slotted Web Connections 7.9.10 Bolted Bracket Connections Other Types of Welded Connection Structures 7.10.1 Eccentrically Braced Frames (EBF) 7.10.2 Dual Systems 7.10.3 Welded Base Plate Details 7.10.4 Vierendeel Truss Systems 7.10.5 Moment Frame Tubular Systems 7.10.6 Welded Connections of Collectors, Ties and Diaphragm Chords 7.10.7 Welded Column Splices 7.10.8 Built-up Moment Frame Members

METALLURGY & WELDING 8.1 Parent Materials 8.1.1 Steels 8.1.2 Chemistry 8.1.3 Tensile/Elongation Properties 8.1.4 Toughness Properties 8.1.5 Lamellar Discontinuities 8.1.6 K-Area Fractures 8.2 Welding 8.2.1 Welding Process 8.2.2 Welding Procedures 8.2.3 Welding Filler Metals 8.2.4 Preheat and Interpass Temperatures 8.2.5 Postheat 8.2.6 Controlled Cooling 8.2.7 Metallurgical Stress Risers 8.2.8 Welding Preparation & Fit-up

REFERENCES

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1. INTRODUCTION 1.1 Purpose There are no modifications to the Guidelines or Commentary of Section 1.1 at this time. 1.2 Scope There are no modifications to the Guidelines or Commentary of Section 1.2 at this time. 1.3 Background Following the January 17, 1994 Northridge, California Earthquake, more than 100 steel buildings with welded moment-resisting frames were found to have experienced beam-to-column connection fractures. The damaged structures cover a wide range of heights ranging from one story to 26 stories; and a wide range of ages spanning from buildings as old as 30 years of age to structures just being erected at the time of the earthquake. The damaged structures are were spread over a large geographical area, including sites that experienced only moderate levels of ground shaking. Although relatively few such buildings were located on sites that experienced the strongest ground shaking, damage to these buildings was quite severe. Discovery of these extensive connection fractures, often with little associated architectural damage to the buildings, was has been alarming. The discovery has also caused some concern that similar, but undiscovered damage may have occurred in other buildings affected by past earthquakes. Indeed, there are now confirmed isolated reports of such damage. In particular, a publicly owned building at Big Bear Lake is known to have been was damaged by the Landers-Big Bear, California sequence of earthquakes, and at least one building, under construction in Oakland, California at the time fo the several buildings were damaged during the 1989 Loma Prieta Earthquake, was reported to have experienced such damage in the San Francisco Bay Area. WSMF construction is used commonly throughout the United States and the world, particularly for mid- and high-rise construction. Prior to the Northridge Earthquake, this type of construction was considered one of the most seismic-resistant structural systems, due to the fact that severe damage to such structures had rarely been reported in past earthquakes and there was no record of earthquakeinduced collapse of such buildings, constructed in accordance with contemporary US practice. However, the widespread severe structural damage which occurred to such structures in the Northridge Earthquake calleds for re-examination of this premise. The basic intent of the earthquake resistive design provisions contained in the building codes is to protect the public safety, however, there is also an intent to control damage. The developers of the building code provisions have explicitly set forth three specific performance goals for buildings designed and constructed to the code provisions (SEAOC - 1990). These are to provide buildings with the capacity to • resist minor earthquake ground motion without damage; Introduction 1-1

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• resist moderate earthquake ground motion without structural damage but possibly some nonstructural damage; and • resist major levels of earthquake ground motion, having an intensity equal to the strongest either experienced or forecast for the building site, without collapse, but possibly with some structural as well as nonstructural damage. In general, WSMF buildings in the Northridge Earthquake met the basic intent of the building codes, to protect life safety. However, the ground shaking intensity experienced by most of these buildings was significantly less than that anticipated by the building codes. Many buildings that experienced moderate intensity ground shaking experienced significant damage that could be viewed as failing to meet the intended performance goals with respect to damage control. Further, some members of the engineering profession (SEAOC - 1995b) and government agencies (Seismic Safety Commission - 1995) have stated that even these performance goals are inadequate for society’s current needs. WSMF buildings are designed to resist earthquake ground shaking based on the assumption that they are capable of extensive yielding and plastic deformation, without loss of strength. The intended plastic deformation is intended to be developed through a combination of consists of plastic rotations developing within the beams, at their connections to the columns, and plastic shear yielding of the column panel zones,. and is tTheoretically these mechanisms should be capable of resulting in benign dissipation of the earthquake energy delivered to the building. Damage is expected to consist of moderate yielding and localized buckling of the steel elements, not brittle fractures. Based on this presumed behavior, building codes require a minimum lateral design strength for WSMF structures that is approximately 1/8 that which would be required for the structure to remain fully elastic. Supplemental provisions within the building code, intended to control the amount of interstory drift sustained by these flexible frame buildings, typically result in structures which are substantially stronger than this minimum requirement and in zones of moderate seismicity, substantial overstrength may be present to accommodate wind and gravity load design conditions. In zones of high seismicity, most such structures designed to minimum code criteria will not start to exhibit plastic behavior until ground motions are experienced that are 1/3 to 1/2 the severity anticipated as a design basis. This design approach has been developed based on historical precedent, the observation of steel building performance in past earthquakes, and limited research that has included laboratory testing of beamcolumn models, albeit with mixed results, and non-linear analytical studies. Observation of damage sustained by buildings in the Northridge Earthquake indicates that contrary to the intended behavior, in some many cases brittle fractures initiated within the connections at very low levels of plastic demand, and in some cases, while the structures remained essentially elastic. Typically, but not always, fractures initiated at, or near, the complete joint penetration (CJP) weld between the beam bottom flange and column flange (Figure 1-1). Once initiated, these fractures progressed along a number of different paths, depending on the individual joint and stress conditions. Figure 1-1 indicates just one of these potential fracture growth patterns. Investigators initially identified a number of factors which may have contributed to the initiation of fractures at the weld root including: notch effects created by the backing bar which was commonly left in place following joint completion; sub-standard welding that included excessive porosity and slag inclusions as well as incomplete fusion; Introduction 1-2

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and potentially, pre-earthquake fractures resulting from initial shrinkage of the highly restrained weld during cool-down. Such problems could be minimized in future construction, with the application of appropriate welding procedures and more careful exercise of quality control during the construction process. However, it is now known that these were not the only causes of the fractures which occurred. Column flange Fused zone Beam flange

Backing bar Fracture

Figure 1-1 - Common Zone of Fracture Initiation in Beam -Column Connection Current production processes for structural steel shapes result in inconsistent strength and deformation capacities for the material in the through-thickness direction. Non-metallic inclusions in the material, together with anisotropic properties introduced by the rolling process can lead to lamellar weakness in the material. Further, the distribution of stress across the girder flange, at the connection to the column is not uniform. Even in connections stiffened by continuity plates across the panel zone, significantly higher stresses tend to occur at the center of the flange, where the column web produces a local stiffness concentration. Large secondary stresses are also induced into the girder flange to column flange joint by kinking of the column flanges resulting from shear deformation of the column panel zone. The dynamic loading experienced by the moment-resisting connections in earthquakes is characterized by high strain tension-compression cycling. Bridge engineers have long recognized that the dynamic loading associated with bridges necessitates different connection details in order to provide improved fatigue resistance, as compared to traditional building design that is subject to “static” loading due to gravity and wind loads. While the nature of the dynamic loads resulting from earthquakes is somewhat different than the high cycle dynamic loads for which fatigue-prone structures are designed, similar detailing may be desirable for buildings subject to seismic loading. In design and construction practice for welded steel bridges, mechanical and metallurgical notches should be avoided because they may be the initiators of fatigue cracking. As fatigue cracks grow under repetitive loading, a critical crack size may be reached whereupon the material toughness (which is a function of temperature) may be unable to resist the onset of brittle (unstable) crack growth. The beam-to-column connections in WSMF buildings are comparable to category C or D bridge details that have a reduced allowable stress range as opposed to category B details for which special metallurgical, inspection and testing requirements are applied. The rapid rate of loading imposed by seismic events, and the complete inelastic range of tension-compression-tension loading applied to these connections is Introduction 1-3

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much more severe than typical bridge loading applications. The mechanical and metallurgical notches or stress risers created by the beam-column weld joints are a logical point for fracture problems to initiate. This, coupled with the tri-axial restraint provided by the beam web and the column flange, is a recipe for brittle fracture. During the Northridge Earthquake, oOnce fractures initiated in beam-column joints, they progressed in a number of different ways. In some cases, the fractures initiated but did not grow, and could not be detected by visual observation. In other cases, In many cases, the fractures progressed completely directly through the thickness of the weld, and if fireproofing was removed, the fractures were evident as a crack through exposed faces of the weld, or the metal just behind the weld (Figure 12a). Other fracture patterns also developed. In some cases, the fracture developed into a surface that resembled a through-thickness failure of the column flange material behind the CJP weld (Figure 1-2b). In these cases, a portion of the column flange remained bonded to the beam flange, but pulled free from the remainder of the column. This fracture pattern has sometimes been termed a “divot” or “nugget” failure. A number of fractures progressed completely through the column flange, along a near horizontal plane that aligns approximately with the beam lower flange (Figure 1-3a). In some cases, these fractures extended into the column web and progressed across the panel zone Figure (1-3b). Investigators have reported some instances where columns fractured entirely across the section.

a. Fracture at Fused Zone

b. Column Flange “Divot” Fracture

Figure 1-2 - Fractures of Beam to Column Joints

a. Fractures through Column Flange

b. Fracture Progresses into Column Web

Figure 1-3 - Column Fractures Introduction 1-4

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Once these fractures have occurred, the beam - column connection has experienced a significant loss of flexural rigidity and capacity. Residual flexural strength and rigidity must be developed through a couple consisting of forces transmitted through the remaining top flange connection and the web bolts. Initial rResearch suggests that residual stiffness is approximately 20% of that of the undamaged connection and that residual strength varies from 10% to 40% of the undamaged capacity, when loading results in tensile stress normal to the fracture plane. When loading produces compression across the fracture plane, much of the original strength and stiffness remain. However, in providing this residual strength and stiffness, the beam shear connections can themselves be subject to failures, consisting of fracturing of the welds of the shear plate to the column, fracturing of supplemental welds to the beam web or fracturing through the weak section of shear plate aligning with the bolt holes (Figure 1-4).

Figure 1-4 - Vertical Fracture through Beam Shear Plate Connection It is now known that these fractures were the result of a number of complex factors that were not well understood either when these connections were first adopted as a standard design approach, or when the damage was discovered immediately following the Northridge earthquake. Engineers had commonly assumed that when these connections were loaded to yield levels, flexural stresses in the beam would be transferred to the column through a force couple comprised of nearly uniform yield level tensile and compressive stresses in the beam flanges. It was similarly assumed that nearly all of the shear stress in the beam was transferred to the column through the shear tab connection to the beam web. In fact, the actual behavior is quite different from this. As a result of local deformations that occur in the column at the location of the beam connection, a significant portion of the shear stress in the beam is actually transferred to the column through the beam flanges. This causes large localized secondary stresses in the beam flanges, both at the toe of the weld access hole and also in the complete joint penetration weld at the face of the column. The presence of the column web behind the column flange tends to locally stiffen the joint of the beam flange to the column flange, further concentrating the distribution of connection stresses and strains. Finally, the presence of the heavy beam and column flange plates, arranged in a “+” shaped pattern at the beam flange to column flange joint produces a condition of very high restraint, which retards the onset of yielding, by raising the effective yield strength of the material, and allowing the development of very large stresses. Introduction 1-5

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The most severe stresses typically occur at the root of the complete joint penetration weld of the beam bottom flange to the column flange. This is precisely the region of this welded joint that is most difficult for the welder to properly complete, as the access to the weld is restricted by the presence of the beam web and the welder often performs this weld while seated on the top flange, in the so-called “wildcat” position. The welder must therefore work from both sides of the beam web, starting and terminating the weld near the center of the joint, a practice that often results in poor fusion and the presence of slag inclusions at this location. These conditions, which are very difficult to detect when the weld backing is left in place, as was the typical practice, are ready-made crack initiators. When this region of the welded joints is subjected to the large concentrated tensile stresses, the weld defects begin to grow into cracks and these cracks can quickly become unstable and propagate as brittle fractures. Once these brittle fractures initiate, they can grow in a variety of patterns, as described above, under the influence of the stress field and the properties of the base and weld metals present at the zone of the fracture. Despite the obvious local strength impairment resulting from these fractures, many damaged buildings did not display overt signs of structural damage, such as permanent drifts or extreme damage to architectural elements. Until news of the discovery of connection fractures in some buildings began to spread through the engineering community, it was relatively common for engineers to perform cursory post-earthquake evaluations of WSMF buildings and declare that they were undamaged. In order to reliably determine if a building has sustained connection damage, it is necessary to remove architectural finishes and fireproofing and perform nondestructive examination including visual inspection and ultrasonic testing careful visual inspection of the welded joints supplemented, in some cases, by nondestructive testing. Even if no damage is found, this is a costly process. Repair of damaged connections is even more costly. A few WSMF buildings have sustained so much connection damage that it has been deemed more practical to demolish the structures rather than to repair them. In the case of one WSMF building, damaged by the Northridge earthquake, repair costs were sufficiently large that the owner elected to demolish rather than replace than building. Immediately following the Northridge Earthquake, a series of tests of beam-column subassemblies were performed at the University of Texas at Austin, under funding provided by the AISC as well as private sources. The test specimens used heavy W14 column sections and deep (W36) beam sections commonly employed in some California construction. Initial specimens were fabricated using the standard prequalified connection specified by the Uniform Building Code (UBC). Section 2211.7.1.2 of UBC-94 {NEHRP-91 Section 10.10.2.3} specified this prequalified connection as follows: “2211.7.1.2 Connection strength. The girder top column connection may be considered to be adequate to develop the flexural strength of the girder if it conforms to the following: 1. the flanges have full penetration butt welds to the columns. 2. the girder web to column connection shall be capable of resisting the girder shear determined for the combination of gravity loads and the seismic shear forces which result from compliance with Section 2211.7.2.1. This connection strength need not exceed that required to develop gravity loads plus 3(Rw/8) times the girder shear resulting from the prescribed seismic forces.

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Where the flexural strength of the girder flanges is greater than 70 percent of the flexural strength of the entire section, (i.e. btf/(d-tf)Fy>0.7ZxFy) the web connection may be made by means of welding or high-strength bolting. For girders not meeting the criteria in the paragraph above, the girder web-to-column connection shall be made by means of welding the web directly or through shear tabs to the column. That welding shall have a strength capable of developing at least 20 percent of the flexural strength of the girder web. The girder shear shall be resisted by means of additional welds or friction-type slip-critical high strength bolts or both. and: 2211.7.2.1 Strength. The panel zone of the joint shall be capable of resisting the shear induced by beam bending moments due to gravity loads plus 1.85 times the prescribed seismic forces, but the shear strength need not exceed that required to develop 0.8ΣMs of the girders framing into the column flanges at the joint...”

In order to investigate the effects that backing bars and weld tabs had on connection performance, these were removed from the specimens prior to testing. Despite these precautions, the test specimens failed at very low levels of plastic loading. Following these tests at the University of Texas at Austin, reviews of literature on historic tests of these connection types indicated a significant failure rate in past tests as well, although these had often been ascribed to poor quality in the specimen fabrication. It was concluded that the prequalified connection, specified by the building code, was fundamentally flawed and should not be used for new construction in the future. In retrospect, this conclusion may have been somewhat premature. More recent testing of connections having configurations similar to those of the prequalified connection, but incorporating tougher weld metals, having backing bars removed from the bottom flange joint, and fabricated with greater care to avoid the defects that can result in crack initiation, have performed better than those initially tested at the University of Texas. However, as a class, when fabricated using currently prevailing construction practice, these connections still do not appear to be capable of consistently developing the levels of ductility presumed by the building codes for service in moment-resisting frames that are subjected to large inelastic demands.When the first test specimens for that series were fabricated, the welder failed to follow the intended welding procedures. Further, no special precautions were taken to assure that the materials incorporated in the work had specified toughness. Some engineers, with knowledge of fracture mechanics, have suggested that if materials with adequate toughness are used, and welding procedures are carefully specified and followed, adequate reliability can be obtained from the traditional connection details. Others believe that the conditions of high triaxial restraint present in the beam flange to column flange joint (Blodgett - 1995) would prevent ductile behavior of these joints regardless of the procedure used to make the welds. Further they point to the important influence of the relative yield and tensile strengths of beam and column materials, and other variables, that can affect connection behavior. To date, there has not been sufficient research conducted to resolve this issue.

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In reaction to the University of Texas tests as well as the widespread damage discovered following the Northridge Earthquake, and the urging of the California Seismic Safety Commission, in September, 1994 the International Conference of Building Officials (ICBO) adopted an emergency code change to the 1994 edition of the Uniform Building Code (UBC-94) {1994 NEHRP Recommended Provisions Section 5.2}. This code change, jointly developed by the Structural Engineers Association of California, AISI and ICBO staff, deleted the prequalified connection and substituted the following in its place: “2211.7.1.2 Connection Strength. Connection configurations utilizing welds or high-strength bolts shall demonstrate, by approved cyclic test results or calculation, the ability to sustain inelastic rotation and develop the strength criteria in Section 2211.7.1.1 considering the effect of steel overstrength and strain hardening.” “2211.7.1.1 Required strength. The girder-to-column connection shall be adequate to develop the lesser of the following: 1. The strength of the girder in flexure. 2. The moment corresponding to development of the panel zone shear strength as determined from formula 11-1.”

Unfortunately, neither the required “inelastic rotation”, or calculation and test procedures are well defined by these code provisions. Design Advisory No. 3 (SAC-1995) included an Interim Recommendation (SEAOC-1995) that attempted to clarify the intent of this code change, and the preferred methods of design in the interim period until additional research could be performed and reliable acceptance criteria for designs re-established. The State of California similarly published a joint Interpretation of Regulations (DSA-OSHPD - 1994) indicating the interpretation of the current code requirements which would be enforced by the state for construction under its control. This applied only to the construction of schools and hospitals in the State of California. The intent of these Interim Guidelines is to supplement these previously published documents and to provide updated recommendations based on the results of the limited directed research performed to date. 1.4 The SAC Joint Venture There are no modifications to the Guidelines or Commentary of Section 1.4 at this time. 1.5 Sponsors There are no modifications to the Guidelines or Commentary of Section 1.5 at this time. 1.6 Summary of Phase 1 Research There are no modifications to the Guidelines or Commentary of Section 1.6 at this time. 1.7 Intent Introduction 1-8

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There are no modifications to the Guidelines or Commentary of Section 1.7 at this time. 1.8 Limitations There are no modifications to the Guidelines or Commentary of Section 1.8 at this time. 1.9 Use of the Guidelines There are no modifications to the Guidelines or Commentary of Section 1.9 at this time.

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12. REFERENCES ATLSS, Fractographic Analysis of Specimens from Failed Moment Connections, (publication pending, title not exact)Fracture Analysis of Failed Moment Frame Weld Joints Produced in FullScale Laboratory Tests and Buildings Damaged in the Northridge Earthquake, SAC95-08, 1995. ATLSS, Testing of Welded “T” Specimens, (publication pending, title not exact), SAC, 1995 A Study of the Effects of Material and Welding Factors on Moment-Frame Weld Joint Performance Using a Small-Scale Tension Specimen. Kauffman, E.J., and Fisher, J.W., SAC9508 1995. Allen J., Personal Correspondence, Test Reports for New Detail, July 30, 1995. Allen J., Partridge, J.E., and Richard, R.M., Stress Distribution in Welded/Bolted Beam to Column Moment Connections. The Allen Company, March, 1995. American Association of State Highway and Transportation Officials, Bridge Welding Code AASHTO/AWS D1.5, 1995. American Institute of Steel Construction, Seismic Provisions for Structural Steel Buildings, April, 1997 American Institute of Steel Construction, Statistical Analysis of Charpy V-notch Toughness For Steel Wide Flange Structural Shapes, July, 1995. American Institute of Steel Construction, Manual of Steel Construction, ASD, Ninth Edition, 1989. American Institute of Steel Construction, Manual of Steel Construction, LRFD, Second Edition, 1998. American Institute of Steel Construction, Load and Resistance Factor Design Specification for Structural Steel Buildings, December 1, 1993. American Institute of Steel Construction, Specification for Structural Joints using ASTM A325 or A490 Bolts. 1985. American Institute of Steel Construction, AISC Northridge Steel Update I, October, 1994. American Welding Society, Guide for Nondestructive Inspection of Welds, AWS B1.10-86, 1986. American Welding Society, Guide for Visual Inspection of Welds, AWS B1.11-88, 1988. American Welding Society, Surface Roughness Guide for Oxygen Cutting, AWS C4.1-77, 1977. American Welding Society, Structural Welding Code - Steel AWS D1.1-94, 1994. References 12-1

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American Welding Society, Structural Welding Code – Steel AWS D1.1-98, 1998 Anderson, J.C., Johnson, R.G., Partridge, J.E., “Post Earthquake Studies of A Damaged Low Rise Office Building” Technical Report: Case Studies of Steel Moment Frame Building Performance in the Northridge Earthquake of January 17, 1994 SAC 95-07. SAC, December, 1995. Anderson, J.C., Filippou, F.C., Dynamic Response Analysis of the 18 Story Canoga Building, SAC, March, 1995. Anderson, J.C., Test Results for Repaired Specimen NSF#1, Report to AISC Steel Advisory Committee, June, 1995. Applied Technology Council, Earthquake Damage Evaluation Data for California ATC-13, Redwood City, CA 1985. Applied Technology Counicl, Procedures for Post Earthquake Safety Evaluations of Buildings ATC-20, Redwood City, CA, 1989. Applied Technology Council, Guidelines for Cyclic Seismic Testing of Components of Steel Structures, ATC-24, Redwood City, CA, 1992. Astaneh-Asl, A. Post-Earthquake Stability of Steel Moment Frames with Damaged Connections. Proceedings of the Third International Workshop on Connections in Steel Structures, University of Trento, Trento, Italy, 1995. Avent, R., “Designing Heat-Straightening Repairs,” National Steel Construction Conference Proceedings, Las Vegas, NV, AISC, 1992. Avent, R., “Engineered Heat Straightening,” National Steel Construction Conference Proceedings, San Antonio, TX, AISC, 1995. Barsom, J. M. and Korvink, S. A. “Through-thickness Properties of Structural Steels”, manuscript submitted to ASCE Journal of Structural Engineering, 1997. Beck, J.L., May, B.S., Polidori, D.C., Vanik, M.W., “Ambient Vibration Surveys of Three SteelFrame Buildings Strongly Shaken by the 1994 Northridge Earthquake”, Analytical and Field Investigations of Buildings Affected by the Northridge Earthquake of January 17, 1994, SAC 9504 Part 2, SAC, December, 1995. Bertero, V.V., and Whittaker, A. and Gilani, A., Testing of Repaired Welded Beam Column AssembliesSeismic Tesing of Full-Scale Steel Beam-Column Assemblies, SAC96-01, publication pending (title not exact), 1995X1996. Blodgett, O., “Evaluation of Beam to Column Connections”, SAC Steel Moment Frame Connection Advisory No. 3, Feb. 1995. References 12-2

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Bonowitz, D, and Youssef, N. “SAC Survey of Steel-Moment Frames Affected by the 1994 Northridge Earthquake”, Surveys and Assessment of Damage to Buildings Affected by the Northridge Earthquake of January 17, 1994 SAC 95-06, SAC, 1995. Building Seismic Safety Council. NEHRP Recommended Provisions for Seismic Regulations for New Buildings -1991 Edition FEMA 222, (Commentary FEMA 223), Washington D.C., January, 1992. Building Seismic Safety Council. NEHRP Recommended Provisions for Seismic Regulations for New Buildings -1994 Edition FEMA 222A, (Commentary FEMA223A), Washington D.C., July, 1995. Building Seismic Safety Council. NEHRP Recommended Provisions for Seismic Regulations for New Buildings and Other Structures. – 1997 Edition, FEMA 302, (Commentary FEMA303), Washington, D.C., February, 1998 Campbell, K.W. and Bazorgnia, Y., “Near Source Attentuation of Peak Horizontal Acceleration from World Wide Accelerogram Records from 1957 - 1993,” Proceedings of the Fifth National Conference on Earthquake Engineering, Chicago, Ill, 1994. Campbell, S., “Modeling of Weld Fractures Using the Drain Programs”, Technical Report: Parametric Analytical Investigations of Ground Motion and Structural Response, Northridge Earthquake of January 17, 1994 SAC95-05. SAC, 1995. Chen, S.J. and Yeh, C.H., Enhancement of Ductility of Steel Beam-to-Column Connections for Seismic Resistance, Department of Construction Engineering, National Taiwan University, May, 1995. Diererlein, G. “Summary of Building Analysis Studies” Analytical and Field Investigations of Buildings Affected by the Northridge Earthquake of January 17, 1994, SAC 95-04 Part 1, SAC, December, 1995 Durkin, M. E., “Inspection, Damage, and Repair of Steel Frame Buildings Following the Northridge Earthquake”, Technical Report: Surveys and Assessment of Damage to Buildings Affected by the Northridge Earthquake of January 17, 1994 SAC 95-06, SAC, December, 1995. Engelhardt, M.D., and Sabol, T.A. Testing of Welded Steel Moment Connections In Response to the Northridge Earthquake, Progress Report to the AISC Advisory Subcommittee on Special Moment Resisting Frame Research, October, 1994. Engelhardt, M. D., Keedong, K.M. Sabol T. A., Ho, L., Kim, H. Uzarski, J. and Abunnasar, H. “Analysis of a Six Story Steel Moment Frame Building in Santa Monica”, Analytical and Field Investigations of Buildings Affected by the Northridge Earthquake of January 17, 1994, SAC 9504 Part 1 SAC, December, 1995.

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Engelhardt, M. D., Keedong, K.M., Uzarski, J., Abunassar, H., Sabol, T.A., Ho, L., and Kim, H. “Parametric Studies on Inelastic Modeling of Steel Moment Frames”, Technical Report: Parametric Analytical Investigations of Ground Motion and Structural Response, Northridge Earthquake of January 17, 1994 SAC95-05. SAC, 1995. Engelhardt, M.D., Sabol, T. A., and Shuey, B.D. Testing of Repair Concepts for Damaged Steel Moment Connections.et. al. Testing of Repaired Welded Beam Column Assemblies, SAC96-01, publication pending (title not exact), 19951996. Englehardt, M.D. Fowler, T.J., and Barnes, C.A., Acoustic Emission Monitoring of Welded Steel Moment Connection Tests.et. al. Accoustic Emission Recordings for Welded Beam Column Assembly Tests, SAC95-08, publication pending (title not exact), 1995. Frank, K.H. “The Physical and Metallurgical Properties Of Structural Steels” State of Art Papers: Metallurgy, Fracture Mechanics, Welding, Moment Connections and Frame System Behavior SAC 95-09. SAC, September, 1996 Fillippou, F.C. “Nonlinear Static and Dynamic Analysis of Canoga Park Towers with FEAPSTRUC”, Analytical and Field Investigations of Buildings Affected by the Northridge Earthquake of January 17, 1994, SAC 95-04 Part 2, SAC., December, 1995. Fisher, J.W., Dexter, R.J., and Kauffman, E.J., “Fracture Mechanics of Welded Structural Steel Connections.” State of Art Papers: Metallurgy, Fracture Mechanics, Welding, Moment Connections and Frame System Behavior SAC 95-09. SAC, September, 1996 Forrel/Elsesser Engineers, Inc., Lawrence Berkeley National Labs Steel Joint Test - Technical Brief, San Francisco, CA, July 17, 1995. Gates, W.E., and Morden, M., “Lessons from Inspection, Evaluation, Repair and Construction of Welded Steel Moment Frames Following the Northridge Earthquake”, Surveys and Assessment of Damage to Buildings Affected by the Northridge Earthquake of January 17, 1994 SAC 95-06 SAC, December, 1995. Gates, W.E. “Interpretation of SAC Survey Data on Damaged Welded Steel Moment Frames Following the Northridge Earthquake”, Surveys and Assessment of Damage to Buildings Affected by the Northridge Earthquake of January 17, 1994 SAC 95-06, SAC, December, 1995. Green, M. “Santa Clarita City Hall; Northridge Earthquake Damage” Technical Report: Case Studies of Steel Moment Frame Building Performance in the Northridge Earthquake of January 17, 1994 SAC 95-07. SAC, December, 1995. Hall, J.F., “Parameter Study of the Response of Moment-Resisting Steel Frame Buildings to Near-Source Ground Motions”, Technical Report: Parametric Analytical Investigations of Ground Motion and Structural Response, Northridge Earthquake of January 17, 1994 SAC9505. SAC, 1995.

References 12-4

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Hajjar, J.F., O’Sullivan D.P., Leon, R. T., Gourley, B.C. “Evaluation of the Damage to the Borax Corporate Headquarters Building As A Result of the Northridge Earthquake”, Technical Report: Case Studies of Steel Moment Frame Building Performance in the Northridge Earthquake of January 17, 1994 SAC 95-07. SAC, December, 1995. Harrison, P.L. and Webster, S.E., Examination of Two Moment Resisting Frame Connectors Utilizing a Cover-Plate Design, British Steel Technical, Swinden Laboratories, Moorgate, Rotherham, 1995. Hart, G.C., Huang, S.C., Lobo, R.F., Van Winkle, M., Jain, A., “Earthquake Response of Strengthened Steel Special moment Resisting Frames” Analytical and Field Investigations of Buildings Affected by the Northridge Earthquake of January 17, 1994, SAC 95-04 Part 1, SAC., December, 1995 Hart, G.C., Huang, S., Lobo, R., and Stewart, J., “Elastic and Inelastic Analysis for Weld Failure Prediction of Two Adjacent Steel Buildings”, ” Analytical and Field Investigations of Buildings Affected by the Northridge Earthquake of January 17, 1994, SAC 95-04 Part 1, SAC, December, 1995. Hart, G.C., Huang, S., Lobo, R., and Stewart, J., “Influence of Vertical Ground Motion on Special Moment-Resisting Frames”, Technical Report: Parametric Analytical Investigations of Ground Motion and Structural Response, Northridge Earthquake of January 17, 1994 SAC9505. SAC, 1995. Heaton, T.H., Hall, J.F., Wald, D.J., and Halling, M.W. “Response of High-Rise and BaseIsolated Buildings to a Hypothetical Mw 7.0 Blind Thrust Earthquake” Science Vol. 26, pp 206211, January, 1995. International Conference of Building Officials, Uniform Building Code UBC-97, Whittier, CA, 1997. International Conference of Building Officials, Uniform Building Code UBC-94. Whittier, CA, 1994. Iwan, W.D., “Drift Demand Spectra for Selected Northridge Sites”, Technical Report: Parametric Analytical Investigations of Ground Motion and Structural Response, Northridge Earthquake of January 17, 1994 SAC95-05. SAC, 1995. Joyner, W.B., and Boore, D.M., “Ground Motion Parameters for Seismic Design,”Bulletin of the Sesimological Society of America, 1994. Kariotis, J. and Eimani, T.J., “Analysis of a Sixteen Story Steel Frame Building at Site 5, for the Northridge Earthquake”, Analytical and Field Investigations of Buildings Affected by the Northridge Earthquake of January 17, 1994, SAC 95-04 Part 2, SAC, December, 1995.

