DG04 - Extended End-Plate Moment Connections

Steel Design Guide Series

ExtendedMoment End-Plate Connections

Steel Design Guide Series

Extended End-Plate Moment Connections Design Guide for Extended End-Plate Moment Connections Thomas M. Murray, PhD, RE. Montague-Betts Professor of Structural Steel Design Virginia Polytechnic Institute and State University Blacksburg, Virginia

A M E R I C A N

I N S T I T U T E

OF

S T E E L

C O N S T R U C T I O N

© 2003 by American Institute of Steel Construction, Inc. All rights reserved. This publication or any part thereof must not be reproduced in any form without permission of the publisher.

Copyright  1990 by American Institute of Steel Construction, Inc.

All rights reserved. This book or any part thereof must not be reproduced in any form without the written permission of the publisher. The information presented in this publication has been prepared in accordance with recognized engineering principles and is for general information only. While it is believed to be accurate, this information should not be used or relied upon for any specific application without competent professional examination and verification of its accuracy, suitablility, and applicability by a licensed professional engineer, designer, or architect. The publication of the material contained herein is not intended as a representation or warranty on the part of the American Institute of Steel Construction or of any other person named herein, that this information is suitable for any general or particular use or of freedom from infringement of any patent or patents. Anyone making use of this information assumes all liability arising from such use. Caution must be exercised when relying upon other specifications and codes developed by other bodies and incorporated by reference herein since such material may be modified or amended from time to time subsequent to the printing of this edition. The Institute bears no responsibility for such material other than to refer to it and incorporate it by reference at the time of the initial publication of this edition. Printed in the United States of America Second Printing: October 2003

© 2003 by American Institute of Steel Construction, Inc. All rights reserved. This publication or any part thereof must not be reproduced in any form without permission of the publisher.

TABLE OF CONTENTS 1. I N T R O D U C T I O N . . . . . . . . . . . . . . . . . . . . . . . . . . . 1.1 Background . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 1.2 Overview of Design Guide . . . . . . . . . . . . . . . . . . 1.3 Brief Literature O v e r v i e w . . . . . . . . . . . . . . . . . . .

1 1

2 2

2. RECOMMENDED DESIGN PROCEDURES . . . 5 2.1 Basis of Design Recommendations . . . . . . . . . . . 5 2.2 Limit States Check L i s t . . . . . . . . . . . . . . . . . . . . 6 3. UNSTIFFENED, EXTENDED END-PLATE CONNECTION D E S I G N . . . . . . . . . . . . . . . . . . . . . 3.1 The Four-Bolt Configuration Design Procedures and E x a m p l e s . . . . . . . . . . . . . . . . . . . 3.1.1 Design Procedures . . . . . . . . . . . . . . . . . . . . 3.1.2 Allowable Stress Design E x a m p l e s . . . . . . . 3.1.3 Load and Resistance Design Example . . . . 3.2 Eight-Bolt Design Procedures and Allowable Stress Design Example . . . . . . . . . . . . . . . . . . . . .

7

7 7

8 11 14

4. STIFFENED, EXTENDED END-PLATE CONNECTION DESIGN . . . . . . . . . . . . . . . . . . . . 17 4.1 Design P r o c e d u r e s . . . . . . . . . . . . . . . . . . . . . . . . . 17

4.2 Design Examples . . . . . . . . . . . . . . . . . . . . . . . . . 18 4.2.1 Allowable Stress Design Examples . . . . . . 18 4.2.2 Load and Resistance Factor Design Examples . . . . . . . . . . . . . . . . . . . . . . . . . . . 21

BIBLIOGRAPHY . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 25 APPENDIX A—ASD NOMENCLATURE, DESIGN AIDS AND QUICK REFERENCE EXAMPLES . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . A.1 ASD Nomenclature . . . . . . . . . . . . . . . . . . . . . . . A.2 ASD Design Aids . . . . . . . . . . . . . . . . . . . . . . . . A.3 ASD Quick Reference Examples . . . . . . . . . . . .

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31 32 34

APPENDIX B—LRFD NOMENCLATURE, DESIGN AIDS AND QUICK REFERENCE EXAMPLES . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 39 B.1 LRFD Nomenclature . . . . . . . . . . . . . . . . . . . . . . 39 B.2 LRFD Design Aids . . . . . . . . . . . . . . . . . . . . . . . 40 B.3 LRFD Quick Reference Examples . . . . . . . . . . . 41

© 2003 by American Institute of Steel Construction, Inc. All rights reserved. This publication or any part thereof must not be reproduced in any form without permission of the publisher.

PREFACE

This booklet was prepared under the direction of the Committee on Research of the American Institute of Steel Construction, Inc. as part of a series of publications on special topics related to fabricated structural steel. Its purpose is to serve as a supplemental reference to the AISC Manual of Steel Construction to assist practicing engineers engaged in building design. The design guidelines suggested by the authors that are outside the scope of the AISC Specifications or Code do not represent an official position of the Institute and are not intended to exclude other design methods and procedures. It is recognized that the design of structures is within the scope of expertise of a competent licensed structural engineer, architect or other licensed professional for the application of principles to a particular structure. The sponsorship of this publication by the American Iron

and Steel Institute is gratefully acknowledged.

The information presented in this publication has been prepared in accordance with recognized engineering principles and is for general information only. While it is believed to be accurate, this information should not be used or relied upon for any specific application without competent professional examination and verification of its accuracy, suitability, and applicability by a licensed professional engineer, designer or architect. The publication of the material contained herein is not intended as a representation or warranty on the part of the American Institute of Steel Construction, Inc. or the American Iron and Steel Institute, or of any other person named herein, that this information is suitable for any general or particular use or of freedom infringement of any patent or patents. Anyone making use of this information assumes all liability arising from such use.

© 2003 by American Institute of Steel Construction, Inc. All rights reserved. This publication or any part thereof must not be reproduced in any form without permission of the publisher.

Chapter 1 INTRODUCTION 1.1 BACKGROUND

figuration are found in the 9th edition AISC Manual of Steel Construction (1989a). As with any connection, end-plate connections have certain advantages and disadvantages. The principal advantages are: (a) The connection is suitable for winter erection in that only field bolting is required. (b) All welding is done in the shop, eliminating field welding associated problems. (c) Without the need for field welding, the erection process is relatively fast. (d) If fabrication is accurate, it is easy to maintain plumbness of the frame. (e) Lower total installed cost for many cases. The principal disadvantages are: (a) The fabrication techniques are somewhat more stringent because of the need for accurate beam length and "squareness" of the beam end. (b) Column out-of-squareness can cause erection difficulties but can be controlled by fabricating the beams ¼ in. to in. short and providing "finger" shims. (c) End plates often warp due to the heat of welding.

The use of moment end-plate connections in multi-story, moment resistant frame construction is becoming more com-

mon because of advancements in design methods and fabrication techniques, both of which have resulted in decreased costs. A typical moment end-plate connection is composed of a steel plate welded to the end of a beam section with attachment to an adjacent member using rows of fully tensioned high-strength bolts. The connection may be between two beams (splice plate connection) or between a beam and a column. End-plate moment connections are classified as either flush or extended with or without stiffeners and further classified depending on the number of bolts at the tension flange. A flush connection is detailed such that the endplate does not appreciably extend beyond the beam flanges and all bolts are located between the beam flanges. An extended end-plate is one which extends beyond the tension flange a sufficient distance to allow the location of bolts other than between the beam flanges. Extended end-plates may be used with or without a stiffener between the end-plate and the beam flange in the plane of the beam web. Flush end-plate connections are typically used in frames subject to light lateral loadings or near inflection points of gable frames. Extended end-plates are used for beam-to-column moment connections. Only extended end-plates are considered in this design guide. Four extended end-plate configurations are shown in Fig. 1.1. The four-bolt unstiffened configuration shown in Fig. 1.1(a) is probably the most commonly used in multi-story frame construction. An allowable stress design (ASD) procedure for this connection is found in the 8th and 9th editions, American Institute of Steel Construction (AISC) Manual of Steel Construction (1980, 1989a) and a load and resistance factor design (LRFD) procedure is found in the AISC Load and Resistance Factor Design Manual of Steel Construction (1986a). Assuming the full beam moment capacity is to be resisted, A325 bolts and a maximum bolt diameter of 1½ in. (maximum practical size because of tightening considerations), this connection is limited because of bolt capacity to use with less than one-half of the available beam sections. The connection strength can be increased by adding a stiffener, Fig. 1.1(b), or increasing the number of bolts per row to four, Fig. 1.1(c). Formal design procedures are not available for the former, and the latter requires a wide column flange. The stiffened A325 eight-bolt connection shown in Fig. 1.1(d) is capable of developing the full moment capacity of most of the available beam sections even if bolt diameter is limited to 1½ in. Design procedures for this con-

Fig. 1.1. Extended end-plate configurations.

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"tee-stub" analogy. All of these methods resulted in design procedures which predict a high degree of "prying action" resulting in large end-plate thicknesses and large bolt diameters. One such method for the four-bolt, extended configuration (Fig. 1.1(a)) is found in the 7th edition of the AISC Manual of Steel Construction (1969). More recently, methods based on refined yield-line analyses have been suggested. A number of configurations have been studied in Europe (Zoetermeijer, 1974, 1981; Packer and Morris, 1977; Mann and Morris, 1979) as well as in the United States (Srouji, 1983; Hendrick et al., 1985; Morrison, 1986). Most of this work has involved flush endplate configurations. Finite element methodology for the analysis of end-plates was first developed by Krishnamurthy (1978, 1981). His exhaustive analytical study of four-bolt, unstiffened, extended end plates (Fig. 1.1(a)), along with a series of experimental investigations, led to the development of a design procedure first published in the 8th edition of the AISC Manual of Steel Construction. More recently, Ahuja (1982) and Ghassemieh (1983) have investigated the stiffened configuration with two rows of two bolts on each side of the tension flange (Fig 1.1(d)). They used regression analysis to develop design equations. Murray and Kukreti (1988) have developed a simplified design procedure using their regression results which appears in the 9th edition AISC Manual of Steel Construction. Bolt Design. Early end-plate design procedures (Douty and McGuire, 1965; Nair et al., 1969; Kato and McGuire, 1973) all involved the calculation of bolt prying forces based on various assumptions. The assumed location of the prying force was at or near the edge of the end-plate. Packer and Morris (1977), Phillips and Packer (1981), Mann and Morris (1979), and Zoetermeijer (1974, 1981) have all included prying action forces in their yield-line based design procedures. The various recommendations range from rather complicated analytical procedures to a simple increase in bolt force over the applied tension (Mann and Morris, 1979). Krishnamurthy (1978a) argues that even though prying action is present, it is overly conservative to assume it to be acting at the edge of the plate as this normally results in thicker than necessary end-plates. His studies describe prying force as a pressure bulb which is formed under the bolt head due to the tensioning of the bolt and shifts towards the edge as the beam flange force increases. For any given loading, the pressure bulb is located somewhere between the edge of the end plate and the bolt head. He states, for service load conditions when the beam flange loads are small, the pressure bulb is closer to the bolt head than to the plate edge, and the plate moments are much smaller than those predicted by prying force formulas. Consequently, in his design procedure for four-bolt, extended, unstiffened end plates (Fig. 1.1(a)), prying forces are ignored, that is, the bolt size is determined directly from the force delivered by the beam flange.

