CIRIA Report 181 - Piled Foundations in Weak Rock

Who we are For almost 40 years CIRIA has managed collaborative research and produced information aimed at providing best practice solutions to industry problems. CIRIA stimulates the exchange of experience across the industry and its clients, and has a reputation for publishing practical, high-quality information.

How you can join CIRIA offers several participation options that have been designed to meet different needs. These include: •

Core Programme membership - for organisations that wish to influence CIRIA's collaboratively funded research programme and obtain early access to the results.

•

Project funding - for organisations that wish to direct funds to specific projects of interest. Project funders influence the direction of the research and obtain early access to the results.

•

New Books Club - popular with organisations that wish to acquire CIRIA publications at special member prices.

•

Construction Productivity Network - for organisations interested in improving their performance and efficiency through sharing and application of knowledge with others.

•

Construction Industry Environmental Forum - provides a focus for the exchange of experience on environmental problems and opportunities.

Where we are To discover how your organisation can benefit from CIRIA's authoritative and practical guidance contact CIRIA by: Post Tel Fax Email

6 Storey's Gate, Westminster, London SW1 P 3AU 0171 222 8891 0171 222 1708 [email protected]

Details are available on CIRIA's website: www.ciria.org.uk

Cover photograph: Piling for motorway bridge over the River White Cart, near Glasgow (courtesy Glasgow City Council). Printed and bound in Great Britain by Multiplex Medway, Walderslade, Kent.

Report 181

London 1999

Piled foundations in weak rock

J A Gannon G G T Masterton W A Wallace D Muir Wood

••

~ _

,horing knowledge. building be" pmctice

6 Storey's Gate, Westminster, London SW1 P 3AU TELEPHONE 0171 2228891 FAX 0171 222 1708 EMAIL [email protected] WEBSITE www.ciria.org.uk

Summary

This report provides guidance on the ground investigation requirements, preliminary and detailed design, and construction of piled foundations in weak rock. An appropriate definition of weak rock is developed and this leads on to consideration of the nature and properties of the material and how it should be investigated, characterised and classified for the purpose of pile design. Information concerning the ground conditions is collected in the geotechnical model. Guidance is given on the choice of pile types for weak rock. Several factors particular to the properties of weak rock and the installation of piles in weak rock will influence the initial selection of pile type and initial geometry of wellconditioned piles. Design of piles in weak rock under axial loads requires initial understanding of the mechanisms of load transfer down the shaft and through the base of the pile which are strongly influenced by the relative stiffnesses of pile and rock materials. Four published design methods are presented which analyse the deformation of the pile and the rock, and emerge with serviceability relationships between pile settlement and axial load. Worked examples are given for each method. The design of pile groups and of piles subjected to lateral loads and moments is considered briefly. The need for effective communication and free flow of information between structural designer, pile designer and piling contractor throughout the process of investigation, design and construction is emphasised and consequent implications for procurement practice are explored. A concluding section gives recommendations for improved practice, some of which should form the basis for future research and development. Five case studies are used to illustrate the main points of the report and the extensive bibliography includes reference to databases of case histories of piles in weak rock.

Piled foundations in weak rock Construction Industry Research and Information Association Report 181

© CIRIA 1999

ISBN 0 86017 494 8

Keywords

Geot~chnics,

piling, foundations, weak rock, pile design methods, pile selection

Reader interest

Classification

Geotechnical engineers,

AVAILABILITY Unrestricted

structural engineers, civil engineers

CONTENT

Best practice guidance

STATUS

Commissioned, committee-guided

USER

Geotechnical and structural engineers

Published by CIRIA. 6 Storey's Gate. Westminster. London SWIP 3AU. All rights reserved. No part of this publication may be reproduced or transmitted in any form or by any means. including photocopying and recording, without the written permission of the copyright-holder, application for which should be addressed to the publisher. Such written permission must also be obtained before any part of this publication is stored in a retrieval system of any nature.

2

CIRIA Report 181

Acknowledgements

This report is the outcome of a research project in CIRIA's ground engineering programme. The Babtie Group Limited undertook the research under contract to CIRIA. The authors of the report were Mr J A Gannon, Mr G G T Masterton, Mr W A Wallace and Professor D Muir Wood. In accordance with CIRIA practice, the project was guided by a steering group, which comprised: Mr R Fernie (Chairman)

K vaerner Cementation Foundations Limited

MrG Doe

AMECPiling

Dr D R Carder

Transport Research Laboratory

Mr J D Findlay

Stent Foundations Limited

Dr S R Hencher

then a/University of Leeds

Dr S M Springman

then a/University of Cambridge

Mr A J Turner

Ove Arup & Partners

Mr S Winter

then a/the Highways Agency

CIRIA's research manager for the project was Mr F M Jardine.