References 12-5

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Krawinkler, H.K., “Systems Behavior of Structural Steel Frames Subjected to Earthquake Ground Motions” State of Art Papers: Metallurgy, Fracture Mechanics, Welding, Moment Connections and Frame System Behavior SAC 95-09. SAC, September, 1996 Krawinkler, H.K., Ali, A.A., Thiel, C.C., Dunlea, J.M., “Analysis of a Damaged 4-Story Building and an Undamaged 2- Story Building”, Analytical and Field Investigations of Buildings Affected by the Northridge Earthquake of January 17, 1994, SAC 95-04 Part 1, SAC, December, 1995. Ksai, K. , and Bleiman, D. “Bolted Brackets for Repair of Damaged Steel Moment Frame Connections,” 7th U.S.-Japan Workshop on the Improvement of Structural Design and Construction Practices: Lessons Learned from Northridge and Kobe, Kobe, Japan, January, 1996 Leon, R. T., “Seismic Performance of Bolted and Riveted Connections” State of Art Papers: Metallurgy, Fracture Mechanics, Welding, Moment Connections and Frame System Behavior SAC 95-09. SAC, September, 1996 Miller, D.K. “Welding of Seismically Resistant Steel Structures” State of Art Papers: Metallurgy, Fracture Mechanics, Welding, Moment Connections and Frame System Behavior SAC 95-09. SAC, September, 1996 Naeim F., DiJulio, R., Benuska, K., Reinhorn, A. M., and Chen, L. “Evaluation of Seismic Performance of an 11 Story Steel Moment Frame Building During the 1994 Northridge Earthquake”, ” Analytical and Field Investigations of Buildings Affected by the Northridge Earthquake of January 17, 1994, SAC 95-04 Part 2 SAC, December, 1995. Newmark, N.M. and Hall W.J., Earthquake Spectra and Design. Earthquake Engineering Research Institute, 1982. NIST and AISC. Modification of Existing Welded Steel Moment Frame Connections for Seismic Resistance. National Institute of Standards and Technology and American Institute of Steel Construction. 1999 Paret, T.F., Sasaki, K.K., “Analysis of a 17 Story Steel Moment Frame Building Damaged by the Northridge Earthquake”, Analytical and Field Investigations of Buildings Affected by the Northridge Earthquake of January 17, 1994, SAC 95-04 Part 2, SAC, December, 1995. Popov, E.P. and Yang, T.S. Steel Seismic Moment Resisting Connections. University of California at Berkeley, May, 1995. Popov, E.P. Blondet, M., Stepanov, L, and Stodjadinovic, B. Full-Scale Beam-Column Connection Tests. et. al. Testing of Repaired Welded Beam Column Assemblies, SAC, publication pending (title not exact), 1995 SAC 96-01. 1996.. SAC, Proceedings of the International Workshop on Steel Moment Frames, October 23-24, 1994 SAC-94-01. Sacramento, CA, December, 1994.

References 12-6

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SAC . Steel Moment Frame Advisory No. 1. September, Sacramento, CA, 1994. SAC . Steel Moment Frame Advisory No. 2. October, Sacramento, CA, 1994. SAC . Steel Moment Frame Advisory No. 3 SAC-95-01, February, Sacramento, CA, 1995. Shonafelt, G.O., and Horn, W.B.. Guidelines for Evaluation and Repair of Damaged Steel Bridge Members, NCHRP Report 271, Transportation Research Board, 1984. Skiles, J.L. and Campbell, H.H., “Why Steel Fractured in the Northridge Earthquake” SAC Advisory No. 1, October, 1994. Seismic Safety Commission, Northridge Earthquake Turning Loss to Gain, Report to the Governor, Sacramento, CA, 1995. Smith Emery Company. Report of Test, July, 1995. Sommerville, P, Graves, R., Chandan, S. Technical Report: Characterization of Ground Motion During the Northridge Earthquake of January 17, 1994, SAC 95-03, SAC, December, 1995. State of California. Division of the State Architect (DSA) and Office of Statewide Health Planning and Development (OSHPD). Interpretation of Regulations Steel Moment Resisting Frames, Sacramento, CA, 1994. Structural Engineers Association of California (SEAOC), Seismology Committee, Recommended Lateral Force Requirements and Commentary, Sacramento, CA. 1990. Structural Engineers Association of California (SEAOC), Seismology Committee, Interim Recommendations for Design of Steel Moment Resisting Connection,. Sacramento, CA, January, 1995. Structural Engineers Association of California (SEAOC), Vision 2000: A Framework for Performance Based Engineering of Buildings, Sacramento, CA, April, 1995. Structural Shape Producers Council, Statistical Analysis of Tensile Data for Wide Flange Structural Shapes, 1994. Thiel, C.C., and Zsutty, T.C., “Earthquake Characteristics and Damage Statistics,” Earthquake Spectra, Volume 3, No. 4., Earthquake Engineering Research Institute, Oakland, Ca. 1987. Tremblay, R., Tchebotarev, N., and Filiatrault, A., “Seismic Performance of RBS Connections for Steel Moment Resisting Frames: Influence of Loading Rate and Floor Slab,” Proceedings of the Second International Conference on the Behavior of Steel Structures in Seismic Area, Kyoto, Japan, August, 1997

References 12-7

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Tsai, K.C. and Popov, E. P. “Seismic Steel Beam-Column Moment Connections” State of Art Papers: Metallurgy, Fracture Mechanics, Welding, Moment Connections and Frame System Behavior SAC 95-09. SAC, September, 1996 Uang, C.M. and Latham, C.T. Cyclic Testing of Full-Scale MNH-SMRF Moment Connections, Structural Systems Research, University of California, San Diego, March, 1995. Tsai, K.C. and Popov, E.P., Steel Beam - Column Joints In Seismic Moment Resisting Frames, Report No. UCB/EERC-88/19, Earthquake Engineering Research Center, University of California, Berkeley, Nov., 1988. Uang, C.M., Yu, Q.S., Sadre, A., Bonowitz, D., Youssef, N. “Performance of a 13 Story Steel Moment-Resisting Frame Damaged in the 1994 Northridge Earthquake”, ” Analytical and Field Investigations of Buildings Affected by the Northridge Earthquake of January 17, 1994, SAC 9504 Part 2 SAC, December, 1995. Uang, C.M. and Bondad, D. Progress Report on Cyclic Testing of Three Repaired UCSD Specimens, SAC, 1995. Uang, C.M. and Lee, C.H. “Seismic Response of Haunch Repaired Steel SMRFs: Analytical Modelling and Case Studies” ” Analytical and Field Investigations of Buildings Affected by the Northridge Earthquake of January 17, 1994, SAC 95-04 Part 2, SAC., December, 1995 Wald, D.J., Heaton, T.H., and Hudnut, K.W., The Slip History of the 1994 Northridge, California, Earthquake Determined from Strong-Motion, Teleseismic, GPS, and Leveling Data, United Sates Geologic Survey, 1995. Watabe, M. Peformance of Wooden Houses and Steel Buildings during the Great Hanshin Earthquake, Architectural Institute of Japan, May, 1995. Youssef, N.F.G, Bonowitz, D., and Gross, J.L., A Survey of Steel Moment-Resisting Frame Buildings Affected by the 1994 Northridge Earthquake, NISTR 5625, Gaithersburg Md, April, 1995.

References 12-8

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3. CLASSIFICATION AND IMPLICATIONS OF DAMAGE 3.1 Summary of Earthquake Damage There are no modifications to the Guidelines or Commentary of Section 3.1 at this time. 3.2 Damage Types There are no modifications to the Guidelines or Commentary of Section 3.2 at this time. 3.2.1 Girder Damage

There are no modifications to the Guidelines or Commentary of Section 3.2.1 at this time. 3.2.2 Column Flange Damage

There are no modifications to the Guidelines or Commentary of Section 3.2.2 at this time. 3.2.3 Weld Damage, Defects and Discontinuities

Six types of weld discontinuities, defects and damage are defined in Table 3-3 and illustrated in Figure 3-4. All apply to the complete joint penetration (CJP) welds between the girder flanges and the column flanges. This category of damage was the most commonly reported type fFollowing the Northridge Earthquake, many instances of W1a and W1b conditions were reported as damage. These conditions, which are detectable only by ultrasonic testing or by removal of weld backing, are now thought more likely to be construction defects than damage. Table 3-3 - Types of Weld Damage, Defects and Discontinuities Type W1 W1a W1b W2 W3 W4 W5

Description Weld root indications Incipient indications -– depth bf/4 Crack through weld metal thickness Fracture at girder interface Fracture at column interface Root indication— non-rejectable Partial crack at weld to column (beam flanges sound) Partial crack at weld to column (beam flange cracked) Crack in Supplemental Weld (beam flanges sound) Crack in Supplemental Weld (beam flange cracked) Fracture through tab at bolt holes Yielding or buckling of tab Damaged, or missing bolts4 Full length fracture of weld to column Fracture, buckle, or yield of continuity plate3 Fracture of continuity plate welds3 Yielding or ductile deformation of web3 Fracture of doubler plate welds3 Partial depth fracture in doubler plate3 Partial depth fracture in web3 Full (or near full) depth fracture in web or doubler plate3 Web buckling3 Fully severed column

Index2dj 4 1 8 8 10 4 10 8 4 8 8 8 6 8 8 01 04 8 8 8 0 4 8 1 8 10 6 6 10 4 4 1 4 4 8 8 6 10

Notes To Table 4-3a: 1. See Figures 3-2 through 3-6 for illustrations of these types of damage. 2. Where multiple damage types have occurred in a single connection, then: a. Sum the damage indices for all types of damage with d=1 and treat as one type. If multiple types still exist; then:

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b. For two types of damage refer to Table 4-3b. If the combination is not present in Table 4-3b and the damage indices for both types are greater than or equal to 4, use 10 as the damage index for the connection. If one is less than 4, use the greater value as the damage index for the connection. c. If three or more types of damage apply and at least one is greater than 4, use an index value of 10, otherwise use the greatest of the applicable individual indices. 3. Panel zone damage should be reflected in the damage index for all moment connections attached to the damaged panel zone within the assembly. 4. Missing or loose bolts may be a result of construction error rather than damage. The condition of the metal around the bolt holes, and the presence of fireproofing or other material in the holes can provide clues to this. Where it is determined that construction error is the cause, the condition should be corrected and a damage index of “0” assigned.

Table 4-3b - Connection Damage Indices for Common Damage Combinations1 Girder, Column or Weld Damage

Shear Tab Damage

Damage Index

Girder, Column or Weld Damage

Shear Tab Damage

Damage Index

G3 or G4

S1a S1b S2a S2b S3 S4 S5 S6 S1a S1b S2a S2b S3 S4 S5 S6 S1a S1b S2a S2b S3 S4 S5 S6

8 10 8 10 10 10 10 10 8 10 8 10 10 10 10 10 8 10 8 10 10 10 10 10

C5

S1a S1b S2a S2b S3 S4 S5 S6 S1a S1b S2a S2b S3 S4 S5 S6

6 10 6 10 10 10 10 10 8 10 8 10 10 10 10 10

C2

C3 or C4

1.

W2, W3, or W4

See Table 4-3a, footnote 2 for combinations other than those contained in this table.

More complete descriptions (including sketches) of the various types of damage are provided in Section 3.1. When the engineer can show by rational analysis that other values for the relative severities of damage are appropriate, these may be substituted for the damage indices provided in Post Earthquake Evaluation 4-7

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the tables. A full reporting of the basis for these different values should be provided to the building official, upon request. Commentary: The connection damage indices provided in Table 4-3 (ranging from 0 to 10) represent judgmental estimates of the relative severities of this damage. An index of 0 indicates no damage and an index of 10 indicates very severe damage. When initially developed, these connection damage indices were conceptualized as estimates of the connection’s lost capacity to reliably participate in the building’s lateral-force-resisting system in future earthquakes (with 0 indicating no loss of capacity and 10 indicating complete loss of capacity). However, due to the limited data available, no direct correlation between these damage indices and the actual residual strength and stiffness of a damaged connection was ever made. They do provide a convenient measure, however, of the extent of damage that various connections in a building have experienced. When FEMA-267 was first published, weld root discontinuities, Type W1a and defects, type W1b, were classified as damage in Table 4-3a with damage indices of 1 and 4, respectively assigned. Recent evidence and investigations, however, suggest strongly that these W1 conditions are not likely to be damage, and also are difficult to reliably detect. As a result, with the publication of Interim Guidelines Advisory No. 2, the damage indices for these conditions has been reduced to a null value, consistent with classifying them as pre-existing conditions, rather than damage. It should be noted that the reduced damage index associated with these conditions is not intended to indicate that these are not a concern with regard to future performance of the building. In particular, type W1b conditions can serve as ready initiators for the types of brittle fractures associated with the other damage types and connections having such conditions are more susceptible to future earthquake-induced damage than connections that do not have these conditions. Correction of these conditions should generally be considered an upgrade or modification, rather than a damage repair. 4.3.5 Step 4— Inspect Connections Adjacent to Damaged Connections

There are no modifications to the Guidelines or Commentary of Section 4.3.5 at this time. 4.3.6 Step 5— Determine Average Damage Index for Each Group

There are no modifications to the Guidelines or Commentary of Section 4.3.6 at this time.

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4.3.7 Step 6— Determine the Probability that the Connections in a Group at a Floor Level Sustained Excessive Damage

There are no modifications to the Guidelines or Commentary of Section 4.3.7 at this time. 4.3.7.1 Some Connections in Group Not Inspected

There are no modifications to the Guidelines or Commentary of Section 4.3.7.1 at this time. 4.3.7.2 All Connections in Group Inspected

There are no modifications to the Guidelines or Commentary of Section 4.3.7.2 at this time. 4.3.8 Step 7— Determine Recommended Recovery Strategies for the Building

There are no modifications to the Guidelines or Commentary of Section 4.3.8 at this time. 4.3.9 Step 8 - Evaluation Report

There are no modifications to the Guidelines or Commentary of Section 4.3.9 at this time. 4.4 Alternative Group Selection for Torsional Response There are no modifications to the Guidelines or Commentary of Section 4.4 at this time. 4.5 Qualified Independent Engineering Review There are no modifications to the Guidelines or Commentary of Section 4.5 at this time. 4.5.1 Timing of Independent Review

There are no modifications to the Guidelines or Commentary of Section 4.5.1 at this time. 4.5.2 Qualifications and Terms of Employment

There are no modifications to the Guidelines or Commentary of Section 4.5.2 at this time. 4.5.3 Scope of Review

There are no modifications to the Guidelines or Commentary of Section 4.5.3 at this time. 4.5.4 Reports

There are no modifications to the Guidelines or Commentary of Section 4.5.4 at this time.

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4.5.5 Responses and Corrective Actions

There are no modifications to the Guidelines or Commentary of Section 4.5.5 at this time. 4.5.6 Distribution of Reports

There are no modifications to the Guidelines or Commentary of Section 4.5.6 at this time. 4.5.7 Engineer of Record

There are no modifications to the Guidelines or Commentary of Section 4.5.7 at this time. 4.5.8 Resolution of Differences

There are no modifications to the Guidelines or Commentary of Section 4.5.8 at this time.

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5. POST-EARTHQUAKE INSPECTION When required by the building official, or recommended by the Interim Guidelines in Chapter 4, post-earthquake inspections of buildings may be conducted in accordance with the Interim Guidelines of this Chapter. In order to determine, with certainty, the actual post-earthquake condition of a building, it is necessary to inspect all elements and their connections. However, it is permissible to select An an appropriate sample (or samples) of WSMF connections should be selected for inspection in accordance with the Chapter 4 Guidelines. These connections, and others deemed appropriate by the engineer, should be subjected to visual inspection (VI) and supplemented by non-destructive testing (NDT) as required by this Chapter. Commentary: The only way to be certain that all damage sustained by a building is detected is to perform complete inspections of every structural element and connection. In most cases, such exhaustive post-earthquake inspections would be both economically impractical and also unnecessary. As recommended by these guidelines, the purpose of post-earthquake inspections is not to detect all damage that has been sustained by a building, but rather, to detect with reasonable certainty, that damage likely to result in a significant degradation in the building’s ability to resist future loading. The connection sampling process, suggested by Chapter 4 of these Interim Guidelines was developed to provide a low probability that damage in buildings that had sustained a substantial reduction in load carrying capacity would be overlooked while avoiding the performance of exhaustive investigations of buildings that have sustained relatively insignificant damage. Where greater certainty in the detection of damage is desired for a building, a more extensive program of inspection can be conducted. For those cases in which it is desired to perform an analytical determination of the residual load carrying capacity of the structure, complete inspections of elements and connections should be performed so that an analytical model of the building can be developed that reasonably represents its post-earthquake condition. 5.1 Connection Types Requiring Inspection 5.1.1 Welded Steel Moment Frame (WSMF) Connections

The inspection of a WSMF connection should start with visual inspection of the welded bottom beam flange to column flange joint and the base materials immediately adjacent to this joint. If damage to this joint is apparent, or suspected, then inspections of that connection should be extended to include the complete joint penetration (CJP) groove welds connecting both top and bottom beam flanges to the column flange, including the backing bar and the weld access holes in the beam web; the shear tab connection, including the bolts, supplemental welds and Post-Earthquake Inspection 5-1

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beam web; the column's web panel zone, including doubler plates; and the continuity plates and continuity plate welds (See Figure 3-1). In addition, where visual inspection indicates potential concealed damage, visual inspection should be supplemented with other methods of nondestructive testing. Commentary: The largest concentration of reported damage following the Northridge Earthquake occurred at the welded joint between the bottom girder flange and column, or in the immediate vicinity of this joint. To a much lesser extent, damage was also observed in some buildings at the joint between the top girder flange and column. If damage at either of these locations is substantial (dj per Chapter 4 greater than 5), then damage is also commonly found in the panel zone or shear tab areas. When originally published,These these Interim Guidelines recommended complete inspection, by visual and NDT assisted means, of all of these potential damage areas for a small representative sample of connections. This practice is was consistent with that followed by most engineers in the Los Angeles area, following the Northridge Earthquake. It requires removal of fireproofing from a relatively large surface of the steel framing, which at most connections will be undamaged. In the time since the Interim Guidelines were first published, extensive investigations have been conducted of the statistical distribution of damage sustained by buildings in the Northridge earthquake, the nature of this damage and the effect of this damage on the future load-carrying capacity of the buildings. These investigations strongly suggest that the W1a and W1b conditions at the weld root are unlikely to be earthquake damage, but rather, conditions of discontinuity and defects from the original construction. Further, studies have shown that NDT methods are generally unreliable in the detection of these conditions. As a result, the current recommendation is not to conduct exhaustive NDT investigations of connections in order to discover hidden damage, as was originally recommended. In a series of analytical investigations of the effect of moment-resisting connection damage on building behavior, it was determined that even if a large number of connections experience fracture at one beam flange to column joint, there is relatively little increase in the probability of global collapse in a future earthquake. Similarly, these investigations indicate that if both the top and bottom beam flange to column joints fracture in a large a number of connections, a very significant increase in the probability of global building collapse occurs. Therefore, to reduce the costs associated with post-earthquake inspections, with the publication of Interim Guidelines Advisory No.2 it is recommended that postearthquake inspections initially be limited to visual inspection of the beam bottom Post-Earthquake Inspection 5-2

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flange to column joint region. If there is evidence of potential damage in this region that is not directly observable by visual means, for example, a gap between the weld backing and column flange, then supplemental investigations of this joint should be conducted using NDT. Similarly, if it is determined that fractures have occurred at the beam bottom flange joint, then inspections of that connection should be extended to encompass the entire connection including the top beam flange joint, the shear tab and column panel zone. This approach was permitted as an alternate, in the original publication of the Interim Guidelines. Some engineers have suggested an alternative approach consisting of visual only inspections, limited to the girder bottom flange to column joint, but for a very large percentage of the total connections in the building. These bottom flange joint connections can be visually inspected with much less fireproofing removed from the framing surfaces. When significant damage is found at the exposed bottom connection, then additional fireproofing is removed to allow full exposure of the connection and inspection of the remaining surfaces. These engineers feel that by inspecting more connections, albeit to a lesser scope than recommended in these Interim Guidelines, their ability to locate the most severe occurrences of damage in a building is enhanced. These engineers use NDT assisted inspection on a very small sample of the total connections exposed to obtain an indication of the likelihood of hidden problems including damage types. If properly executed, such an approach can provide sufficient information to evaluate the post-earthquake condition of a building and to make appropriate occupancy, structural repair and/or modification decisions. It is important that the visual inspector be highly trained and that visual inspections be carefully performed, preferably by a structural engineer. Casual observation may miss clues that hidden damage exists. If, as a result of the partial visual inspection, there is any reason to believe that damage exists at a connection (such as small gaps between the CJP weld backing and column face), then complete inspection of the suspected connection, in accordance with the recommendations of these Interim Guidelines should be performed. If this approach is followed, it is recommended that a significantly larger sample of connections than otherwise recommended by these Interim Guidelines, perhaps nearly all of the connections, be inspected. 5.1.2 Gravity Connections

There are no modifications to the Guidelines or Commentary of Section 5.1.2 at this time. 5.1.3 Other Connection Types

There are no modifications to the Guidelines or Commentary of Section 5.1.3 at this time. Post-Earthquake Inspection 5-3

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5.2 Preparation 5.2.1 Preliminary Document Review and Evaluation 5.2.1.1 Document Collection and Review

There are no modifications to the Guidelines or Commentary of Section 5.2.1.1 at this time. 5.2.1.2 Preliminary Building Walk-Through.

There are no modifications to the Guidelines or Commentary of Section 5.2.1.2 at this time. 5.2.1.3 Structural Analysis

There are no modifications to the Guidelines or Commentary of Section 5.2.1.3 at this time. 5.2.1.4 Vertical Plumbness Check

There are no modifications to the Guidelines or Commentary of Section 5.2.1.4 at this time. 5.2.2 Connection Exposure

Pre-inspection activities to expose and prepare a connection for inspection should include the local removal of suspended ceiling panels or (as applicable) local demolition of permanent ceiling finish to access the connection; and cleaning of sufficient fireproofing from the beam and column surfaces to allow visual observation of the area to be inspected. If initial inspections are to be limited to the beam bottom flange to column joint and the surrounding material, fireproofing should be removed from the connection as indicated in Figure 5-1a. Removal of fireproofing need only be sufficient to permit observation of the surfaces of base and weld metals. Wire brushing and cleaning to remove all particles of fireproofing material is not necessary unless ultrasonic testing of the joint area is to be conducted. In the event that damage is found at the bottom beam flange to column joint, then additional fireproofing should be removed, as indicated in Figure 51b, to expose the column panel zone, the column flange, continuity plates, beam web and flanges. The extent of the removal of fireproofing should be sufficient to allow adequate inspection of the surfaces to be inspected. Figure 5-1b suggests a pattern that will allow both visual and NDT inspection of the top and bottom beam flange to column joints, the beam web and shear connection, column panel zone and continuity plates, and column flanges in the areas of highest expected demands. The maximum extent of the removal of fireproofing need not be greater than a distance equal to the beam depth "d" into the beam span to expose evidence of any yielding.

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Exposed surfaces

6”

6” 6” Fireproofing

Figure 5-1a Recommended Zone for Fireproofing Removal for Initial Inspections

6”

12” 6” Fireproofing

Figure 5-1b Recommended Zone for Removal of Fireproofing for Complete Inspections Commentary: If inspection is to be limited to visual observation of the surfaces of the base metal and welds, cleaning of fireproofing need only be sufficient to expose these surfaces. However, if ultrasonic testing is to be performed, the surface over which the scanning will be performed must be free Cleaning of weld areas and removal of mill scale and weld spatter. Such cleaning should be done with care, preferably using a power wire brush, to ensure a clean surface that does not affect the accuracy of ultrasonic testing. The resulting surface finish should be clean, free of mill scale, rust and foreign matter. The use of a chisel should be avoided to preclude scratching the steel surfaces which could be mistaken for yield lines. Sprayed-on fireproofing on WSMFs erected prior to about 19801970 is likely to contain asbestos and should be handled according to Post-Earthquake Inspection 5-5

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applicable standards for the removal of hazardous materials. Health hazards associated with asbestos were recognized by industry in the late 1960s and by 1969, most commercial production of asbestos containing materials had ceased. In April, 1973, the federal government formally prohibited the production of asbestos containing materials with the adoption of the National Emission Standards for Hazardous Air Pollutants. Allowing for shelf life of materials produced prior to that date, it should be considered possible that buildings constructed prior to 1975 contain some asbestos hazards. To preclude physical exposure to hazardous materials and working conditions in such buildings, the structural engineer should require by contractual agreement with the building owner, prior to the start of the inspection program, that the building owner deliver to the structural engineer for his/her review and files a laboratory certificate that confirms the absence of asbestos in structural steel fireproofing, local pipe insulation, ceiling tiles, and drywall joint compound. The pattern of fireproofing removal indicated in Figure 5-1 is adequate to allow visual and UT inspection of the top and bottom girder flange to column joints, the beam web and shear connection and the column panel zone. As discussed in the commentary to Section 5.1.1, some engineers prefer to initially inspect only the bottom beam flange to column joint. In such cases, the initial removal of fireproofing can be more limited than indicated in the figure. If after initial inspection, damage at a connection is suspected, then full removal, as indicated in the figure, should be performed to allow inspection of all areas of the connection. 5.3 Inspection Program 5.3.1 Visual Inspection (VI)

There are no modifications to the Guidelines or Commentary of Section 5.3.1 at this time. 5.3.1.1 Top Flange

There are no modifications to the Guidelines or Commentary of Section 5.3.1.1 at this time. 5.3.1.2 Bottom Flange

There are no modifications to the Guidelines or Commentary of Section 5.3.1.2 at this time. 5.3.1.3 Column and Continuity Plates

There are no modifications to the Guidelines or Commentary of Section 5.3.1.3 at this time.

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5.3.1.4 Beam Web Shear Connection

There are no modifications to the Guidelines or Commentary of Section 5.3.1.4 at this time. 5.3.2 Nondestructive Testing (NDT)

NDT should may be used to supplement the visual inspection of connections selected in accordance with the Interim Guidelines of Chapter 4. The testing agency and NDT personnel performing this work should conform to the qualifications indicated in Chapter 11 of these Interim Guidelines. The following NDT techniques should may be used at the top and bottom of each connection, where accessible, to supplement visual inspection: These techniques should be used whenever visual inspection indicates the potential for damage that is not directly observable. a) Magnetic particle testing (MT) of the beam flange to column flange weld surfaces may be used to confirm the presence of suspected surface cracks based on visual evidence. Where fractures are evident from visual inspection, MT should be used to confirm the lateral extent of the fracture.All surfaces which were visually inspected should be tested using the magnetic particle technique. Commentary: The color of powder should be selected to achieve maximum contrast to the base and weld metal under examination. The test may be further enhanced by applying a white coating made specifically for MT or by applying penetrant developer prior to the MT examination. This background coating should be allowed to thoroughly dry before performing the MT. b) Ultrasonic testing (UT) may be used to detect the presence of hidden fractures, where visual inspection reveals the potential for such fractures. of all faces at the beam flange welds and adjacent column flanges (extending at least 3 inches above and below the location of the CJP weld, along the face of the column, but not less than 1-1/2 times the column flange thickness). Commentary: The purpose of UT is to 1) locate and describe the extent of internal defects not visible on the surface and 2) to determine the extent of cracks observed visually and by MT. These guidelines recommend the use of visual inspection as the primary tool for detecting earthquake damage (See commentary to Sec. 5..1.1). UT can be a useful technique for confirmation of the presence of suspected fractures at the beam flange to column flange joints. Visual evidence that may suggest the need for such testing could include apparent separation of the base of the weld backing from the face of the column. Requirements and acceptance criteria for NDT should be as given in AWS D1.1-98 Sections 6 and 8. Acceptance or rejection of planar weld discontinuity (cracks, slag inclusion, or lack of fusion), including root indications, should, as a minimum, be consistent with AWS Discontinuities Severity Class designations of cracks and defects per Table 8.26.2 of AWS D1.1-98 for Static Post-Earthquake Inspection 5-7

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Structures. Beam flange welds should be tested as "tension welds" per AWS D1.1 Table 8.15.3, Note 3. Backing bars need not be removed prior to performing UT. Commentary: The value of UT for locating small discontinuities at the root of beam flange to column flange welds when the backing is left in place is not universally accepted. The reliability of this technique is particularly questionable at the center of the joint, where the beam web obscures the signal. There have been a number of reported instances of UT detected indications which were not found upon removal of the backing, and similarly, there have been reported instances of defects which were missed by UT examination but were evident upon removal of the backing. The smaller the defect, the less likely it is that UT alone will reliably detect its presence. Despite the potential inaccuracies of this technique, it is the only method currently available, short of removal of the backing, to find subsurface damage in the welds. It is also the most reliable method for finding lamellar problems in the column flange (type C4 and C5 damage) opposite the girder flange. Removal of weld backing at these connections results in a significant cost increase that is probably not warranted unless UT indicates widespread, significant defects and/or damage in the building. The proper scanning techniques, beam angle(s) and transducer sizes should be used as specified in the written UT procedure contained in the Written Practice, prepared in accordance with Section 5.3.3 of these Interim Guidelines. The acceptance standard should be that specified in the original contract documents, but in no case should it be less than the acceptance criteria of AWS D1.1, Chapter 8, for Statically Loaded Structures. The base metal should be scanned with UT for cracks. Cracks which have propagated to the surface of the weld or beam and column base metal will probably have been detected by visual inspection and magnetic particle tests performed earlier. The purpose of ultrasonic testing of the base metal is to: 1. Locate and describe the extent of internal indications not apparent on the surface and, 2. Determine the extent of cracks found visually and by magnetic particle test. Commentary: Liquid dye penetrant testing (PT) may be used where MT is precluded due to geometrical conditions or restricted access. Note that more stringent requirements for surface preparation are required for PT than MT, per AWS D1.1. If practical, NDT should be performed across the full width of the bottom beam flange joint. However, if there are no discontinuity signals from UT of Post-Earthquake Inspection 5-8

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accessible faces on one side of the bottom flange weld, obstructions on the other side of the connection need not be removed for testing of the bottom flange weld. Slabs, flooring and roofing need not be removed to permit NDT of the top flange joint unless there is significant visible damage at the bottom beam flange, adjacent column flange, column web, or shear connection. Unless such damage is present, NDT of the top flange should be performed as permitted, without local removal of the diaphragms or perimeter wall obstructions. It should be noted that UT is not 100% effective in locating discontinuities and defects in CJP beam flange to column flange welds. The ability of UT to reliably detect such defects is very dependent on the skill of the operator and the care taken in the inspection. Even under perfect conditions, it is difficult to obtain reliable readings of conditions at the center of the beam flange to column flange connection as return signals are obscured by the presence of the beam web. If backing is left in place on the welds, UT becomes even less reliable. There have been a number of reported instances in which UT indicated apparent defects, that were found not to exist upon removal of the backing. Similarly, UT has failed in some cases to locate defects that were later discovered upon removal of the backing. Additional information on UT may be found in AWS B1.10. 5.3.3 Inspector Qualification 5.3.4 Post-Earthquake Field Inspection Report

There are no modifications to the Guidelines or Commentary of Section 5.3.4 at this time. 5.3.5 Written Report

There are no modifications to the Guidelines or Commentary of Section 5.3.5 at this time.