(d) End-plates are subject to lamellar tearing in the region of the top flange tension weld. (e) The bolts are in tension, which can result in prying forces. A number of designers and fabricators in the United States have successfully used moment end-plate connections for building frames up to 30 stories in height. It is believed that, in spite of the several disadvantages, moment end-plate connections can provide economic solutions for rigid frame construction. Because very little research has been conducted on the low cycle fatigue strength of end-plate connections, their use is not presently recommended in areas of high seismic activity.

1.2 OVERVIEW OF DESIGN GUIDE The intent of this guide is to present complete design procedures and examples for extended moment end-plate connections suitable for fully restrained (or continuous frame) construction. Chapter 2 presents the basic design procedures for the end-plate configurations shown in Figs, 1.1(a), (c) and (d). Chapter 3 contains ASD and LRFD design examples for the four-bolt unstiffened configuration shown in Fig. 1.1 (a) and the eight-bolt unstiffened configuration shown in Fig. 1.1(c). Chapter 4 contains ASD and LRFD examples for the eight-bolt stiffened configuration shown in Fig. 1.1 (d). Appendix A includes allowable stress design (ASD) nomenclature, several design aids and quick reference examples. Appendix B is similar to Appendix A except it is for load and resistance factor design (LRFD). The quick reference examples serve as a guide for designers who are thoroughly familiar with moment end-plate design. The following section is a brief review of available literature for background purposes.

1.3 BRIEF LITERATURE OVERVIEW End Plate Design. Research starting in the early 1950s and continuing to the present has resulted in refined design procedures for both flush and extended end-plate connections. The earlier design methods were based on statics and simple assumptions concerning prying forces. These methods resulted in thick end-plates and large diameter bolts. Other studies have been based on yield-line theory. The more recent studies have used the finite element method and regression analysis to develop design equations. Accurate solutions can be developed using the latter technique; however, the procedure is time consuming and the resulting design equations usually involve terms to odd powers which virtually eliminates "structural feel" from the design. Early attempts (prior to about 1975) to develop design criteria for moment end-plate connections were based on the

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the beam flange through the column flange and fillet. If the stress at this critical section exceeds the yield stress of the column material, a column web stiffener is required opposite the beam tension and compression flanges. For the case of end-plate moment connections, the width of the stress pattern at the critical section may be considerably wider due to the insertion of the end plate into the load path. Hendrick and Murray (1983) conducted a number of column compression region tests using both stiffened and unstiffened end plates and concluded that the slope of the stress path through the end plate can be taken as 1:1 and that in the column as 3:1. This recommendation is also found in Hendrick and Murray (1984) and in the AISC LRFD manual (1986a). Hendrick's recommendations, except for the 3:1 slope, are also found in AISC Engineering for Steel Construction (1984), where 2½:1 is used. Newlin and Chen (1971) recommend that an interaction equation be used to check combined web yielding strength and web buckling. Possibly anticipating resistance to such form, they also provided a simple check for web buckling. This latter provision was adopted by AISC in their 1978 specification revision.

Kennedy et al. (1981) have presented a method for calculation of prying forces as a function of plate "thickness" relative to applied load. They identified three types of end-plate behavior. The first type is characterized by the absence of plastic hinges in the end plate. These end-plates are said to be "thick." Under low loading conditions all end plates fall into this category. The upper limit of this behavior occurs at a load which causes flexural yielding in the end-plate at the beam flange. Once this load is exceeded, a plastic hinge is formed at the flange and the end-plate is said to be of "intermediate" thickness. As the load is increased, a second plastic hinge forms at the bolt lines. At this load, the

end-plate is considered to be a "thin" plate. Further, they consider bolt force to be the sum of a portion of the flange force plus prying force and identify three stages of prying action corresponding to the three phases of end-plate behavior. For "thick" plates, the prying force is assumed to be zero. When the end plate is considered as "thin," the prying force is at its maximum. For "intermediate" plates, the prying force is somewhere between zero and the maximum value. They suggest that for ideal design, the end-plate should be "thick" under service loads, "intermediate" under factored loads and function as a "thin" plate at ultimate loads. Srouji (1983), Hendrick et al. (1985) and Morrison (1986) have modified the Kennedy et al. (1981) approach for use with two- and four-bolt flush end plates; four-bolt, stiffened extended end plates; and extended end plates with multiple bolt rows below the tension flange. Each researcher has presented experimental evidence to verify the prediction equations. Ahuja (1982) and Ghassemieh (1983) have presented finite element/regression analysis equations to predict bolt forces above the pretension level for eight-bolt, stiffened, extended end plates. Ahuja's results are based on elastic material properties, but Ghassemieh's results include inelastic material properties. Both authors limit the use of their results to A36 steel and A325 bolts. Beam-to-End-Plate Weld Design. Griffiths (1984) suggests that either full penetration welds or fillet welds sufficient to develop the beam flange in tension be used to connect the end plate to the beam. This recommendation holds even if the full capacity of the beam is not being utilized because of the large local deformations that occur along the end plate. Column Side Design. Relative to end-plate research, the amount of effort devoted to the column side of end-plate moment connections is quite limited. Only a few papers have been published which suggest design guidelines for the three column side failure modes: column web yielding, column web buckling and column flange bending failure. The critical section for column web yielding is at the toe of the column web fillet. For design of welded connections, the present AISC Manual (1989a) criteria is based on a load path which is assumed to vary linearly on a 2½:1 slope from

Witteveen et al. (1982) found three modes of failure for bending of the column flange. The first mode prevails when the column flange is thick when compared with bolt diameter. The second failure mode is when the stiffnesses of the bolts and flange are such that prying forces can develop because yield lines form in the flange near the fillet, causing both the flange and the bolts to fail. The third failure mode occurs when yield lines form in the flange near both the bolts and the fillet. Design procedures for each failure mode are presented as well as test results to verify the analytical work.

Mann and Morris (1979) present complete design procedures for the column side of end-plate connections. The recommendations are based primarily on the work of Packer and Morris (1977). However, only the case when the column flange is much less stiff than the end plate is considered. Three possible failure modes were found to exist. If the flange is very stiff, there are no prying forces and the failure occurs when the bolts rupture. The second failure mode occurs when the column flange is less stiff, which results in a combination of bolt fracture and flange yielding near the column web. The third failure mode is characterized by yield lines forming and causing double curvature in the flange plate. Provisions to estimate the column flange capacity for each of the failure modes are provided. If the first failure mode governs, the total bolt force is equal to the applied flange force. For the second failure mode, prying forces are accounted for by limiting bolt capacity to 80% of tensile capacity. Mann and Morris do not provide methods to estimate prying forces if the third failure mode governs. Granstrom (1980) extended tee-hanger results to include column flanges. The procedure to determine the required column flange thickness is the same as that used for tee-hanger

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flange thickness except that an effective column flange length is used. Granstrom does not consider prying action effects. Hendrick and Murray (1983) conducted a limited series of tests to evaluate the methods suggested by Mann and Morris (1979), Granstrom (1980) and Witteveen et al. (1982) for use with North American rolled sections. They concluded that the method proposed by Mann and Morris (1979) is the most suitable for the evaluation of unstiffened column flanges in the tension region of four-bolt, unstiffened end-plate connections. They also modified the Krishnamurthy (1978a) procedure for end plates by introducing an effective column flange length to obtain the same results as found with the Mann and Morris equations. Finally, they developed the "rule of thumb" found in the AISC Engineering for Steel Construction manual (1984) which states that, under certain limitations, the column flange is adequate if its thickness is greater than the required bolt diameter from the Krishnamurthy end-plate design procedure. All of his work applies only to A36 steel. Curtis (1985) has proposed design rules for column flange strength in the tension region of eight-bolt, stiffened endplate connections. His method is based on the Ghassemieh

(1983) end-plate design procedure with an effective column flange length and is therefore limited to A36 steel. Curtis and Murray (1989) have modified both the Hendrick and Murray (1983) and Curtis (1985) recommendations to ensure adequate column flange stiffness for use in fully restrained (continuous) construction. Procedures for the design of column web stiffeners to prevent web yielding or buckling have been suggested by Hendrick and Murray (1984) and have the same form as for welded beam-to-column connections in the 1989 AISC ASD Specification. Mann and Morris (1979) have presented methods to estimate the resistance of column flanges stiffened using standard web stiffeners. Zoetemeijer (1974) and Moore and Sims (1986) have recommended the use of "flange washer plate stiffeners." They have also provided design rules for the fourbolt unstiffened end-plate configuration. Curtis (1985) reported extensive analytical (yield-line) and experimental work on washer flange stiffening at both four-bolt unstiffened and eight-bolt stiffened, extended end plates. Some of the literature cited was used to develop the design procedures presented in the following chapter.

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Chapter 2 RECOMMENDED DESIGN PROCEDURES 2.1 BASIS OF DESIGN RECOMMENDATIONS

(RCSC, 1985). This Commentary states: "Connections of the type... in which some of the bolts lose a part of their clamping force due to applied tension suffer no overall loss of frictional resistance. The bolt tension produced by the moment is coupled with a compensating compressive force on the other side of the axis of bending." Thus, the frictional resistance of the connection remains unchanged. If very high shear forces exist, a bearing type connection may be necessary. In this case, the tension bolts must be designed with a shear-tension interaction equation. It is noted that shear is rarely a major concern in the design of moment end-plate connections. 6. It is assumed that the width of the end plate which is effective in resisting the applied beam moment is not greater than the beam flange width plus 1 in. This assumption is based on engineering judgment and is not part of any of the referenced end-plate design procedures. Further, the writer is unaware of any endplate connection tests conducted with end-plates substantially greater in width than the connected beam flange. 7. The gage of the tension bolts (horizontal distance between vertical bolt lines) should not exceed the beam tension flange width, again based on engineering judgment. 8. Beam web to end-plate welds in the vicinity of the tension bolts are designed to develop 0.6 of the beam web. This weld strength is recommended even if the full moment capacity of the beam is not required for frame strength. 9. Only the web to end-plate weld between the mid-depth of the beam and the inside side face of the beam compression flange or between the inner row of tension bolts plus two bolt diameters and the inside face of the beam compression flange, whichever is smaller, may be used to resist the beam shear. This assumption is based on the author's opinion. Literature was not found to substantiate or contradict this assumption. Column web stiffeners are expensive to fabricate and can interfere with weak axis column framing. Therefore, it is recommended that they be avoided whenever possible. If the