CIRIA gratefully acknowledges funding from the Highways Agency through the Transport Research Laboratory and contributions in kind from across industry. The authors and CIRIA are grateful for the help given to this project, not only by funders and members of the steering group, but also by many individuals and organisations, including: Mr N J Wharmby

AMECPiling

Mr R P Allwright

AMEC Piling

Mr C Harding

Babtie Group Limited

Mr A D Barley

Keller Colcrete Limited

Mr C A Raison

Keller Foundations Limited

Mr C R Heath

May Gurney (Technical Services) Limited

Dr J P Seidel

University of Monash

Mr J Whitworth

then a/Soletanche Limited

The authors also wish to thank Heather Cook for patient hours of typing and Allan Rice for preparation of figures.

CIRIA Report 181

3

4

CIRIA Report 181

Contents

SUMMARY ........................................................................................... 2 ACKNOWLEDGEMENTS .................................................................... 3 CONTENTS .......................................................................................... 5 LIST OF FIGURES ............................................................................... 7 LIST OF TABLES ................................................................................. 9 NOTATION ......................................................................................... 10 ABBREVIATIONS .............................................................................. 12

1

2

3

4

CIRIA Report 181

INTRODUCTION ........................................................................... 13 1.1

Background ................................................................................................... 13

1.2

Scope ............................................................................................................ 13

THE GEOTECHNICAL MODEL ................................................... 17 2.1

Definition of weak rock ............................................................................... 17

2.2

Nature of weak rock .................................................................................... 19

2.3

Properties of weak rock ............................................................................... 20

2.4

Investigation of weak rock ........................................................................... 25

2.5

Classification of weak rock ......................................................................... 33

2.6

Geotechnical model and design parameters ................................................. 36

PILE SELECTION AND INSTALLATION ..................................... 38 3.1

Choice of pile type ....................................................................................... 38

3.2

Installation and behaviour ............................................................................ 46

3.3

Guidance on sizing the piles ........................................................................ 50

DESIGN OF PILES IN WEAK ROCK ........................................... 53 4.1

Load transfer ................................................................................................ 53

4.2

Design methods: axial loads ........................................................................ 64

4.3

Piles subjected to lateral loads and moments ............................................... 78

4.4

Group effects ............................................................................................... 79

4.5

Design review .............................................................................................. 81

5

5

6

7

PROCUREMENT AND PERFORMANCE ..................................... 82 5.1

Communication between designer and constructor ..................................... 83

5.2

Specification ................................................................................................ 85

5.3

Control during installation ........................................................................... 85

5.4

Assessing compliance with performance criteria ........................................ 86

5.5

Monitoring and observation during installation ........................................... 87

CASE STUDIES ........................................................................... 89 6.1

Case" Study 1: determining design parameters for piles in mudstone .......... 90

6.2

Case Study 2: application of the geotechnical model- variability .............. 93

6.3

Case Study 3: design. construction and performance of driven H-piles in weak mudstone ........................................................................................ 96

6.4

Case Study 4: design of piles in layered Coal Measures sequence .............. 99

6.5

Case Study 5: procurement of bored piles in weak rock ........................... 102

RECOMMENDATIONS FOR IMPROVING PRACTICE .............. 105 7.1

Practical improvements ............................................................................. 105

7.2

Research and development ........................................................................ 106

7.3

Dual approach ............................................................................................ 107

REFERENCES ................................................................................. 108

A1 GEOMECHANICS CLASSIFICATION OF JOINTED ROCK MASSES (BIENIAWSKI. 1987) .................................................. 117 A2 PILE TYPES USED IN WEAK ROCK ......................................... 119 A3 CALCULATION EXAMPLES ...................................................... 120 A3.1 Calculation using method of Williams. Johnston and Donald (1980) ........ 122 A3.2 Calculation using method of Rowe and Armitage (1987) ......................... 125 A3.3 Calculation using method of Kulhawy and Carter (1992) ......................... 129 A3.4 Fleming (1992) .......................................................................................... 132