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6. POST-EARTHQUAKE REPAIR AND MODIFICATION 6.1 Scope There are no modifications to the Guidelines or Commentary of Section 6.1 at this time. 6.2 Shoring There are no modifications to the Guidelines or Commentary of Section 6.2 at this time. 6.3 Repair Details There are no modifications to the Guidelines or Commentary of Section 6.3 at this time. 6.4 Preparation There are no modifications to the Guidelines or Commentary of Section 6.4 at this time. 6.5 Execution There are no modifications to the Guidelines or Commentary of Section 6.5 at this time. 6.6 STRUCTURAL MODIFICATION 6.6.1 Definition of Modification

There are no modifications to the Guidelines or Commentary of Section 6.6.1 at this time. 6.6.2 Damaged vs. Undamaged Connections

There are no modifications to the Guidelines or Commentary of Section 6.6.2 at this time. 6.6.3 Criteria

Connection modification intended to permit inelastic frame behavior should be proportioned so that the required plastic deformation of the frame may be accommodated through the development of plastic hinges at pre-determined locations within the girder spans, as indicated in Figure 6-12 Figure 6.6.3-1. Beam-column connections should be designed with sufficient strength (through the use of cover plates, haunches, side plates, etc.) to force development of the plastic hinge away from the column face. This condition may also be attained through local weakening of the beam section, at the desired location for plastic hinge formation. All elements of the connection should have adequate strength to develop the forces resulting from the

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formation of the plastic hinge at the predetermined location, together with forces resulting from gravity loads.

h

Undeformed frame

Deformed frame shape

Plastic Hinges

drift angle - θ

L’ L

Figure 6-12 Figure 6.6.3-1 - Desired Plastic Frame Behavior Commentary: Nonlinear deformation of frame structures is typically accommodated through the development of inelastic flexural or shear strains within discrete regions of the structure. At large inelastic strains these regions can develop into plastic hinges, which can accommodate significant concentrated rotations at constant (or nearly constant) load through yielding at tensile fibers and buckling at compressive fibers. If a sufficient number of plastic hinges develop in a frame, a mechanism is formed and the frame can deform laterally in a plastic manner. This behavior is accompanied by significant energy dissipation, particularly if a number of members are involved in the plastic behavior, as well as substantial local damage to the highly strained elements. The formation of hinges in columns, as opposed to beams, is undesirable, as this results in the formation of weak story mechanisms with relatively few elements participating, and consequently little energy dissipation occurring. In addition, such mechanisms also result in local damage to critical gravity load bearing elements. The prescriptive connection contained in the UBC and NEHRP Recommended Provisions prior to the Northridge Earthquake was based on the assumed development of plastic hinge zones within the beams at adjacent to the face of the column, or within the column panel zone itself. If the plastic hinge develops in the column panel zone, the resulting column deformation results in very large secondary stresses on the beam flange to column flange joint, a condition which can contribute to brittle failure. If the plastic hinge forms in the beam, at the face of the column, this can result in very large through-thickness strain demands on Post-earthquake Repair and Modification 6-2

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the column flange material and large inelastic strain demands on the weld metal and surrounding heat affected zones stress and strain demands on the welded beam flange to column flange joint. These conditions can also lead to brittle joint failure. Although ongoing research may reveal conditions of material properties, design and detailing configurations that permit connections with yielding occurring at the column face to perform reliably, for the present, it is recommended In order to achieve more reliable performance, it is recommended that the connection of the beam to the column be modified to be sufficiently strong to force the inelastic action (plastic hinge) away from the column face. Plastic hinges in steel beams have finite length, typically on the order of half the beam depth. Therefore, the location for the plastic hinge should be shifted at least that distance away from the face of the column. When this is done, the flexural demands on the columns are increased. Care must be taken to assure that weak column conditions are not inadvertently created by local strengthening of the connections. It should be noted that connection modifications of the type described above, while believed to be effective in preventing brittle connection fractures, will not prevent structural damage from occurring. Brittle connection fractures are undesirable because they result in a substantial reduction in the lateral-forceresisting strength of the structure which, in extreme cases, can result in instability and collapse. Connections modified as described in these Interim Guidelines should experience many fewer such brittle fractures than unmodified connections. However, the formation of a plastic hinge within the span of a beam is not a completely benign event. Beams which have formed such hinges may exhibit large buckling and yielding deformation, damage which typically must be repaired. The cost of such repairs could be comparable to the costs incurred in repairing fracture damage experienced in the Northridge Earthquake. The primary difference is that life safety protection will be significantly enhanced and most structures that have experienced such plastic deformation damage should continue to be safe for occupancy while repairs are made. If the types of damage described above are unacceptable for a given building, then alternative methods of structural modification should be considered that will reduce the plastic deformation demands on the structure during a strong earthquake. Appropriate methods of achieving such goals include the installation of supplemental braced frames, energy dissipation systems, and similar systematic modifications of the building’s basic lateral force resisting system. It is important to recognize that in frames with relatively short bays, the flexural hinging indicated in Figure 6.6.3-1 may not be able to form. If the effective flexural length (L’in the figure) of beams in a frame becomes too short, then the beams or girders will yield in shear before zones of flexural plasticity

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can form, resulting in an inelastic behavior that is more like that of an eccentrically braced frame than that of a moment frame. This behavior may inadvertently occur in frames in which relatively large strengthened connections, such as haunches, cover plates or side plates have been used on beams with relatively short spans. This behavior is illustrated in Figure 6.6.3-2. The guidelines contained in this section are intended to address the design of flexurally dominated moment resisting frames. When utilizing these guidelines, it is important to confirm that the configuration of the structure is such that the presumed flexural hinging can actually occur. It is possible that shear yielding of frame beams, such as that schematically illustrated in Figure 6.6.3-2 may be a desirable behavior mode. However, to date, there has not been enough research conducted into the behavior of such frames to develop recommended design guidelines. If modifications to an existing frame result in such a configuration designers should consider referring to the code requirements for eccentrically braced frames. Particular care should be taken to brace the shear link of such beams against lateral-torsional buckling and also to adequately stiffen the webs to avoid local buckling following shear plastification. Shear Link

Shear Link

Figure 6.6.3-2 Shear Yielding Dominated Behavior of Short Bay Frames 6.6.4 Strength and Stiffness 6.6.4.1 Strength

When these Interim Guidelines require determination of the strength of a framing element or component, this shall be calculated in accordance with the criteria contained in UBC-94, Section

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2211.4.2 {NEHRP-91 Section 10.2, except that the factor φshould be taken as 1.0}, restated as follows: 2211.4.1 Member strength. Where this section requires that the strength of the member be developed, the following shall be used: Flexure Shear Axial compression Axial tension Connectors Full Penetration welds Partial Penetration welds Bolts and fillet welds

Ms = Z F y Vs = 0.55 Fy d t Psc = 1.7 Fa A Pst = Fy A Fy A 1.7 allowable (see commentary) 1.7 allowable

Alternatively, the criteria contained in the 1997 edition of the AISC Seismic Provisions for Structural Steel Buildings (AISC, 1997) may be used. Commentary: At the time the Interim Guidelines were first published, they were based on the 1994 edition of the Uniform Building Code and the 1994 edition of the NEHRP Provisions. In the time since that initial publication, more recent editions of both documents have been published, and codes based on these documents have been adopted by some jurisdictions. In addition, the American Institute of Steel Construction has adopted a major revision to its Seismic Provisions for Structural Steel Buildings (AISC Seismic Provisions), largely incorporating, with some modification, the recommendations contained in the Interim Guidelines. This updated edition of the AISC Seismic Provisions has been incorporated by reference into the 1997 edition of the NEHRP Provisions and has also been adopted by some jurisdictions as an amendment to the model building codes. Structural upgrades designed to comply with the requirements of the 1997 AISC Seismic Provisions may be deemed to comply with the intent of these Interim Guidelines. Where reference is made herein to the requirements of the 1994 Uniform Building Code or 1994 NERHP Provisions, the parallel provisions of the 1997 editions may be used instead, and should be used in those jurisdictions that have adopted codes based on these updated standards. Partial penetration welds are not recommended for tension applications in critical connections resisting seismic induced stresses. The geometry of partial penetration welds creates a notch-like condition that can initiate brittle fracture under conditions of high tensile strain. Many WSMF structures are constructed with concrete floor slabs that are provided with positive shear attachment between the slab and the top flanges of the girders of the moment-resisting frames. Although not generally accounted for in the design of the frames, the resulting composite action can increase the

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effective strength of the girder significantly, particularly at sections where curvature of the girder places the top flange into compression. Although this effect is directly accounted for in the design of composite systems, it is typically neglected in the design of systems classified as moment resisting steel frames. The increased girder flexural strength caused by this composite action can result in a number of effects including the unintentional creation of weak column strong beam and weak panel zone conditions. In addition, this composite effect has the potential to reduce the effectiveness of reduced section or “dog-bone” type connection assemblies. Unfortunately, very little laboratory testing of large scale connection assemblies with slabs in place has been performed to date and as a result, these effects are not well quantified. In keeping with typical contemporary design practice, the design formulae provided in these Guidelines neglect the strengthening effects of composite action. Designers should, however, be alert to the fact that these composite effects do exist. Similar, and perhaps more severe, effects may also exist where steel beams support masonry or concrete walls. 6.6.4.2 Stiffness

Calculation of frame stiffness for the purpose of determining interstory drift under the influence of the design lateral forces should be based on the properties of the bare steel frame, neglecting the effects of composite action with floor slabs. The stiffening effects of connection reinforcements (e.g.: haunches, side plates, etc.) may be considered in the calculation of overall frame stiffness and drift demands. When reduced beam section connections are utilized, the reduction in overall frame stiffness, due to local reductions in girder cross section, should be considered. Commentary: For design purposes, frame stiffness is typically calculated considering only the behavior of the bare frame, neglecting the stiffening effects of slabs, gravity framing, and architectural elements such as walls. The resulting calculation of building stiffness and period typically underestimates the actual properties, substantially. Although this approach can result in unconservative estimates of design force levels, it typically produces conservative estimates of interstory drift demands. Since the design of most moment-resisting frames are controlled by considerations of drift, this approach is considered preferable to methods that would have the potential to over-estimate building stiffness. Also, many of the elements that provide additional stiffness may be subject to rapid degradation under severe cyclic lateral deformation, so that the bare frame stiffness may provide a reasonable estimate of the effective stiffness under long duration ground shaking response. Notwithstanding the above, designers should be alert to the fact that unintentional stiffness introduced by walls and other non-structural elements can

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significantly alter the behavior of the structure in response to ground shaking. Of particular concern, if these elements are not uniformly distributed throughout the structure, or isolated from its response, they can cause soft stories and torsional irregularities, conditions known to result in poor behavior. 6.6.5 Plastic Rotation Capacity

The plastic rotation capacity of modified connections should reflect realistic estimates of the required level of plastic rotation demand. In the absence of detailed calculations of rotation demand, connections should be shown to be capable of developing a minimum plastic rotation capacity on the order of 0.025 to 0.030 radian. The demand may be lower when braced frames, supplemental damping, base isolation, or other elements are introduced into the moment frame system, to control its lateral deformation; when the design ground motion is relatively low in the range of predominant periods for the structure; and when the frame is sufficiently strong and stiff. As used in these Guidelines, plastic rotation is defined as the plastic chord rotation angle. The plastic chord rotation angle is calculated using the rotated coordinate system shown in Fig. 6.6.51 as the plastic deflection of the beam or girder, at the point of inflection (usually at the center of its span,) ∆CI, divided by the distance between the center of the beam span and the centerline of the panel zone of the beam column connection, LCL. This convention is illustrated in Figure 6.6.51. It is important to note that this definition of plastic rotation is somewhat different than the plastic rotation that would actually occur within a discrete plastic hinge in a frame model similar to that shown in Figure 6.6.3-1. These two quantities are related to each other, however, and if one of them is known, the other may be calculated from Eq. 6.6.5-1. cL

LCL

Plastic hinge

Beam span center line

cL

∆CL

lh

θp =

∆ CL

LCL

Figure 6.6.5-1 Calculation of Plastic Rotation Angle

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θ p = θ ph where: θp θph LCL lh

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( LCL − lh ) LCL

(6.6.5-1)

is the plastic chord angle rotation, as used in these Guidelines is the plastic rotation, at the location of a discrete hinge is the distance from the center of the beam span to the center of the beam-column assembly panel zone is the assumed location of the discrete plastic hinge relative to the center of the beam-column assembly panel zone

If calculations are performed to determine the required connection plastic rotation capacity, the capacity should be taken somewhat greater than the calculated deformation demand, due to the high variability and uncertainty inherent in predictions of inelastic seismic response. Until better guidelines become available, a required plastic rotation capacity on the order of 0.005 radians greater than the demand calculated for the design basis earthquake (or if greater conservatism is desired - the maximum capable considered earthquake) is recommended. Rotation demand calculations should consider the effect of plastic hinge location within the beam span, as indicated in Figure 6-12 Figure 6.6.3-1, on plastic rotation demand. Calculations should be performed to the same level of detail specified for nonlinear dynamic analysis for base isolated structures in UBC-94 Section 1655 {NEHRP-94 Section 2.6.4.4}. Ground motion time histories utilized for these nonlinear analyses should satisfy the scaling requirements of UBC-94 Section 1655.4.2 {NEHRP-94 Section 2.6.4.4} except that instead of the base isolated period, TI, the structure period, T, calculated in accordance with UBC-94 Section 1628 {NEHRP-94 Section 2.3.3.1} should be used. Commentary. When the Interim Guidelines were first published, the plastic rotation was defined as that rotation that would occur at a discrete plastic hinge, similar to the definition of θph. in Eq. 6.6.5-1, above. In subsequent testing of prototype connection assemblies, it was found that it is often very difficult to determine the value of this rotation parameter from test data, since actual plastic hinges do not occur at discrete points in the assembly and because some amount of plasticity also occurs in the panel zone of many assemblies. The plastic chord angle rotation, introduced in Interim Guidelines Advisory No. 1, may more readily be obtained from test data and also more closely relates to the drift experienced by a frame during earthquake response. Traditionally, structural engineers have calculated demand in moment frames by sizing the members for strength and drift using code forces (either equivalent static or reduced dynamic forces) and then "developing the strength of the members." Since 1988, "developing the strength" has been accomplished by prescriptive means. It was assumed that the prescribed connections would be strong enough so that the girder would yield (in bending), or the panel zone

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would yield (in shear) in a nearly perfectly plastic manner producing the plastic rotations necessary to dissipate the energy of the earthquake. It is now known that the prescriptive connection is often incapable of behaving in this manner. In the 1994 Northridge earthquake, many moment-frame connections fractured with little evidence of plastic hinging of the girders or yielding of the column panel zones. Testing of moment frame connections both prior to and subsequent to the earthquake suggests that the standard welded flange-bolted web connection is unable to reliably provide plastic rotations beyond about 0.005 radian for all ranges of girder depths and often fails below that level. Thus, for frames designed for code forces and for the code drift limits, new connection configurations must be developed to reliably accommodate such rotation without brittle fracture. In order to develop reasonable estimates of the plastic rotation demands on a frame’s connections, it is necessary to perform inelastic time history analyses. For regular structures, approximations of the plastic rotation demands can be obtained from linear elastic analyses. Analytical research (Newmark and Hall 1982) suggests that for structures having the dynamic characteristics of most WSMF buildings, and for the ground motions typical of western US earthquakes, the total frame deflections obtained from an unreduced (no R or Rw factor) dynamic analysis provide an approximate estimate of those which would be experienced by the inelastic structure. For the typical spectra contained in the building code, this would indicate expected drift ratios on the order of 1%. The drift demands in a real structure, responding inelastically, tend to concentrate in a few stories, rather than being uniformly distributed throughout the structure’s height. Therefore, it is reasonable to expect typical drift demands in individual stories on the order of 1.5% to 2% of the story height. As a rough approximation, the drift demand may be equated to the joint rotation demand, yielding expected rotation demands on the order of perhaps 2%. Since there is considerable variation in ground motion intensity and spectra, as well as the inelastic response of buildings to these ground motions, conservatism in selection of an appropriate connection rotation demand is warranted. In recent testing of large scale subassemblies incorporating modified connection details, conducted by SAC and others, when the connection design was able to achieve a plastic rotation demand of 0.025 radians or more for several cycles, the ultimate failure of the subassembly generally did not occur in the connection, but rather in the members themselves. Therefore, the stated connection capacity criteria would appear to result in connections capable of providing reliable performance. It should be noted that the connection assembly capacity criteria for the modification of existing buildings, recommended by these Interim Guidelines, is Post-earthquake Repair and Modification 6-9

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somewhat reduced compared to that recommended for new buildings (Chapter 7). This is typical of approaches normally taken for existing structures. For new buildings, these Interim Guidelines discourage building-specific calculation of required plastic rotation capacity for connections and instead, encourage the development of highly ductile connection designs. For existing buildings, such an approach may lead to modification designs that are excessively costly, as well as the modification of structures which do not require such modification. Consequently, an approach which permits the development of semi-ductile connection designs, with sufficient plastic rotation capacity to withstand the expected demands from a design earthquake is adopted. It should be understood that buildings modified to this reduced criteria will not have the same reliability as new buildings, designed in accordance with the recommendations of Chapter 7. The criteria of Chapter 7 could be applied to existing buildings, if superior reliability is desired. When performing inelastic frame analysis, in order to determine the required connection plastic rotation capacity, it is important to accurately account for the locations at which the plastic hinges will occur. Simplified models, which represent the hinge as occurring at the face of the column, maywill underestimate the plastic rotation demand. This problem becomes more severe as the column spacing, L, becomes shorter and the distance between plastic hinges, L’, a greater portion of the total beam span. Eq. 6.6.5-1 may be used to convert calculated values of plastic rotation at a hinge remotely located from the column, to the chord angle rotation, used for the definition of acceptance criteria contained in these Guidelines. In extreme cases, the girder will not form plastic hinges at all, but instead, will develop a shear yield, similar to an eccentric braced frame. 6.6.6

Connection Qualification and Design

Modified girder-column connections may be qualified by testing or designed using calculations. Qualification by testing is the preferred approach. Preliminary designs of connections to be qualified by test may be obtained using the calculation procedures of Section 6.6.6.3. The procedures of that section may also be used to calibrate previous tests of similar connection configurations to slightly different applications, by extrapolation. Extrapolation of test results should be limited to connections of elements having similar geometries and material specifications as the tested connections. Designs based on calculation alone should be subject to qualified independent third party review. Commentary: Because of the cost of testing, use of calculations for interpolation or extrapolation of test results is desirable. How much extrapolation should be accepted is a difficult decision. As additional testing is done, more information may be available on what constitutes "conservative" testing conditions, thereby

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allowing easier decisions relative to extrapolating tests to actual conditions which are likely to be less demanding than the tests. For example, it is hypothesized that connections of shallower, thinner flanged members are likely to be more reliable than similar connections consisting of deeper, thicker flanged members. Thus, it may be possible to test the largest assemblages of similar details and extrapolate to the smaller member sizes? - at least within comparable member group families. However, there is evidence to suggest that extrapolation of test results to assemblies using members of reduced size is not always conservative. In a recent series of tests of cover plated connections, conducted at the University of California at San Diego, a connection assembly that produced acceptable results for one family of beam sizes, W24, did not behave acceptably when the beam depth was reduced significantly to W18. In that project, the change in relative flexibilities of the members and connection elements resulted in a shift in the basic behavior of the assembly and initiation of a failure mode that was not observed in the specimens with larger member sizes. In order to minimize the possibility of such occurrences, when extrapolation of test results is performed, it should be done with a basic understanding of the behavior of the assembly, and the likely effects of changes to the assembly configuration on this behavior. Test results should not be extrapolated to assembly configurations that are expected to behave differently than the tested configuration. Extrapolation or interpolation of results with differences in welding procedures, details or material properties is even more difficult. 6.6.6.1 Qualification Test Protocol

There are no modifications to the Guidelines or Commentary of Section 6.6.6.1 at this time. 6.6.6.2 Acceptance Criteria

The minimum acceptance criteria for connection qualification for specimens tested in accordance with these Interim Guidelines should be as follows: a) The connection should develop beam plastic rotations as indicated in Section 6.6.5, for at least one complete cycle. b) The connection should develop a minimum strength equal to 80% of the plastic strength of the girder, calculated using minimum specified yield strength Fy, throughout the loading history required to achieve the required plastic rotation capacity, as indicated in a), above. c) The connection should exhibit ductile behavior throughout the loading history. A specimen that exhibits a brittle limit state (e.g. complete flange fracture, column cracking, through-thickness failures of the column flange, fractures in welds subject to

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tension, shear tab cracking, etc. ) prior to reaching the required plastic rotation shall be considered unsuccessful. d) Throughout the loading history, until the required plastic rotation is achieved, the connection should be judged capable of supporting dead and live loads required by the building code. In those specimens where axial load is applied during the testing, the specimen should be capable of supporting the applied load throughout the loading history. The evaluation of the test specimen’s performance should consistently reflect the relevant limit states. For example, the maximum reported moment and the moment at the maximum plastic rotation are unlikely to be the same. It would be inappropriate to evaluate the connection using the maximum moment and the maximum plastic rotation in a way that implies that they occurred simultaneously. In a similar fashion, the maximum demand on the connection should be evaluated using the maximum moment, not the moment at the maximum plastic rotation unless the behavior of the connection indicated that this limit state produced a more critical condition in the connection. Commentary: Many connection configurations will be able to withstand plastic rotations on the order of 0.025 radians or more, but will have sustained significant damage and degradation of stiffness and strength in achieving this deformation. The intent of the acceptance criteria presented in this Section is to assure that when connections experience the required plastic rotation demand, they will still have significant remaining ability to participate in the structure’s lateral load resisting system. In evaluating the performance of specimens during testing, it is important to distinguish between brittle behavior and ductile behavior. It is not uncommon for small cracks to develop in specimens after relatively few cycles of inelastic deformation. In some cases these initial cracks will rapidly lead to ultimate failure of the specimen and in other cases they will remain stable, perhaps growing slowly with repeated cycles, and may or may not participate in the ultimate failure mode. The development of minor cracks in a specimen, prior to achievement of the target plastic rotation demand should not be cause for rejection of the design if the cracks remain stable during repeated cycling. Similarly, the occurrence of brittle fracture at inelastic rotations significantly in excess of the target plastic rotation should not be cause for rejection of the design. 6.6.6.3 Calculations

There are no modifications to the Guidelines or Commentary of Section 6.6.6.3 at this time.

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6.6.6.3.1 Material Strength Properties In the absence of project specific material property information (for example, mill test reports), the values listed in Table 6-3 Table 6.6.6.3.1-1 should be used to determine the strength of steel shape and plate for purposes of calculation. The permissible strength for weld metal should be taken in accordance with the building code. Table 6-3Table 6.6.6.3-1 - Properties for Use in Connection Modification Design Material Fy (ksi) Fy m (ksi) Fu (ksi) 1 1 A36 Beam 36 Dual Certified Beam Axial, Flexural 50 65 min. Shape Group 1 552 Shape Group 2 582 Shape Group 3 572 Shape Group 4 542 Through-Thickness Note 3 A572 Column/Beam Axial, Flexural 50 65 min. Shape Group 1 582 Shape Group 2 582 Shape Group 3 572 Shape Group 4 572 Shape Group 5 552 Through-Thickness Note 3 A992 Structural Shape1 Use same values as for A572, Gr. 50 Notes: 1. See Commentary 2. Based on coupons from web. For thick flanges, the Fy flange is approximately 0.95 Fy web. 3. See Commentary

Commentary: Table 6-3, Note 1 - The material properties for steel nominally designated on the construction documents as ASTM A36 can be highly variable and in recent years, steel meeting the specified requirements for both ASTM A36 and A572 has routinely been incorporated in projects calling for A36 steel. Consequently, unless project specific data is available to indicate the actual strength of material incorporated into the project, the properties for ASTM A572 steel should be assumed when ASTM A36 is indicated on the drawings, and the assumption of a higher yield stress results in a more severe design condition. The ASTM A992 specification was specifically developed by the steel industry in response to expressed concerns of the design community with regard to the permissible variation in chemistry and mechanical properties of structural steel under the A36 and A572 specifications. This new specification, which was adopted in late 1998, is very similar to ASTM A572, except that it includes somewhat more restrictive limits on chemistry and on the permissible variation in Post-earthquake Repair and Modification 6-13

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yield and ultimate tensile stress, as well as the ratio of yield to tensile strength. At this time, no statistical data base is available to estimate the actual distribution of properties of material produced to this specification. However, the properties are likely to be very similar, albeit with less statistical scatter, to those of material recently produced under ASTM A572, Grade 50. Table 6-3Table 6.6.6.3-1, Note 3 - In the period immediately following the Northridge earthquake, the Seismology Committee of the Structural Engineers Association of California and the International Conference of Building Officials issued Interim Recommendation No. 2 (SEAOC-1995) to provide guidance on the design of moment resisting steel frame connections. Interim Recommendation No. 2 included a recommendation that the through-thickness stress demand on column flanges be limited to a value of 40 ksi, applied to the projected area of beam flange attachment. This value was selected somewhat arbitrarily, to ensure that through-thickness yielding did not initiate in the column flanges of momentresisting connections and because it was consistent with the successful tests of assemblies with cover plates conducted at the University of Texas at Austin (Engelhardt and Sabol - 1994), rather than being the result of a demonstrated through-thickness capacity of typical column flange material. Despite the somewhat arbitrary nature of the selection of this value, its use often controls the overall design of a connection assembly including the selection of cover plate thickness, haunch depth, and similar parameters. It would seem to be important to prevent the inelastic behavior of connections from being controlled by through-thickness yielding of the column flanges. This is because it would be necessary to develop very large local ductilities in the column flange material in order to accommodate even modest plastic rotation demands on the assembly. However, extensive investigation of the throughthickness behavior of column flanges in a “T” joint configuration reveals that neither yielding, nor through-thickness failure are likely to occur in these connections. Barsom and Korvink (1997) conducted a statistical survey of available data on the tensile strength of rolled shape material in the throughthickness direction. These tests were generally conducted on small diameter coupons, extracted from flange material of heavy shapes. The data indicates that both the yield stress and ultimate tensile strength of this material in the throughthickness direction is comparable to that of the material in the direction parallel to rolling. However, it does indicate somewhat greater scatter, with a number of reported values where the through-thickness strength was higher, as well as lower than that in the longitudinal direction. Review of this data indicates with high confidence that for small diameter coupons, the yield and ultimate tensile values of the material in a through-thickness direction will exceed 90% and 80% respectively of the comparable values in the longitudinal direction. theThe causes for through-thickness failures of column flanges (types C2, C4, and C5), observed

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both in buildings damaged by the Northridge Earthquake and in some test specimens, are not well understood. They are thought to be a function of the metallurgy and “purity” of the steel; conditions of loading including the presence of axial load and rate of loading application; conditions of tri-axial restraint; conditions of local hardening and embrittlement within the weld’s heat affected zone; stress concentrations induced by the presence of backing bars and defects at the root of beam flange to column flange welds; and by the relationship of the connection components as they may affect flange bending stresses and flange curvature induced by panel zone yielding. Given the many complex factors which can affect the through-thickness strength of the column flange, determination of a reliable basis upon which to set permissible design stresses will require significant research. Such research is currently being conducted under the SAC phase II program. While this statistical distribution suggests the likelihood that the throughthickness strength of column flanges could be less than the flexural strength of attached beam elements, testing of more than 40 specimens at Lehigh University indicates that this is not the case. In these tests, high strength plates, representing beam flanges and having a yield strength of 100 ksi were welded to the face of A572, Grade 50 and A913, Grade 50 column shapes, to simulate the portion of a beam-column assembly at the beam flange. These specimens were placed in a universal testing machine and loaded to produce high throughthickness tensile stresses in the column flange material. The tests simulated a wide range of conditions, representing different weld metals as well and also to include eccentrically applied loading. In 40 of 41 specimens tested, the assembly strength was limited by tensile failure of the high strength beam flange plate as opposed to the column flange material. In the one failure that occurred within the column flange material, fracture initiated at the root of a low-toughness weld, at root defects that were intentionally introduced to initiate such a fracture. The behavior illustrated by this test series is consistent with mechanics of materials theory. At the joint of the beam flange to column flange, the material is very highly restrained. As a result of this, both the yield strength and ultimate tensile strength of the material in this region is significantly elevated. Under these conditions, failure is unlikely to occur unless a large flaw is present that can lead to unstable crack propagation and brittle fracture. In light of this evidence, Interim Guidelines Advisory No. 2 deletes any requirement for evaluation of through-thickness flange stress in columns. Interim Recommendation No. 2 (SEAOC-1995) included a value of 40 ksi, applied to the projected area of beam flange attachment, for the throughthickness strength to be used in calculations. This value was selected because it was consistent with the successful tests of cover plated assemblies conducted at

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the University of Texas at Austin (Engelhardt and Sabol - 1994). However, because of the probable influence of all the factors noted above, this value can only be considered to reflect the specific conditions of those tests and specimens. Although reduced stresses at the column face produced acceptable results in the University of Texas tests, the key to that success was more likely the result of forcing the plastic hinge away from the column than reduction of the throughthickness stress by the cover plates. Reduction of through-thickness column flange stress to ever lower levels by the use of thicker cover plates is not recommended, since such cover plates will result in ever higher forces on the face of the column flange as well as larger weldments with potential for enlarged heat affected zones, higher residual stresses and conditions of restraint. Since the initial publication of the Interim Guidelines, a significant number of tests have been performed on reduced beam section connections (See section 7.5.3), most of which employed beam flanges which were welded directly to the column flanges using improved welding techniques, but without reinforcement plates. No through-thickness failures occurred in these tests despite the fact that calculated through-thickness stresses at the root of the beam flange to column flange joint ranged as high as 58 ksi. The successful performance of these welded joints is most probably due to the shifting of the yield area of the assembly away from the column flange and into the beam span. Based on the indications of the above described tests, and noting the undesirability of over reinforcing connections, it is now suggested that a maximum through-thickness stress of 0.9Fyc may be appropriate for use with connections that shift the hinging away from the column face. Notwithstanding this recommendation, engineers are still cautioned to carefully consider the through-thickness issue when these other previously listed conditions which are thought to be involved in this type of failure are prevalent. Notwithstanding all of the above, successful tests using cover plates and other measures of moving hinges (and coincidentally reducing through-thickness stress) continue to be performed. In the interim, structural engineers choosing to utilize connections relying on through-thickness strength should recognize that despite the successful testing, connections relying on through-thickness strength can not be considered to be fully reliable until the influence of the other parameters discussed above can be fully understood. A high amount of structural redundancy is recommended for frames employing connections which rely on through-thickness strength of the column flange. 6.6.6.3.2 Determine Plastic Hinge Location The desired location for the formation of plastic hinges should be determined as a basic parameter for the calculations. For beams with gravity loads representing a small portion of the Post-earthquake Repair and Modification 6-16

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total flexural demand, the location of the plastic hinge may be assumed to occur as indicated in Table 6.6.6.3.2-1 and illustrated in Figure 6.6.6.3.2-1, at a distance equal to 1/3 of the beam depth from the edge of the reinforced connection (or start of the weakened beam section), unless specific test data for the connection indicates that a different value is appropriate. Refer to Figure 6-13. Table 6.6.6.3.2-1 Plastic Hinge Location - Strengthened Connections Connection Type

Reference Section

Hinge Location “sh”

Sect. 7.9.1

d/4 beyond end of cover plates

Haunches

Sect. 7.9.3, 7.9.4

d/3 beyond toe of haunch

Vertical Ribs

Sect. 7.9.2

d/3 beyond toe of ribs

Plastic hinge s h=

Edge of reinforced connection

d/4

Connection reinforcement

sh = d/3

Edge of reinforced connection

Beam depth - d

Cover plates

L’

L

Figure 6-13 Figure 6.6.6.3.2-1 - Location of Plastic Hinge Commentary: The suggested locations for the plastic hinge, at a distance d/3 away from the end of the reinforced section indicated in Table 6.6.6.3.2-1 and Figure 6.6.6.3.2-1 are is based on the observed behavior of test specimens, with no significant gravity load present. If significant gravity load is present, this can shift the locations of the plastic hinges, and in the extreme case, even change the form of the collapse mechanism. If flexural demand on the girder due to gravity load is less than about 30% of the girder plastic capacity, this effect can safely be neglected, and the plastic hinge locations taken as indicated. If gravity demands significantly exceed this level then plastic analysis of the girder should be

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performed to determine the appropriate hinge locations. Note that in zones of high seismicity (UBC Zones 3 and 4, and NEHRP Map Areas 6 and 7) gravity loading on the girders of earthquake resisting frames typically has a very small effect. 6.6.6.3.3 Determine Probable Plastic Moment at Hinges The probable value of the plastic moment, Mpr, at the location of the plastic hinges should be determined from the equation: M pr = 0.95aZ b Fya M pr = 1.1Z b Fya where: α

(6-1) (6.6.6.3.3-1)

is a coefficient that accounts for the effects of strain hardening and modeling uncertainty, taken as: 1.1

when qualification testing is performed or calculations are correlated with previous qualification testing

1.3

when design is based on calculations, alone.