The recommended design procedures in Chapter 3 for the four- and eight-bolt unstiffened end-plate configurations, Figs, 1.1(a) and (c), are based on the work of Krishnamurthy (1978a), "A Fresh Look at Bolted End-Plate Behavior and Design," and the procedures in the ASD and LRFD AISC manuals (1980, 1986a, 1989a). Column side design for the four-bolt configuration is based on the work of Hendrick and Murray (1984), "Column Web Compression Strength at EndPlate Connections," and Curtis and Murray (1989), "Column Flange Strength at Moment End-Plate Connections." The eight-bolt stiffened end plate, Fig. 1.1(d), design procedures in Chapter 4 are based on the works of Ghassemeih (1983), "Inelastic Finite Element Analysis of Stiffened EndPlate Moment Connections," and Murray and Kukreti (1988), "Design of 8-bolt Stiffened Moment End Plates," and the procedures in the 9th edition ASD AISC Manual of Steel Construction (1989a). Column side design procedures for this configuration are based on the previously cited works of Hendrick and Murray (1984) and Curtis and Murray (1989). In addition, the following assumptions or conditions are inherent to the design procedures: 1. All bolts are tightened to a tension not less than that given in the AISC ASD and LRFD specifications. 2. The design procedures for the 8-bolt, stiffened configuration (Fig. 1.1(d)) are valid for use with A325 bolts. A490 bolts should not be used in this configuration. 3. Only static loading is permitted. Temperature, wind and snow loadings are considered static loadings (AISC, 1986, 1989). The design procedures should not be used, pending further research, when seismic loading is a major design consideration. 4. The smallest possible bolt pitch (distance from face of beam flange to centerline of nearer bolt) generally results in the most economical connection. The recommended minimum pitch dimension is bolt diameter plus ½ in. However, many fabricators prefer to use a standard pitch dimension, usually 2 in., for all bolt diameters. 5. End-plate connections can be designed to resist shear force at the interface of the end-plate and column flange using either "slip critical" or "bearing" assumptions. If slip critical (type "SC") criteria are used, all bolts at the interface can be assumed to resist the shear force and shear/tension interaction can be ignored as explained in the Commentary on "Specification for Structural Joints Using ASTM A325 or A490 Bolts"

need for a stiffener is marginal, it may be more economical to increase the column size rather than install stiffeners. If column web stiffeners are required because of inadequate column flange bending strength or stiffness, increasing the effective length of the column flange may eliminate the need for stiffening. This can be accomplished by increasing the

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tension bolt pitch or by switching from a two row configuration, Figs, 1.1(a), (b) or (c), to a four row configuration, Fig. 1.1(d). Alternately, column flange washer plates (loose plates with holes, placed on the column flange opposite the end-plate and connected with the end-plate connection tension bolts) may be used. This approach is widely used in Europe (Mann and Morris, 1979; Zoetemeijer, 1981; Moore and Sims, 1986) and has been studied in the United States (Curtis, 1985), but final design recommendations have not been formulated at this writing.

4. Failure of bolt, or slip of bolt in slip critical connections, due to shear at the interface between the end

plate and column flange. 5. Plate bearing failure of end-plate or column flange at bolts. 6. Rupture of beam tension flange to end-plate welds or beam web tension region to end-plate welds. 7. Shear yielding of beam web to end-plate weld or of beam web base metal. 8. Column web yielding opposite either the tension or compression flanges of the connected beam. 9. Column web buckling opposite the compression flange of the connected beam. 10. Column flange yielding in the vicinity of the tension bolts. As with flexural yielding of the end plate, this state in itself is not limiting but results in rapid increases in tension bolt forces and excessive rotation. 11. Column web stiffener failure due to yielding, local buckling or weld failure. 12. Column flange stiffener failure due to yielding or weld failure. 13. Excessive rotation (flexibility) at the connection due to end-plate and/or flange bending. 14. Column panel zone failure due to yielding or web plate buckling.

2.2 LIMIT STATES CHECK LIST Limit states (or failure modes) for moment end-plate beamto-column connections are: 1. Flexural yielding of the end-plate material near the tension flange bolts. This state in itself is not limiting, but yielding results in rapid increases in tension bolt forces and excessive rotation. 2. Shear yielding of the end-plate material. This limit state is not usually observed, but shear in combination with bending can result in reduced flexural capacity and stiffness. 3. Bolt rupture because of direct load and prying force effects. This limit state is obviously a brittle failure mode and is the most critical limit state in an endplate connection.

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Chapter 3 UNSTIFFENED, EXTENDED END-PLATE CONNECTION DESIGN 3.1 THE FOUR-BOLT CONFIGURATION DESIGN PROCEDURES AND EXAMPLES

The term was originally defined and values tabulated in the AISC ASD manual. The same values were printed in the AISC LRFD manual. However, to account for the differences in weak axis bending strength between the AISC ASD and LRFD specifications, the original values of must be increased by (0.90/0.75) = 1.20 for use in LRFD. Further, the values printed in both manuals are for cases where the end-plate and beam material have the same yield strengths, which is generally not the case except for A36 steel. Values of for various combinations of beam and end-plate material are found in Tables A.2 and A.3 for ASD use and in Tables B.2 and B.3 for LRFD use. Tables A.2 and B.2 are for A325 bolts and Tables A.3 and B.3 are for A490 bolts. Values of for hot-rolled beam sections are found in Table A.4.

3.1.1 Design procedures The following design recommendations for the extended, four-bolt, unstiffened, beam-to-column, end-plate connection shown in Fig. 3.1 are based on the works of Krishnamurthy (1978a), "A Fresh Look at Bolted End-Plate Behavior and Design"; Hendrick and Murray (1984), "Column Web Compression Strength at End-Plate Connections"; and Curtis and Murray (1989), "Column Flange Strength at Moment EndPlate Connections." The basic procedures for end-plate and bolt design are also found in the AISC ASD Manual of Steel Construction (1989a) and the LRFD Manual of Steel Construction (1986a). In Krishnamurthy's design procedure, prying action forces are considered negligible and the tension flange force is considered to be distributed equally to the four tension bolts. Possible local yielding of the tension flange and tension area of the web is neglected. The required end-plate thickness is determined using the tee-stub analogy with the effective critical moment in the end plate given by

in ASD

(3.1a)

in LRFD

(3.1b)

or

with unfactored beam flange force, kips factored beam flange force, kips a constant depending on the plate material yield stress, the bolt material and the design method (ASD or LRFD) beam flange width, in. effective end-plate width, in. (not more than

1 in.) 2 area of beam tension flange, in. 2 web area, clear of flanges, in. effective pitch, in.

distance from center line of bolt to nearer surface of the tension flange, in. + ½ in. is generally enough to provide wrench clearance; 2 in. is a common fabricator standard) fillet weld throat size or reinforcement of groove weld, in. nominal bolt diameter, in.

Fig. 3.1. Four-bolt unstiffened end-plate connection geometry.

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The required end-plate thickness, from

, is then determined in ASD

with column flange thickness, in. required column flange thickness, in.

(3.2a)

or in LRFD

The required column flange thickness is determined using Equation 3.2 modified as follows:

(3.2b)

with

in ASD the allowable bending stress for the end-plate material (0.75 times the specified yield stress), ksi specified yield stress of the end-plate material, ksi

in LRFD

(3.5b)

with

The column side limit states are to be checked as follows: 1. To prevent column web yielding at either the beam tension or compression flanges in ASD

(3.5a)

or

effective column flange length, in. 2.5c vertical spacing between rows of tension bolts, in.

(3.3a)

or in LRFD (3.3b) and or 3.1b with 1.0;

with factored beam flange force equal to times the beam flange force when the flange force is due to live and dead loads only, or by when the flange force is due to live and dead loads in conjunction with wind force, kips specified yield stress of column material, ksi column web thickness, in. beam flange thickness, in. distance from outer face of flange to web toe of column fillet, in. end-plate thickness, in. leg size of fillet weld or reinforcement weld, in. 1.0

with the column section

If the selected criterion is not satisfied, standard colstiffeners can be used to increase the flexural strength of the column flanges. 4. To prevent column web shear yielding within the connection, column web reinforcement is required if in ASD (3.6a)

or in LRFD (3.6b) with

connected beam end moments, ft-kips,

connected beam factored end moments, ftkips, and planar area of the column connection, in.2 In the above equations, the effect of column shear has been conservatively ignored. The following examples illustrate the above design procedures for four-bolt, unstiffened extended end-plate connections. Examples 3.1 and 3.2 use the ASD format and Example 3.3 uses the LRFD format. For these examples, the beam top flange is in tension and moment reversal is not a consideration.

(3.4a)

or

in LRFD

distance, in.

umn flange to web stiffeners or flange washer plate

If inequality 3.3 is not satisfied, column web stiffeners, capable of resisting a force equal to the difference between the left and right sides of the inequality, must be provided. 2. To prevent column web buckling at the beam compression flange in ASD

are calculated using Equations 3.1a or for ASD and 1.36 for LRFD; and

(3.4b)

with column web depth clear of fillets, in. 0.90 If inequality 3.4 is not satisfied, column web stiffeners are required at the beam compression flange. 3. To prevent column flange yielding in the tension region of the connection, the following must be satisfied assuming A36 material even if the column material yield stress is higher:

3.1.2 Allowable stress design examples EXAMPLE 3.1. Use ASD procedures to design a beam-tocolumn end-plate connection for a moment of 200 ft-kips and a shear of 40 kips. The beam is a W24x55 and the column is a W14x159. A36 steel is used for all members and

8

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plates. Bolts are ASTM A325. The end plate is to be shop

welded to the beam using E70XX electrodes. W24x55

Check bolt bearing on end plate since it is less thick than the column flange. Assume, conservatively, that the compression side bolts resist all of the shear.

W14x159

A. Bolt design, A325-SC bolts The beam tension flange force,

Check end-plate shear

is

End-Plate Selection The tension force per bolt, B, is then

From Table A.1, try diameter bolts (allowable capacity is 26.5 kips). Assuming A325-SC bolts, the single shear

C. Weld design, E70XX electrode

capacity from Table A.1 is 10.5 kips. The number of bolts

i. Beam flanges to end-plate welds: The flange weld must develop the force in the beam flange.

required to resist the applied shear is then

For E70XX electrodes the capacity of a fillet weld is

1-in. long

Bolt Selection Use A325-SC bolts fully tightened, 4 at the tension beam flange and 2 at the compression beam flange.

Use ½-in. fillet welds at both beam flanges. Note minimum weld size from the AISC ASD Specification is ¼ in., which

could be used at the beam compression flange if desired. B. End-plate design, A36 steel Try edge distance = 1¼ in. gage, g = 5½ in. pitch, Required end-plate width is 1¼ + 5½ + 1¼ = 8 in. Effective end-plate width must be less than beam flange width plus 1 in.

Determine

ii. Beam web to end-plate weld: Minimum size fillet weld is ¼ in. Required weld to develop the bending stress in the beam web near the tension bolts is

Use fillet weld both sides of beam web from inside face of beam flange to centerline of inside bolt holes plus two bolt diameters. The applied shear (40 kips) is to be resisted by weld

from Equation 3.1a:

between mid-depth of the beam and the inside face of the compression flange or between the inner row of tension bolts plus two bolt diameters and the inside face of the compression flange, whichever is minimum. By inspection the former governs for this example.