A4 LOAD TEST DATA ..................................................................... 135 A4.1 Summary of bored pile load tests on Categories 1 and 2 intermediate geomaterials .......................................................................... 135 A4.2 Summary of shaft and end resistance values used in design and/or verified by load testing of piles ...................................................... 138

6

CIRIA Report 181

Figures

CIRIA Report 181

1.1

Flowchart for the design and construction of piles in weak rock ........................... 15

2.1

Flowchart for the process of assembling the geotechnical model ......................... 16

2.2

Classifications of rock material strength ............................................................... 18

2.3

Modulus ratio ranges (values shown above as diagonal lines) for some Trias rock ................................................................................................................. 19

2.4

Influence of water content on measured uniaxial compressive strength of synthetic rock specimens ........................................................................................ 23

2.5

Depths of exploration .............................................................................................. 27

2.6

Correlations between "mass" and unconfined compressive strengths and penetration resistance of weak rock ........................................................................ 29

2.7

Comparison of initial elastic moduli values determined from pressuremeter tests with values deduced from pile tests ................................................................ 30

2.8

Example of correlation between point load and uniaxial compressive strength test results .................................................................................................. 33

2.9

Correlation between the in-situ modulus of deformability, Em, and RMR value ............................................................................................................... ;35

-----"""---..

3.1

Flowchart for the process of selecting the pile ....................................................... 39

3.2

Classification of piling systems .............................................................................. 41

3.3

Applicability of excavation method as a function of rock mass rating value or rock mass quality index ...................................................................................... 43

3.4

Excavatability of rock related to compressive strength and fracture spacing ...... .44

3.5

Comparison of total applied load versus displacement behaviour for piles with grooved sockets and non-grooved sockets .................................................... .47

3.6

Influence of rock structure on behaviour of displacement piles ............................ 49

4.1

Flowchart for the pile design process ..................................................................... 52

4.2

Idealised displacement behaviour of a drilled pier in rock .................................... 55

4.3

Development of subsequent shear plane ................................................................ 56

4.4

Design curves for unit side resistance and adhesion factor versus strength .......... 57

4.5

Dimensions of rough shaft for definition of roughness factor, RF ........................ 59

4.6

Side resistance reduction factor reflecting ratio of stiffness of rock mass and intact rock ......................................................................................................... 59

4.7

Adhesion factor as a function of rock strength ....................................................... 60

4.8

Simplified design charts for adhesion factor for Melbourne Mudstone ................ 61

4.9

Link between SPT N value and pile shaft resistance: tsu

4.10

Field test results for end-bearing pressure qb on class IV sandstone ..................... 63

4.11

Elastic settlement influence factor as a function of embedment ratio and modular ratio .................................................................................................... 68

4.12

Elastic load distribution as a function of embedment ratio .................................... 68

4.13

Allowance for nonlinearity in development of side resistance .............................. 69

=KsN ............................ 63

7

--............

8

4.14

Allowance for nonlinearity in development of base resistance ............................. 70

4.15

Design curves relating bearing capacity factor, Ns = qb/Em, to settlement ratio for different embedment ratios ....................................................................... 70

4.16

Design chart for a complete socketed pier ............................................................. 71

4.17

Elastic analysis of complete socketed pile (Poisson's ratio: 0.15 for pile, 0.3 for rock): (a) settlement factor, (b) proportion of load carried at base ............ 72

4.18

Predicted axial behaviour of a socketed shaft foundation in rock (L = 1.8m, D = 0.5m, Em = 100 MPa, Oc = 5 MPa, Poisson's ratio = 0.25 for rock) .............. 73

4.19

Interface strength properties deduced from field tests correlated to uniaxial compressive strength of intact rock: (a) cohesion, (b) composite friction and dilation .............................................................................................................. 75

4.20

Performance of pile sockets, 0.75 m diameter, designed for maximum settlement of 10 mm at working load of 4 MN ...................................................... 77

5.1

Flowchart for the process of procurement .............................................................. 84

CIRIAReport181

Tables

CIRIA Report 181

2.1

Examples of processes in the formation of some UK weak rocks ..................... 21

2.2

Engineering properties of some British mudrocks ............................................ 22

2.3

Typical engineering properties of Sherwood Sandstone (Nottingham Castle formation) .......................................................................................................... 22