Fya

is the actual yield stress of the material, as identified from mill test reports. Where mill test data for the project is not traceable to specific framing elements, the average of mill test data for the project for the given shape may be used. When mill test data for the project is not available, the value of Fym, from table 6-3Table 6.6.6.3-1 may be used.

Zb

is the plastic modulus of the section

Commentary: The 1.10.95 factor, in equation 6.6.6.3.3-1, is used to adjust account for two effects. First, it is intended to account for the typical difference between the yield stress in the beam web, where coupons for mill certification tests are normally extracted, andto the value in the beam flange. Beam flanges, being comprised of thicker material, typically have somewhat lower yield strengths than do beam web material. Second, it is intended toThe factor of 1.1 recommended to account for strain hardening, or other sources of strength above yield, and agrees fairly well with available test results. It should be noted that the 1.1 factor could underestimate the over-strength where significant flange buckling does not act as the gradual limit on the connection. Nevertheless, the 1.1 factor seems a reasonable expectation of over-strength considering the complexities involved.

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Connection designs that result in excessive strength in the girder connection relative to the column or excessive demands on the column panel zone are not expected to produce superior performance. There is a careful balance that must be maintained between developing connections that provide for an appropriate allowance for girder overstrength and those that arbitrarily increase connection demand in the quest for a “conservative” connection design. The factors suggested above were chosen in an attempt to achieve this balance, and arbitrary increases in these values are not recommended. When the Interim Guidelines were first published, Eq. 6.6.6.3.3-1 included a coefficient, α, intended to account both for the effects of strain hardening and also for modeling uncertainty when connection designs were based on calculations as opposed to a specific program of qualification testing. The intent of this modeling uncertainty factor was twofold: to provide additional conservatism in the design when specific test data for a representative connection was not available, and also as an inducement to encourage projects to undertake connection qualification testing programs. After the Interim Guidelines had been in use for some time, it became apparent that this approach was not an effective inducement for projects to perform qualification testing, and also that the use of an overly large value for the α coefficient often resulted in excessively large connection reinforcing elements (cover plates, e.g.) and other design features that did not appear conducive to good connection behavior. Consequently, it was decided to remove this modeling uncertainty factor from the calculation of the probable strength of an assembly. 6.6.6.3.4 Determine Beam Shear The shear in the beam at the location of the plastic hinge should be determined. A free body diagram of that portion of the beam located between plastic hinges is a useful tool for obtaining the shear at each plastic hinge. Figure 6-14Figure 6.6.3.4-1 provides an example of such a calculation.

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Plastic hinge

P

Note: if 2Mpr /L’is less then the gravity shear in the free body (in this case P/2 + wL’/2), then the plastic hinge location will shift and L’ must be adjusted, accordingly

L’ sh L

P

VA

w Mpr

“A”

Vp

Mpr

L’

taking the sum of moments about “A” = 0 Vp ={Mpr + Mpr + P L’/2 + wL’2/2}/L’

Figure 6-14 Figure 6.6.3.4-1 - Sample Calculation of Shear at Plastic Hinge 6.6.6.3.5 Determine Strength Demands on Connection In order to complete the design of the connection, including sizing the various plates and joining welds which make up the connection, it is necessary to determine the shear and flexural strength demands at each critical section. These demands may be calculated by taking a free body of that portion of the connection assembly located between the critical section and the plastic hinge. Figure 6-15 Figure 6.6.3.5-1 demonstrates this procedure for two critical sections, for the beam shown in Figure 6-14Figure 6.6.3.4-1.

Plastic hinge

Plastic hinge

Mpr

Mf

Vp

dc

x

Vp x+dc/2

Mf=Mpr +Vpx Critical Section at Column Face

Mpr

Mc

Mc=Mpr +Vp(x+dc/2) Critical Section at Column Centerline

Figure 6-15 Figure 6.6.3.5-1 - Calculation of Demands at Critical Sections Post-earthquake Repair and Modification 6-20

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Commentary: Each unique connection configuration may have different critical sections. The vertical plane that passes through the joint between the beam flanges and column (if such joining occurs) will typically define at least one such critical section, used for designing the joint of the beam flanges to the column, as well as evaluating shear demands on the column panel zone. A second critical section occurs at the center line of the column. Moments calculated at this point are used to check weak beam - strong column conditions. Other critical sections should be selected as appropriate to the connection configuration. 6.6.6.3.6 Check for Strong Column - Weak Beam Condition Buildings which form sidesway mechanisms through the formation of plastic hinges in the beams can dissipate more energy than buildings that develop mechanisms consisting primarily of plastic hinges in the columns. Therefore, if an existing building’s original design was such that hinging would occur in the beams rather than the columns, care should be taken not to alter this behavior with the addition of connection reinforcement. To determine if the desired strong column - weak beam condition exists, the connection assembly should be checked to determine if the following equation is satisfied:

∑Z where:

c

(Fyc − f a )

∑M

c

> 1.0

(6.6.6.3.6-12)

Zc is the plastic modulus of the column section above and below the connection Fyc is the minimum specified yield stress for the column above and below fa is the axial load in the column above and below ΣMc is the moment calculated at the center of the column in accordance with Section 6.6.6.3.5 sum of the column moments at the top and bottom of the panel zone, respectively, resulting from the development of the probable beam plastic moments, Mpr, within each beam in the connection. Commentary: Equation 6.6.6.3.6-12 is based on the building code provisions for strong column - weak beam design. The building code provisions for evaluating strong column - weak beam conditions presume that the flexural stiffness of the columns above and below the beam are approximately equal, that the beams will yield at the face of the column, and that the depth of the columns and beams are small relative to their respective span lengths. This permits the code to use a relatively simple equation to evaluate strong column - weak beam conditions in which the sum of the flexural capacities of columns at a connection are compared to the sums of the flexural capacities in the beams. The first publication of the Interim Guidelines took this same approach, except that the definition of ΣMc was modified to explicitly recognize that because flexural hinging of the beams would occur at a location removed from the face of the column, the moments delivered by the beams to the connection would be larger than the plastic moment strength of the beam. In this equation, ΣMc was taken as the sum of the moments at the Post-earthquake Repair and Modification 6-21

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center of the column, calculated in accordance with the procedures of Sect. 6.6.3.5. This simplified approach is not always appropriate. If non-symmetrical connection configurations are used, such as a haunch on only the bottom side of the beam, this can result in an uneven distribution of stiffness between the two column segments, and premature yielding of the column, either above, or below, the beam-column connection. Also, it was determined that for connection configurations in which the panel zone depth represents a significant fraction of the total column height, such as can occur in some haunched and side-plated connections, the definition of ΣMc contained in the initial printing of the Guidelines could lead to excessive conservatism in determining whether or not a strong column - weak beam condition exists in a structure. Consequently, Interim Guidelines Advisory No. 1 adopted the current definition of ΣMc for use in this evaluation. This definition requires that the moments in the column, at the top and bottom of the panel zone be determined for the condition when a plastic hinge has formed at all beams in the connection. Figure 6.6.6.3.6-1 illustrates a method for determining this quantity. In such cases, When evaluation indicates that a strong column - weak beam condition does not exist, a plastic analysis should be considered to determine if an undesirable story mechanism is likely to form in the building. assumed point of zero moment

ht

Vc

Vp

Vc =

∑ [M

M ct = Vc ht

Mct

dp

Mpr Vf

(

pr

]

(

+ V p ( L − L ' ) / 2) − V f hb + d p / 2

)

hb + d p + ht

)

M cb = Vc + V f hb

∑

M c = M ct + M cb

Mpr Mcb

hb

Vp

Vc+Vf (L-L’)/2

Note: The quantities Mpr, Vp, L, and L’are as previously identified. Vf is the incremental shear distributed to the column at the floor level. Other quantities are as shown.

Figure 6.6.6.3.6-1 Calculation of Column Moment for Strong Column Evaluation

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6.6.6.3.7 Check Column Panel Zone The adequacy of the shear strength of the column panel zone should be checked. For this purpose, the term 0.8ΣMf should be substituted for the term 0.8ΣMs in UBC-94 Section 2211.7.2.1 {0.9ΣφbMp in NEHRP-91 Section 10.10.3.1} repeated below for convenience of reference. Mf is the calculated moment at the face of the column, when the beam mechanism forms, calculated as indicated in Section 6.6.6.3.5, above. In addition, it is recommended not to use the alternative design criteria indicated in UBC-94 Section 2211.7.2.1 (NEHRP-91 Sect. 10.10.3.1), permitting panel zone shear strength to be proportioned for the shear induced by bending moments from gravity loads plus 1.85 times the prescribed seismic forces. For convenience of reference, UBC-94 Section 2211.7.2.1 is reproduced below, edited, to indicate the recommended application: 2211.7.2.1 Strength (edited). The panel zone of the joint shall be capable of resisting the shear induced by beam bending moments due to gravity loads plus 1.85 times the prescribed seismic forces, but the shear strength need not exceed that required to develop 0.8ΣMs 0.8ΣMf of the girders framing into the column flanges at the joint. The joint panel zone shear strength may be obtained from the following formula:

 3b c t c f 2  V = 0.55Fy d c t 1 +  dbdct  

(11-1)

where: bc = width of column flange db = the depth of the beam (including haunches or cover plates) dc = the depth of the column t = the total thickness of the panel zone including doubler plates tcf = the thickness of the column flange

Commentary: The effect of panel zone shear yielding on connection behavior is not well understood. In the past, panel zone shear yielding has been viewed as a benign mechanism that permits overall frame ductility demands to be accommodated while minimizing the extent of inelastic behavior required of the beam and beam flange to column flange joint. The criteria permitting panel zone shear strength to be proportioned for the shears resulting from moments due to gravity loads plus 1.85 times the design seismic forces was adopted by the code specifically to encourage designs with weak panel zones. However, during recent testing of large scale connection assemblies with weak panel zones, it has been noted that in order to accommodate the large shear deformations that occur in the panel zone, extreme “kinking” deformations were induced into the column flanges at the beam flange to column flange welded joint. While this did not lead to premature joint failure in all cases, it is believed to have contributed to such premature failures in at least some of the specimens. The recommendations of this section are intended to result in stronger panel zones than previously Post-earthquake Repair and Modification 6-23

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permitted by the code, thereby avoiding potential failures due to this kinking action on the column flanges. 6.6.7

Modification Details

There are no modifications to the Guidelines or Commentary of Section 6.6.7 at this time. 6.6.7.1 Haunch at Bottom Flange

Figure 6-166.6.7.1-1 illustrates the basic configuration for a connection modification consisting of the addition of a welded haunch at the bottom beam flange. Several tests of such a modification were conducted by Uang under the SAC phase I project (Uang, 1995). Following that work, additional research on the feasibility of improving connection performance with welded haunches was conducted under a project that was jointly sponsored by NIST and AISC (NIST, 1998). As indicated in the report of that work, the haunched modification improves connection performance by altering the basic behavior of the connection. In essence, the haunch creates a prop type support, beneath the beam bottom flange. This both reduces the effective flexural stresses in the beam at the face of the support, and also greatly reduces the shear that must be transmitted to the column through the beam. Laboratory tests indicate this modification can be effective when the existing low-toughness welds between the beam bottom flange and column are left in place, however, more reliable performance is obtained when the top welds are modified. A complete procedure for the design of this modification may be found in NIST, 1998. two alternative configurations of this detail that have been tested (Uang - 1995). The basic concept is to reinforce the connection with the provision of a triangular haunch at the bottom flange. The intended behavior of both configurations is to shift the plastic hinge from the face of the column and to reduce the demand on the CJP weld by increasing the effective depth of the section. In one test, shown on the left of Figure 6-16, the joint between the girder bottom flange and column was cut free, to simulate a condition which might occur if the bottom joint had been damaged, but not repaired. In a second tested configuration, the bottom flange joint was repaired and the top flange was replaced with a locally thickened plate, similar to the detail shown in Figure 6-9. Design Issues: This approach developed acceptable levels of plastic rotation. Acceptable levels of connection strength were also maintained during large inelastic deformations of the plastic hinge. This approach does not require that the top flange be modified, or slab disturbed, unless other conditions require repair of the top flange, as in the detail on the left of Figure 6-16. The bottom flange is generally far more accessible than the top flange because a slab does not have to be removed. In addition, the haunch can be installed at perimeter frames without removal of the exterior building cladding. There did not appear to be any appreciable degradation in performance when the bottom beam flange was not re-welded to the face of the column. Eliminating this additional welding should help reduce the cost of the repair. Performance is dependent on properly executed complete joint penetration welds at the column face and at the attachment of the haunch to the girder bottom flange. The joint can be subject to through-thickness flaws in the column flange; however, this connection may not be as sensitive Post-earthquake Repair and Modification 6-24

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to this potential problem because of the significant increase in the effective depth of the beam section which can be achieved. Welding of the bottom haunch requires overhead welding. The skewed groove welds of the haunch flanges to the girder and column flanges may be difficult to execute. Experimental Results: This approach developed excellent levels of plastic rotation. In Specimen 1, the bottom flange CJP weld was damaged in a prior test but was not repaired: only the bottom haunch was added. During the test of specimen 1, a slowly growing crack developed at the underside of the top flange-web intersection, perhaps exacerbated by significant local buckling of the top flange. Some of the buckling may be attributed to lateral torsional buckling that occurred because the bottom flange was not restrained by a CJP weld. A significant portion of the flexural strength was lost during the cycles of large plastic rotation. In the second specimen, the bottom girder flange weld was intact during the haunch testing, and its performance was significantly improved compared with the first specimen. The test was stopped when significant local buckling led to a slowly growing crack at the beam flange and web intersection. At this time, it appears that repairing damaged bottom flange welds in this configuration can produce better performance. Acceptable levels of flexural strength were maintained during large inelastic deformations of the plastic hinge for both specimens. As reported in NIST, 1998, a total of 9 beam-column connection tests incorporating bottom haunch modifications of preNorthridge connections have been tested in the laboratory, including two dynamic tests. Most of the connection assemblies tested resisted in excess of 0.02 radians of imposed plastic rotation. However, for those specimens in which the existing low-toughness weld was left in place at the beam top flange, without modification, connection behavior was generally limited by fractures generating at these welds at relatively low plastic rotations. It may be expected that enhanced performance can be obtained by replacing or reinforcing these welds as part of the modification.

Figure 6-166.6.7.1-1 - Bottom Haunch Connection Modification

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Quantitative Results: No. of specimens tested: 29 Girder Size: W30 x 99 Column Size: W14 x 176 Plastic Rotation achievedSpecimen 1 UCSD-1R: 0.04 radian (w/o bottom flange weld) Specimen 2 UCSD-3R:0.05 radian (with bottom flange weld) Specimen UCSD-4R: 0.014 radian (dynamic- limited by test setup) Speciemn UCSD-5R: 0.015 radian (dynamic- limited by test setup) Girder Size: W36x150 Column Size: W14x257 Plastic Rotation achieved Specimen UCB-RN2: 0.014 radian (no modification of top weld) Specimen UTA-1R: 0.019 radian (partial modification of top weld) Specimen UTA-1RB: 0.028 radian (modified top weld) Girder Size: W36x150 Column Size: W14x455 Plastic Rotation achievedSpecment UTA-NSF4: 0.015 radian (no modification of top weld) Girder Size: W18x86 Column Size: W24x279 Plastic Rotation achievedSpecimen SFCCC-8: 0.035 radian (cover plated top flange) 6.6.7.2 Top and Bottom Haunch

There are no modifications to the Guidelines or Commentary of Section 6.6.7.2 at this time. 6.6.7.3 Cover Plate Sections

Figure 6.6.7.3-1 Figure 6-18 illustrates the basic configurations of cover plate connections. The assumption behind the cover plate is that it reduces the applied stress demand on the weld at the column flange and shifts the plastic hinge away from the column face. Only the connection with cover plates on the top of the top flange has been tested. There are no quantitative results for cover plates on the bottom side of the top flange, such as might be used in repair. It is likely that thicker plates would be required where the plates are installed on the underside of the top flange. The implications of this deviation from the tested configuration should be considered.

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d

Near and Far Sides

d/2, typical Top &Bottom

Top &Bottom

Figure 6-18 Figure 6.6.7.3-1 - Cover Plate Connection Modification Design Issues: Following the Northridge earthquake, the University of Texas at Austin conducted a program of research, under private funding, to develop a modified connection configuration for a specific project. Following a series of unsuccessful tests on various types of connections,Approximately eight connections similar to that shown in Figure 6-18Figure 6.7.3-1 have been were tested (Engelhardt & Sabol - 1994), and have demonstrated the ability to achieve acceptable levels of plastic rotation provided that the beam flange to column flange welding wasis correctly executed and through-thickness problems in the column flange wereare avoided. Due to the significant publicity that followed these successful tests, as well as the economy of these connections relative to some other alternatives, cover plated connections quickly became the predominant configuration used in the design of new buildings. As a result, a number of qualification tests have now been performed on different variations of cover plated connections, covering a wide range of member sizes ranging from W16 to W36 beams, as part of the design process for individual building projects. The results of these tests have been somewhat mixed, with a significant number of failures reported. Although this connection type appears to be significantly more reliable than the typical pre-Northridge connection, it should be expected that some connections in buildings incorporating this detail may still be subjected to earthquake initiated fracture damage. Designers should consider using alternative connection types, unless highly redundant framing systems are employed. The option with the top flange cover plate located on top of the flange can be used on perimeter frames where access to the outer side of the beam is restricted by existing building cladding. The option with the cover plate for the top flange located beneath the flange can be installed without requiring modification of the slab. In the figures shown, the bottom cover plate is rectangular, and sized slightly wider than the beam flange to allow downhand fillet welding of the joint between the two plates. Some configurations using triangular plates at the bottom flange, similar to the top flange have also been tested.

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Designers using this detail are cautioned to be mindful of not making cover plates so thick that excessively large welds of the beam flange combination to column flange result. As the cover plates increase in size, the weld size must also increase. Larger welds invariably result in greater shrinkage stresses and increased potential for cracking prior to actual loading. In addition, larger welds will lead to larger heat affected zones in the column flange, a potentially brittle area. Performance is dependent on properly executed girder flange welds. The joint can be subject to through-thickness failures in the column flange. Access to the top of the top flange requires demolition of the existing slab. Access to the bottom of the top flange requires overhead welding and may be problematic for perimeter frames. Costs are greater than those associated with approaches that concentrate modifications on the bottom flange Experimental Results: Six of eight connections tested by the University of Texas at Austin were able to achieve plastic rotations of at least 0.025 radians, or better. These tests were performed using heavy column sections which forced nearly all of the plastic deformation into the beam plastic hinge; very little column panel zone deformation occurred. Strength loss at the extreme levels of plastic rotation did not reduce the flexural capacity to less than the plastic moment capacity of the section based on minimum specified yield strength. One specimen achieved plastic rotations of 0.015 radians when a brittle fracture of the CJP weld (type W2 failure) occurred. This may partially be the result of a weld that was not executed in conformance with the specified welding procedure specification. The second unsuccessful test specimen achieved plastic rotations of 0.005 radian when a section of the column flange pulled out (type C2 failure). The successful tests were terminated either when twisting of the specimen threatened to damage the test setup or the maximum stroke of the loading ram was achieved. Since the completion of that testing, a number of additional tests have been performed. Data for 18 tests, including those performed by Engelhardt and referenced above, are in the public domain. At least 12 other tests have been performed on behalf of private parties, however, the data from these tests are not available. Some of those tests exhibited premature fractures. Quantitative Results: No. of specimens tested: 18 Girder Size: W21 x 68 to W36 x 150 Column Size: W12 x 106 to W14 x 455, and 426 Plastic Rotation achieved6 13 Specimens : >.025 radian to 0.05 radian 13 Specimens: 0.005 < θp < 0.0250.015 radian (W2 failure) 12 Specimens: 0.005 radian (C2 failure) 6.6.7.4 Upstanding Ribs

There are no modifications to the Guidelines or Commentary of Section 6.6.7.4 at this time.

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6.6.7.5 Side-Plate Connections

There are no modifications to the Guidelines or Commentary of Section 6.6.7.5 at this time. 6.6.7.6 Bolted Brackets

Heavy bolted brackets, incorporating high strength bolts, may be added to existing welded connections to provide an alternative load path for transfer of stress between the beams and columns. To be compatible with existing welded connections, the brackets must have sufficient strength and rigidity to transfer beam stresses with negligible deformation. Pre-tensioning of the bolts or threaded rods attaching the brackets to the column flanges and use of welds or slipcritical connections between the brackets and beam flanges can help to minimize deformation under load. Reinforcement of the column flanges may be required to prevent local yielding and excessive deformation of these elements. Two alternative configurations, which may be used either to repair an existing damaged, welded connection or to reinforce an existing undamaged connection are illustrated in Figure 6.6.7.6-1. The developer of these connections offers the brackets in the form of proprietary steel castings. Several tests of these alternative connections have been performed on specimens with beams ranging in size from W16 to W36 sections and with large plastic rotations successfully achieved. Under a project jointly funded by NIST and AISC, the use of a single bracket at the bottom flange of the beam was investigated. It was determined that significant improvement in connection behavior could be obtained by placing a bracket at the bottom beam flange and by replacing existing low-toughness welds at the top flange with tougher material. NIST, 1998 provides a recommended design procedure for such connection modifications. Design Issues: The concept of bolted bracket connections is similar to that of the riveted “wind connections” commonly installed in steel frame buildings in the early twentieth century. The primary difference is that the riveted wind connections were typically limited in strength either by flexural yielding of outstanding flanges of the brackets, or shear and tension on the rivets, rather than by flexural hinging of the connected framing. Since the old-style wind connections could not typically develop the flexural strength of the girders and also could be quite flexible, they would be classified either as partial strength or partially restrained connections. Following the Northridge earthquake, the concept of designing such connections with high strength bolts and heavy plates, to behave as fully restrained connections, was developed and tested by a private party who has applied for patents on the concept of using steel castings for this purpose. Bolted bracket connections can be installed in an existing building without field welding. Since this reduces the risk of construction-induced fire, brackets may be installed with somewhat less demolition of existing architectural features than with welded connections. In addition, the quality assurance issues related to field welding are eliminated. However, the fabrication of the brackets themselves does require quality assurance. Quality assurance is also required for operations related to the drilling of bolt holes for installation of bolts, surface preparation of faying surfaces and for installation and tensioning of the bolts themselves.

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Steel casting

SAC 99-01 High tensile threaded rod

Pipe Plate Bolts

Bolts

WARNING: The information presented in this figure is PROPRIETARY. US and Foreign Patents have been applied for. Use of this information is strictly prohibited except as authorized in writing by the developer. Violators shall be prosecuted in accordance with US and Foreign Patent Intellectual Property Laws.

Figure 6.6.7.6-1 Bolted Bracket Modification Bolted brackets can be used to repair damaged connections. If damage is limited to the beam flange to column flange welds, the damaged welds should be dressed out by grinding. Any existing fractures in base metal should be repaired as indicated in Section 6.3, in order to restore the strength of the damaged members and also to prevent growth of the fractures under applied stress. Since repairs to base metal typically require cutting and welding, this reduces somewhat the advantages cited above, with regard to avoidance of field welding. Experimental Results: A series of tests on several different configurations of proprietary heavy bolted bracket connections have been performed at Lehigh University (Ksai & Bleiman, 1996) to qualify these connections for use in repair and modification applications. To test repair applications, brackets were placed only on the bottom beam flange to simulate installations on a connection where the bottom flange weld in the original connection had failed. In these specimens, bottom flange welds were not installed, to approximate the condition of a fully fractured weld. The top flange welds of these specimens were made with electrodes rated for notch toughness, to preclude premature failure of the specimens at the top flange. For specimens in which brackets were placed at both the top and bottom beam flanges, both welds were omitted. Acceptable plastic rotations were achieved for each of the specimens tested. No testing has yet been performed to determine the effectiveness of bolted brackets when applied to an existing undamaged connection with full penetration beam flange to column flange welds with low toughness or significant defects or discontinuities. Quantitative Results: No. of specimens tested: 8 Girder Size: W16x40 and W36x150 Column Size: W12x65 and W14x425 Plastic Rotation achieved - 0.05 radians - 0.07 radians

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7. NEW CONSTRUCTION 7.1 Scope This Chapter presents interim design guidelines for new welded steel moment frames (WSMFs) intended to resist seismic demands through inelastic behavior. The criteria apply to all SMRF structures designed for earthquake resistance and those IMRF and OMRF structures located in Uniform Building Code (UBC) Seismic Zones 3 and 4 {National Earthquake Hazards Reduction Program (NEHRP) Map Areas 6 and 7} or assigned to 1997 NEHRP Seismic Design Categories D, E, or F. Light, single-story buildings, the design of which is governed by wind, need not consider these Interim Guidelines. Frames with bolted connections, either fully restrained (type FR) or partially restrained (type PR), are beyond the scope of this document. However, the acceptance criteria for connections may be applied to type FR bolted connections as well. Commentary: Observation of damage experienced by WSMF buildings in the Northridge Earthquake and subsequent laboratory testing of large scale beamcolumn assemblies has demonstrated that the standard details for WSMF connections commonly used in the past are not capable of providing reliable service in the post-elastic range. Therefore, structures which are expected to experience significant post-elastic demands from design earthquakes, or for which highly reliable seismic performance is desired, should be designed using the Interim Guidelines presented herein. In order to determine if a structure will experience significant inelastic behavior in a design earthquake, it is necessary to perform strength checks of the frame components for the combination of dead and live loads expected to be present, together with the full earthquake load. Except for structures with special performance goals, or structures located within the near field (within 10 kilometers) of known active earthquake faults, the full earthquake load may be taken as the minimum design earthquake load specified in the building code, but calculated using a lateral force reduction coefficient (Rw or R) of unity. If all components of the structure and its connections have adequate strength to resist these loads, or nearly so, then the structure may be considered to be able to resist the design earthquake, elastically. Design of frames to remain elastic under unreduced (Rw {R} taken as unity) earthquake forces may not be an overly oppressive requirement, particularly in more moderate seismic zones. Most frame designs are currently controlled by drift considerations and have substantially more strength than the minimum specified for design by the building code. As part of the SAC Phase 1 research, a number of modern frame buildings designed with large lateral force reduction coefficients (Rw = 12, {R = 8}) were evaluated for unreduced forces calculated New Construction 7-1

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using the standard building code spectra. It was determined that despite the nominally large lateral force reduction coefficients used in the original design, the maximum computed demands from the dynamic analyses were only on the order of 2 to 3 times those which would cause yielding of the real structures (Krawinkler, et. al. - 1995; Uang, et. al. - 1995; Engelhardt, et. al. - 1995, Hart, et. al. - 1995; Kariotis and Eimani - 1995). Therefore, it is not unreasonable to expect that OMRF structures (nominally designed with a lateral force reduction coefficient Rw = 6 {R = 4.5}) could resist the design earthquakes with near elastic behavior. Regardless of these considerations, better seismic performance can be expected by designing structures with greater ductility rather than less, and engineers are not encouraged to design structures for elastic behavior using brittle or unreliable details.. For structures designed to meet special performance goals, and buildings located within the near field of major active faults, full earthquake loads calculated in accordance with the above procedure may not be adequate. For such structures, the full earthquake load should be determined using a site specific ground motion characterization and a suitable analysis procedure. Recent research (Heaton, et. al. - 1995) suggests that the elastic response spectrum technique, typically used for determining seismic forces for structural design, may not provide an adequate indication of the true earthquake demands produced by the large impulsive ground motions common in the near field of large earthquake events. Further, this research indicates that frame structures, subjected to such impulsive ground motions can experience very large drifts, and potential collapse. In an attempt to address this, both the 1997 edition of the Uniform Building Code and the 1997 edition of the NEHRP Provisions specify design ground motions for structures located close to major active faults that are substantially more severe than those contained in earlier codes. While the more severe ground motion criteria contained in these newer provisions are probably adequate for the design of most structures, analytical studies conducted by SAC confirm that even structures designed to these criteria can experience very large drift demands, and potentially collapse, if the dynamic characteristics of the impulsive loading and those of the structure are matched. Direct nonlinear time history analysis, using an appropriate ground motion representation would be one method of more accurately determining the demands on structures located in the near field. Additional research on these effects is required. As an alternative to use of the criteria contained in these Interim Guidelines, OMRF structures in zones of high seismicity (UBC seismic zones 3 and 4 and NEHRP map areas 6 and 7) and OMRF structures assigned to 1997 NEHRP Seismic Design Categories D, E or F, may be designed for the connections to remain elastic (Rw or R taken as 1.0) while the beams and columns are designed using the standard lateral force reduction coefficients specified by the building

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code. Although this is an acceptable approach, it may result in much larger connections than would be obtained by following these Interim Guidelines. The use of partially restrained connections may be an attractive and economical alternative to the design of frames with fully restrained connections. However, the design of frames with partially restrained connections is beyond the scope of this document. The AISC is currently working on development of practical design guidelines for frames with partially restrained connections. 7.2 General - Welded Steel Frame Design Criteria 7.2.1 Criteria

Welded Steel Moment Frame (WSMF) systems should, as a minimum, be designed for the provisions of the prevailing building code and these Interim Guidelines. Special MomentResisting Frames (SMRF)s and Ordinary Moment-Resisting Frames (OMRF)s with FR connections, should additionally be designed in accordance with either the 1997 edition of the AISC Seismic Provisions for Structural Steel Buildings (AISC, 1997) or the emergency code change to the 1994 UBC {NEHRP-1994}, restated as follows: 2211.7.1.1. Required Strength {NEHRP-1994 Section 5.2, revision to Ref. 8.2c of Ref. 5.3} The girder-to-column connections shall be adequate to develop the lesser of the following: 1.

The strength of the girder in flexure.

2.

The moment corresponding to development of the panel zone shear strength as determined by Formula (11-1).

2211.7.1.3-2 Connection Strength Connection configurations utilizing welds and high strength bolts shall demonstrate, by approved cyclic test results or calculation, the ability to sustain inelastic rotations and to develop the strength criteria in Section 2211.7.1.1 considering the effects of steel overstrength and strain hardening.