Determine

from Equation 3.2a:

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Use ¼-in. fillet weld (minimum size for ¾-in thick plate) both sides of beam web below tension bolt region.

column is a W14x90 A572 Gr50 steel. Only the column side limit states need to be checked. ASD procedures apply.

iii. Check beam web yielding

D. Check column side limit states and design stiffeners if necessary, A36 steel.

i. Check column web yielding using inequality 3.3a, 50 ksi:

i. Check column web yielding using inequality 3.3a:

Therefore, stiffeners are not required to prevent column web yielding. Therefore, stiffeners are not required opposite the beam tension and compression flanges to prevent column web yielding.

ii. Check column web buckling using inequality 3.4a:

Therefore, web stiffeners are not required opposite the beam compression flange to prevent column web buckling.

iii. Check column flange bending: The required column flange thickness is determined using Equation 3.2(a) with the previously discussed modifications.

Therefore, neither column web or column flange stiffeners are required for this design.

iv. Check column web yielding using inequality 3.6a, 50 ksi:

Therefore, column web reinforcement is not required. Final design details are shown in Fig. 3.2. EXAMPLE 3.2. Using the data, bolt design and end plate from Example 3.1, determine if stiffeners are required if the

Fig. 3.2. Final design details, Example 3.1.

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ii. Check column web buckling using inequality 3.4a,

=

Try 2PL ½ x 4 x 0'-7

50 ksi:

Use ¾-in. x ¾-in. clips to clear column web fillets.

Column flange to stiffener weld: Therefore, stiffeners are not required to prevent column web buckling.

Minimum weld is ¼ in. Use Column web to stiffener weld:

iii. Check column flange bending: From Example 3.1, = 8.14 in., and from similar calculations = 1.66 in., =1.326, = 57.29 in.-kips and = 1.25 in. Note that this check is made assuming the column material is A36 steel. Since is greater than 0.710 in., a stiffener is required opposite the beam tension flange. Because of the expense and possibility of interference with weak axis framing, the use of column web stiffeners is not recommended. Possible solutions for this example are: (1) to use an 8-bolt stiffened connection (Chapter 4) which increases the effective column flange length, (2) to increase the column flange thickness by using a heavier column or (3) to increase the bolt pitch which also increases the effective column flange length. If the third change is made, a thicker end-plate may be required. Obviously, the suggested changes require additional expense; however, the resulting connection may be more economical because column web stiffeners are eliminated. If changes are not practical, the following procedure can be used to determine stiffener size. Curtis and Murray (1989) do not provide recommendations for designing stiffeners when the column flange is inadequate. Assuming that only force in excess of what the unstiffened column flange can resist need be resisted by the stiffener, the capacity of the unstiffened column flange is first computed by rearranging Equation 3.2a and then 3.1a:

fillet weld both sides.

Minimum weld is To simplify detailing, use let weld both sides. Check shear stress in stiffener base metal.

fil-

Stiffener Selection Use 2PL ½x4x0'-7 with fillet welds all around. iv. Check column web yielding using inequality 3.6a, 50 ksi:

Therefore, column web reinforcement is not required. Final design details are shown in Fig. 3.3.

3.1.3 Load and resistance factor design example EXAMPLE 3.3. Using LRFD procedures, design a beamto-column end-plate connection for a factored moment of 260 ft-kips, an unfactored shear of 40 kips and a factored shear of 52 kips. The beam is a W24x55 and the column is a W14x90. A36 steel is to be used for all members and plates. Bolts are A325. The end plate is to be shop welded to the beam using E70XX electrodes.

Thus, the stiffeners will be designed for the unfactored beam flange force less the capacity of the column flange: With an allowable stress of Stiffeners do not need to be full depth of the column web if only one beam is connected to the column at a given elevation. Since the stiffener is in tension, local buckling is not a limit state and AISC ASD Specification width and thickness rules do not apply; however, good engineering practice requires the stiffener to be proportioned to match the beam flange.

A. Bolt design, A325-SC bolts The factored beam tension flange force,

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is

The factored tension force per bolt,

B. End-plate design, A36 steel Try edge distance = 1¼ in. gage = 5½ in. pitch,

is then

From Table B.1, try diameter bolts (design strength is 40.6 kips). Assuming A325-SC bolts, the single shear design strength from Table B.1 is 10.2 kips. The number of bolts required to resist the applied shear (unfactored) is then

Required end-plate width is 1¼ + 5½ + 1¼ = 8 in. Effective end-plate width must be less than beam flange width plus 1 in.

Bolt Selection

Determine

from Equation 3.1b:

Use A325-SC bolts fully tightened, 4 at the tension beam flange and 2 at the compression beam flange.

Determine

from Equation 3.2b:

Check bolt bearing on end-plate (note column flange thickness is larger and, conservatively, only the compression side bolts are considered).

in. Check end-plate shear:

End-Plate Selection

C. Weld design, E70XX electrode i. Beam flanges to end-plate welds: Flange weld must develop the force in beam flange. For E70XX electrodes the capacity of a 1-in. long fillet weld is

Use fillet welds at beam tension flange and minimum weld size at beam compression flange. From the AISC LRFD Specification minimum weld size is ¼ in.

Fig. 3.3. Final design details, Example 3.2.

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ii. Beam web to end-plate weld: Minimum size fillet weld is ¼ in. Required weld to develop maximum bending stress in web near tension bolts is

Use fillet weld both sides of beam web from inside face of beam flange to centerline of inside bolt holes plus two bolt diameters. The factored shear (52 kips) is to be resisted by weld between mid-depth of the beam and the inside face of the compression flange or between the inner row of tension bolts plus two bolt diameters and the inside face of compression flange, whichever is minimum. By inspection the former governs for this example.

Therefore, a stiffener is required opposite the beam tension flange. As was previously discussed, because of the expense and possibility of interference with weak axis framing, the

use of column web stiffeners is not recommended. Possible solutions for this example are: (1) to use an 8-bolt, stiffened end-plate (Chapter 4) which increases the effective column flange length, (2) to increase the column flange thickness by using a heavier column or (3) to increase the bolt pitch which increases the effective column flange length and decreases the required column flange thickness. If the third change is made, a thicker end-plate may be required. Obviously, any change requires additional expense; however, the resulting connection may be more economical if the column web stiffeners are eliminated. If changes are not practical, the following procedure can be used to determine stiffener size. Assuming only force in excess of what the unstiffened column flange can resist need be resisted by the stiffener, the capacity of the unstiffened column flange is first computed.

Use ¼-in. fillet weld (minimum size for ¾-in. thick plate) both sides of beam web below tension bolt region.

iii. Check beam web yielding:

D. Check column side limit states and design stiffeners if necessary, A36 steel i. Check column web yielding using Inequality 3.3b: Thus, the stiffener will be designed for Therefore, stiffeners are not required opposite the beam tension and compression flanges to prevent column web yielding.

The required stiffener area is then

ii. Check column web buckling using Inequality 3.4b:

Stiffeners do not need to be full depth of the column web if only one beam is connected to the column at a given elevation. Since the stiffener is in tension, local buckling is not a limit

Therefore, web stiffeners are not required opposite the beam compression flange to prevent column web buckling.

do not apply; however, good engineering practice requires the stiffener to be proportioned to be compatible with the beam flange. Assume ¾-in. "clip" to clear column web fillets.

state and AISC LRFD specification width and thickness rules

iii. Check column flange bending: The required column flange thickness is determined using Equation 3.2b with the modifications that resulted in Equation 3.5b.

Column flange to stiffener weld:

Minimum weld is ¼ in. Use

fillet weld both sides.

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Column web to stiffener weld: By inspection 8 bolts are required. The force per bolt, B, is then

Minimum weld is To simplify detailing, use fillet weld both sides. Check shear force in stiffener base metal using AISC specification Equation J5-3 (length along flange governs).

From Table A.1, try 1-in. diameter bolts (allowable capacity is 34.6 kips). Assuming A325-SC bolts, the single shear capacity from Table A.1 is 13.7 kips. The number of bolts required to resist the applied shear is then

Stiffener Selection Use 2PL ½x4x0'-7 with fillet welds.

Bolt Selection Use 12 1-in. diameter

iv. Check column web yielding using inequality 3.6b:

A325-SC bolts fully tightened, 8 at beam tension

flange and 4 at beam compression flange. Therefore, column web reinforcement is not required. Final design details are shown in Fig. 3.4.

3.2 EIGHT-BOLT DESIGN PROCEDURES

AND ALLOWABLE STRESS DESIGN EXAMPLE The design procedures for unstiffened extended end-plates in the AISC Manuals imply that the end-plate configuration shown in Fig. 1.1(c) can be designed using the work of Krishnamurthy (1978a). The work of Hendrick and Murray (1984) can be used to evaluate column web yielding and buckling. Column flange bending strength requires special consideration. A suggested approach is given in the following ASD example. Only slight modifications are required for LRFD design (see Example 3.3). EXAMPLE 3.4. Design a beam-to-column end-plate connection for a moment of 700 ft-kips and a shear of 90 kips using ASD procedures. The beam is a W33x118 and the column is a W14x311. All material is A36. Bolts are A325 and are

limited to 1-in. diameter. E70XX electrodes will be used for all welding. The beam top flange is in tension and moment reversal is not a consideration.

A. Bolt design, A325-SC bolts The beam tension flange force,

is

Fig. 3.4. Final design details, Example 3.3.

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Use

B. End-plate design, A36 steel Try edge distance = 1¼ in.

gages = 6 in. and 12 in. (inside and outside bolts) pitch,

flange or between the inner row of tension bolts plus two bolt diameters and the inside face compression flange, whichever is minimum. By inspection the former governs for this example.

Required end-plate width is 1¼ + 3 + 6 + 3 + 1¼ = 14½ in. (Note column flange width is 16¼ in.) Effective end-plate width must be less than beam flange width plus 1 in.

Determine

fillet weld both sides of beam web from inside

of beam flange to centerline of bolt holes plus two bolt diameters. The applied shear is to be resisted by weld between middepth of the beam and the inside face of the compression

from Equation 3.1a:

Use fillet weld (minimum size for thick plate) both sides of beam web below tension bolt region.

iii. Check beam web yielding

D. Check column side limit states and design stiffeners if necessary, A36 steel i. Check column web yielding using inequality 3.3a:

Check bolt bearing on end-plate (note column flange thickness is larger and, conservatively, only the compression side bolts are considered).