2.4

Engineering properties of Mercia Mudstone .................................................... 23

2.5

RQD and its relationship to other rock mass measurements ............................. 25

2.6

Range of shear modulus values measured in weak rocks by the self-boring pressuremeter .................................................................................. 31

3.1

Factors governing the choice of pile type ......................................................... .40

3.2

Construction-related design assumptions for bored piles in Melbourne Mudstone ........................................................................................................... 48

3.3

Allowable stresses in pile materials ................................................................... 50

4.1

Values of X in expression for adhesion factor (equation (4.4) ........................... 58

4.2

Correlations between

4.3

Roughness clarification ..................................................................................... 58

4.4

Maximum bearing pressures for specified displacements for footings on sandstone ...................................................................................................... 62

4.5

Key attributes of four pile design methods ........................................................ 80

A 1.1

Classification parameters and their ratings ...................................................... 117

AI.2

Rating adjustment for joint orientations .......................................................... 118

AI.3

Rock mass classes determined from total ratings ............................................ 118

A 1.4

Meaning of rock mass classes ......................................................................... 118

A2.1

Pile types used in weak rocks .......................................................................... 119

A3.1

Summary of results of example calculations ................................................... 121

A4.1

Summary of pile load tests on Category 1 and 2 intermediate geomaterials .................................................................................................... 135

A4.2

Shaft and end resistance values for some pile load tests ................................. 138

t

su

.and a proposed by various authors ........................... 58 c

9

Notation

a Ab As b B C Cu

C d D

E E Eb Ee Ej Em Emd E, E' Icu

IE

Iy IT

F F I , F z, F3 G Gj Gm Go Gu, Id Ip

lR 15 (50) Ig j

K KE Ks L Ls Lt M, Ms n N Ns N60 Pa q qb qba qbe qbm

10

factor linking cohesion to uniaxial compressive strength area of pile base area of pile shaft factor linking friction-dilation expression (tan tamp) to uniaxial compressive strength width of foundation cohesion undrained cohesion constant used in deriving Hiley formula drop of pile hammer diameter of pile energy (hammer weight times drop) Young's modulus modulus of deformability of rock below base of pile Young's modulus of concrete intact rock modulus modulus of deformability for rock mass design value of rock mass modulus modulus of deformability of rock around pile shaft drained Young's modulus characteristic compressive strength of concrete partial factor used in design method of Rowe and Armitage characteristic yield stress of steel partial factor used in design method of Rowe and Armitage ultimate safety factor lumped safety factors shear modulus initial shear modulus shear modulus of rock mass dynamic shear modulus unload-reload shear modulus settlement factor plasticity index relative rigidity point load strength settlement influence factor ratio of rock mass modulus to intact modulus (Hobbs' rock mass factor) normal stiffness of pile/rock interface factor controlling load transfer for calculation of elastic shortening of pile ratio of shaft resistance to SPT N value pile socket length in weak rock socket length total arc distance down length of socket wall profile modulus ratio: Em/ae initial slope of shaft response function porosity standard penetration test value elastic bearing capacity factor standard penetration test value corrected for 60% free-fall hammer energy atmospheric pressure stress base stress allowable base stress elastic base stress allowable maximum base stress

CIRIA Report 181

!::.r !::.rm rs R Rs RJ -R6 SU

S

a (3 Ydry Ysat

Yrn

S T]

A v Vr

CIRIA Report 181

plastic correction to base stress pessimistic estimate of base stress base pressure to be used as normalising pressure average axial pile stress at surface of weak rock ultimate base stress total load allowable pile load base load allowable base load elastic estimate of base load ultimate base capacity design axial load total elastic load shaft load allowable shaft load elastic estimate of shaft load ultimate shaft capacity total pile load working load original cavity (socket) radius average height of asperities radial movement nominal socket radius pile load resistance group settlement ratio composite factors (Kulhawy and Carter) undrained shear strength pile set per blow of hammer field shear wave velocity laboratory shear wave velocity water content weight of pile hammer adhesion factor modifying factor linking shaft resistance and rock quality dry density saturated density partial material factor elastic settlement factor (Kulhawy and Carter) efficiency of energy transfer coefficient fracture frequency Poisson's ratio Poisson's ratio for rock settlement allowable settlement design settlement elastic shortening of pile uniaxial compressive strength radial stress increase initial radial stress effective vertical stress shaft stress elastic shaft stress plastic correction to shaft stress shaft resistance design shaft resistance ultimate shaft resistance angle of friction along shaft of pile multiplier in expression for adhesion factor: (4.4) angle of dilation along shaft of pile