Commentary: At the time the Interim Guidelines were first published, they were based on the 1994 edition of the Uniform Building Code and the 1994 edition of the NEHRP Provisions. In the time since that initial publication, more recent editions of both documents have been published, and codes based on these documents have been adopted by some jurisdictions. In addition, the American Institute of Steel Construction has adopted a major revision to its Seismic Provisions for Structural Steel Buildings (AISC Seismic Provisions), largely incorporating, with some modification, the recommendations contained in the Interim Guidelines. This updated edition of the AISC Seismic Provisions has been incorporated by reference into the 1997 edition of the NEHRP Provisions and has also been adopted by some jurisdictions as an amendment to the model building codes. SMRF and OMRF systems that are designed to comply with the requirements of the 1997 AISC Seismic Provisions may be deemed to comply with the intent of these Interim Guidelines. Where reference is made herein to the New Construction 7-3

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requirements of the 1994 Uniform Building Code or 1994 NERHP Provisions, the parallel provisions of the 1997 editions may be used instead, and should be used in those jurisdictions that have adopted codes based on these updated standards. The 1997 edition NEHRP Provisions and AISC Seismic Provisions introduce a new structural system termed an Intermediate Moment Resisting Frame (IMRF). Provisions for IMRF structures include somewhat more restrictive detailing and design requirements than those for OMRF structures, and less than those for SMRF structures. The intent was to provide a system that would be more economical than SMRF structures yet have better inelastic response capability than OMRF structures. The SAC project is currently conducting research to determine if the provisions for the new IMRF system are adequate, but has not developed a position on this at this time. At this time, no recommendations are made to change the minimum lateral forces, drift limitations or strength calculations which determine member sizing and overall performance of moment frame systems, except as recommended in Sections 7.2.2, 7.2.3 and 7.2.4. The design of joints and connections is discussed in Section 7.3. The UBC permits OMRF structures with FR connections, designed for 3/8Rw times the earthquake forces otherwise required, to be designed without conforming to Section 2211.7.1. However, this is not recommended. 7.2.2 Strength and Stiffness 7.2.2.1 Strength

When these Interim Guidelines require determination of the strength of a framing element or component, this shall be calculated in accordance with the criteria contained in UBC-94, Section 2211.4.2 {NEHRP-91 Section 10.2, except that the factor φshould be taken as 1.0}, restated as follows: 2211.4.1 Member strength. Where this section requires that the strength of the member be developed, the following shall be used: Flexure Shear Axial compression Axial tension Connectors Full Penetration welds Partial Penetration welds Bolts and fillet welds

Ms = Z F y Vs = 0.55 Fy d t Psc = 1.7 Fa A Pst = Fy A Fy A 1.7 allowable (see commentary) 1.7 allowable

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Commentary: Partial penetration welds are not recommended for tension applications in critical connections resisting seismic-induced stresses. The geometry of partial penetration welds creates a notch-like condition that can initiate brittle fracture under conditions of high tensile strain. Many WSMF structures are constructed with concrete floor slabs that are provided with positive shear attachment between the slab and the top flanges of the girders of the moment-resisting frames. Although not generally accounted for in the design of the frames, the resulting composite action can increase the effective strength of the girder significantly, particularly at sections where curvature of the girder places the top flange into compression. Although this effect is directly accounted for in the design of composite systems, it is typically neglected in the design of systems classified as moment resisting steel frames. The increased girder flexural strength caused by this composite action can result in a number of effects including the unintentional creation of weak column strong beam and weak panel zone conditions. In addition, this composite effect has the potential to reduce the effectiveness of reduced section or “dog-bone” type connection assemblies. Unfortunately, very little laboratory testing of large scale connection assemblies with slabs in place has been performed to date and as a result, these effects are not well quantified. In keeping with typical contemporary design practice, the design formulae provided in these Guidelines neglect the strengthening effects of composite action. Designers should, however, be alert to the fact that these composite effects do exist. 7.2.2.2 Stiffness

Calculation of frame stiffness for the purpose of determining interstory drift under the influence of the design lateral forces should be based on the properties of the bare steel frame, neglecting the effects of composite action with floor slabs. The stiffening effects of connection reinforcements (e.g.: haunches, side plates, etc.) may be considered in the calculation of overall frame stiffness and drift demands. When reduced beam section connections are utilized, the reduction in overall frame stiffness, due to local reductions in girder cross section, should be considered. Commentary: For design purposes, frame stiffness is typically calculated considering only the behavior of the bare frame, neglecting the stiffening effects of slabs, gravity framing, and architectural elements. The resulting calculation of building stiffness and period typically underestimates the actual properties, substantially. Although this approach can result in unconservative estimates of design force levels, it typically produces conservative estimates of interstory drift demands. Since the design of most moment-resisting frames are controlled by considerations of drift, this approach is considered preferable to methods that would have the potential to over-estimate building stiffness. Also, many of the

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elements that provide additional stiffness may be subject to rapid degradation under severe cyclic lateral deformation, so that the bare frame stiffness may provide a reasonable estimate of the effective stiffness under long duration ground shaking response. Notwithstanding the above, designers should be alert to the fact that unintentional stiffness introduced by walls and other non-structural elements can significantly alter the behavior of the structure in response to ground shaking. Of particular concern, if these elements are not uniformly distributed throughout the structure, or isolated from its response, they can cause soft stories and torsional irregularities, conditions known to result in poor behavior. 7.2.3 Configuration

Frames should be proportioned so that the required plastic deformation of the frame can may be accommodated through the development of plastic hinges at pre-determined locations within the girder spans, as indicated in Figure 7-1Figure 7.2.3-1. Beam-column connections should be designed with sufficient strength (through the use of cover plates, haunches, side plates, etc.) to force development of the plastic hinge away from the column face. This condition may also be attained through local weakening of the beam section at the desired location for plastic hinge formation.

h

Undeformed frame

Deformed frame shape

Plastic Hinges

drift angle - θ

L’ L

Figure 7-1 Figure 7.2.3-1 - Desired Plastic Frame Behavior Commentary: Nonlinear deformation of frame structures is typically accommodated through the development of inelastic flexural or shear strains within discrete regions of the structure. At large inelastic strains these regions can develop into plastic hinges, which can accommodate significant concentrated rotations at constant (or nearly constant) load through yielding at tensile and compressive fibers and by buckling at compressive fibers. If a sufficient number of plastic hinges develop in a frame, a mechanism is formed and the frame can New Construction 7-6

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deform laterally in a plastic manner. This behavior is accompanied by significant energy dissipation, particularly if a number of members are involved in the plastic behavior, as well as substantial local damage to the highly strained elements. The formation of hinges in columns, as opposed to beams, is undesirable, as this results in the formation of weak story mechanisms with relatively few elements participating, so called “story mechanisms” and consequently little energy dissipation occurring. In addition, such mechanisms also result in local damage to critical gravity load bearing elements. The prescriptive connection contained in the UBC and NEHRP Recommended Provisions prior to the Northridge Earthquake was based on the assumed development of plastic hinge zones within the beams at adjacent to the face of the column, or within the column panel zone itself. If the plastic hinge develops in the column panel zone, the resulting column deformation results in very large secondary stresses on the beam flange to column flange joint, a condition which can contribute to brittle failure. If the plastic hinge forms in the beam, at the face of the column, this can result in very large through-thickness strain demands on the column flange material and large inelastic strain demands on the weld metal and surrounding heat affected zones. These conditions can also lead to brittle joint failure. Although ongoing research may reveal conditions of material properties, design and detailing configurations that permit connections with yielding occurring at the column face to perform reliably, for the present it is recommended In order to achieve more reliable performance, it is recommended that the connection of the beam to the column be configured to force the inelastic action (plastic hinge) away from the column face. This can be done either by local reinforcement of the connection, or locally reducing the cross section of the beam at a distance away from the connection. Plastic hinges in steel beams have finite length, typically on the order of half the beam depth. Therefore, the location for the plastic hinge should be shifted at least that distance away from the face of the column. When this is done through reinforcement of the connection, the flexural demands on the columns, for a given beam size, are increased. Care must be taken to assure that weak column conditions are not inadvertently created by local strengthening of the connections. It should be noted that some professionals and researchers believe that configurations which permit plastic hinging to occur adjacent to the column face may still provide reliable service under some conditions. These conditions may include limitations on the size of the connected sections, the use of base and weld metals with adequate notch toughness, joint detailing that minimizes notch effects, and appropriate control of the relative strength of the beam and column materials. Sufficient research has not been performed to date either to confirm these suggestions or define the conditions in which they are valid. Research however does indicate that reliable performance can be attained if the plastic

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hinge is shifted away from the column face, as suggested above. Consequently, these Interim Guidelines make a general recommendation that this approach be taken. Additional research should be performed to determine the acceptability of other approaches. It should also be noted that reinforced connection (or reduced beam section) configurations of the type described above, while believed to be effective in preventing brittle connection fractures, will not prevent structural damage from occurring. Brittle connection fractures are undesirable because they result in a substantial reduction in the lateral-force-resisting strength of the structure which, in extreme cases, can result in instability and collapse. Connections configured as described in these Interim Guidelines should experience many fewer such brittle fractures than unmodified connections. However, the formation of a plastic hinge within the span of a beam is not a completely benign event. Beams which have formed such hinges may, if plastic rotations are large, exhibit significantlarge buckling and yielding deformation, damage which typically must be repaired. The cost of such repairs could be comparable to the costs incurred in repairing fracture damage experienced in the Northridge Earthquake. The primary difference is that life safety protection will be significantly enhanced and most structures that have experienced such plastic deformation damage should continue to be safe for occupancy while repairs are made. If the types of damage described above are unacceptable for a given building, then alternative structural systems should be considered that will reduce the plastic deformation demands on the structure during a strong earthquake. Appropriate methods of achieving such goals include the installation of supplemental braced frames, energy dissipation systems, base isolation systems and similar structural systems. Framing systems incorporating partially restrained connections may also be quite effective in resisting large earthquake induced deformation with limited damage. It is important to recognize that in frames with relatively short bays, the flexural hinging indicated in Figure 7.2.3-1 may not be able to form. If the effective flexural length (L’in the figure) of beams in a frame becomes too short, then the beams or girders will yield in shear before zones of flexural plasticity can form, resulting in an inelastic behavior that is more like that of an eccentrically braced frame than that of a moment frame. This behavior may inadvertently occur in frames in which relatively large strengthened connections, such as haunches, cover plates or side plates have been used on beams with relatively short spans. This behavior is illustrated in Figure 7.2.3-2. The guidelines contained in this section are intended to address the design of flexurally dominated moment resisting frames. When utilizing these guidelines, it is important to confirm that the configuration of the structure is such that the New Construction 7-8

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presumed flexural hinging can actually occur. It is possible that shear yielding of frame beams, such as that schematically illustrated in Figure 7.2.3-2 may be a desirable behavior mode. However, to date, there has not been enough research conducted into the behavior of such frames to develop recommended design guidelines. Designers wishing to utilize such configurations should refer to the code requirements for eccentrically braced frames. Particular care should be taken to brace the shear link of such beams against lateral-torsional buckling and also to adequately stiffen the webs to avoid local buckling following shear plastification. Shear Link

Shear Link

Figure 7.2.3-2 Shear Yielding Dominated Behavior of Short Bay Frames 7.2.4 Plastic Rotation Capacity

The plastic rotation capacity of tested connection assemblies should reflect realistic estimates of the total (elastic and plastic) drift likely to be induced in the frame by earthquake ground shaking, and the geometric configuration of the frame. For frames of typical configuration, and for ground shaking of the levels anticipated by the building code, a minimum plastic rotation capacity of 0.03 radian is recommended. As used in these Guidelines, plastic rotation is defined as the plastic chord rotation angle. The plastic chord rotation angle is calculated using the rotated coordinate system shown in Fig. 7.2.4-1 as the plastic deflection of the beam or girder, at its point of inflection (usually the mid-span,) ∆CI, divided by the distance between this mid-span point and the centerline of the panel zone of the beam column connection, LCL. This convention is illustrated in Figure 7.2.4-1.

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LCL

Plastic hinge

Beam span center line

cL

∆CL

lh

θp =

∆ CL

LCL

Figure 7.2.4-1 Calculation of Plastic Rotation Angle It is important to note that this definition of plastic rotation is somewhat different than the plastic rotation that would actually occur within a discrete plastic hinge in a frame model similar to that shown in Figure 7.2.3-1. These two quantities are related to each other, however, and if one of them is known, the other may be calculated from Eq. 7.2.4-1.When the configuration of a frame is such that the ratio L/L’is greater than 1.25, the plastic rotation demand should be taken as follows:

θ p = θ ph where: θp θph LCL lh

( LCL − lh ) LCL

(7.2.4-1)

is the plastic chord angle rotation, as used in these Guidelines is the plastic rotation, at the location of a discrete hinge is the distance from the center of the beam-column assembly panel zone to the center of the beam span is the location of the discrete plastic hinge relative to the center of the beam-column assembly panel zone θ = 0.025(1 + ( L − L' ) L' )

(7-1)

where: L is the center to center spacing of columns, and L’is the center to center spacing of plastic hinges in the bay under consideration The indicated rotation demands may be reduced when positive means, such as the use of base isolation or energy dissipation devices, are introduced into the design to control the building’s response. When such measures are taken, nonlinear dynamic analyses should be performed and the connection demands taken as 0.005 radians greater than the plastic rotation demands calculated in the analyses. The nonlinear analyses should conform to the criteria specified in UBC-94 Section 1655 {NEHRP-94 Section 2.6.4.2} for nonlinear dynamic analysis of base New Construction 7-10

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isolated structures. Ground motion time histories utilized for these nonlinear analyses should satisfy the scaling requirements of UBC-94 Section 1655.4.2 {NEHRP-94 Section 2.6.4.4}, except that if the building is not base isolated, the structure period T, calculated in accordance with UBC-94 Section 1628 {NEHRP-94 Section 2.3.3.1} should be substituted for TI. When using methods of nonlinear analysis to establish the plastic rotation demands on frame connections, the analysis results should not be scaled by the factor Rw (R) or RWi (Ri ), as otherwise permitted by the building code. Commentary: When the Interim Guidelines were first published, the plastic rotation was defined as that rotation that would occur at a discrete plastic hinge, similar to the definition of θph. in Eq. 7.2.4-1, above. In subsequent testing of prototype connection assemblies, it was found that it is often very difficult to determine the value of this rotation parameter from test data, since actual plastic hinges do not occur at discrete points in the assembly and because some amount of plasticity also occurs in the panel zone of many assemblies. The plastic chord angle rotation, introduced in this advisory, may more readily be obtained from test data and also more closely relates to the drift experienced by a frame during earthquake response. This change in the definition of plastic rotation does not result in any significant change in the acceptance criteria for beam-column assembly qualification testing. When the Interim Guidelines were first published, they recommended an acceptance criteria given by Eq. 7.2.4-2, below:

 L − L'  θ p = 0.0251 +   L' 

(7.2.4-2)

For typical beam-column assemblies in which the plastic hinge forms relatively close to the face of the column, perhaps within a length of 1/2 the beam depth, this typically resulted in a plastic rotation demand of 0.03 radians, as currently measured. Traditionally, engineers have calculated demand in moment frames by sizing the members for strength and drift using code forces (either equivalent static or reduced dynamic forces) and then "developing the strength of the members." Since 1988, "developing the strength" has been accomplished by prescriptive means based on a review of testing of moment frame connections to that date. It was assumed that the prescribed connections would be strong enough that the beam or girder would yield (in bending), or the panel zone would yield (in shear) in a nearly perfectly plastic manner producing the plastic rotations necessary to dissipate the energy of the earthquake.

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A realistic estimate of the interstory drift demand for most structures and most earthquakes is on the order of 0.015 to 0.025 times the story height for WSMF structures designed to code allowable drift limits. In such frames, a portion of the drift will be due to elastic deformations of the frame, while the balance must be provided by inelastic rotations of the beam plastic hinges, by yielding of the column panel zone, or by a combination of the two. In the 1994 Northridge Earthquake, many moment-frame connections fractured with little evidence of plastic hinging of the beams or yielding of the column panel zones. Testing of moment frame connections both prior to and subsequent to the earthquake suggests that the standard, pre-Northridge, welded flange-bolted web connection is unable to reliably provide plastic rotations beyond about 0.005 radian for all ranges of beam depths and often fails below that level. Since the elastic contribution to drift may approach 0.01 radian, the necessary inelastic contributions will exceed the capability of the standard connection in many cases. For frames designed for code forces and for the code drift, the necessary plastic rotational demand may be expected to be on the order of 0.02 radian or more and new connection configurations should be developed to accommodate such rotation without brittle fracture. The recommended plastic rotation connection demand of 0.03 radians was selected both to provide a comfortable margin against the demands actually expected in most cases and because in recent testing of connection assemblies, specimens capable of achieving this demand behaved in a ductile manner through the formation of plastic hinges. For a given building design, and known earthquake hazard, it is possible to more accurately estimate plastic rotation demands on frame connections. This requires the use of nonlinear analysis techniques. Analysis software capable of performing such analyses is becoming more available and many design offices will have the ability to perform such analyses and develop more accurate estimates of inelastic demands for specific building designs. However, when performing such analyses, care should be taken to evaluate building response for multiple earthquake time histories, representative of realistic ground motions for sites having similar geologic characteristics and proximity to faults as the actual building site. Relatively minor differences in the ground motion time history used as input in such an analysis can significantly alter the results. Since there is significant uncertainty involved in any ground motion estimate, it is recommended that analysis not be used to justify the design of structures with non-ductile connections, unless positive measures such as the use of base isolation or energy dissipation devices are taken to provide reliable behavior of the structure.

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It has been pointed out that it is not only the total plastic rotation demand that is important to connection and frame performance, but also the connection mechanism (for example - panel zone yielding, girder flange yielding/buckling, etc.) and hysteretic loading history. These are matters for further study in the continuing research on connection and joint performance. 7.2.5 Redundancy

The frame system should be designed and arranged to incorporate as many moment-resisting connections as is reasonable into the moment frame. Commentary: Early moment frame designs were highly redundant and nearly every column was designed to participate in the lateral-force-resisting system. In an attempt to produce economical designs, recent practice often yieldedproduced designs which utilized only a few large columns and beams in a small proportion of the building’s frames for lateral resistance, with the balance of the building columns designed not considered or designed to participate in lateral resistance. This practice led to the need for large welds at the connections and to reliance on only a few connections for the lateral stability of the building. The resulting large framing elements and connections are believed to have exacerbated the poor performance of the pre-Northridge connection. Further, if only a few framing elements are available to resist lateral demands, then failure of only a few connections has the potential to result in a significant loss of earthquakeresisting strength. Together, these effects are not beneficial to building performance. The importance of redundancy to building performance can not be overemphasized. Even connections designed and constructed according to the improved procedures recommended by these Interim Guidelines will have some potential, albeit greatly reduced, for brittle failures. As the number of individual beams and columns incorporated into the lateral-force-resisting system is increased, the consequences of isolated connection failures are significantly reduceds. Further, as more framing elements are activated in the building’s response to earthquake ground motion, the building develops greater potential for energy absorption and dissipation, and greater ability to limit control earthquake-induced deformations to acceptable levels. Incorporation of more of the building framing into the lateral-force-resisting system will lead to smaller members and therefore an anticipated increase in the reliability of individual connections. It will almost certainly lead to improved overall system reliability. Further, recent studies conducted by designers indicate that under some conditions, redundant framing systems can be constructed as economically as non-redundant systems. In these studies, the additional costs incurred in making a greater number of field-welded moment-resisting New Construction 7-13

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connections in the more redundant frame were balanced by a reduced total tonnage of steel in the lateral-force-resisting systems, and sometimes reduced foundation costs as well. In order to codify the need for more redundant structural systems, the 1997 Uniform Building Code has specifically adopted a reliability coefficient, ρx, tied to the redundancy of framing present in the building. This coefficient, with values varying from 1.0 for highly redundant structures to 1.5 for non-redundant structures, is applied to the design earthquake forces, E, in the load combination equations, and has the effect of requiring more conservative design force levels for structures with nonredundant systems. The Building Seismic Safety Council’s Provisions Update Committee has also approved a proposal to include such a coefficient in the1997 NEHRP Provisions also includes such a coefficient. The formulation of this coefficient and its application are very similar in both the 1997 Uniform Building Code and 1997 NEHRP Provisions. As proposed contained in the 1997 NEHRP Provisions, the reliability coefficient is given by the equation: ρ = 2−

20 rmax Ax

(7.25-1)

where:

rmax x =

the ratio of the design story shear resisted by the single element

carrying the most shear force in the story to the total story shear, for a given direction of loading. For moment frames,

rmax x

is taken as the

maximum of the sum of the shears in any two adjacent columns in a moment frame divided by the story shear. For columns common to two bays with moment resisting connections on opposite sides at the level under consideration, 70% of the shear in that column may be used in the column shear summation. Ax = the floor area in square feet of the diaphragm level immediately above the story. The 1997 UBC and NEHRP Provisions also require that structures utilizing moment resisting frames as the primary lateral force resisting system be proportioned such that they qualify for a maximum value of ρx of 1.25. Structures located within a few kilometers of major active faults must be configured so as to qualify for a maximum value of ρx of 1.1.

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The most redundant moment-resisting frame systems are distributed frames in which all beam-column connections are detailed to be moment resisting. In these types of structures, half of the moment-resisting connections will be to the minor axis of the column which will typically result in weak column/strong beam framing. The 1994 UBC requirements limit the portion of the building design lateral forces that can be resisted by relative number of weak column/strong beam connections in the moment frame system. This limitation was adopted to avoid the design of frames likely to develop story mechanisms as opposed to concern about the adequacy of moment-resisting connections to the minor axis of columns. However, the limited research data available on such connections suggests that they do not behave well. There is a divergence of opinion among structural engineers on the desirability of frames in which all beam-column connections are made momentresisting, including those of beams framing to the minor axis of columns. Use of such systems as a means of satisfying the redundancy recommendations of these Interim Guidelines requires careful consideration by the structural engineer. Limited testing in the past has indicated that moment connections made to the minor axis of wide flange columns are subject to the same types of fracture damage experienced by major axis connections. As of this time, there has not been sufficient research to suggest methods of making reliable connections to the column minor axis. 7.2.6 System Performance

There are no modifications to the Guidelines or Commentary of Section 7.2.6 at this time. 7.2.7 Special Systems

There are no modifications to the Guidelines or Commentary of Section 7.2.7 at this time. 7.3 Connection Design & Qualification Procedures - General 7.3.1 Connection Performance Intent

The intent of connection design should be to force the plastic hinge away from the face of the column to a pre-determined location within the beam span. This may be accomplished by local reinforcement of the connection itself (cover plates, haunches, side plates, etc.) or by local reductions of the beam section (drilled holes, trimmed flanges, etc.). All elements of the connection should have adequate strength to develop the forces resulting from the formation of the plastic hinge at the predetermined location, together with forces resulting from gravity loads. Section 7.5.2 outlines a design procedure for reinforced connection designs. Section 7.5.3 provides a design procedure for reduced section connections.

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7.3.2 Qualification by Testing

There are no modifications to the Guidelines or Commentary of Section 7.3.2 at this time. 7.3.3 Design by Calculation

There are no modifications to the Guidelines or Commentary of Section 7.3.3 at this time. 7.4 Guidelines for Connection Qualification by Testing 7.4.1 Testing Protocol

There are no modifications to the Guidelines or Commentary of Section 7.4.1 at this time. 7.4.2 Acceptance Criteria

The minimum acceptance criteria for connection qualification for specimens tested in accordance with these Interim Guidelines should be as follows: a)

The connection should develop beam plastic rotations as indicated in Section 7.2.4, for at least one complete cycle.

b)

The connection should develop a minimum strength equal to the plastic strength of the girder, calculated using minimum specified yield strength Fy, tThroughout the loading history required to achieve the required plastic rotation capacity, as indicated in a), above, the connection should develop a minimum moment at the column face as follows: i)

For strengthened connections, the minimum moment at the column face should be equal to the plastic moment of the girder, calculated using the minimum specified yield strength, Fy of the girder. If the load limiting mechanism in the test is buckling of the girder flanges, the engineer, upon consideration of the effect of strength degradation on the structure, may consider a minimum of 80% of the nominal strength as acceptable.

ii)

For reduced section connection designs, the minimum moment at the column face should be equal to the moment corresponding to development of the nominal plastic moment of the reduced section at the reduced section, calculated using the minimum specified yield strength, Fy of the girder, and the plastic section modulus for the reduced section. The moment at the column face should not be less than 80% of the nominal plastic moment capacity of the unreduced girder section.

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c)

The connection should exhibit ductile behavior throughout the loading history. A specimen that exhibits a brittle limit state (e.g. complete flange fracture, column cracking, through-thickness failures of the column flange, fractures in welds subject to tension, shear tab cracking, etc. ) prior to reaching the required plastic rotation should be considered unsuccessful.

d)

Throughout the loading history, until the required plastic rotation is achieved, the connection should be judged capable of supporting dead and live loads required by the building code. In those specimens where axial load is applied during the testing, the specimen should be capable of supporting the applied load throughout the loading history.

The evaluation of the test specimen’s performance should consistently reflect the relevant limit states. For example, the maximum reported moment and the moment at the maximum plastic rotation are unlikely to be the same. It would be inappropriate to evaluate the connection using the maximum moment and the maximum plastic rotation in a way that implies that they occurred simultaneously. In a similar fashion, the maximum demand on the connection should be evaluated using the maximum moment, not the moment at the maximum plastic rotation unless the behavior of the connection indicated that this limit state produced a more critical condition in the connection. Commentary: While the testing of all connection geometries and member combinations in any given building might be desirable, it would not be very practical nor necessary. Test specimens should replicate, within the limitations associated with test specimen simplification, the fabrication and welding procedures, connection geometry and member size, and potential modes of failure. If the testing is done in a manner consistent with other testing programs, reasonable comparisons can be made. On the other hand, testing is expensive and it is difficult to realistically test the beam-column connection using actual boundary conditions and earthquake loading histories and rates. It was suggested in Interim Recommendation No. 2 by the SEAOC Seismology Committee that three tested specimens be the minimum for qualification of a connection. Further consideration has led to the recognition that while three tests may be desirable, the actual testing program selected should consider the conditions of the project. Since the purpose of the testing program is to "qualify the connection", and since it is not practical for a given project to do enough tests to be statistically meaningful considering random factors such as material, welder skills, and other variables, arguments can be made for fewer tests of identical specimens, and concentration on testing specimens which represent the range of different properties which may occur in the project. Once a connection is qualified, that is, once it has been confirmed that the connection can work, monitoring of actual materials and quality control to assure emulation of the tested design becomes most important. New Construction 7-17

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Because of the cost of testing, use of calculations for interpolation or extrapolation of test results is desirable. How much extrapolation should be accepted is a difficult decision. As additional testing is done, more information may be available on what constitutes "conservative" testing conditions, thereby allowing easier decisions relative to extrapolating tests to actual conditions which are likely to be less demanding than the tests. For example, it is hypothesized that connections of shallower, thinner flanged members are likely to be more reliable than similar connections consisting of deeper, thicker flanged members. Thus, it may be possible to test the largest assemblages of similar details and extrapolate to the smaller member sizes - at least within comparable member group families. However, there is evidence to suggest that extrapolation of test results to assemblies using members of reduced size is not always conservative. In a recent series of tests of cover plated connections, conducted at the University of California at San Diego, a connection assembly that produced acceptable results for one family of beam sizes, W24, did not behave acceptably when the beam depth was reduced significantly, to W18. In that project, the change in relative flexibilities of the members and connection elements resulted in a shift in the basic behavior of the assembly and initiation of a failure mode that was not observed in the specimens with larger member sizes. In order to minimize the possibility of such occurrences, when extrapolation of test results is performed, it should be done with a basic understanding of the behavior of the assembly, and the likely effects of changes to the assembly configuration on this behavior. Test results should not be extrapolated to assembly configurations that are expected to behave differently than the tested configuration. Extrapolation or interpolation of results with differences in welding procedures, details or material properties is even more difficult. 7.5 Guidelines for Connection Design by Calculation In conditions where it has been determined that design of connections by calculation is sufficient, or when calculations are used for interpolation or extrapolation, the following guidelines should be used. 7.5.1 Material Strength Properties

In the absence of project specific material property information, the values listed in Table 7-1 Table 7.5.1-1 should be used to determine the strength of steel shape and plate for purposes of calculation. The permissible strength for weld metal should be taken in accordance with the building code. Additional information on material properties may be found in the Interim Guidelines of Chapter 8.

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Table 7-1Table 7.5.1-1 - Properties for Use in Connection Design Material A36

Fy (ksi) 36

Fy m (ksi) use values for Dual Certified

Fu (ksi) 58

Dual Certified Beam Axial, Flexural3 50 65 min. Shape Group 1 551 Shape Group 2 581 Shape Group 3 571 Shape Group 4 541 Through-Thickness Note 5 A572 Column/Beam Axial, Flexural3 50 65 min. Shape Group 1 581 Shape Group 2 581 Shape Group 3 571 Shape Group 4 571 Shape Group 5 551 , Through-Thickness Note 5 A9922 Use same values as ASTM A572 A913-50 Axial, Flexural 50 581 65 min. , Through-thickness Note 5 A913--— 65 Axial, Flexural 65 751(4) 80 min. Through-thickness Note 5 Notes: 1. Based on coupons from web. For thick flanges, the Fy flange is approximately 0.95 Fy web. 2. See Commentary 3. Values based on (SSPC-1994) 4. ASTM A913, Grade 65 material is not recommended for use in the beams of moment resisting frames 5. See Commentary

Commentary: Table7.5.1-1 Note 2 - The ASTM A992 specification was specifically developed by the steel industry in response to expressed concerns of the design community with regard to the permissible variation in chemistry and mechanical properties of structural steel under the A36 and A572 specifications. This new specification, which was adopted in late 1998, is very similar to ASTM A572, except that it includes somewhat more restrictive limits on chemistry and on the permissible variation in yield and ultimate tensile stress, as well as the ratio of yield to tensile strength. At this time, no statistical data base is available to estimate the actual distribution of properties of material produced to this specification. However, the properties are likely to be very similar, albeit with less statistical scatter, to those of material recently produced under ASTM A572, Grade 50.