Therefore, stiffeners are not required opposite the beam tension and compression flanges to prevent column web yielding.

ii. Check column web buckling using inequality 3.4a:

End-Plate Selection

iii. Check column flange bending Design procedures are not available to assess the column flange bending strength for this bolt pattern. However, the strength can be evaluated if a small triangular stiffener between the column flange and the column web is used since this pattern is similar to that of the eight-bolt stiffened endplate discussed in Chapter 4. When this approach is used

C. Weld design, E70XX electrode i. Beam flanges to end-plate welds: By inspection, fillet welds will be impractical; therefore, use full penetration groove weld with reinforcement at beam tension flange. Use minimum weld at beam compression flange.

the column web is equivalent to the beam flange and the column flange is equivalent to the end-plate. Since test data is not available, it is recommended that the effective column

ii. Beam web to end-plate weld Minimum size of fillet weld is The required weld to develop the bending stress in the beam web near the tension bolts is

flange length (equivalent to the end-plate width) be taken as that recommended for the four-bolt configuration (Curtis and Murray, 1989), e.g., 2.5c. With reference to Chapter 4, for details of the design procedure, the column flange for this example is now checked. (See ASD nomenclature for definition of terms.) Details are shown in Fig. 3.5.

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The equivalent gage and pitch are

From Equation 4.4a and the equivalent beam flange and end-plate thicknesses are

and from Equation 4.5a The effective end-plate width is equal to 2.5c or

The column flange-to-web stiffener should be approximately equal to the beam flange thickness (0.740 in.) and extend beyond the outside row of bolts, thus use a rectangular plate ¾ in. x 7 in. x 7 in. Since all limitations given in Chapter

Since only 6 bolts are assumed effective, the capacity of the stiffened flange is 6 x 106.4 = 638.4 kips which is greater than the applied beam flange force of 261.5 kips and the stiffened column flange is adequate. Conservatively, the stiffener to flange and web welds will be designed for the applied beam flange force. Assuming a 1½ in. "clip" to clear the column fillet, the required fillet weld size is

4 are satisfied, the simplified method, Equation 4.4a, can be used to determine the adequacy of the stiffened column flange. From Equation 4.7a with

Use fillet welds both sides of stiffener. (Full penetration groove welds are not practical at this location.)

Stiffener Selection Use 2PL ¾ x 7 x 0'-7 with fillet welds.

iv. Check column web yielding using inequality 3.6a:

Therefore, column web reinforcement is not required. Final design details are shown in Fig. 3.5.

Fig. 3.5. Final design details, Example 3.4.

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Chapter 4 STIFFENED, EXTENDED END-PLATE CONNECTION DESIGN 4.1 DESIGN PROCEDURES The following ASD design recommendations for the extended, four-bolt, stiffened, beam-to-column, end-plate connection shown in Fig. 4.1 are based on the works of Murray and Kukreti (1988), "End-Plate Moment Connections— Their Use and Misuse," Hendrick and Murray (1984), "Column Web Compression Strength at End-Plate Connections," and Curtis and Murray (1989), "Column Flange Strength at Moment End-Plate Connections." The basic procedures for end-plate and bolt design are also found in the 9th ed. AISC ASD Manual of Steel Construction. Murray and Kukreti (1988) present two methods for determining end-plate thickness and bolt diameter. Both methods are limited to use for A36 end-plate steel and A325 bolts and both include bolt prying action effects. The first method is a series of equations developed from regression analyses of data generated by the finite element method. The finite element model included both second order geometry effects and inelastic plate and bolt material properties. With this method, the required end-plate thickness is the larger of and determined from (see Fig. 4.1 for definition of terms): in ASD

(4.1a)

in ASD

(4.2a)

in LRFD

(4.1b)

in LRFD

(4.2b)

with = minimum bolt tension as given in AISC specifications and reproduced here for A325 bolts in Tables A.1 and B.1. Equation 4.3a includes a factor of safety of 2.0. Equation 4.3b does not include a resistance factor, thus the specified minimum tensile strength of the bolt material must be used to determine the required bolt diameter. In the application of Equations 4.1, 4.2 and 4.3, a preliminary bolt diameter is selected assuming that 6.8 of the 8 tension bolts are effective. This ratio must often be decreased

or

The regression-based Equations 4.1 are stiffness criteria which control end-plate flexibility for use in Type I construction. Equations 4.2 are strength criteria which limit maximum strain on the end-plate. Both ASD equations include a factor of safety of 1.67 and both LRFD equations include a resistance factor of 0.9. Ultimate bolt force including prying action effects is estimated from

Fig. 4.1. Eight-bolt stiffened end-plate connection geometry.

17

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if Equation 4.3 shows the selected bolt diameter to be inadequate. The second method is a simplified approach which was formulated because of the difficulty in using Equations 4.1, 4.2 and 4.3, except for completely computerized designs. The

The following limitations, in addition to those given in Chapter 2, apply to the simplified method: 1. The connected beam section must be hot-rolled and included in the "Allowable Stress Design Selection Table" in the AISC ASD Manual. 2. The vertical pitch, from the face of the beam tension flange to the first row of bolts must not exceed 2½ in. The recommended minimum pitch is bolt diameter plus ½ in. 3. The vertical spacing between bolt rows, must not exceed 4. The horizontal gage, g, must be between 5½ and 7½ in. 5. Bolt diameter must not be less than ¾ in. nor greater than 1½ in.

method was developed by first generating end-plate and bolt sizes using the above equations for all hot-rolled A36 steel beam sections at various moment levels. An effective number of bolts was then determined for each connection and a conservative lower bound of six bolts established. Next, it was assumed that plate thickness could be established from tee-stub analogy bending (see Fig. 4.2), that is, in ASD

(4.4a)

in LRFD

(4.4b)

or

with force per bolt based on six effective bolts and an effective pitch. From the generated designs it was determined that in ASD

(4.5a)

in LRFD

(4.5b)

The recommendations of Hendrick and Murray (1984) can be used to check column web yielding at either the beam tension or compression flanges (inequality 3.3) and column web buckling at the beam compression flange (inequality 3.4). Since Type I construction is assumed for this connection, a stiff column flange is required. Thus, unless the column

or

flange is considerably thicker than the end-plate, flange to web stiffeners are required. If effective flange length effects are neglected, the behavior of the column flange is identical to that of the end-plate and, therefore, the column flange must be at least as thick as the end-plate, and the column stiffener must be as thick as the beam flange. Further, the stiffener to flange weld must be sufficient to develop the strength of the full thickness of the stiffener plate. If the column flange is substantially thicker than the endplate (1.5-2 times), stiffeners may not be necessary. Based on the work of Curtis and Murray (1989), such an unstiffened flange can be evaluated using Equation 3.5 with

The required end-plate thickness is then determined from in ASD

(4.6a)

in LRFD

(4.6b)

or

with

in ASD

(4.7a)

(4.8)

The referenced work included only A36 steel. Therefore, it is recommended that if the column material yield strength is greater than 36 ksi, the column flange strength be checked assuming A36 steel is being used. Column web shear strength should be checked using inequality 3.6.

4.2 DESIGN EXAMPLES 4.2.1 Allowable stress design examples The following three examples demonstrate the use of the above ASD procedures. Example 4.1 uses the simplified design method, Equation 4.6a. Example 4.2 uses the more exact design method, Equations 4.1a, 4.2a and 4.3a. Example 4.3 demonstrates the ASD procedures for checking the column side of the connection. For all examples, the beam top flange is in tension and moment reversal does not occur.

Fig. 4.2. Tee-stub analogy moments.

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and the equivalent tee-stub analogy moment from Equation 4.4a

EXAMPLE 4.1. Using the ASD procedures, design a beamto-column end-plate connection for a moment of 700 ft-kips and a shear force of 90 kips. The beam is a W33x118 and the column is a W14x311. All material is A36 steel. Bolts are A325. E70XX electrodes will be used for all welding. Use Equation 4.6a to determine end-plate thickness and

The required section modulus is then

assume only 6 bolts are effective. And the required end-plate thickness from Equation 4.6a is A. Bolt design, A325-SC bolts The beam tension flange force, Check bolt bearing on end-plate (note that (1) column flange thickness is larger and (2) conservatively only the compression side bolts are considered).

Assuming 6 bolts effective, the force per bolt is From Table A.1, try diameter bolts (allowable capacity is 43.7 kips). Assuming A325-SC bolts, the single shear capacity from Table A.1 is 17.4 kips. The number of bolts required to resist the applied shear is then

End-Plate Selection

C. Weld design, E70XX electrodes i. Beam flanges to end-plate welds: By inspection, the fillet welds will be impractical. Use full penetration groove weld with reinforcement at beam tension flange and fillet weld (minimum for.1¼-in. plate at beam compression flange).

Bolt Selection Use

diameter

A325-SC bolts fully tightened, 8 at beam tension flange and 2 at compression flange.

ii. Beam web to end-plate weld: Minimum size fillet weld is Conservatively, the required weld to develop the bending stress in the beam web near the tension bolts is

(Note if the four bolt unstiffened configuration shown in Fig. 1.1(a) is used, the required bolt diameter is

B. End-plate design, A36 steel Try edge distance = 1½ in. gage g = 6 in. pitch pitch between bolt rows stiffener thickness

Use

fillet weld both sides of beam web from inside

face of beam flange to centerline of innermost bolt holes plus two bolt diameters. The applied shear is to be resisted by weld between the minimum of the mid-depth of the beam and the compression flange or the inner row of tension bolts plus two bolt diameters and the compression flange. By inspection the former governs for this example.

Note that all of the specified limitations for the simplified method are satisfied. Minimum end-plate width is

Effective end-plate width must be less than or equal to the beam flange width plus 1 in., e.g.

12.48 in. Use 12½ in. end-plate width and Determine effective pitch from Equation 4.5a, Use fillet weld (minimum size for 1¼-in. thick plate) both sides of beam web below tension bolt region.

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iii. Check beam web shear yielding:

Thus, diameter A325-SC bolts are satisfactory. Since the end-plate thickness and bolt diameter are the same as in Example 4.1, the number of bolts required to resist the shear force is the same and bolt bearing is adequate. Hence, the final design using the regression based Equations 4.1a, 4.2a and 4.3a is identical to that obtained using the split-tee analogy method, Equation 4.6a. Column side limit states are checked in Example 4.3.

Column side limit states are checked in Example 4.3. EXAMPLE 4.2. For the conditions of Example 4.1, determine required end-plate thickness and bolt diameter using Equations 4.1a, 4.2a and 4.3a. ASD procedures apply. A. Trial bolt size, A325-SC bolts From Example 4.1, the flange force is 261.5 kips. A trial bolt size is selected assuming 6.8 bolts are effective.

From Table A.1, try ity is 43.7 kips).

EXAMPLE 4.3. Using the data, bolt design and end-plate from Example 4.1, determine if stiffeners are required if the column is a W14x311 A36 steel. Only the column side limit states need to be checked. ASD procedures apply.

diameter bolts (allowable capac-

B. End-plate design, A36 steel Try: edge distance = 1½ in. gage g = 6 in. pitch pitch between bolt rows stiffener thickness

i. Check column web yielding using inequality 3.3a, A36 steel:

From Example 4.1, use 12½-in. wide end-plate with = 12.48 in. Determine the required end-plate thickness from the stiffness criterion, Equation 4.1a.

Therefore, stiffeners are not required opposite the beam tension and compression flanges to prevent column web yielding. ii. Check column web buckling using inequality 3.4a, A36 steel:

Determine the required end-plate thickness from the strength criterion, Equation 4.2a.