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Table 7.5.1-1 Note 5 -In the period immediately following the Northridge earthquake, the Seismology Committee of the Structural Engineers Association of California and the International Conference of Building Officials issued Interim Recommendation No. 2 (SEAOC-1995) to provide guidance on the design of moment resisting steel frame connections. Interim Recommendation No. 2 included a recommendation that the through-thickness stress demand on column flanges be limited to a value of 40 ksi, applied to the projected area of beam flange attachment. This value was selected somewhat arbitrarily, to ensure that through-thickness yielding did not initiate in the column flanges of momentresisting connections and because it was consistent with the successful tests of assemblies with cover plates conducted at the University of Texas at Austin (Engelhardt and Sabol - 1994), rather than being the result of a demonstrated through-thickness capacity of typical column flange material. Despite the somewhat arbitrary nature of the selection of this value, its use often controls the overall design of a connection assembly including the selection of cover plate thickness, haunch depth, and similar parameters. It would seem to be important to prevent the inelastic behavior of connections from being controlled by through-thickness yielding of the column flanges. This is because it would be necessary to develop very large local ductilities in the column flange material in order to accommodate even modest plastic rotation demands on the assembly. However, extensive investigation of the throughthickness behavior of column flanges in a “T” joint configuration reveals that neither yielding, nor through-thickness failure are likely to occur in these connections. Barsom and Korvink (1997) conducted a statistical survey of available data on the tensile strength of rolled shape material in the throughthickness direction. These tests were generally conducted on small diameter coupons, extracted from flange material of heavy shapes. The data indicates that both the yield stress and ultimate tensile strength of this material in the throughthickness direction is comparable to that of the material in the direction parallel to rolling. However, it does indicate somewhat greater scatter, with a number of reported values where the through-thickness strength was higher, as well as lower than that in the longitudinal direction. Review of this data indicates with high confidence that for small diameter coupons, the yield and ultimate tensile values of the material in a through-thickness direction will exceed 90% and 80% respectively of the comparable values in the longitudinal direction. the actual The causes for through-thickness failures of column flanges (types C2, C4, and C5), observed both in buildings damaged by the Northridge Earthquake and in some test specimens, are not well understood. They are thought to be a function of the metallurgy and “purity” of the steel; conditions of loading including the presence of axial load and rate of loading application; conditions of tri-axial restraint; conditions of local hardening and embrittlement within the weld’s heat affected zone; stress concentrations induced by the presence of backing bars and

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defects at the root of beam flange to column flange welds; and by the relationship of the connection components as they may affect flange bending stresses and flange curvature induced by panel zone yielding. Given the many complex factors which can affect the through-thickness strength of the column flange, determination of a reliable basis upon which to set permissible design stresses will require significant research. Such research is currently being conducted under the SAC phase II program. While this statistical distribution suggests the likelihood that the throughthickness strength of column flanges could be less than the flexural strength of attached beam elements, testing of more than 40 specimens at Lehigh University indicates that this is not the case. In these tests, high strength plates, representing beam flanges and having a yield strength of 100 ksi were welded to the face of A572, Grade 50 and A913, Grade 50 and 65 column shapes, to simulate the portion of a beam-column assembly at the beam flange. These specimens were placed in a universal testing machine and loaded to produce high through-thickness tensile stresses in the column flange material. The tests simulated a wide range of conditions, representing different weld metals as well and also to include eccentrically applied loading. In 40 of 41 specimens tested, the assembly strength was limited by tensile failure of the high strength beam flange plate as opposed to the column flange material. In the one failure that occurred within the column flange material, fracture initiated at the root of a lowtoughness weld, at root defects that were intentionally introduced to initiate such a fracture. The behavior illustrated by this test series is consistent with mechanics of materials theory. At the joint of the beam flange to column flange, the material is very highly restrained. As a result of this, both the yield strength and ultimate tensile strength of the material in this region is significantly elevated. Under these conditions, failure is unlikely to occur unless a large flaw is present that can lead to unstable crack propagation and brittle fracture. In light of this evidence, Interim Guidelines Advisory No. 2 deletes any requirement for evaluation of through-thickness flange stress in columns. Interim Recommendation No. 2 (SEAOC-1995) included a value of 40 ksi, applied to the projected area of beam flange attachment, for the throughthickness strength to be used in calculations. This value was selected because it was consistent with the successful tests of assemblies with cover plates conducted at the University of Texas at Austin (Engelhardt and Sabol - 1994). However, because of the probable influence of all the factors noted above, this value can only be considered to reflect the specific conditions of those tests and specimens. Although reduced stresses at the column face produced acceptable results in the University of Texas tests, the key to that success was more likely the result of New Construction 7-21

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forcing the plastic hinge away from the column than reduction of the throughthickness stress by the cover plates. Reduction of through-thickness column flange stress to ever lower levels by the use of thicker cover plates is not recommended, since such cover plates will result in ever higher forces on the face of the column flange as well as larger weldments with potential for enlarged heat affected zones, higher residual stresses and conditions of restraint. Since the initial publication of the Interim Guidelines, a significant number of tests have been performed on reduced beam section connections (See section 7.5.3), most of which employed beam flanges which were welded directly to the column flanges using improved welding techniques, but without reinforcement plates. No through-thickness failures occurred in these tests despite the fact that calculated through-thickness stresses at the root of the beam flange to column flange joint ranged as high as 58 ksi. The successful performance of these welded joints is most probably due to the shifting of the yield area of the assembly away from the column flange and into the beam span. Based on the indications of the above described tests, and noting the undesirability of over reinforcing connections, it is now suggested that a maximum through-thickness stress of 0.9Fyc may be appropriate for use with connections that shift the hinging away from the column face. Notwithstanding this recommendation, engineers are still cautioned to carefully consider the through-thickness issue when these other previously listed conditions which are thought to be involved in this type of failure are prevalent. Connections relying on through-thickness strength can not be considered to be fully reliable until the influence of the other parameters discussed above can be fully understood. A high amount of structural redundancy is recommended for frames employing connections which rely on through-thickness strength of the column flange. Notwithstanding all of the above, successful tests using cover plates and other measures of moving hinges (and coincidentally reducing through-thickness stress) continue to be performed. In the interim, engineers choosing to utilize connections relying on through-thickness strength should recognize that despite the successful testing, connections relying on through-thickness strength can not be considered to be fully reliable until the influence of the other parameters discussed above can be fully understood. A high amount of structural redundancy is recommended for frames employing connections which rely on through-thickness strength of the column flange. 7.5.2 Design Procedure - Strengthened Connections

The following procedure may be followed to size the various elements of strengthened connection assemblies that are intended to promote formation of plastic hinges within the beam span by providing a reinforced beam section at the face of the column. Section 7.5.3 provides a

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modified procedure recommended for use in the design of connection assemblies using reduced beam sections to promote similar inelastic behavior. Begin by selecting Select a connection configuration, such as one of those indicated in Sections 7.9.1, 7.9.2, 7.9.3, 7.9.4, or 7.9.5, that will permit the formation of a plastic hinge within the beam span, away from the face of the column, when the frame is subjected to gravity and lateral loads. Then proceed as described in the following sections. The following procedure should be followed to size the various elements of the connection assembly: 7.5.2.1 Determine Plastic Hinge Locations

For beams with gravity loads representing a small portion of the total flexural demand, the location of the plastic hinge may be assumed to occur as indicated in Table 7.5.2.1-1 at a distance equal to 1/3 of the beam depth from the edge of the reinforced connection (or start of the reduced beam section), unless specific test data for the connection indicates that a different location value is more appropriate. Refer to Figure 7-2Figure 7.5.2.1-1. Table 7.5.2.1-1 Plastic Hinge Location - Strengthened Connections Connection Type

Reference Section

Hinge Location “sh”

Sect. 7.9.1

d/4 beyond end of cover plates

Haunches

Sect. 7.9.3, 7.9.4

d/3 beyond toe of haunch

Vertical Ribs

Sect. 7.9.2

d/3 beyond toe of ribs

Plastic hinge sh=

Edge of reinforced connection

d/4

Connection reinforcement

sh= d/3

Edge of reinforced connection

Beam depth - d

Cover plates

L’

L

Figure 7-2 Figure 7.5.2.1-1 - Location of Plastic Hinge Commentary: The suggested locations for the plastic hinge, at a distance d/3 away from the end of the reinforced section (or beginning of reduced section) indicated in Table 7.5.2.1-1 and Figure 7.5.2.1-1 are is based on the observed behavior of test specimens, with no significant gravity load present. If significant

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gravity load is present, this can shift the locations of the plastic hinges, and in the extreme case, even change the form of the collapse mechanism. If flexural demand on the girder due to gravity load is less than about 30% of the girder plastic capacity, this effect can safely be neglected, and the plastic hinge locations taken as indicated. If gravity demands significantly exceed this level, then plastic analysis of the girder should be performed to determine the appropriate hinge locations. In zones of high seismicity (UBC Zones 3 and 4, and NEHRP Map Areas 6 and 7), gravity loading on the girders of earthquake resisting frames typically has a very small effect, unless tributary areas for gravity loads are large. 7.5.2.2 Determine Probable Plastic Moment at Hinges

Determine the probable value of the plastic moment, Mpr, at the location of the plastic hinges as: M pr = βM p = βZ b Fy where: ß

Zb

(7.5.2.2-12)

is a coefficient that adjusts the nominal plastic moment to the estimated hinge moment based on the mean yield stress of the beam material and the estimated strain hardening. A value of 1.2 should be taken for β for ASTM A572, A992 and A913 steels. When designs are based upon calculations alone, an additional factor is recommended to account for uncertainty. In the absence of adequate testing of the type described above, ß should be taken as 1.4 for ASTM A572 and for A913, Grades 50 and 65 steels. Where adequate testing has been performed ß should be permitted to be taken as 1.2 for these materials. is the plastic modulus of the section

Commentary: In order to compute β, the expected yield strength, strain hardening and an appropriate uncertainty factor need to be determined. The following assumed strengths are recommended: Expected Yield:

The expected yield strength, for purposes of computing (Mpr) may be taken as: Fye = 0.95 Fym

(7.5.2.2-2-3)

The 0.95 factor is used to adjust the yield stress in the beam web, where coupons for mill certification tests are normally extracted, to the value in the beam flange. Beam flanges, being comprised of thicker material, typically have somewhat lower yield strengths than do beam web material.

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Fy m for various steels are as shown in Table 7-1 Table 7.5.1-1, based on a survey of web coupon tensile tests (Steel Shape Producers Council - 1994). The engineer is cautioned that there is no upper limit on the yield point for ASTM A36 steel and consequently, dual-certification steel having properties consistent with ASTM A572, Grade 50 is routinely supplied when ASTM A36 is specified. Consequently, it is the recommendation here that the design of connections be based on an assumption of Grade 50 properties, even when A36 steel is specified for beams. It should be noted that at least one producer offers A36 steel with a maximum yield point of 50 ksi in shape sizes ranging up to W 24x62. Refer to the commentary to Section 8.1.3 for further discussion of steel strength issues. Strain Hardening: A factor of 1.1 is recommended for use with the mean yield stress in the foregoing table when calculating the probable plastic moment capacity Mpr.. The 1.1 factor for strain hardening, or other sources of strength above yield, agrees fairly well with available test results. The 1.1 factor could underestimate the over-strength where significant flange buckling does not act as a gradual limit on the beam strength. Nevertheless, the 1.1 factor seems a reasonable expectation of over-strength considering the complexities involved. Modeling Uncertainty: Where a design is based on approved cyclic testing, the modeling uncertainty may be taken as 1.0, otherwise the recommended value is 1.2. When the Interim Guidelines were first published, the β coefficient included a 1.2 factor to account for modeling uncertainty when connection designs were based on calculations as opposed to a specific program of qualification testing. The intent of this factor was twofold: to provide additional conservatism in the design when specific test data for a representative connection was not available and also as an inducement to encourage projects to undertake connection qualification testing programs. After the Interim Guidelines had been in use for some time, it became apparent that this approach was not an effective inducement for projects to perform qualification testing, and also that the use of an overly large value for the β coefficient often resulted in excessively large connection reinforcing elements (cover plates, e.g.) and other design features that did not appear conducive to good connection behavior. Consequently, it was decided to remove this modeling uncertainty factor from the calculation of β. In summary, for Grade 50 steel, we have: β = [0.95 (54 ksi to 58 ksi)/50 ksi] (1.1) 1.2) = 1.35 t0 1.45, say 1.4 β = [0.95 (54 ksi to 59 ksi)/50 ksi] (1.1) = 1.13 to 1.21, say 1.2

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7.5.2.3 Determine Shear at the Plastic Hinge

The shear at the plastic hinge should be determined by statics, considering gravity loads acting on the beam. A free body diagram of that portion of the beam between plastic hinges, is a useful tool for obtaining the shear at each plastic hinge. Figure 7-3 Figure 7.5.2.3-1 provides an example of such a calculation. For the purposes of such calculations, gravity load should be based on the load combinations required by the building code in use. L/2

Plastic hinge

P

Note: Gravity loads can effect the location of the plastic hinges. If 2Mpr /L’ is less then the gravity shear in the free body (in this case P/2 + wL’/2), then the plastic hinge location will shift significantly and L’must be adjusted, accordingly

w

L’ sh L

P w

VA

Mpr “A”

Vp

Mpr

L’

taking the sum of moments about “A” = 0 Vp ={Mpr + Mpr + P L’/2 + wL’2/2}/L’

Figure 7-3 Figure 7.5.2.3-1- Sample Calculation of Shear at Plastic Hinge Commentary: The UBC gives no specific guidance on the load combinations to use with strength level calculations while the NEHRP Recommended Provisions do specify load factors for the various dead, live and earthquake components of load. For designs performed in accordance with the UBC, it is customary to use unfactored gravity loads when checking the strength of elements. 7.5.2.4 Determine Strength Demands at Each Critical Section

In order to complete the design of the connection, including sizing the various plates and joining welds which make up the connection, it is necessary to determine the shear and flexural strength demands at each critical section. These demands may be calculated by taking a free body of that portion of the connection assembly located between the critical section and the plastic hinge. Figure 7-4 Figure 7.5.2.4-1 demonstrates this procedure for two critical sections, for the beam shown in Figure 7-3 Figure 7.5.2.3-1.

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Plastic hinge

Plastic hinge

Mpr

Mf

Mpr

Mc

Vp

dc

x

Vp x+dc/2

Mf =Mpr +Vpx

Mc=Mpr +Vp(x+dc/2)

Critical Section at Column Face

Critical Section at Column Centerline

Figure 7-4 Figure 7.5.2.4-1 - Calculation of Demands at Critical Sections Commentary: Each unique connection configuration may have different critical sections. The vertical plane that passes through the joint between the beam flanges and column (if such joining occurs) will typically define at least one such critical section, used for designing the joint of the beam flanges to the column, as well as evaluating shear demands on the column panel zone. A second critical section occurs at the center line of the column. Moments calculated at this point are used to check strong column - weak beam conditions. Other critical sections should be selected as appropriate. 7.5.2.5 Check for Strong Column - Weak Beam Condition

When required by the building code, the connection assembly should be checked to determine if strong column - weak beam conditions are satisfied. In lieu of UBC-94 equation 11-3.1 {NEHRP-91 equation 10-3}, the following equation should be used:

∑Z where:

c

(Fyc − f a )

∑M

c

> 1.0

(7.5.2.5-1-4)

Zc is the plastic modulus of the column section above and below the connection Fyc is the minimum specified yield stress for the column above and below fa is the axial load in the column above and below ΣMc is the moment calculated at the center of the column in accordance with Section 7.5.2.4 sum of the column moments at the top and bottom of the panel zone, respectively, resulting from the development of the probable beam plastic moments, Mpr, within each beam in the connection.

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Commentary: The building code provisions for evaluating strong column - weak beam conditions presume that the flexural stiffness of the columns above and below the beam are approximately equal, that the beams will yield at the face of the column, and that the depth of the columns and beams are small relative to their respective span lengths. This permits the code to use a relatively simple equation to evaluate strong column - weak beam conditions in which the sum of the flexural capacities of columns at a connection are compared against the sums of the flexural capacities in the beams. The first publication of the Interim Guidelines took this same approach, except that the definition of ΣMc was modified to explicitly recognize that because flexural hinging of the beams would occur at a location removed from the face of the column, the moments delivered by the beams to the connection would be larger than the plastic moment strength of the beam. In this equation, ΣMc was taken as the sum of the moments at the center of the column, calculated in accordance with the procedures of Sect. 7.5.2.4. assumed point of zero moment

ht

Vc

Vp

Vc =

∑ [M

M ct = Vc ht

Mct

dp

Mpr Vf

(

pr

]

(

)

+ V p ( L − L ' ) / 2) − V f hb + d p / 2 hb + d p + ht

)

M cb = Vc + V f hb

∑

M c = M ct + M cb

Mpr Mcb

hb

Vp

Vc+Vf (L-L’)/2

Note: The quantities Mpr, Vp, L, and L’are as previously identified. Vf is the incremental shear distributed to the column at the floor level. Other quantities are as shown.

Figure 7.5.2.5-1 Calculation of Column Moment for Strong Column Evaluation This simplified approach is not always appropriate. If non-symmetrical connection configurations are used, such as a haunch on only the bottom side of the beam, this can result in an uneven distribution of stiffness between the two column segments, and premature yielding of the column, either above, or below, the beam-column connection. Also, it was determined that for connection configurations in which the panel zone depth represents a significant fraction of New Construction 7-28

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the total column height, such as can occur in some haunched and side-plated connections, the definition of ΣMc contained in the initial printing of the Guidelines could lead to excessive conservatism in determining whether or not a strong column - weak beam condition exists in a structure. Consequently, Interim Guidelines Advisory No. 1 adopted the current definition of ΣMc for use in this evaluation. This definition requires that the moments in the column, at the top and bottom of the panel zone be determined for the condition when a plastic hinge has formed at all beams in the connection. Figure 7.5.2.5-1 illustrates a method for estimating this quantity. 7.5.2.6 Check Column Panel Zone

The adequacy of the shear strength of the column panel zone should be checked. For this purpose, the term 0.8ΣMf should be substituted for the term 0.8ΣMs in UBC-94 Section 2211.7.2.1 {0.9ΣφbMp in NEHRP-91 Section 10.10.3.1}, repeated below for convenience of reference. Mf is the calculated moment at the face of the column, when the beam mechanism forms, calculated as indicated in Section 7.5.2.4 above. In addition, it is recommended that the alternative design criteria indicated in UBC-94 Section 2211.7.2.1 (NEHRP-91 Sect. 10.10.3.1), permitting panel zone shear strength to be proportioned for the shear induced by bending moments from gravity loads plus 1.85 times the prescribed seismic forces, not be used. For convenience of reference, UBC-94 Section 2211.7.2.1 is reproduced below, edited, to indicate the recommended application. 2211.7.2.1 Strength (edited). The panel zone of the joint shall be capable of resisting the shear induced by beam bending moments due to gravity loads plus 1.85 times the prescribed seismic forces, but the shear strength need not exceed that required to develop 0.8ΣMs 0.8ΣMf of the girders framing into the column flanges at the joint. The joint panel zone shear strength may be obtained from the following formula:

 3b c t c f 2  V = 0.55Fy d c t 1 +  dbdct  

(11-1)

where: bc = width of column flange db = the depth of the beam (including any haunches or cover plates) dc = the depth of the column t = the total thickness of the panel zone including doubler plates tcf = the thickness of the column flange

Commentary: The effect of panel zone shear yielding on connection behavior is not well understood. In the past, panel zone shear yielding has been viewed as a benign, or even beneficial mechanism that permits overall frame ductility demands to be accommodated while minimizing the extent of inelastic behavior required of the beam and beam flange to column flange joint. The criteria

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permitting panel zone shear strength to be proportioned for the shears resulting from moments due to gravity loads plus 1.85 times the design seismic forces was adopted by the code specifically to permit designs with somewhat weak panel zones. However, during recent testing of large scale connection assemblies with weak panel zones, it has been noted that in order to accommodate the large shear deformations that occur in the panel zone, extreme “kinking” deformations were induced into the column flanges at the beam flange to column flange welded joint. While this did not lead to premature joint failure in all cases, it is believed to have contributed to such premature failures in at least some of the specimens. The recommendations of this section are intended to result in stronger panel zones than previously permitted by the code, thereby avoiding potential failures due to this kinking action on the column flanges. 7.5.3 Design Procedure - Reduced Beam Section Connections

The following procedure may be followed to size the various elements of reduced beam section (RBS) assemblies with circular curved reductions in beam flanges, such as shown in Figure 7.5.3-1., such as those indicated in Section 7.9.6 indicates other configurations for such connections, however, the circular curved configuration shown in Figure 7.5.3-1 is currently preferred. RBS assemblies are intended to promote the formation of plastic hinges within the beam span by developing a segment of the beam with locally reduced section properties and strength. Begin by selecting an RBS configuration, such as one of those indicated in Figure 7.5.31, that will permit the formation of a plastic hinge within the reduced section of the beam. Of the configurations shown in the figure, the circular curved configuration is preferred. 2

2

4a + l R = radius of cut = 8a

a b f a c l Figure 7.5.3-1 Geometry of Reduced Beam Section

Commentary: Connection assemblies in which inelastic behavior is shifted away from the column face through development of a segment of the beam with intentionally reduced properties, so-called reduced beam section (RBS) or “dogbone” connections, appear to have the potential to provide an economical New Construction 7-30

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solution to the WMSF connection problem. These recommendations are based on limited design configurations that have successfully been tested ing that has been conducted of these types of connections to date. While a A large number of RBS tests have been conducted, these tests have not included the effects of floor slabs or loading rates approximating those that would be produced by a building’s response to earthquake ground motionsincluding some tests of assemblies with floor slabs present. Extensive additional testing of RBS connections, intended to explore these and other factors relevant to connection performance, are currently planned under funding provided by NIST and the SAC phase II program. In the interim, designers specifying RBS connections may wish to consider provision of details to minimize the participation of the slab in the flexural behavior of the beam at the reduced section. The criteria presented in this section are partially based on a draft procedure developed by AISC (Iwankiw, 1996).

Circular

Straight

Reduced Section

Drilled Constant

Tapered

Drilled Tapered

Figure 7.5.3-2 Alternative Reduced-Beam Section Patterns Figure 7.5.3-1 Reduced Beam Section Patterns Several alternative configurations of RBS connections have also been tested to date. As indicated in Figures 7.5.3-21 and 7.9.6-1, these include constant section, tapered section, curved section, and drilled hole patterns. It appears that several of these configurations are more desirable than others. In particular, the drilled hole section patterns have been subject to tensile failure across the reduced net section of the flange through the drill holes. A few RBS tests utilizing straight or tapered cuts have failed within the reduced section at plastic rotation demands less than recommended by these Guidelines. In all of these cases, the failure occurred at locations at which there was a change in direction of the cuts in the beam flange, resulting in a geometric stress riser or notch effect. It is also reported that one of these tests failed at the beam flange continuity plate - to column flange joint. There have been no reported failures of RBS connection New Construction 7-31

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assemblies employing the circular curved flange cuts, and therefore, this is the pattern recommended in these Guidelines. This would appear, therefore, to be a more desirable configuration, although some successful tests have been performed using the straight and tapered configurations. It is important that the pattern of any cuts made in the flange be proportioned so as to avoid sharp cut corners. All corners should be rounded to minimize notch effects and in addition, cut edges should be cut or ground in the direction of the flange length to have a surface roughness meeting the requirements of AWS C4.1-77 class 4, or smootherroughness value less than or equal to 1,000, as defined in ANSI/ASME B46.1. Concerns have been raised by some engineers over the strength reduction inherent in the RBS. Clearly, code requirements for strength, considering gravity loads and gravity loads in combination with wind, seismic and other loads must be met. For higher seismic zones, beam sizes are typically governed by elastic stiffness considerations (drift control) and this must be addressed. Also, for seismic loads, the Building Codes typically require that connections for Special Moment Resisting Frames must develop the “strength” or the “plastic bending moment” of the beam. There may be a problem of semantics where these requirements are applied to a system using RBS connections. Is the RBS part of the connection or is it part of the beam, the strength of which must be developed by the connection? Clearly, the latter interpretation should be applied. Notwithstanding the above, it must be kept in mind that, although unstated, and typically not quantified, there is inherent in design practice an implied relationship between the elastic behavior that we analyze and the inelastic behavior which the building is expected to experience. Elastic drift limitations commonly used are considered to be related to the anticipated inelastic drifts and ultimate lateral stability of the framed structure in at least an intuitively predictable manner. It can be shown that RBS’s such as those that have been tested will reduce the elastic stiffness (increase the drift) on the order of 5%. However, because of the reduction in strength, the effect on the inelastic drift may be more significant. Thus, it seems prudent to require that the RBS maintain a reasonably high proportion of the frame inelastic strength. For the connections tested to date, the inelastic strength of the RBS section has been in the range of 70% of that of the full section. However, the moment demand at the face of the column, corresponding to development of this reduced section strength, is likely to be in the range of 85% to 90% of the strength of the full beam. This seems to be quite reasonably high considering the accuracy of other seismic design assumptions. Although the use of RBS designs tends to reduce the total strength demand on the beam flange - to - column flange connection, relative to strengthened New Construction 7-32

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connections, designs utilizing RBS configurations should continue to follow the recommendations for beam flange continuity plates, weld metal and base metal notch toughness recommended by the Interim Guidelines for strengthened connections. 7.5.3.1 Determine Reduced Section and Plastic Hinge Locations

Beam depth - d

The reduced beam section should be located at a sufficient distance from the face of the column flange (dimension “c” in Figures 7.5.3-1 and 7.5.3.1-1) to avoid significant inelastic behavior of the material at the beam flange - to - column flange joint. Based on testing performed to date, it appears that a value of “c” on the order of ½ to ¾ of the beam width, bf, is sufficient. d/4 (where “d” is the beam depth) is sufficient. The total length of the reduced section of beam flange (dimension “l” in Figures 7.5.3-1 and 7.5.3.1-1) should be on the order of 0.65d to 0.85d, where d is the beam depth.3d/4 to d. The location of the plastic hinge, sh,, may be taken as ½ the length of the cut-out, l.indicated in Table 7.5.3.1-1, unless test data indicates a more appropriate value should be used. When tapered configurations are utilized, the slope of the tapered cut in the beam flange should be arranged such that the variation of the plastic section modulus, Zx, within the reduced section approximates the moment gradient in the beam during the condition when plastic hinges have formed within the reduced beam sections at both ends.

l

Plastic hinge

reduced section

L’

sh c

L

Figure 7.5.3.1-1 Critical Dimensions - RBS Assemblies 7.5.3.2 Determine Strength and Probable Plastic Moment in RBS

The RBS may be proportioned to meet the following criteria:

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1. The section at the RBS should be sufficient to satisfy the strength criteria specified by the building code for Dead, Live, Seismic, Snow, Wind, and other applicable design forces. 2. The elastic stiffness of the frame, considering the effects of the RBS, should be sufficient to meet the drift requirements specified by the code, under the design seismic and other forces. 3. The expected stress in the beam flange - to - column flange weld, under the application of gravity forces and that seismic force that results in development of the probable plastic moment of the reduced section at both ends of the beam, should be less than or equal to the strength of the weld, as indicated in Section 7.2.2 of the Interim Guidelines. 4. The expected through-thickness stress on the face of the column flange, calculated as Mf/Sc, under the application of gravity forces and that seismic force that results in development of the probable plastic moment of the reduced section at both ends of the beam, should be less than or equal to the values indicated in Section 7.5.1, where Mf is the moment at the face of the column flange, calculated as indicated in Section 7.5.2.4, and Sc is the elastic section modulus of the beam at the connection considering weld reinforcement, bolt holes, reinforcing plates, etc. The maximum moment at the face of the column should be in the range of 85 percent to 100 percent of the beam’s expected plastic moment capacity. The depth of cut-out, a, should be selected to be less than or equal to bf/4. The plastic section modulus of the RBS may be calculated from the equation:

(

Z RBS = Z x − bR t f d − t f

)

(7.5.3.2-1)

where: ZRBS Zx bR tf d

is the plastic section modulus of the reduced beam section is the plastic modulus of the unreduced section is the total width of material removed from the beam flange is the thickness of the beam flange is the depth of the beam

The probable plastic moment, Mpr, at the RBS shall be calculated from the equation:

M pr = Z RBS βFy where: ZRBS β

is the plastic section modulus of the reduced beam section is as defined in Section 7.5.2.2 New Construction 7-34

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The strength demand on the beam flange - to - column flange weld and on the face of the column may be determined by following the procedures of Section .7.5.2.3 and 7.5.2.4 of the Interim Guidelines, using the value of Mpr determined in accordance with Eq. 7.5.3.2-2. Commentary: Initial design procedures for RBS connections published by SAC recommended that sufficient reduction of the beam flange be made to maintain flexural stresses in the beam, at the column face, below the anticipated throughthickness yield strength of the column flange material. Since the publication of those recommendations, extensive testing of RBS connections has been conducted, both with and without composite slabs. The testing conducted to date on RBS specimens This testing has typically been for configurations that would result in somewhat larger strength demands at the face of the column flange than suggested by the criteria originally published by SAC. contained in this Advisory. Typically, the tested specimens had reductions in the beam flange area on the order of 35% to 45% and produced moments at the face of the column that resulted in stresses on the weld and column as large as large as 90 to 100% of the expected material strength of the beam, which is often somewhat in excess of the through-thickness yield strength of the column material. The specimens in these tests all developed acceptable levels of inelastic deformation. Recent studies conducted for SAC at Lehigh University confirm that the significant conditions of restraint that exist at the beam flange to column flange joint results in substantially elevated column through-thickness strength, negating a need to reduce flexural stresses below the anticipated column yield strength. In view of this evidence, SAC has elected to adopt design recommendations consistent with configurations that were successfully tested. The criteria contained in this Advisory suggest that these demands be reduced to a level which would maintain weld stresses within their normally specified values and through-thickness column flange stresses at the same levels recommended for strengthened connections. This may require the beam flanges to be reduced by as much as 50% or more for some frame configurations, or that supplemental reinforcement such as cover plates or vertical ribs be provided in addition to the reduced section. This approach was taken to maintain consistency with the criteria recommended for strengthened connections and with the knowledge that the factors affecting the performance of these connections are not yet fully understood. 7.5.3.3 Strong Column - Weak Beam Condition

The adequacy of the design to meet strong column - weak beam conditions should be checked in accordance with the procedures of Section 7.5.2.5

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7.5.3.4 Column Panel Zone

The adequacy of the column panel zone should be checked in accordance with the procedures of Section 7.5.2.6. 7.5.3.5 Lateral Bracing

The reduced section of the beam flanges should be provided with adequate lateral support to prevent lateral-torsional buckling of the section. Lateral braces should be located within a distance equal to 1/2 the beam depth from the expected location of plastic hinging, but should not be located within the reduced section of the flanges. Commentary: Unbraced compression flanges of beams are subject to lateraltorsional buckling, when subjected to large flexural stresses , such as occur in the plastic hinges of beams reduced sections of RBS connections during response to strong ground motion. To prevent such behavior lateral-torsional buckling, it is recommended that both flanges of beams be provided with lateral support. Section 9.8 of the 1997 AISC Seismic Specification requires such bracing in general, and specifically states as follows: “Both flanges of beams shall be laterally supported directly or indirectly. The unbraced length between lateral supports shall not exceed 2500ry/Fy. In addition, lateral supports shall be placed near concentrated forces, changes in cross section and other locations where analysis indicates that a plastic hinge will form during inelastic deformations of the SMF.” Adequate lateral support of the top flanges of beams supporting concrete filled metal deck or formed slabs can usually be obtained through the normal welded attachments of the deck to the beam or through shear studs. Lateral support of beam flanges can also be provided through the connections of transverse framing members or by provision of special lateral braces, attached directly to the flanges. Such attachments should not be made within the reduced section of the beam flange as the welding or bolting required to make such attachments can lead to premature fracturing in these regions of high plastic demands. For beams in moment-resisting frames, it has traditionally been assumed that the direct attachment of the beam flanges to the columns provided sufficient lateral support of both beam flanges to accommodate the plastic hinges anticipated to develop in these frames at the beam-column connection. However, connection configurations like the RBS, developed following the Northridge earthquake, are intended to promote formation of these plastic hinges at some distance from the beam-column interface. This brings to question the adequacy of the beam flange to column flange attachments to provide the necessary lateral

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support at the plastic hinge. While this issue is pertinent for any connection configuration that promotes plastic hinge formation remote from the beamcolumn interface, RBS connections could be more susceptible to lateral-torsion buckling at the plastic hinge because the reductions in the beam flange used to achieve plastic hinge formation also locally reduce the torsional resistance of the section. For that reason, FEMA-267a recommended provision of lateral bracing adjacent to the reduced beam section. Provision of lateral bracing does result in some additional cost. Therefore, SAC has engaged in specific investigations to evaluate the effect of lateral bracing both on the hysteretic behavior of individual connections as well as overall frame response to large lateral displacements. Until these investigations have concluded SAC continues to recommend provision of lateral bracing for RBS connections. It should be noted that Section 9.8 of the 1997 AISC Seismic Specification states: “If members with Reduced Beam Sections, tested in accordance with Appendix S are used, the placement of lateral support for the member shall be consistent with that used in the tests.” Most testing of RBS specimens performed as part of the SAC project have consisted of single beams cantilevered off a column to simulate the exterior connection in a multi-bay moment-resisting frame. The beams have generally been braced at the end of the cantilever length, typically located about 100 inches from the face of the column. For the ASTM A572, Grade 50, W36x150 sections typically tested, this results in a nominal length between lateral supports that is comparable to 2500ry/Fy. The appropriate design strength for lateral bracing of compression elements has long been a matter of debate. Most engineers have applied “rules of thumb” that suggest that the bracing element should be able to resist a small portion, perhaps on the order of 2% to 6% of the compressive force in the element being braced, applied normal to the line of action of the compression. A recent successful test of an RBS specimen conducted at the University of Texas at Austin incorporated lateral bracing with a strength equal to 6% of the nominal compressive yield force in the reduced section. 7.5.3.6 Welded Attachments

Headed studs for composite floor construction should not be placed on the beam flange between the face of the column and the extreme end of the RBS, as indicated in Figure 7.5.3.6-1. Other welded attachments should also be excluded from these regions of the beam.

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welded attachment prohibited welded attachment permitted

Reduced beam section

Figure 7.5.3.6-1 Welded Attachments to RBS Beams Commentary: There are two basic reasons for omitting headed studs in the region between the reduced beam section and the column. The first of these is that composite action of the slab and beam can effectively counteract the reduction in beam section properties achieved by the cutouts in the top beam flange. By omitting shear studs in the end region of the beam, this composite behavior is neutralized, protecting the effectiveness of the section reduction. The second reason is that the portion of the beam at the reduced section is expected to experience large cyclic inelastic strains. If welded attachments are made to the beam in this region, the potential for low-cycle fatigue of the beam, under these large cyclic inelastic strains is greatly increased. For this same reason, other welded attachments should also be excluded from this region. 7.6 Metallurgy and Welding There are no modifications to the Guidelines or Commentary of Section 7.6 at this time. 7.7 Quality Control/Quality Assurance There are no modifications to the Guidelines or Commentary of Section 7.7 at this time. 7.8 Guidelines on Other Connection Design Issues There are no modifications to the Guidelines or Commentary of Section 7.8 at this time.

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7.8.1 Design of Panel Zones

No current recommendations are made to supplement or modify the UBC-1994 {NEHRP-91} provisions for the design of panel zones, other than as indicated in Section 7.5.2.6, above. Panel zone demands should be calculated in accordance with Section 7.5.2.6. As with other elements of the connection, available panel zone strength should be computed using minimum specified yield stress for the material, except when the panel zone strength is used as a limit on the required connection strength, in which case Fym should be used. Where connection design for two-sided connection assemblies is relying on test data for onesided connection assemblies, consideration should be given to maintaining the level of panel zone deformation in the design to a level consistent with that of the test, or at least assume that the panel zone must remain elastic, under the maximum expected shear demands. Commentary: At present, no changes are recommended to the code requirements governing the design of panel zones, other than in the calculation of the demand. As indicated in Section 7.5.2.6, it is recommended that the formulation for panel zone demand contained in the UBC, based on 1.85 times the prescribed seismic forces, not be utilized. This formulation, which is not contained in either the AISC Seismic Provisions or the NEHRP Provisions, is felt to lead to the design of panel zones that are excessively flexible and weak in shear. There is evidence that panel zone yielding may contribute to the plastic rotation capability of a connection. However, there is also concern and some evidence that if the deformation is excessive, a kink will develop in the column flange at the joint with the beam flange and, if the local curvature induced in the beam and column flanges is significant, can contribute to failure of the joint. This would suggest that greater conservatism in column panel zone design may be warranted. In addition to the influence of the deformation of the panel zone on the connection performance, it should be recognized that the use of doubler plates and especially the welding associated with them is likely to be detrimental to the connection performance. It is recommended that the Engineer consider use of column sizes which will not require addition of doubler plates, where practical. 7.8.2 Design of Web Connections to Column Flanges

Specific modifications to the code requirements for design of shear connections are not made at this time. It should be noted that the emergency code change to the UBC-94 {NEHRP-94} deleted the former requirements for supplemental web welds on shear connections. This is felt to be appropriate since these welds can apparently contribute to the potential for shear tab failure at large induced rotations. When designing shear connections for moment-resisting assemblies, the designer should calculate shear demands on the web connection in accordance with Section 7.5.2.4, above. For

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connection designs based on tested configurations, the web connection design should be consistent with the conditions in the tested assemblies. Commentary: Some engineers consider that it is desirable to develop as much bending strength in the web as possible. Additionally, it has been observed in some laboratory testing that pre-mature slip of the bolted web connection can result in large secondary flexural stresses in the beam flanges and the welded joints to the column flange. However, there is some evidence to suggest that if flange connections should fail, welding of shear tabs to the beam web may promote tearing of the tab weld to the column flange or the tab itself through the bolt holes, and some have suggested that welding be avoided and that web connections should incorporate horizontally slotted holes to limit the moment which can be developed in the shear tab, thereby protecting its ability to resist gravity loads on the beam in the event of flexural connection failure. Some recent finite element studies of typical connections by Goel, Popov and others have suggested that even when the shear tab is welded, shear demands at the connections tend to be resisted by a diagonal tension type behavior in the web that tends to result in much of the shear being resisted by the flanges. Investigation of these effects is continuing. 7.8.3 Design of Continuity Plates

There are no modifications to the Guidelines or Commentary of Section 7.8.3 at this time. 7.8.4 Design of Weak Column and Weak Way Connections

There are no modifications to the Guidelines or Commentary of Section 7.8.4 at this time. 7.9 Moment Frame Connections for Consideration in New Construction There are no modifications to the Guidelines or Commentary of Section 7.9 at this time. 7.9.1 Cover Plate Connections

Figure 7-5 Figure 7.9.1-1 illustrates the basic configuration of cover plated connections. Short cover plates are added to the top and bottom flanges of the beam with fillet welds adequate to transfer the cover plate forces to the beam flanges. The bottom flange cover plate is shop welded to the column flange and the beam bottom flange is field welded to the column flange and to the cover plate. The top flange and the top flange cover plate are both field welded to the column flange with a common weld. The web connection may be either welded or high strength (slip critical) bolted. Limited testing of these connections (Engelhardt & Sabol - 1994), (Tsai & Popov -1988) has been performed. More than 30 tests of such connections have been performed, with data on at least 18 of these tests available in the public domain. New Construction 7-40

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A variation of this concept which has been tested successfully very recently (Forell/Elsesser Engineers -1995), uses cover plates sized to take the full flange force, without direct welding of the beam flanges themselves to the column. In this version of the detail, the cover plate provides a cross sectional area at the column face about 1.7 times that of the beam flange area. In the detail which has been tested, a welded shear tab is used, and is designed to resist a significant portion of the plastic bending strength of the beam web.

T&B

Figure 7-5 Figure 7.9.1-1 - Cover Plate Connection Design Issues: Following the Northridge earthquake, the University of Texas at Austin conducted a program of research, under private funding, to develop a modified connection configuration for a specific project. Following a series of unsuccessful tests on various types of connections, approximatelyApproximately eight connections similar to that shown in Figure 7-5 Figure 7.9.1-1 were have been recently tested (Engelhardt & Sabol - 1994), and they have demonstrated the ability to achieve acceptable levels of plastic rotation provided that the beam flange to column flange welding wasis correctly executed and through-thickness problems in the column flange were are avoided. This configuration is relatively economical, compared to some other reinforced configurations, and has limited architectural impact. As a result of these factors, and the significant publicity that followed the first successful tests of these connections, cover plated connections quickly became the predominant configuration used in the design of new buildings. As a result, a number of qualification tests have now been performed on different variations of cover plated connections, covering a wide range of member sizes ranging from W16 to W36 beams, as part of the design process for individual building projects. The results of these tests have been somewhat mixed, with a significant number of failures reported. Although this connection type appears to be significantly more reliable than the typical pre-Northridge connection, it should be expected that some connections in buildings incorporating this detail may still be subjected to earthquake initiated fracture damage. Designers should consider using alternative connection types, unless highly redundant framing systems are employed.

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Six of eight connections tested by the University of Texas at Austin were able to achieve plastic rotations of at least 0.025 radians, or better. Strength loss at the extreme levels of plastic rotation did not reduce the flexural capacity to less than the plastic moment capacity of the section based on minimum specified yield strength. One specimen achieved plastic rotations of 0.015 radians when a brittle fracture of the CJP weld (type W2 failure) occurred. This may partially be the result of a weld that was not executed in conformance with the specified welding procedure specification. The second unsuccessful test specimen achieved plastic rotations of 0.005 radian when a section of the column flange (type C2 failure) occurred. A similar failure occurred in recent testing by Popov of a specimen with cover plates having a somewhat modified plan shape. Quantitative Results: No. of specimens tested: 18 Girder Size: W21 x 68 to W36 x 150 Column Size: W12 x 106 to W14 x 455 Plastic Rotation achieved6 13 Specimens : >0.025 radian 1 3 Specimens: 0.015 0.005 < θp < 0.025 radian 1 2 Specimens: 0.005 radian Although apparently more reliable than the former prescriptive connection, this configuration is subject to some of the same flaws including dependence dependent on properly executed beam flange to column flange welds, and through-thickness behavior of the column flange. Further these effects are somewhat exacerbated as the added effective thickness of the beam flange results in a much larger groove weld at the joint, and therefore potentially more severe problems with brittle heat affected zones and lamellar defects in the column. Indeed, a significant percentage of connections of this configuration have failed to produce the desired amount of plastic rotation. One of the issues that must be faced by designers utilizing cover plated connections is the sequence of operations used to attach the cover plate and beam flange to the column. In one approach, the bottom cover plate is shop welded to the column, and then used as the backing for the weld of the beam bottom flange to the column flange. This approach has the advantage of providing an erection seat and also results in a somewhat reduced amount of field welding for this joint. A second approach is to attach the cover plate to the beam flange, and then weld it to the column, in the field, as an integral part of the beam flange. There are tradeoffs to both approaches. The latter approach results in a relatively large field weld at the bottom flange with large heat input required into the column and beam. If this operation is not performed with proper preheat and control of the heat input, it can potentially result in an enlarged and brittle heat affected zone in both members. The first approach results in reduced heat input and therefore, somewhat minimized potential for this effect. However, proper control of preheat and heat input remains as important in either case, as improper procedures can still result in brittle conditions in the heat affected zone. Further, the detail in which the cover plate is shop welded to the column can lead to a notch effect for the column flange at the seam between the beam

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flange and cover plate. This is effect is illustrated in Figure 7.9.1-2. At least one specimen employing this detail developed a premature fracture across the column flange that has been related to this notch effect. This effect has been confirmed by recent fracture mechanics modeling of this condition conducted by Deierlein. When developing cover plated connection details, designers should attempt to minimize the total thickness of beam flange and cover plate, so as to reduce the size of the complete joint penetration weld of these combined elements to the column flange. For some frame configurations and member sizes, this combined thickness and the resulting CJP weld size can approach or even exceed the thickness of the column flange. While there is no specific criteria in the AWS or AISC specifications that would suggest such weldments should not be made, judgementally they would not appear to be desirable from either a constructability or performance perspective. As a rough guideline, it is recommended that for connections in which both the beam flange and cover plate are welded to the column flange, the combined thickness of these elements should not exceed twice the thickness of the beam flange nor 100% of the thickness of the column flange. For cover plated connections in which only the cover plate is welded to the column flange, the same thickness limits should be applied to the cover plate.

seam acts as notch beam bottom flange column flange (in tension)

cover plate

Figure 7.9.1-2 Notch Effect at Cover Plated Connections 7.9.2 Flange Rib Connections

There are no modifications to the Guidelines or Commentary of Section 7.9.2 at this time.

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7.9.3 Bottom Haunch Connections

Figure 7.9.3-1 7-7 indicates the configuration of a connection with a haunch at the bottom beam flange.several potential configurations for single, haunched beam-column connections. As with the cover plated and ribbed connections, the intent is to shift the plastic hinge away from the column face and to reduce the demand on the CJP weld by increasing the depth of the section. To date, the configuration incorporating the triangular haunch has been subjected to limited testing. Testing of configurations incorporating the straight haunch are currently planned, but have not yet been performed. Several tests of this connection type were conducted by Uang under the SAC phase I project (Uang, 1995). Following that work, additional research on the feasibility of improving connection performance with welded haunches was conducted under a project that was jointly sponsored by NIST and AISC (NIST, 1998). That project was primarily focused on the problem of upgrading connections in existing buildings. As indicated in the report of that work, the haunched modification improves connection performance by altering the basic behavior of the connection. In essence, the haunch creates a prop type support, beneath the beam bottom flange. This both reduces the effective flexural stresses in the beam at the face of the support, and also greatly reduces the shear that must be transmitted to the column through the beam. A complete procedure for the design of this modification may be found in NIST, 1998. Figure 7-7 - Bottom Haunch Connection Modification

Figure 7.9.3-1 Bottom Haunch Connecction Two Nine tests are known to have been performed to date, both successfully all intended to replicate the condition of an existing connection that has been upgraded. Except for those specimens in which existing vulnerable welded joints were left in place at the top flange, these connections generally achieved large plastic rotations. Several dynamic tests have also been New Construction 7-44

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successfully conducted, although only moderate plastic deformation demands could be imposed due to limitations of the laboratory equipment. Both tests were conducted in a repair/modification configuration. In one test, a portion of the girder top flange, adjacent to the column, was replaced with a thicker plate. In addition, the bottom flange and haunch were both welded to the column. This specimen developed a plastic hinge within the beam span, outside the haunched area and behaved acceptably. A second specimen did not have a thickened top flange and the bottom girder flange was not welded to the column. Plastic behavior in this specimen occurred outside the haunch at the bottom flange and adjacent to the column face at the top flange. Failure initiated in the girder at the juncture between the top flange and web, possibly contributed to by buckling of the flange as well as lateral torsional buckling of the section. Fracture progressed slowly along the top fillet of the girder and eventually, traveled into the flange itself. Design Issues: The haunch can be attached to the girder in the shop, reducing field erection costs. Weld sizes are smaller than in cover plated connections. The top flange is free of obstructions. Quantitative Results: No. of specimens tested: 92 Girder Size: W30 x 99 Column Size: W14 x 176 Plastic Rotation achievedSpecimen 1 UCSD-1R:0.04 radian (w/o bottom flange weld and reinforced top flange) Specimen 2 UCSD-3R:0.05 radian (with bottom flange weld and reinforced top flange) Specimen UCSD-4R: 0.014 radian (dynamic- limited by test setup) Specimen UCSD-5R: 0.015 radian (dynamic- limited by test setup) Girder Size: W36x150 Column Size: W14x257 Plastic Rotation achieved Specimen UCB-RN2: 0.014 radian (no modification of top weld) Specimen UTA-1R: 0.019 radian (partial modification of top weld) Specimen UTA-1RB: 0.028 radian (modified top weld) Girder Size: W36x150 Column Size: W14x455 Plastic Rotation achievedSpecimen UTA-NSF4: 0.015 radian (no modification of top weld) Girder Size: W18x86 Column Size: W24x279 Plastic Rotation achievedSpecimen SFCCC-8: 0.035 radian (cover plated top flange) New Construction 7-45

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Performance is dependent on properly executed complete joint penetration welds at the column face. The joint can be subject to through-thickness flaws in the column flange; however, this connection may not be as sensitive to this potential problem because of the significant increase in the effective depth of the beam section which can be achieved. Welding of the bottom haunch requires overhead welding when relatively shallow haunches are used. The skewed groove welds of the haunch flanges to the girder and column flanges may be difficult to execute. The increased depth of the beam, resulting from the haunch may have undesirable impact on architectural design. Unless the top flange is prevented from buckling at the face of the column, performance may not be adequate. For configurations incorporating straight haunches, the haunch must be long, in order to adequately develop stress into the haunch, through the web. This tends to increase demands at the column face. Additional testing of all these configurations is recommended. 7.9.4 Top and Bottom Haunch Connections

There are no modifications to the Guidelines or Commentary of Section 7.9.4 at this time. 7.9.5 Side-Plate Connections

There are no modifications to the Guidelines or Commentary of Section 7.9.5 at this time. 7.9.6 Reduced Beam Section Connections

In this connection, the cross section of the beam is intentionally reduced within a segment, to produce an intended plastic hinge zone or fuse, located within the beam span, away from the column face. Several ways of performing this cross section reduction have been proposed. One method includes removal of a portion of the flanges, symmetrical about the beam centerline, in a so-called “dog bone” profile. Care should be taken with this approach to provide for smoothly contoured transitions to avoid the creation of stress risers which could initiate fracture. It has also been proposed to create the reduced section of beam by drilling a series of holes in the beam flanges. Figure 7-11 Figure 7.9.6-1 illustrates both concepts. The most successful configurations have used reduced sections formed with circular cuts. Configurations which taper the reduced section, through the use of unsymmetrical cut-outs, or variable size holes, to balance the cross section and the flexural demand have also been tested with success. Testing of this concept was first performed by a private party, and US patents were applied for and granted. These patents have now been released. Limited testing of both “dog-bone” and drilled hole configurations have been performed in Taiwan (Chen and Yeh - 1995). The American Institute of Steel Construction is currently performing additional tests of this configuration (Smith-Emery - 1995), however the full results of this testing are not yet available. has performed successful testing of 4 linearly tapered RBS connections. In the time since the first publication of the Interim Guidelines, a number of tests have been successfully conducted of RBS connections with circular curved cut-outs, including investigations and at the University of Texas at Austin, has successfully tested 4 circular curved RBS specimens. Others, including Popov at the New Construction 7-46

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University of California at Berkeley, and Texas A&M University., have also tested circular curved RBS connections with success. When this connection type was first proposed, There is a concern was expressed that the presence of a concrete slab at the beam top flange would tend to limit the effectiveness of the reduced section of that flange, particularly when loading places the top flange into compression. It may be possible to mitigate this effect with proper detailing of the slab. Limited testing of RBS specimens with composite slabs has recently been successfully conducted at Ecole Polytechnic, in Montreal, Canada. In these tests, shear studs were omitted from the portion of the top flange having a reduced section, in order to minimize the influence of the slab on flexural hinging. In addition, a 1 inch wide gap was placed in the slab, around the column, to reduce the influence of the slab on the connection at the column face. More recently, both the University of Texas at Austin and Texas A&M University have conducted successful tests of RBS connections with slabs and without such gaps present between the slab and column. This most recent testing suggests that the presence of the slab actually enhances connection behavior by retarding buckling of the top flange in compression and delaying strength degradation effects commonly observed in specimens tested without slabs. Design Issues: This connection type is potentially the most economical of the several types which have been suggested. The reliability of this connection type is dependent on the quality of the complete joint penetration weld of the beam to column flange, and the through-thickness behavior of the column flange. If the slab is not appropriately detailed, it may inhibit the intended “fuse” behavior of the reduced section beam segment. It is not clear at this time whether it would be necessary to use larger beams with this detail to attain the same overall system strength and stiffness obtained with other configurations. In limited testing conducted to date of the unsymmetrical “dog-bone” configuration (Smith-Emery - 1995), the plastic hinging which occurred at the reduced section was less prone to buckling of the flanges than in some of the other configurations which have been tested, due to the very compact nature of the flange in the region of the plastic hinge. However, the tendency for lateral-torsional buckling is significantly increased suggesting the need for lateral bracing of the beam flanges, near the reduced section. Experimental Results: A number of researchers have performed tests on RBS specimens to date. Most tests have utilized the ATC-24 loading protocol, which is similar to the protocol described in Section 7.4.1 of the Interim Guidelines. Testing employed at Ecole Polytechnic, in Montreal, Canada utilized a series of different testing protocols including the ATC-24 procedure and a dynamic excitation simulating the response of a connection in a building to an actual earthquake accelerogram (Tremblay, et. al., 1997). This research included two tests of connections with composite floor slabs. All of the reported tests with circular flange cuts have performed acceptably, however, the dynamic tests at Ecole Polytechnic only imposed 0.025 radians of plastic rotation on the assembly.

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Straight

Tapered

Circular Reduced Section

Drilled Constant

Drilled Tapered

Figure 7-11 7.9.6-1 - Reduced Beam Section Connection Quantitative Results: No. of specimens tested: 219 published (without slabs)2 Girder Size: W21 x 62W30 x 99 thru W 36 x 194 Column Size: W14x120W14 x 176 thru W 14 x 426, W24 x 229 Plastic Rotation achieved:- 0.03 radian Straight: - 0.02 radian Tapered - 0.027 - 0.045 radian Circular - 0.03 - 0.04 radian No. of specimens tested: 42published (with slabs) Girder Size: W21 x 44 to W36 x 150 Column Size: W14 x 90 to W14x257 Plastic Rotation achieved: 0.03-0.05 radians (ATC-24 loading protocol) 0.025 radians (earthquake simulation – limited by laboratory setup, no failure observed) 7.9.7 Slip - Friction Energy Dissipating Connection

There are no modifications to the Guidelines or Commentary of Section 7.9.7 at this time. 7.9.8 Column-Tree Connection

There are no modifications to the Guidelines or Commentary of Section 7.9.8 at this time. 7.9.9 Proprietary Slotted Web Connections

In the former prescriptive connection, in which the beam flanges were welded directly to the column flanges, beam flexural stress was transferred into the column web through the combined

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action of direct tension across the column flange, opposite the column web, and through flexure of the column flange. This stress transfer mechanism and its resulting beam flange prying moment results in a large stress concentration at the center of the beam flange, opposite the column web. Recent research (Allen, et. al. - 1995) indicates that the provision of continuity plates within the column panel zone reduces this stress concentration somewhat, but not completely. The intent of the proprietary slotted web connections is to further reduce this stress concentration and to achieve a uniform distribution of flexural stress across the beam flange at the connection, and also, to promote local buckling of the beam flanges under compressive loads to limit the amount of demand on the beam flange to column flange weld. Claimed assets for this connection include elimination of the vertical beam shear in the beam flange welds, elimination of the beam lateral torsional buckling mode, and the participation of the beam web in resisting its portion of the beam moment. A number of different configurations for this connection type have been developed and tested. Figure 7.9.9-17-14 indicates one such configuration for this connection type that has been successfully tested and which has been used in both new and retrofit steel moment-resisting frames. In this configuration, slots are cut into the beam web, extending from the weld access hole adjacent to the top and bottom flanges, and extending along the beam axis a sufficient length to alleviate the stress concentration effects at the beam flange to column flange weld. The beam web is welded to the column flange. vertical plates are placed between the column flanges, opposite the edges of the top and bottom beam flanges to stiffen the outstanding column flanges and draw flexural stress away from the center of the beam flange. Horizontal plates are placed between these vertical plates and the column web to transfer shear stresses to the panel zone. The web itself is softened with the cutting of a vertical slot in the column web, opposite the beam flange. High fidelity finite element models were utilized to confirm that a nearly uniform distribution of stress occurs across the beam flange.

Slot, typ. NOTICE OF CONFIDENTIAL INFORMATION: WARNING: The information presented in this figure is PROPRIETARY. US patents have been granted and Foreign Patents have been applied for. Use of this information is strictly prohibited except as authorized in writing by the developer. Violators shall be prosecuted in accordance with US and Foreign Patent Intellectual Property Laws.

Figure 7.9.9-17-14 - Proprietary Slotted Web Connection New Construction 7-49

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Design Issues: This detail is potentially quite economical, entailing somewhat more shop fabrication than the former prescriptive connection, but similar levels of field erection work. Contrary to the recommendations contained in these Interim Guidelines, this connection does not shift the location of plastic hinging away from the column face. However, two a number of connections employing details similar to that shown in Figure 7-147.9.9-1 have recently been tested successfully (Allen. - 1995). The connection detail is sensitive to the quality of welding employed in the critical welds, including those between the beam and column flanges., and between the vertical and horizontal plates and the column elements. It has been reported that one specimen, with a known defect in the beam flange to column flange weld was informally tested and failed at low levels of loading. The detail is also sensitive to the balance in stiffness of the various plates and flanges. For configurations other than those tested, detailed finite element analyses may be necessary to confirm that the desired uniform stress distribution is achieved. The developer of this detail indicates that for certain column profiles, it may be possible to omit the vertical slots in the column web and still achieve the desired uniform beam flange stress distribution. This detail may also be sensitive to the toughness of the column base metal at the region of the fillet between the web and flanges. In heavy shapes produced by some rolling processes the metal in this region may have substantially reduced toughness properties relative to the balance of the section. This condition, coupled with local stress concentrations induced by the slot in the web may have the potential to initiate premature fracture. The developer believes that it is essential to perform detailed analyses of the connection configuration, in order to avoid such problems. Popov tested one specimen incorporating a locally softened web, but without the vertical and horizontal stiffener plates contained in the detail shown in Figure 7-14. That specimen failed by brittle fracture through the column flange which progressed into the holes cut into the web. The stress patterns induced in that specimen, however, were significantly different than those which occur in the detail shown in the figure. Quantitative Results: Number of specimens tested: 2 Girder Size: W 27x94 Column Size: W 14x176 Plastic Rotation Achieved: Specimen 1: 0.025 radian Specimen 2: 0.030 radian Quantitative data on connection testing may be obtained from the licensor. 7.9.10 Bolted Bracket Connections

Framing connections employing bolted or riveted brackets have been used in structural steel construction since its inception. Early connections of this type were often quite flexible, and also had limited strength compared to the members they were connecting, resulting in partially restrained type framing. However, it is possible to construct heavy bolted brackets employing New Construction 7-50

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high strength bolts to develop fully restrained moment connections. Pretensioing of the bolts or threaded rods attaching the brackets to the column flanges and use of slip-critical connections between the brackets and beam flanges can help to provide the rigidity required to obtain fully restrained behavior. Reinforcement of the column flanges may be required to prevent local yielding and excessive deformation of these elements, as well. Two alternative configurations that have been tested recently are illustrated in Figure 7.9.10-1. The developer of these configurations offers the brackets in the form of proprietary steel castings. Several tests of these alternative connections have been performed on specimens with beams ranging in size from W16 to W36 sections and with large plastic rotations successfully achieved. Design Issues: The concept of bolted bracket connections is similar to that of the riveted “wind connections” commonly installed in steel frame buildings in the early twentieth century. The primary difference is that the riveted wind connections were typically limited in strength either by flexural yielding of outstanding flanges of the brackets, or shear and tension on the rivets, rather than by flexural hinging of the connected framing. Since the old-style wind connections could not typically develop the flexural strength of the girders and also could be quite flexible, they would be classified either as partial strength or partially restrained connections. Following the Northridge earthquake, the concept of designing such connections with high strength bolts and heavy plates, to behave as fully restrained connections, was developed and tested by a private party who has applied for patents on the concept of using steel castings for this purpose. Bracket

High tensile threaded rod

Pipe Plate

Bolts

WARNING: The information presented in this figure is PROPRIETARY. US and Foreign Patents have been applied for. Use of this information is strictly prohibited except as authorized in writing by the developer. Violators shall be prosecuted in accordance with US and Foreign Patent Intellectual Property Laws.

Figure 7.9.10-1 Bolted Bracket Connections Bolted connections offer a number of potential advantages over welded connections. Since no field welding is required for these connections, they are inherently less labor intensive during New Construction 7-51

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erection, and also less dependent on the technique of individual welders for successful performance. However, quality assurance should be provided for installation and tensioning of the bolts, as well as correction of any problems with fit-up due to fabrication tolerances. Experimental Results: A series of tests on several different configurations of proprietary heavy bolted bracket connections have been performed at Lehigh University (Ksai & Bleiman, 1996) to qualify these connections for use in repair and modification applications. To test repair applications, brackets were placed only on the bottom beam flange to simulate installations on a connection where the bottom flange weld in the original connection had failed. In these specimens, bottom flange welds were not installed, to approximate the condition of a fully fractured weld. The top flange welds of these specimens were made with electrodes rated for notch toughness, to preclude premature failure of the specimens at the top flange. For specimens in which brackets were placed at both the top and bottom beam flanges, both welds were omitted. Acceptable plastic rotations were achieved for each of the specimens tested. Quantitative Results: No. of specimens tested: 8 Girder Size: W16x40 and W36x150 Column Size: W12x65 and W14x425 Plastic Rotation achieved - 0.05 radians - 0.07 radians 7.10 Other Types of Welded Connection Structures There are no modifications to the Guidelines or Commentary of Section 7.10 at this time. 7.10.1 Eccentrically Braced Frames (EBF)

There are no modifications to the Guidelines or Commentary of Section 7.10.1 at this time. 7.10.2 Dual Systems

There are no modifications to the Guidelines or Commentary of Section 7.10.2 at this time. 7.10.3 Welded Base Plate Details

There are no modifications to the Guidelines or Commentary of Section 7.10.3 at this time. 7.10.4 Vierendeel Truss Systems

There are no modifications to the Guidelines or Commentary of Section 7.10.4 at this time. 7.10.5 Moment Frame Tubular Systems

There are no modifications to the Guidelines or Commentary of Section 7.10.5 at this time.

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7.10.6 Welded Connections of Collectors, Ties and Diaphragm Chords

There are no modifications to the Guidelines or Commentary of Section 7.10.6 at this time. 7.10.7 Welded Column Splices

There are no modifications to the Guidelines or Commentary of Section 7.10.7 at this time. 7.10.8 Built-up Moment Frame Members

There are no modifications to the Guidelines or Commentary of Section 7.10.8 at this time.