Therefore column web stiffeners are not required to prevent column web buckling.

Check adequacy of 4.3a.

iii. Check column flange bending, A36 steel: Since the column flange is significantly thicker than the endplate, column flange stiffeners may not be required. The unstiffened column flange can be investigated using Equation 3.2a with appropriate modifications. From Curtis and Murray (1989), the effective column flange length, which is equivalent to the end-plate width in Equation 3.2, is

diameter bolts using Equation

The ultimate bolt force must be less than the tensile strength of the bolt which is twice the allowable capacity given in Table A.1, that is

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iv. Check column web yielding using inequality 3.6a, A36 steel:

Therefore, column web reinforcement is not required. Final design details are shown in Fig. 4.3.

The required flange thickness from Equation 3.2 is

4.2.2 Load and resistance factor design examples The following three examples demonstrate the use of the LRFD procedures. Example 4.4 uses the simplified design method, Equation 4.6b. Example 4.5 uses the more exact method, Equations 4.1b, 4.2b and 4.3b. Example 4.6 demonstrates the LRFD procedures for checking the column side of the connection. For all examples, the beam top flange is in tension and moment reversal does not occur.

Therefore, column web stiffeners are not required for this example.

EXAMPLE 4.4. Using the LRFD procedures, design a beamto-column end-plate connection for a factored moment of 1050 ft-kips, an unfactored shear force of 90 kips and a factored shear force of 135 kips. The beam is a W33x118 and the column is a W14x311. All material is A36 steel. Bolts are A325. E70XX electrodes will be used for all welding. Use Equation 4.6b to determine end-plate thickness and assume only 6 bolts are effective.

A. Bolt design, A325-SC bolts The beam tension flange force,

is

Assuming 6 bolts effective, the force per bolt is

From Table B.1, try diameter bolts (design tension load is 67.1 kips). Assuming A325-SC bolts, the single shear capacity from Table B.1 is 16.9 kips. The number of bolts required to resist the applied shear is then 90 / 16.9 = 5.3.

Bolt Selection Use diameter A325-SC bolts fully tightened, 8 at beam tension flange and 2 at beam compression flange.

(Note if the four bolt unstiffened configuration shown in Fig. 1.1(a) is used, the required bolt diameter is

Fig. 4.3. Final design details for eight-bolt stiffened end-plate examples.

21 © 2003 by American Institute of Steel Construction, Inc. All rights reserved. This publication or any part thereof must not be reproduced in any form without permission of the publisher.

The applied shear is to be resisted by weld between the minimum of the mid-depth of the beam and the compression flange or the inner row of tension bolts plus two bolt diameters and the compression flange. By inspection the former governs for this example.

B. End-plate design, A36 steel From Example 4.1 use:

gage g = 6 in. pitch pitch between bolt rows stiffener thickness end-plate width = 12½ in. effective end-plate width Note that all of the specified limitations for the simplified method are satisfied. Determine effective pitch from Equation 4.5b.

Use fillet weld (minimum size for 1¼-in. thick plate) both sides of beam web below tension bolt region.

iii. Check beam web yielding:

Column side limit states are checked in Example 4.6.

and the equivalent tee-stub analogy moment from Equation 4.4b

The required section modulus is then

EXAMPLE 4.5. For the conditions of Example 4.4, determine required end-plate thicknesses and bolt diameter using Equations 4.1b, 4.2b and 4.3b. LRFD procedures apply.

And the required end-plate thickness from Equation 4.6b is

A. Trial bolt size, A325-SC bolts From Example 4.4, the factored flange force is 392.3 kips. A trial bolt size is selected assuming 6.8 bolts are effective.

From Table B.1, try ity is 67.1 kips).

Check bolt bearing on end-plate (note column flange thickness is larger and conservatively only the compression side bolts are considered).

diameter bolts (allowable capac-

B. End-plate design, A36 steel From Example 4.1 use:

End-Plate Selection

C. Weld design, E70XX electrode i. Beam flanges to end-plate welds: By inspection, fillet welds will be impractical. Use full penetration groove weld with reinforcement at beam tension flange and fillet weld (minimum for 1¼-in. plate at beam compression flange).

Determine the required end-plate thickness from the stiffness criterion, Equation 4.1b.

ii. Beam web to end-plate weld: Minimum size fillet weld is Required weld to develop maximum bending stress in web near tension bolts is Determine the required end-plate thickness from the strength criterion, Equation 4.2b. Use fillet weld both sides of beam web from inside face of beam flange to centerline of innermost bolt holes plus two bolt diameters.

22 © 2003 by American Institute of Steel Construction, Inc. All rights reserved. This publication or any part thereof must not be reproduced in any form without permission of the publisher.

ii. Check column web buckling using inequality 3.4b, A36 steel:

Check adequacy of 4.3b.

diameter bolts using Equation Therefore column web stiffeners are not required to prevent column web buckling.

iii. Check column flange bending, , A36 steel: Since the column flange is significantly thicker than the endplate, column flange stiffeners may not be required. The unstiffened column flange can be investigated using Equation 3.2b with appropriate modifications. From Curtis and Murray (1989), the effective column flange length, which is equivalent to the end-plate width in Equation 3.2b, is

The ultimate bolt force must be less than the tensile strength of the bolt which is the design tension capacity given in Table B.1, that is

Thus, diameter A325-SC bolts are satisfactory. Since the end-plate thickness and bolt diameter are the same as in Example 4.4, the number of bolts required to resist the shear force is the same and bolt bearing is adequate. Hence, the final design using the regression based Equations 4.1b, 4.2b and 4.3b is identical to that obtained using the split-tee analogy method, Equation 4.6b. Column side limit states are

checked in Example 4.6. EXAMPLE 4.6. Using the data, bolt design and end-plate from Example 4.4, determine if stiffeners are required if the column is a W14x311 A36 steel. Only the column side limit states need to be checked. LRFD procedures apply.

The required flange thickness from Equation 3.2b is

Therefore, column web stiffeners are not required for this example.

iv. Check column web yielding using inequality 3.6b, A36 steel:

i. Check column web yielding using inequality 3.3b, A36 steel:

Therefore, stiffeners are not required opposite the beam tension and compression flanges to prevent column web yielding.

Therefore, column web reinforcement is not required. Final design details are the same as for the ASD Example 4.3 and are shown in Fig. 4.3.

23 © 2003 by American Institute of Steel Construction, Inc. All rights reserved. This publication or any part thereof must not be reproduced in any form without permission of the publisher.

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Chaudhari, V. N. (1975), "Finite Element Analysis of Multiple Bolt Row and Flush Type Connections," Unpublished Thesis, Auburn University, Auburn, AL, March 1975.

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26 © 2003 by American Institute of Steel Construction, Inc. All rights reserved. This publication or any part thereof must not be reproduced in any form without permission of the publisher.

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Ioannides, S. A. and T. S. Tarpy (1979), "Finite Element Analysis of Unstiffened Beam-to-Column End-Plate Connections," Proceedings of the Third International Conference in Australia on Finite Element Methods, The University of New South Wales, New South Wales, Australia, July 1979, pp. 49-63.

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27 © 2003 by American Institute of Steel Construction, Inc. All rights reserved. This publication or any part thereof must not be reproduced in any form without permission of the publisher.

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28 © 2003 by American Institute of Steel Construction, Inc. All rights reserved. This publication or any part thereof must not be reproduced in any form without permission of the publisher.

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Moment Connections," Master of Science Thesis, School of Civil Engineering and Environmental Science, University of Oklahoma, Norman, OK, 1986.

Newlin, D. E. and W. E. Chen (1971), "Strength and Stability of Column Web in Welded Beam-to-Column Connections," Fritz Engineering Laboratory, Report No. 333.14, Lehigh University, Bethlehem, PA, May 1971.

Munse, W. H., K. S. Peterson and E. Chesson (1959), "Strength in Tension of Rivets and High Strength Bolts," Journal of the Structural Division, ASCE, Vol. 85, No. ST3, Part I, March 1959, pp. 7-28.

Onderdonk, A. B., R. P. Lathrop and J. Goel (1964), "EndPlate Connections in Plastically Designed Structures," Engineering Journal, AISC, Vol. 1, No. 1, Jan. 1964, pp. 24-27.

Murray, T. M. (1984), "Beam-to-Column End-Plate Connections—Column Design Considerations," 36th Annual AISC National Conference, Tampa Bay, PL, March 29-30, 1984.

Oswalt, R. E. (1978), "Two-Dimensional Finite Element Analysis of the Effects of Bolt Heads and Welds in Steel EndPlate Connections," Unpublished thesis, Vanderbilt University, Nashville, TN, May 1978.

Murray, T. M. (1985), "Bolted End-Plate Moment Connections in the USA," Symposium Proceedings, International Symposium on Structural Steel Design and Construction, Singapore, July 3-4, 1985, pp. 43-57.

Packer, J. A. (1975), "A Study of the Tension Region of Plastically Designed, Bolted Beam-to-Column Connections," Ph.D, thesis presented to the University of Manchester, at Manchester, England, in 1975.

Murray, T. M. (1988), "Recent Developments for the Design of Moment End-Plate Connections," Journal of Constructional Steel Research, Elsevier Applied Science Publishers, October 1988, pp. 133-162.

Packer, J. A. and L. J. Morris (1975), "Behavior and Design of Haunched Steel Portal Frame Knees," Research Report

M002, Simon Engineering Laboratories, University of Manchester, England, 1975.

Murray, T. M. and A. Kukreti (1984), "Large Capacity Moment End-Plate Connections for Offshore Construction," Offshore Engineering, Vol. 4, Pentach Press, London, 1984.

Packer, J. A. and L. J. Morris (1977), "A Limit State Design Method for the Tension Region of Bolted Beam-to-Column Connections," The Structural Engineer, Vol. 5, No. 10, London, October 1977, pp. 446-58.

Murray, T. M. and A. Kukreti (1985), "Design of 8-Bolt Stiffened End-Plate Moment Connections," Papers, Third Conference on Steel Developments, Australian Institute of Steel Construction, Melbourne, Australia, May 20-22, 1985, pp. 145-149.

Packer, J. A. and L. J. Morris (1978), Discussion of "A Limit

State Design Method for the Tension Region of Bolted Beamto-Column Connections," The Structural Engineer, Vol. 56A, No. 8, London, Aug. 1978, pp. 217-223.

Murray, T. M. and A. Kukreti (1988), "Design of 8-bolt

29 © 2003 by American Institute of Steel Construction, Inc. All rights reserved. This publication or any part thereof must not be reproduced in any form without permission of the publisher.

Patel, K. V. (1981), Analysis of Steel Beam-to-Column Moment Connections," Ph.D. Dissertation submitted to Purdue University, Purdue University, West Lafayette, IN, May 1981.

Report 6-60-13, Stevin Laboratory, Delft University of Technology, Delft, Netherlands, 1969.

Struik, J. H. A., A. O. Oyeledum and J. W. Fisher (1973), "Bolt Tension Control with a Direct Tension Indicator," Engineering Journal, AISC, Vol. 10, No. 1, 1st Quarter, 1973.