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8. METALLURGY & WELDING 8.1 Parent Materials 8.1.1 Steels

Designers should specify materials which are readily available for building construction and which will provide suitable ductility and weldability for seismic applications. Structural steels which may be used in the lateral-force-resisting systems for structures designed for seismic resistance without special qualification include those contained in Table 8.1.1-1. Refer to the applicable ASTM reference standard for detailed information. Table 8.1.1-1 - Structural Steel Prequalified for Use in Seismic Lateral-Force-Resisting Systems ASTM Specification ASTM A36 ASTM A283 Grade D ASTM A500 (Grades B & C) ASTM A501 ASTM A572 (Grades 42 & 50) ASTM A588 ASTM A9921 Notes: 1- See Commentary

Description Carbon Structural Steel Low and Intermediate Tensile Strength Carbon Steel Plates Cold-Formed Welded & Seamless Carbon Steel Structural Tubing in Rounds & Shapes Hot-Formed Welded & Seamless Carbon Steel Structural Tubing High-Strength Low-Alloy Columbium-Vanadium Steels of Structural Quality High-Strength Low-Alloy Structural Steel (weathering steel) Steel for Structural Shapes for Use in Building Framing

Structural steels which may be used in the lateral-force-resisting systems of structures designed for seismic resistance with special permission of the building official are those listed in Table 8.1.1-2. Steel meeting these specifications has not been demonstrated to have adequate weldability or ductility for general purpose application in seismic-force-resisting systems, although it may well possess such characteristics. In order to demonstrate the acceptability of these materials for such use in WSMF construction it is recommended that connections be qualified by test, in accordance with the guidelines of Chapter 7. The test specimens should be fabricated out of the steel using those welding procedures proposed for use in the actual work. Table 8.1.1-2 - Non-prequalified Structural Steel ASTM Specification Description ASTM A242 High-Strength Low-Alloy Structural Steel ASTM A709 Structural Steel for Bridges ASTM A913 High-Strength Low-Alloy Steel Shapes of Structural Quality, Produced by Quenching & Self-Tempering Process Metallurgy & Welding 8-1

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Commentary: Many WSMF structures designed in the last 10 years incorporated ASTM A36 steel for the beams and ASTM A572 grade 50 steel for the columns. This provided an economical way to design structures for the strong column weak beam provisions contained in the building code. Recent studies conducted by the Structural Shape Producers Council (SSPC), however, indicate that material produced to the A36 specification has wide variation in strength properties with actual yield strengths that often exceed 50 ksi. This wide variation makes prediction of connection and frame behavior difficult. Some have postulated that one of the contributing causes to damage experienced in the Northridge earthquake was inadvertent pairing of overly strong beams with average strength columns. The AISC and SSPC have been working for several years to develop a new specification for structural steel that would have both minimum and maximum yield values defined and provide for a margin between maximum yield and minimum ultimate tensile stress. AISC recently submitted such a specification, for a material with 50 ksi specified yield strength, to ASTM for development into a standard specification. ASTM formally adopted the new specification for structural shapes, with a yield strength of 50 ksi, under designation A992 in 1998 and It is anticipated that domestic mills will begin have begun producing structural wide flange shapes to this specification. within a few years and that eventually, this new material will replace A36 as the standard structural material for incorporation into lateral-force-resisting systems. Since the formal approval of the A992 specification by ASTM occurred after publication of the 1997 editions of the building codes and the AISC Seismic Specification, it is not listed in any of these documents as a prequalified material for use in lateral force resisting systems. Neither is it listed as prequalified in AWS D1.1-98. However, all steel that complies with the ASTM-992 specification will also meet the requirements of ASTM A572, Grade 50 and should therefore be permissible for any application for which the A572 material is approved. See also, the commentary to Section 8.2.2. Under certain circumstances it may be desirable to specify steels that are not recognized under the UBC for use in lateral-force-resisting systems. For instance, ASTM A709 might be specified if the designer wanted to place limits on toughness for fracture-critical applications. In addition, designers may wish to begin incorporating ASTM A913, Grade 65 steel, as well as other higher strength materials, into projects, in order to again be able to economically design for strong column - weak beam conditions. Designers should be aware, however, that these alternative steel materials may not be readily available. It is also important when using such non-prequalified steel materials, that precautions be taken to ensure adequate weldability of the material and that it has sufficient ductility to perform under the severe loadings produced by earthquakes. The Metallurgy & Welding 8-2

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cyclic test program recommended by these Interim Guidelines for qualification of connection designs, by test, is believed to be an adequate approach to qualify alternative steel material for such use as well. Note that ASTM A709 steel, although not listed in the building code as prequalified for use in lateral-force-resisting systems, actually meets all of the requirements for ASTM A36 and ASTM A572. Consequently, special qualification of the use of this steel should not be required. Although the 1994 editions of the Uniform Building Code and the NEHRP Provisions do not prequalify the use of ASTM A913 steel in lateral force resisting systems, the pending 1997 edition of the UBC does prequalify its use. Both the 1997 NEHRP Provisions and the AISC Seismic Provisions prequalify the use of this steel in elements that do not undergo significant yielding, for example, the columns of moment-resisting frames designed to meet strong column - weak beam criteria. Consequently, special approval of the Building Official should no longer be required as a pre-condition of the use of material conforming to this specification, at least for columns. 8.1.2 Chemistry

There are no modifications to the Guidelines of Section 8.1.2 at this time. Commentary: Some concern has been expressed with respect to the movement in the steel producing industry of utilizing more recycled steel in its processes. This results in added trace elements not limited by current specifications. Although these have not been shown quantitatively to be detrimental to the performance of welding on the above steels, a the new A992specification for structural steel proposed by AISC does place more control on these trace elements. Mill test reports now include elements not limited in some or all of the specifications. They include copper, columbium, chromium, nickel, molybdenum, silicon and vanadium. The analysis and reporting of an expanded set of elements should be possible, and could be beneficial in the preparation of welding procedure specifications (WPSs) by the welding engineer if critical welding parameters are required. Modern spectrographs used by the mills are capable of automated analyses. When required by the engineer, a request for special supplemental requests should be noted in the contract documents. 8.1.3 Tensile/Elongation Properties

Mechanical property test specimens are taken from rolled shapes or plates at the rolling mill in the manner and location prescribed by ASTM A6 and ASTM A370. Table 8-3 Table 8.1.3-1 gives the basic mechanical requirements for commonly used structural steels. Properties specified, and controlled by the mills, in current practice include minimum yield strength or yield point, ultimate

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tensile strength and minimum elongation. However, there can be considerable variability in the actual properties of steel meeting these specifications. SSPC, in cooperation with SEAOC, has collected statistical data on the strength characteristics of two grades (ASTM A36 and ASTM A572 Grade 50) of structural steels, based on mill test reports from selected domestic producers for the 1992 production year. Data were also collected for "Dual Grade" material that was certified by the producers as complying with both ASTM A36 and ASTM A572 Grade 50. Table 8-4 Table 8.1.3-2 summarizes these results as well as data provided by a single producer for ASTM A913 material. Unless special precautions are taken to limit the actual strength of material incorporated into the work to defined levels, new material specified as ASTM A36 should be assumed to be the dual grade for connection demand calculations, whenever the assumption of a higher strength will result in a more conservative design condition. Table 8-3 Table 8.1.3-1 - Typical Tensile Requirements for Structural Shapes Minimum Yield Ultimate Tensile Minimum Elongation Minimum Elongation Strength or Yield Strength, Ksi % % Point, Ksi in 2 inches in 8 inches A36 36 Min. 58-801 212 20 4 A242 42 Min.. 63 MIN. 213 18 A572, Gr. 42 42 Min. 60 Min. 24 20 A572, GR50 50 Min. 65 MIN. 212 18 A588 50 Min. 70 MIN. 213 18 A709, GR36 36 Min. 58-80 212 20 A709, GR50 50 Min. 65 MIN. 21 18 A913, GR50 50 Min. 65 MIN. 21 18 A913, GR65 65 Min. 80 MIN. 17 15 A992 50 Min. – 65 Max. 65 MIN 21 18 Notes: 1No maximum for shapes greater than 426 lb./ft. 2Minimum is 19% for shapes greater than 426 lb. /ft. 3No limit for Shape Groups 1, 2 and 3.Minimum is 18% for shapes greater than 426 lb./ft. 4. Minimum is 50 ksi for Shape Groups 1 and 2, 46 ksi for Shape Group 3, and 42 ksi for Shape Groups 4 and 5. ASTM

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Table 8-4 Table 8.1.3-2 - Statistics for Structural Shapes1,2 Statistic

A 36

Dual GRADE

A572 GR50

A913 GR65

Mean Minimum Maximum Standard Deviation [ s ] Mean + 1 s

Yield Point (ksi) 49.2 36.0 72.4 4.9 54.1

55.2 50.0 71.1 3.7 58.9

57.6 50.0 79.5 5.1 62.7

75.3 68.2 84.1 4.0 79.3

Mean Minimum Maximum Standard Deviation [ s ] Mean + 1 s

Tensile Strength (ksi) 68.5 58.0 88.5 4.6 73.1

73.2 65.0 80.0 3.3 76.5

75.6 65.0 104.0 6.2 81.8

89.7 83.4 99.6 3.5 93.2

Mean Minimum Maximum Standard Deviation [ s ] Mean + 1 s Mean - 1 s

Yield/Tensile Ratio 0.72 0.51 0.93 0.06 0.78 0.66

0.75 0.65 0.92 0.04 0.79 0.71

0.76 0.62 0.95 0.05 0.81 0.71

0.84 0.75 0.90 0.03 0.87 0.81

1: The data presented for ASTM A36, “Dual Grade” and ASTM A572 Grade 50 were included as part of the SSPC study (SSPC-1994). The data for ASTM A913 were derived from a single producer and may not be available from all producers. 2. Statistical Data on the distribution of strength properties for material meeting ASTM A992 are not presently available. Pending the development of such statistics, it should be assumed that A992 material will have similar properties to ASTM A572, Gr. 50 material.

Commentary: The data given in Table 8-4 Table 8.1.3-2 for A36 and A572 Grade 50 is somewhat weighted by the lighter, Group 1 shapes that will not ordinarily be used in WSMF applications. Excluding Group 1 shapes and combining the Dual Grade and A572 Grade 50 data results in a mean yield strength of 48 ksi for A36 and 57 ksi for A572 Grade 50 steel. It should also be noted that approximately 50% of the material actually incorporated in a project will have yield strengths that exceed these mean values. For the design of facilities with stringent requirements for limiting post-earthquake damage, consideration of more conservative estimates of the actual yield strength may be warranted. Until recently, In wide flange sections the tensile test coupons in wide flange sections are currently were taken from the web. The amount of reduction rolling, finish rolling temperatures and cooling conditions affect the tensile and impact Metallurgy & Welding 8-5

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properties in different areas of the member. Typically, the web exhibits about five percent higher strength than the flanges due to faster cooling. In 1998 ASTM A6 was revised to specify that coupons be taken from the flange of wide flange shapes. Design professionals should be aware of the variation in actual properties permitted by the ASTM specifications. This is especially important for yield strength. Yield strengths for ASTM A36 material have consistently increased over the last 15 years so that several grades of steel may have the same properties or reversed properties, with respect to beams and columns, from those the designer intended. Investigations of structures damaged by the Northridge earthquake found some WSMF connections in which beam yield strength exceeded column yield strength despite the opposite intent of the designer. As an example of the variations which can be found, Table 8-5 Table 8.1.3-2 presents the variation in material properties found within a single building affected by the Northridge earthquake. Properties shown include measured yield strength (Fya,), measured tensile strength (Fua ) and Charpy V-Notch energy rating (CVN). Table 8-5Table 8.1.3-2 - Sample Steel Properties from a Building Affected by the Northridge Earthquake Shape

Fya1 ksi

Fua, ksi

CVN, ft-lb.

W36 X 182

38.0

69.3

18

W36 X 230

49.3

71.7

195

Note 1 - ASTM A36 material was specified for both structures.

The practice of dual certification of A36 and A572, Grade 50 can result in mean yield strengths that are fifty percent higher than the specified yield of A36. Since there is no practical way to discern whether dual grade steel will be supplied, unless direct purchase of steel from specific suppliers is made, in the absence of such procurement practices, the prudent action for determining connection requirements, where higher strengths could be detrimental to the design, would be to assume the dual grade material whenever A36 or A572 Grade 50 is specified. In the period since the initial publication of the Interim Guidelines, several researchers and engineers engaged in connection assembly prototype testing have reported that tensile tests on coupons extracted from steel members used in the prototype tests resulted in lower yield strength than reported on the mill test report furnished with the material, and in a few cases lower yield strength than would be permitted by the applicable ASTM specification. This led to some confusion and concern, as to how mill test reports should be interpreted. Metallurgy & Welding 8-6

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The variation of the measured yield strength of coupons reported by researchers engaged in connection prototype testing, as compared to that indicated on the mill test reports, is not unusual and should be expected. These variations are the result of a series of factors including inconsistencies between the testing procedures employed as well as normal variation in the material itself. The following paragraphs describe the basis for the strengths reported by producers on mill certificates, as well as the factors that could cause independent investigators to determine different strengths for the same material. Mill tests of mechanical properties of steel are performed in accordance with the requirements of ASTM specifications A6 and A370. ASTM A6 had historically required that test specimens for rolled W shapes be taken from the webs of the shapes, but recently was revised to require testing from the flanges of wide flange shapes with 6 inch or wider flanges. A minimum of two tests must be made for each heat of steel, although additional tests are required if shapes of significantly different thickness are cast from the same heat. Coupon size and shape is specified based on the thickness of the material. The size of the coupon used to test material strength can effect the indicated value. Under ASTM A6, material that is between 3/4 inches thick and 4 inches thick can either be tested in full thickness “straps” or in smaller 1/2” diameter round specimens. In thick material, the yield strength will vary through the thickness, as a result of cooling rate effects. The material at the core of the section cools most slowly, has larger grain size and consequently lower strength. If full-thickness specimens are used, as is the practice in most mills, the recorded yield strength will be an average of the relatively stronger material at the edges of the thickness and the lower yield material at the center. Many independent laboratories will use the smaller 1/2” round specimens, and sometimes even sub-sized 1/4” round specimens for tensile testing, due to limitations of their testing equipment. Use of these smaller specimens for thick material will result in testing only of the lower yield strength material at the center of the thickness. ASTM A370 specifies the actual protocol for tensile testing including the loading rate and method of reporting test data. Strain rate can affect the strength and elongation values obtained for material. High strain rates result in elevated strength and reduced ductility. Under ASTM A370, yield values may be determined using any convenient strain rate, but not more than 1/16 inch per inch, per minute which corresponds to a maximum loading rate of approximately 30 ksi per second. Once the yield value is determined, continued testing to obtain ultimate tensile values can proceed at a more rapid rate, not to exceed 1/2 inch per inch per minute. Under ASTM A370, there are two different ways in which the yield property for structural steel can be measured and reported. These include yield point and yield strength. These are illustrated in Figure 8.1.3-1. The yield point is the peak Metallurgy & Welding 8-7

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stress that occurs at the limit of the elastic range, while the yield strength is a somewhat lower value, typically measured at a specified offset or elongation under load. Although a number of methods are available to determine yield point, the so-called “drop of the beam” method is most commonly used for structural steel. In this method the load at which a momentary drop-off in applied loading occurs is recorded, and then converted to units of stress to obtain the yield point. Yield strength may also be determined by several methods, but is most commonly determined using the offset method. In this method, the stress strain diagram for the test is drawn, as indicated in Figure 8.1.3-1. A specified offset, typically 0.2% strain for structural steel, is laid off on the abscissa of the curve and a line is drawn from this offset, parallel to the slope of the elastic portion of the test. The stress at the intersection of this offset line with the stressstrain curve is taken as the yield strength.

σ

Yield Point Yield Strength

ε

Offset

Figure 8.1.3-1 Typical Stress - Strain Curve for Structural Steel The material specifications for structural steels typically specify minimum values for yield point but do not control yield strength. The SSPC has reported that actual practice among the mills varies, with some mills reporting yield strength and others reporting yield point. This practice is permissible as yield strength will always be a somewhat lower value than yield point, resulting in a somewhat conservative demonstration that the material meets specified requirements. However, this does mean that there is inconsistency between the values reported by the various mills on certification reports. Similarly, the procedures followed by independent testing laboratories may be different than those followed by the mill, particularly with regard to strain rate and the location at which a coupon is obtained. Metallurgy & Welding 8-8

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Under ASTM A6, coupons for tensile tests had historically been obtained from the webs of structural shapes. However, most engineers and researchers engaged in connection testing have preferred to extract material specimens from the flanges of the shape, since this is more representative of the flexural strength of the section. Coupons removed from the web of a rolled shape tend to exhibit somewhat higher strength properties than do coupons removed from the flanges, due to the extra amount of working the thinner web material typically experiences during the rolling process and also because the thinner material cools more rapidly after rolling, resulting in finer grain size. Given these differences in testing practice, as well as the normal variation that can occur along the length of an individual member and between different members rolled from the same heat, the reported differences in strength obtained by independent laboratories, as compared to that reported on the mill test reports, should not be surprising. It is worth noting that following the recognition of these differences in testing procedure, the SSPC in coordination with AISC and ASTM developed and proposed a revision to the A6 specification to require test specimens to be taken from the flanges of rolled shapes when the flanges are 6 inches or more wide. It is anticipated that mills will begin to alter practice to conform to a revised specification in early 1997 This has since become the standard practice. The discovery of the somewhat varied practice for reporting material strength calls into question both the validity of statistics on the yield strength of structural steel obtained from the SSPC study, and its relevance to the determination of the expected strength of the material for use in design calculations. Although the yield point is the quantity controlled by the ASTM material specifications, it has little relevance to the plastic moment capacity of a beam section. Plastic section capacity is more closely related to the stress along the lower yield plateau of the typical stress-strain curve for structural steel. This strength may often be somewhat lower than that determined by the offset drop-of-the-beam method. Since the database of material test reports on which the SSPC study was based appears to contain test data based on both the offset and drop-of-the-beam methods, it is difficult to place great significance in the statistics derived from it and to draw a direct parallel between this data and the expected flexural strength of rolled shapes. It would appear that the statistics reported in the SSPC study provide estimates of the probable material strength that are somewhat high. Thus, the recommended design strengths presented in Tables 6.6.6.3-1 and 7.5.11 of the Interim Guidelines would appear to be conservative with regard to design of welds, panel zones and other elements with demands limited by the beam yield strength. Under the phase II program of investigation, SAC, together with the shape producers, is engaged in additional study of the statistical distribution of yield strength of various materials produced by the mills. This study is intended to provide an improved understanding of the statistical distribution of the lower Metallurgy & Welding 8-9

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yield plateau strength of material extracted from section flanges, measured in a consistent manner. In addition, it will provide correlation with yield strengths determined by other methods such that the data provided on mill test certificates can be properly interpreted and utilized. In addition, the possibility of revising the ASTM specifications to provide for more consistent reporting of strength data as well as the reporting of strength statistics that are directly useful in the design process will be evaluated. In the interim period, the data reported in Table 8-1.32, extracted from the SSPC study, remain the best currently available information. 8.1.4 Toughness Properties

There are no modifications to the Guidelines or Commentary of Section 8.1.4 at this time. 8.1.5 Lamellar Discontinuities

There are no modifications to the Guidelines or Commentary of Section 8.1.5 at this time. 8.1.6 K-Area Fractures

Recently, there have beenIn the period 1995-96 there were several reports of fractures initiating in the webs of column sections during the fabrication process, as flange continuity plates and/or doubler plates were welded into the sections. This fracturing typically initiated in the region near the fillet between the flange and web. This region has been commonly termed the “k-area” because the AISC Manual of Steel Construction indicates the dimension of the fillet between the web and flange with the symbol “k”. The k-area may be considered to extend from mid-point of the radius of the fillet into the web, approximately 1 to 1-1/2 inches beyond the point of tangency between the fillet and web. The fractures typically extended into, and sometimes across, the webs of the columns in a characteristic “half-moon” or “smiley face” pattern. Investigations of materials extracted from fractured members have indicated that the material in this region of the shapes had elevated yield strength, high yield/tensile ratio, high hardness and very low toughness, on the order of a few foot-pounds at 70oF. Material with these properties can behave in a brittle manner. Fracture can be induced by thermal stresses from the welding process or by subsequent weld shrinkage, as apparently occurred in the reported cases. There have been no reported cases of inservice k-line fracture from externally applied loading, as in beam-column connections, although such a possibility is perceived to exist under large inelastic demand. It appears that this local embrittling of sections can be attributed to the rotary straightening process used by some mills to bring the rolled shapes within the permissible tolerances under ASTM A6. The straightening process results in local cold working of the sections, which strain hardens the material. The amount of cold working that occurs depends on the initial straightness of the section and consequently, the extent that mechanical properties are effected is likely to vary along the length of a member. The actual process used to straighten the section can also affect the amount of local cold working that occurs.

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Engineers can reduce the potential for weld-induced fracture in the k-area by avoiding welding within the k-area region. This can be accomplished by detailing doubler plates and continuity plates such that they do not contact the section in this region. The use of large corner clips on beam flange continuity plates can permit this. Selection of column sections with thicker webs, to eliminate the need for doubler plates; the use of fillet welds rather than full penetration groove welds to attach doubler plates to columns, when acceptable for stress transfer; and detailing of column web doubler plates such that they are offset from the face of the column web can also help to avoid these fabrication-induced fracture problems. Commentary: It appears that detailing and fabrication practice can be adjusted to reduce the potential for k-area fracture during fabrication. However, the acceptability of having low-toughness material in the k-area region for service is a question that remains. It is not clear at this time what percentage of the material incorporated in projects is adversely affected, or even if a problem with regard to serviceability exists. SAC recently placed a public call, asking for reports of fabrication-induced fractures at the k-area, but only received limited response. However, in one of the projects that did report this problem, a significant number of columns were affected. This may have been contributed to by the detailing and fabrication practices applied on that project. Other than detailing structures to minimize the use of doubler plates, and to avoid large weldments in the potentially sensitive k-area of the shape, it is not clear at this time, what approach, if any, engineers should take with regard to this issue. There are several methods available to identify possible low notch toughness in structural carbon steels, including Charpy V-Notch testing and hardness testing of samples extracted from the members. However, both of these approaches are quite costly for application as a routine measure on projects and the need for such measures has not yet been established. Following publication of advisories on the k-line problem by AISC, and the publication of similar advisory information in FEMA-267a,reports on this problem diminished. It is not clear whether this is due to revised detailing practice on the part of engineers and fabricators, revised mill rolling practice, or a combination of both. SAC, AISC and SSPC are continuing to research this issue in order to identify if a significant problem exists, and if it does, to determine its basic causes, and to develop appropriate recommendations for mill, design, detailing, and fabrication practices to mitigate the problem. 8.2 Welding 8.2.1 Welding Process

There are no modifications to the Guidelines or Commentary of Section 8.2.1 at this time.

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8.2.2 Welding Procedures

Welding should be performed within the parameters established by the electrode manufacturer and the Welding Procedure Specification (WPS), required under AWS D1.1. Commentary: A welding procedure specification identifies all the important parameters for making a welded joint including the material specifications of the base and filler metals, joint geometry, welding process, requirements for pre- and post-weld heat treatment, welding position, electrical characteristics, voltage, amperage, and travel speed. Two types of welding procedure specifications are recognized by AWS D1.1. These are prequalified procedures and qualified-bytest procedures. Prequalified procedures are those for which the important parameters are specified within the D1.1 specification. If a prequalified procedure is to be used for a joint, all of the variables for the joint must fall within the limits indicated in the D1.1 specification for the specific procedure. If one or more variables are outside the limits specified for the prequalified procedures, then the fabricator must demonstrate the adequacy of the proposed procedure through a series of tests and submit documentation (procedure qualification records) demonstrating that acceptable properties were obtained. Regardless of whether or not a prequalified or qualified-by-test procedure is employed, the fabricator should prepare a welding procedure specification, which should be submitted to the engineer of record for review and be maintained at the work location for reference by the welders and inspectors. The following information is presented to help the engineer understand some of the issues surrounding the parameters controlled by the welding procedure specification. For example, the position (if applicable), electrode diameter, amperage or wire feed speed range, voltage range, travel speed range and electrode stickout (e.g. all passes, 0.072 in. diameter, 248 to 302 amps, 19 to 23 volts, 6 to 10 inches/minute travel speed, 170 to 245 inches/minute wire feed speed, 1/2" to 1" electrode stickout) should be established. This information is generally submitted by the fabricator as part of the Welding Procedure Specification. Its importance in producing a high quality weld is essential. The following information is presented to help the engineer understand some of the issues surrounding these parameters. The amperage, voltage, travel speed, electrical stickout and wire feed speed are functions of each electrode. If prequalified WPSs are utilized, these parameters must be in compliance with the AWS D1.1 requirements. For FCAW and SMAW, the parameters required for an individual electrode vary from manufacturer to manufacturer. Therefore, for these processes, it is essential that the fabricator/erector utilize parameters that are within the range of recommended operation published by the filler metal manufacturer. Alternately, the fabricator/erector could qualify the welding procedure by test in accordance

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with the provisions of AWS D1.1 and base the WPS parameters on the test results. For submerged arc welding, the AWS D1.1 code provides specific amperage limitations since the solid steel electrodes used by this process operate essentially the same regardless of manufacture. The filler metal manufacturer’s guideline should supply data on amperage or wire feed speed, voltage, polarity, and electrical stickout. The guidelines will not, however, include information on travel speed which is a function of the joint detail. The contractor should select a balanced combination of parameters, including travel speed, that will ensure that the code mandated weld-bead sizes (width and height) are not exceeded. Recently, ASTM approved a new material specification for structural steel shape, ASTM A992. This specification is very similar to the ASTM A572, Grade 50 specification except that it includes additional limitations on yield and tensile strengths and chemical composition. Although material conforming to A992 is expected to have very similar welding characteristics to A572 material, it was adopted too late to be included as a prequalified base material in AWS D1.1-98. Although the D1 committee has evaluated A992 and has taken measures to incorporate it as a prequalified material in AWS D1.1-2000, technically, under AWS D1.1-98, welded joints made with this material should follow qualified-bytest procedures. In reality, structural steel conforming to ASTM A992 may actually have somewhat better weldability than material conforming to the A572 specification. This is because A992 includes limits on carbon equivalent, precluding the delivery of steels where all alloys simultaneously approach the maximum specified limits. Therefore, it should be permissible to utilize prequalified procedures for joint with base metal conforming to this specification. 8.2.3 Welding Filler Metals

There are no modifications to the Guidelines of Section 8.2.3 at this time. Commentary: Currently, there are no notch toughness requirements for weld metal used in welding ASTM A 36 or A 572, Grade 50, steel in AWS D1.1. This topic has been extensively discussed by the Welding Group at the Joint SAC/AISC/AISI/NIST Invitational Workshop on September 8 and 9, 1994, and by all participants of the SAC Invitational Workshop on October 28 and 29, 1994. The topic was also considered by the AWS Presidential Task Group, which decided that additional research was required to determine the need for toughness in weld metal. There is general agreement that adding a toughness requirement for filler metal would be desirable and easily achievable. Most filler metals are fairly tough, but some will not achieve even a modest requirement such as 5 ft-lb. at + 70? F. What is not in unanimous agreement is what level of toughness should be required. The recommendation from the Joint Workshop was Metallurgy & Welding 8-13

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20 ft-lb. at -20? F per Charpy V-Notch [CVN] testing. The recommendation from the SAC Workshop was 20 ft-lb. at 30? F lower than the Lowest Ambient Service Temperature (LAST) and not above 0? F. The AWS Presidential Task Group provided an interim recommendation for different toughness values depending on the climatic zone, referenced to ASTM A709. Specifically, the recommendation was for 20 ft-lb. at temperatures of 70 degrees F for Zone 1, 40 degrees F for Zone 2, and 10 degrees F for Zone 3. The AWS also suggested toughness values for base metals used in these applications. Some fractured surfaces in the Northridge and Kobe Earthquakes revealed evidence of improper use of electrodes and welding procedures. Prominent among the misuses were high production deposition rates. Pass widths of up to 11/2 inches and pass heights of 1/2 inch were common. The kind of heat input associated with such large passes promotes grain growth in the HAZ and attendant low notch toughness. In evaluation of welds in buildings affected by the Northridge earthquake, the parameters found to be most likely to result in damage-susceptible welds included root gap, access capability, electrode diameter, stick-out, pass thickness, pass width, travel speed, wire feed rate, current and voltage were found to be the significant problems in evaluation of welds in buildings affected by the Northridge earthquake. Welding electrodes for common welding processes include: AWS A5.20: AWS A5.29: AWS A5.1: AWS A5.5: AWS A5.17: AWS A5.23: AWS A5.25:

Carbon Steel Electrodes for FCAW Low Alloy Steel Electrodes for FCAW Carbon Steel Electrodes for SMAW Low Alloy Steel Covered Arc Welding Electrodes (for SMAW) Carbon Steel Electrodes and Fluxes for SAW Low Alloy Steel Electrodes and Fluxes for SAW Carbon and Low Alloy Steel Electrodes and Fluxes for Electroslag Welding

In flux cored arc welding, one would expect the use of electrodes that meet either AWS A5.20 or AWS A5.29 provided they meet the toughness requirements specified below. Except to the extent that one requires Charpy V-Notch toughness and minimum yield strength, the filler metal classification is typically selected by the Fabricator. Compatibility between different filler metals must be confirmed by the Fabricator, particularly when SMAW and FCAW-SS processes are mixed. Generally speaking, SMAW-type filler metals may not be applied to FCAW-SS type filler metals (e.g. when a weld has been partially removed) while FCAW-type filler metals may be applied to SMAW-type filler metals. This recommendation considers the use of aluminum as a killing agent in FCAW-SS electrodes that can

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be incorporated into the SMAW filler metal with a reduction in impact toughness properties. As an aid to the engineer, the following interpretation of filler metal classifications is provided below: E1X2X3T4X5 For electrodes specified under AWS A5.20 (e.g. E71T1) 1 2 3 4 5 6 E X X T X X For electrodes specified under AWS A5.29 (e.g. E70TGK2) E1XX7X8X9X10 For electrodes specified under AWS A5.1 or AWS A5.5. (e.g. E7018) NOTES: 1.

Indicates an electrode.

2.

Indicates minimum tensile strength of deposited weld metal (in tens of ksi, e.g., 7 = 70 ksi).

3.

Indicates primary welding position for which the electrode is designed (0 = flat and horizontal and 1 = all positions).

4.

Indicates a flux cored electrode. Absence of a letter indicates a "stick" electrode for SMAW.

5.

Describes usability and performance capabilities. For our purposes, it conveys whether or not Charpy V-Notch toughness is required (1, 5, 6 and 8 have impact strength requirements while 2, 4, 7, 10 and 11 do not). A "G" signifies that the properties are not defined by AWS and are to be agreed upon between the manufacturer and the specifier. Impact strength is specified in terms of the number of foot-pounds at a given temperature (e.g., 20 ft-lb. at 0 degrees F). Note that for electrodes specified under AWS A5.20, the format for usage is "T-X".

6.

Designates the chemical composition of deposited metal for electrodes specified under AWS A5.29. Note that there is no equivalent format for chemical composition for electrodes specified under AWS A5.20.

7.

The first two digits (or three digits in a five digit number) designate the minimum tensile strength in ksi.

8.

The third digit (or fourth digit in a five digit number) indicates the primary welding position for which the electrode is designed (1 = all positions, 2 = flat position and fillet welds in the horizontal position, 4 = vertical welding with downward progression and for other positions.)

9.

The last two digits, taken together, indicate the type of current with which the electrode can be used and the type of covering on the electrode. Metallurgy & Welding 8-15

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Indicates a suffix (e.g., A1, A2, B1, etc.) designating the chemical composition of the deposited metal.

Electrode Diameter: (See AWS D1.1 Section 4.14.1.2) The issue of maximum electrode diameter has not been studied sufficiently to determine whether or not electrode diameter is a critical variable. Recent tests have produced modified frame joints with acceptable test results using the previous standard-of-practice 0.120 in. diameter wire. The use of smaller diameter electrodes will slow the rate of deposition (as measured by volume) but will not, in and of itself, produce an acceptable weld. The following lists the maximum allowable electrode diameters for prequalified FCAW WPS’s according to D1.1: • • • • •

Horizontal, complete or partial penetration welds: 1/8 inch (0.125")* Vertical, complete or partial penetration welds: 5/64 inch (0.078") Horizontal, fillet welds: 1/8 inch (0.125") Vertical, fillet welds: 5/64 inch (0.078") Overhead, reinforcing fillet welds: 5/64 inch (0.078") * This value is not part of D1.1-94, but will be part of D1.1-96.

For a given electrode diameter, there is an optimum range of weld bead sizes that may be deposited. Weld bead sizes that are outside the acceptable size range (either too large or too small) may result in unacceptable weld quality. The D1.1 code controls both maximum electrode diameters and maximum bead sizes (width and thickness). Prequalified WPS’s are required to meet these code requirements. Further restrictions on suitable electrode diameters are not recommended. Low-hydrogen electrodes. Low hydrogen electrodes should be used to minimize the risk of hydrogen assisted cracking (HAC) when conditions of high restraint and the potential for high hardness microstructures exist. Hydrogen assisted cracking can occur in the heat affected zone or weld metal whenever sufficient concentrations of diffusible hydrogen and sufficient stresses are present along with a hard microstructure at a temperature between 100 C and –100 C. Hydrogen is soluble in steel at high temperatures and is introduced into the weld pool from a variety of sources including but not limited to: moisture from coating or core ingredients, drawing lubricants, hydrogenous compounds on the base material, and moisture from the atmosphere. At the present time, the term “low hydrogen” is not well defined by AWS. The degree of hydrogen control required to reduce the risk of hydrogen assisted cracking will depend on the material being welded, level of restraint, preheat/interpass temperature, and heat input level. When a controlled level of diffusible hydrogen is required, electrodes can be purchased with a supplemental designator that indicates a diffusible hydrogen concentration below 16, 8, or 4 ml

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H2/100g in the weld metal can be maintained (H16, H8, and H4 respectively) under most welding conditions . The diffusible hydrogen potential (measured in ml/100g deposited weld metal) will depend on the type of consumable, welding process, plate/joint cleanliness, and atmospheric conditions in the area of welding. Some consumables may absorb moisture after exposure to the atmosphere. Depending on the type of consumable, this may result in a significant increase in the weld metal diffusible hydrogen concentration. In situations where control of diffusible hydrogen concentrations is important, the manufacturer should be consulted for advice on proper storage and handling conditions required to limit moisture absorption. Hydrogen assisted cracking may be avoided through the selection and maintenance of an adequate preheat /interpass temperature and/or minimum heat input. Depending on the type of steel and restraint level, a trade-off between an economic preheat/interpass temperature and the diffusible hydrogen potential of a given process exists. There have been several empirical approaches developed to determine safe preheat levels for a given application that include consideration of carbon equivalent, restraint level, electrode type, and preheat. When followed, the guidelines for preheat that have been established in AWS D1.1 and D1.5 are generally sufficient to reduce the risk of hydrogen assisted cracking in most mild steel weldments. Hydrogen assisted cracking will typically occur up to 72 hours after completion of welding. For the strength of materials currently used in moment frame construction, inspection of completed welds should be conducted no sooner than 24 hours following weld completion. 8.2.4 Preheat and Interpass Temperatures

There are no modifications to the Guidelines or Commentary of Section 8.2.4 at this time. 8.2.5 Postheat

There are no modifications to the Guidelines or Commentary of Section 8.2.5 at this time. 8.2.6 Controlled Cooling

There are no modifications to the Guidelines or Commentary of Section 8.2.6 at this time. 8.2.7 Metallurgical Stress Risers

There are no modifications to the Guidelines or Commentary of Section 8.2.7 at this time.

Metallurgy & Welding 8-17

Interim Guidelines Advisory No. 2

SAC 99-01

8.2.8 Welding Preparation & Fit-up

There are no modifications to the Guidelines or Commentary of Section 8.2.8 at this time.

Metallurgy & Welding 8-18