Patel, K. V. and W. F. Chen (1984), "Nonlinear Analysis of Steel Beam-to-Column Moment Connections," Journal of Structural Engineering, ASCE, Vol. 110, No. 8, August 1984.

Surtees, J. O. and A. P. Mann (1970), "End-Plate Connections in Plastically Designed Structures," Proceedings Conference of Joints in Structures, Vol. 1, Paper 5, University of Sheffield, England, July, 1970.

Phillips, J. and J. A. Packer (1981), "The Effect of Plate Thickness on Flush End-Plate Connections," Joints in Structural Steelwork, John Wiley & Sons, London-Toronto, 1981, pp. 6.77-6.92.

Tall, L., Ed. (1974), Structural Steel Design, 2nd Ed., Ronald Press, New York, 1974.

Punch, S. H. (1976), "A Finite Element Analysis of Column Behavior in a Bolted Beam-to-Column End-Plate Moment Connection," Master of Science Thesis, Vanderbilt University, Nashville, TN, 1976.

Tarpy, T. S. and S. D. Lindsey (1975), "Experimental Test Results on End-Plate Moment Connections," Report No. CEAISC-1, Dept, of Civil Engineering, Vanderbilt University, Nashville, TN, June 1975.

RCSC (1985), "Specifications for Structural Joints Using ASTM A325 or A490 Bolts," Research Council on Structural Connections of the Engineering Foundation, November 13, 1985.

Tarpy, T. S. and J. W. Cardinal (1981), "Behavior of SemiRigid Beam-to-Column End-Plate Connections," Joints in Structural Steelwork, John Wiley & Sons, London-Toronto, 1981, pp. 2.3-2.25.

Salmon, C. G. and J. E. Johnson (1971), Steel Structures, Design and Behavior, Intext Educational Publishers, New York, 1971.

Van Bercum, J. Th., F. S. K. Bijlaard and P. Zoetemeijer (1978), "Design Rules for Bolted Beam-to-Column Connections," (in Dutch) Staalbouwkundig Genootschap, P.O.B.

Salmon, C. G. and J. E. Johnson (1980), Steel Structures, 2nd Ed., Harper & Row, New York, 1980.

20714, 3301 JA, Rotterdam, 1978. Van Douwen, A. A. (1981), "Design for Economy in Bolted and Welded Connections," Joints in Structural Steelwork, John Wiley & Sons, London-Toronto, 1981, pp. 5.18-5.35.

Saxena, K. (1978), Photoelastic and Finite Element Analysis of Steel Bolted Tee Hangers, Unpublished Thesis, Vanderbilt University, Nashville, TN, May 1978.

Witteveen, J., J. W. B. Stark, F. S. K. Bijlaard and P. Zoetemeijer (1982), "Welded and Bolted Beam-to-Column Connections," Journal of the Structural Division, ASCE, Vol. 108, No. ST2, Proc. Paper 16873, February 1982, 433-455.

Schutz, F. W, Jr. (1959), "Strength of Moment Connections Using High Tensile Strength Bolts," Proceedings of the National Engineering Conference, American Institute of Steel Construction, 1959, pp. 98-110.

Sherbourne, A. N. (1961), "Bolted Beam-to-Column Connections," The Structural Engineer, Vol. 39, London, England, June 1961.

Zoetemeijer, P. (1974), "A Design Method for the Tension Side of Statically Loaded, Bolted Beam-to-Column Connections," Heron, Vol. 20, No. 1, Delft University, Delft, Netherlands, 1974, pp. 1-59.

Srouji, R. (1983), "Yield-Line Analysis of End-Plate Connections with Bolt Force Predictions," Master of Science Thesis, School of Civil Engineering and Environmental Science, University of Oklahoma, Norman, OK, 1983.

Zoetemeijer, P. (1981), "Semi-Rigid Bolted Beam-to-Column Connections with Stiffened Column Flanges and Flush EndPlates," Joints in Structural Steelwork, John Wiley & Sons, London-Toronto, 1981, pp. 2.99-2.118.

SSRC (1986), Connections Bibliography, Task Group 25,

Zoetemeijer, P. (1981a), "Bolted Connections with Flush End-Plates and Haunched Beams, Tests and Computations," Stevin Laboratory, Delft University of Technology, 1981.

Structural Stability Research Council, September 1986. Sterling, G. H. and J. W. Fisher (1966), "A440 Steel Joints Connected by A490 Bolts," Journal of the Structural Division, ASCE, Proc. Paper 4845, No. ST3 (1966).

Zoetemeijer, P. and M. H. Kolstein (1975), "Bolted Beamto-Column Connections with Flush End-Plates, Tests and Computation Methods" (in Dutch), Report 6-75-20, Stevin Laboratory, Delft University of Technology, 1975.

Struik, J. H. A. and J. DeBack (1969), "Tests on Bolted TStubs with Respect to Bolted Beam-to-Column Connections,"

30

© 2003 by American Institute of Steel Construction, Inc. All rights reserved. This publication or any part thereof must not be reproduced in any form without permission of the publisher.

Appendix A ASD NOMENCLATURE, DESIGN AIDS, AND QUICK REFERENCE EXAMPLES A.1 ASD NOMENCLATURE planar area of column connection, in.2 area of beam tension flange, in. 2 gross area of plate, in.2 2 column stiffener area, in. 2 area of beam web, in. beam or column flange width, in. effective end-plate width, in. (not more than + 1 in.) effective column flange length, in. unfactored tension force per bolt, kip s allowable tension load in bolt, kips ultimate bolt force including prying action effects, kips vertical spacing between rows of tension bolts, in. a constant depending on the plate material yield stress, the bolt material and the design method

specified yield stress of the end-plate material, ksi specified yield stress of column material, ksi horizontal spacing between vertical bolt lines, in. distance from outer face of flange to web toe of fillet, in. the column section distance, in. unfactored effective end-plate moment, in.-kips connected beam end moments, ft-kips required number of bolts to resist beam shear effective pitch, in. pitch, distance from center line of bolt to nearer surface of the tension flange, in. + ½ in. is generally enough to provide wrench clearance) factored beam flange force equal to times the beam flange force when the flange force is due to live and dead loads only, or by when the flange force is due to live and dead loads in conjunction with wind or earthquake forces, kips effective pitch, in. minimum bolt tension, kips required end-plate elastic section modulus, in.3 beam flange thickness, in. column flange thickness, in. required column flange thickness, in. end-plate thickness, in. required end-plate thickness from stiffness criterion, in. required end-plate thickness from strength criterion, in. end-plate to beam tension flange stiffener thickness (approximately equal in thickness to that of the beam web), in. beam web thickness, in. column web thickness, in. column web depth clear of fillets, in. single shear capacity of bolt, kips leg size of fillet weld or reinforcement weld, in.

depth of beam or column section, in. nominal bolt diameter, in. column web depth clear of fillets, in. required fillet weld throat size, sixteenths edge distance, in. computed bearing stress, ksi computed shear stress, ksi average yield stress for beam and end-plate materials, ksi allowable bending stress for the end-plate material (0.75 times the specified yield stress), ksi allowable bending stress for column flange material (0.75 times the specified yield stress), ksi allowable tensile stress for bolt material, ksi specified minimum tensile strength for bolt material, ksi capacity of unstiffened column flange to resist applied force, kips unfactored beam flange force, kips allowable bearing stress, ksi specified minimum tensile strength, ksi allowable shear stress, ksi

31 © 2003 by American Institute of Steel Construction, Inc. All rights reserved. This publication or any part thereof must not be reproduced in any form without permission of the publisher.

A.2 ASD DESIGN AIDS Table A.1. Allowable Tension and Single Shear Loads for A325 and A490 Bolts (ASD Method) a) A325 Bolts

Diameter (in.) Tension (kips)

26.5

34.6

43.7

54.0

65.3

77.7

Shear A325-SC (kips)

5.4

7.7

10.5

13.7

17.4

21.5

26.0

30.9

Shear A325-N (kips)

6.4

9.3

12.6

16.5

20.9

25.8

31.2

37.1

Shear A325-X (kips)

9.2

13.3

18.0

23.6

29.8

36.8

44.5

53.0

28

39

51

56

71

85

32.5

42.4

53.7

66.3

80.2

95.4

13.2

17.3

21.9

27.0

32.7

38.9

12.4

16.8

22.0

27.8

34.4

41.6

49.5

13.5

Minimum Bolt Tension (kips)

19

19.4

103

b) A490 Bolts

Diameter (in.) Tension (kips)

16.6

Shear A490-SC (kips)

6.7

Shear A490-N (kips)

8.6

23.9 9.7

Shear A490-X (kips)

12.3

17.7

24.1

31.4

39.8

Minimum Bolt Tension (kips)

24

35

49

64

80

49.1 102

59.4

121

70.7 148

All values from AISC ASD Manual (1980, 1989).

Table A.2. ASD Values of for A325 Bolts

Table A.3. ASD Values of for A490 Bolts

32 © 2003 by American Institute of Steel Construction, Inc. All rights reserved. This publication or any part thereof must not be reproduced in any form without permission of the publisher.

Table A.4. Values of Section

Section

W36 x 359 x328 x300 x280 x260 x245 x230 x256 x 232 x210 x194 x182 x170 x160 x150 x135

0.899 0.903 0.887 0.882 0.850 0.835 0.818 0.648 0.644 0.588 0.587 0.579 0.573 0.554 0.530 0.463

W33x354 x318 x291 x 263 x 241 x 221 x 201 x169 x152 x141 x130 x118

0.925 0.926 0.913 0.909 0.853 0.829 0.807 0.667 0.612 0.583 0.541 0.492

W30 x 235 x211 x191 x173 x148 x132 x124 x116 x108 x 99

0.961 0.905 0.887 0.861 0.672 0.606 0.590 0.558 0.516 0.476

Section

W27x217 x194 x178 x161 x146 x129 x114 x102 x 94 x 84

1.003 0.986 0.909 0.902 0.885 0.710 0.646 0.635 0.597 0.545

W24 x 176 x162 x146 x131 x117 x104 x103 x 94 x 84 x 76 x 68 x 62 x 55

1.021 0.994 0.959 0.904 0.877 0.848 0.711 0.683 0.655 0.616 0.560 0.428 0.397

W21 x 166 x147 x132 x122 x111 x101 x 93 x 83 x 73 x 68 x 62 x 57 x 50 x 44

1.140 1.011 1.002 1.003 0.994 0.995 0.683 0.686 0.683 0.667 0.641 0.532 0.465 0.423

Section

W18x143 x130 x119 x106 x 97 x 86 x 76 x 71 x 65 x 60 x 55 x 50 x 46 x 40 x 35

1.204 1.186 1.082 1.059 1.076 1.056 1.048 0.741 0.751 0.751 0.722 0.714 0.604 0.595 0.504

W16x100 x 89 x 77 x 67 x 57 x 50 x 45 x 40 x 36 x 31 x 26

1.170 1.152 1.146 1.149 0.789 0.781 0.768 0.772 0.679 0.589 0.506

W14x120 x109 x 99 x 90 x 82 x 74 x 68 x 61 x 53 x 48 x 43 x 38 x 34 x 30 x 26 x 22

1.855 1.899 1.859 1.860 1.348 1.394 1.382 1.364 1.141 1.115 1.103 0.861 0.824 0.734 0.633 0.557

W12x87 x79 x72 x65 x58 x53 x50 x45 x 40 x35 x30 x26 x22 x19 x16 x14

1.748 1.732 1.720 1.706 1.631 1.527 1.281 1.266 1.281 0.992 0.963 0.936 0.575 0.520 0.419 0.390

W10x60 x 54 x49 x45 x39 x33 x 30 x 26 x22 x19 x17 x15 x12

1.842 1.882 1.859 1.603 1.516 1.348 1.045 1.033 0.913 0.672 0.583 0.497 0.463

W 8x35 x31 x 28 x 24 x21 x18 x15 x13 x10

1.796 1.711 1.495 1.487 1.127 1.007 0.690 0.593 0.635

W 6x25 1.580 x20 x15

1.545 1.238

x16 1.148 x12 0.890 x 9 0.911 W 5x19 x16

1.867 1.748

W 4x13

1.442

33 © 2003 by American Institute of Steel Construction, Inc. All rights reserved. This publication or any part thereof must not be reproduced in any form without permission of the publisher.

A.3 ASD QUICK REFERENCE EXAMPLES EXAMPLE A.1 (Same as Example 3.1) 4-bolt Unstiffened End-Plate

ii. Check bolt bearing, end-plate controls, compression bolts:

Beam W24x55 A36 steel

iii. Check end-plate shear:

Column W14x159 A36 steel

C. End-plate weld design, E70XX electrodes

i. Beam flanges to end-plate weld:

A. Bolt design, A325-SC bolts i. Tension: ii. Beam web to end-plate weld:

ii. Shear,

B. End-plate design, A36 steel

iii. Check beam web yielding

i. Bending, Equation 3.1a:

D. Column side, A36 steel and E70XX electrodes

i. Check column web yielding, inequality 3.3a,

34 © 2003 by American Institute of Steel Construction, Inc. All rights reserved. This publication or any part thereof must not be reproduced in any form without permission of the publisher.

ii. Check column web buckling, inequality 3.4a,

A. Column side i. Check column web yielding, inequality 3.3a:

iii. Check column flange bending,

ii. Check column web buckling, inequality 3.4a:

iii. Check column flange bending,

Calculations to be made with Column web stiffeners are not required. iv. Check column web yielding, inequality 3.6a:

Column web reinforcement is not required. E. Final details: Design stiffeners and welds for

Column flange to stiffener weld, E70XX electrodes:

Column web to stiffener weld, E70XX electrodes:

Check shear stress in stiffener base metal, A36 steel:

EXAMPLE A.2 (Same as Example 3.2) Data is same as Example A.1, except Column W14x90 A572 Gr 50 steel

iv. Check column web yielding inequality 3.6(a),

35 © 2003 by American Institute of Steel Construction, Inc. All rights reserved. This publication or any part thereof must not be reproduced in any form without permission of the publisher.

B. Final details:

ii. Shear,

B. End-plate design, A36 steel

i. Bending, Equation 4.4a:

ii. Check bolt bearing, end-plate controls, compression bolts:

C. End-plate weld design, E70XX electrodes

EXAMPLE A.3 (Same as Examples 4.1 and 4.3) 8-Bolt Stiffened End-Plate Simplified procedure

i. Beam flanges to end-plate weld: Use full penetration groove weld with

reinforcement.

ii. Beam web to end-plate weld:

Beam W33x118 A36 steel

Column W14x311 A36 steel

iii. Check beam web shear yielding: A. Bolt design, A325-SC bolts i. Tension:

D. Column side, A36 steel and E70XX electrodes

i. Check column web yielding, inequality 3.3a,

36

© 2003 by American Institute of Steel Construction, Inc. All rights reserved. This publication or any part thereof must not be reproduced in any form without permission of the publisher.

Column web reinforcement is not required.

ii. Check column web buckling, inequality 3.4a,

D. Final details:

iii. Check column flange bending,

Column web stiffeners are not required. iv. Check column web yielding, inequality 3.6a:

37 © 2003 by American Institute of Steel Construction, Inc. All rights reserved. This publication or any part thereof must not be reproduced in any form without permission of the publisher.

Appendix B LRFD NOMENCLATURE, DESIGN AIDS, AND QUICK REFERENCE EXAMPLES B.1 LRFD NOMENCLATURE = = = = = = = = = = = =

planar area of the column connection, in.2 area of beam tension flange, in.2 gross area of plate, in. 2 column stiffener area, in.2 area of beam web, in.2 beam or column flange width, in. effective end-plate width, in. (not more than + 1 in.) effective column flange length, in. design tension capacity of bolt, kips factored tension force per bolt; ultimate bolt force including prying action effects, kips vertical spacing between rows of tension bolts, in. a constant depending on the plate material yield stress, the bolt material and the design method.

= horizontal spacing between vertical bolt lines, in. = distance from outer face of flange to web toe of fillet, in. = the column section distance, in. = factored effective end-plate moment, in.-kips = factored beam moment, in.-kips = connected beam factored end moments, ft-kips = required number of bolts to resist beam shear = effective pitch, in. = pitch, distance from center line of bolt to nearer surface of the tension flange, in. + ½ in. is generally enough to provide wrench clearance.) = effective pitch, in. = minimum bolt tension, kips = beam flange thickness, in. = column flange thickness, in. = required column flange thickness, in.

= depth of beam or column section, in. = nominal bolt diameter, in. = column web depth clear of fillets, in.

= end-plate thickness, in. = required end-plate thickness from stiffness criterion, in. = required end-plate thickness from strength criterion, in. = end-plate to beam tension flange stiffener thickness, in. = beam web thickness, in. = column web thickness, in. = column web depth clear of fillets, in. = single shear bolt design strength, kips = factored shear force, kips = leg size of fillet weld or reinforcement weld, in. = required end-plate plastic section modulus, in.3

= required fillet weld throat size, sixteenths = edge distance, in. = average yield stress for beam and end-plate materials, ksi = 0.75 of end-plate material), ksi = ASD allowable tensile stress for bolt material, ksi = specified minimum tensile strength for bolt material, ksi = capacity of unstiffened column flange to resist applied force, kips = factored beam flange force, kips = specified minimum tensile strength, ksi = specified yield stress of the end-plate material, ksi = specified yield stress of column material, ksi

= resistance factor

39 © 2003 by American Institute of Steel Construction, Inc. All rights reserved. This publication or any part thereof must not be reproduced in any form without permission of the publisher.

B.2 LRFD DESIGN AIDS Table B.1. Design Tension and Single Shear Strengths for A325 and A490 Bolts (LRFD Method) a) A325 Bolts

Diameter (in.) Tension (kips)

29.8

20.7

5.22

53.0

67.1

82.8

100.2

119.3

10.2

13.4

16.9

20.9

25.2

30.0

Shear A325-N (kips)

10.8

15.5

21.1

27.6

34.9

43.1

52.1

62.0

Shear A325-X (kips)

14.4

20.7

28.1

36.8

46.5

57.4

69.5

82.7

Minimum Bolt Tension (kips)

19

28

39

51

56

71

85

66.3

83.9

103.5

125.3

149.1 37.1

Shear A325-SC (kips)

7.51

40.6

103

b) A490 Bolts

Diameter (in.) Tension (kips)

25.9 6.44

37.3

50.7 12.6

16.5

20.9

25.8

31.2

Shear A490-N (kips)

13.5

19.4

26.4

34.5

43.6

53.8

65.1

77.5

Shear A490-X (kips)

17.9

25.8

35.2

45.9

58.2

71.8

86.9

103.4

Minimum Bolt Tension (kips)

24

35

49

64

80

Shear A490-SC (kips)

9.28

102

121

148

All values from AISC LRFD Manual (1986).

Table B.2. LRFD Values of for A325 Bolts

Table B.3. LRFD Values of for A490 Bolts

40

© 2003 by American Institute of Steel Construction, Inc. All rights reserved. This publication or any part thereof must not be reproduced in any form without permission of the publisher.

B.3 LRFD QUICK REFERENCE EXAMPLES EXAMPLE B.1 (Same as Example 3.3) 4-bolt Unstiffened End-Plate

ii. Check bolt bearing, end-plate controls, compression bolts:

Beam W24x55 A36 steel in. Check end-plate shear:

Column W14x90 A36 steel C. End-plate weld design, E70XX electrodes i. Beam flanges to end-plate weld:

A. Bolt design, A325-SC bolts

ii. Beam web to end-plate weld:

i. Tension:

ii. Shear,

B. End-plate design, A36 steel iii. Check beam web yielding

i. Bending, Equation 3.1b:

D. Column side, A36 steel and E70XX electrodes i. Check column web yielding, inequality 3.3b,

ii. Check column web buckling, inequality 3.4b,

41 © 2003 by American Institute of Steel Construction, Inc. All rights reserved. This publication or any part thereof must not be reproduced in any form without permission of the publisher.

EXAMPLE B.2 (Same as Examples 4.4 and 4.6) 8-Bolt Stiffened End-Plate Simplified procedure

iii. Check column flange bending,

Beam W33x118 A36 steel

Column W14x311 A36 steel

Design stiffeners and welds for

Column flange to stiffener weld, E70XX electrodes:

A. Bolt design, A325-SC bolts i. Tension:

Column web to stiffener weld, E70XX electrodes:

Check shear force in stiffener base metal, A36 steel:

ii. Shear,

iv. Check column web yielding, inequality 3.6(b):

B. End-plate design, A36 steel

E. Final details: i. Bending, Equation 4.4b:

ii. Check bolt bearing, end-plate controls, compression bolts:

42 © 2003 by American Institute of Steel Construction, Inc. All rights reserved. This publication or any part thereof must not be reproduced in any form without permission of the publisher.

C. End-plate weld design, E70XX electrodes i. Beam flanges to end-plate weld: Use full penetration groove weld with

reinforcement.

ii. Beam web to end-plate weld:

Column web stiffeners are not required.

iv. Check column web yielding, inequality 3.6b:

Column web reinforcement is not required.

D. Final details:

iii. Check beam web shear yielding:

D. Column side, A36 steel and E70XX electrodes i. Check column web yielding, inequality 3.3b,

ii. Check column web buckling, inequality 3.4b,

iii. Check column flange bending,

43 © 2003 by American Institute of Steel Construction, Inc. All rights reserved. This publication or any part thereof must not be reproduced in any form without permission of the publisher.

DESIGN GUIDE SERIES American Institute of Steel Construction, Inc. One East Wacker Drive, Suite 3100 Chicago, Illinois 60601-2001

Pub. No. D 8 0 4 (5M194) © 2003 by American Institute of Steel Construction, Inc. All rights reserved. This publication or any part thereof must not be reproduced in any form without permission of the publisher.