11

Abbreviations

12

BS

British Standard

DVL

design verification load

FHW A

(US) Federal Highway Administration

IGM

inteimediate geomaterial

ISRM

International Society for Rock Mechanics

NGI

Norwegian Geotechnical Institute

RF

roughness factor

RMR

rock mass rating

RQD

rock quality designation

SCR

solid core recovery

SID

socket inspection device

SLS

serviceability limit state

SPT

standard penetration test

SWL

specified working load

ULS

ultimate limit state

CIRIA Report 181

1

Introduction

1.1

BACKGROUND In the preface to Piles in Weak Rock, M J Tomlinson (I977a) wrote: "very little research or observation has been undertaken on the problem of piles terminated in rocks". Despite the advances that have since been made in the geotechnical investigation of weak rocks and in the design and construction of piles, piled foundations in weak rocks still present particular difficulties to the pile designer and constructor: •

the geological controls on their formation are not given sufficient attention

•

they are difficult to investigate, sample and test

•

their behaviour is not well understood their properties may be significantly modified by the pile installation process.

There is a perception within the piling industry that design methods are too conservative and that assessments of design parameters are too pessimistic. This is in part because of inadequate understanding of pile behaviour and in part because the lack of a flexible and robust procurement process leads to uneconomic design and contractual disputes during construction. This report summarises current knowledge and practice and provides recommendations for improving this practice through appropriate geotechnical modelling, pile selection, pile design and procurement. Suggestions for future research are also given. The report emphasises that economy can only be achieved if a well-controlled process of designing and constructing a pile in weak rock can be followed.

1.2

SCOPE This report's guidance is drawn from international publications and the expert opinion of a steering group and industry. The text highlights potential piling problems and suggests methods by which these may be investigated and dealt with. The report is for geotechnical and structural engineers working in investigation, design and construction, and research. The coverage of the report is the design and construction of vertical and inclined bearing piles in which foundation loads are predominantly axial (compression and tension). Weak rock is defined in Section 2 by reference to geological characteristics, geotechnical properties and engineering behaviour. These are brought together to show the types of geological materials that are being considered so that a practising engineer engaged in pile design and construction can understand better the nature, properties and behaviour of weak rock.

CIRIA Report 181

13

The report guides the user through the several stages (see Figure 1.1) that make up the process of engineering of a piled foundation:

•

construction of a geotechnical model of the piled zone

• • • • • •

selection of appropriate pile type and installation method pile design and predictive analysis of behaviour procurement performance feedback reappraisal of design if appropriate.

Key points in the process of investigation, design and construction are illustrated by five case studies. The report draws on worldwide published experience of piling in weak rocks. The application of the report, however, is intended primarily for the UK.

14

CIRIA Report 181

:>----l_ piles not needed

0)

ASSEMBLE THE THE GEOTECHNICAL MODEL

revise model

CARRY OUT ANALYSIS OF STRUCTURE • structure loads, displacements • idealised pile loads, displacements

b) revise design

SELECT PILE TYPE, APPROXIMATE DIMENSIONS AND LAYOUT soil-structu re interaction

.-------~--------,

c)

CARRY OUT DESIGN CHECKS TO ARRIVE AT DETAILED DESIGN

PREDICT PILE INSTALLATION CONDITIONS AND BEHAVIOUR

d)

PREPARE SPECIFICATION (INCLUDING PROCEDURES FOR REVIEW)

INSTALL

test failureunforseen conditions

test failure

test pass

a) b) c) d)

See See See See

section section section section

2 3 4 5

ACCEPT

Figure 1.1 Flowchart for the design and construction of piles in weak rock

CIRIA Report 181

15

0

notes: • general arrangement of structure • structure loads • serviceability limits • desk study • preliminary ground investigation

RECOGNISE NEED FOR A PILED FOUNDATION AND PRESENCE OF WEAK ROCK

~ CARRY OUT MAIN GROUND INVESTIGATION

CONSIDER NATURE OF WEAK ROCK: • variable rockhead • variable weathering • variable structure • voids

""C

l..-

C

e

e

Q)

L...~

Q)

III

Q) ""C Q)

U

e Q) c>""C ·iii_ C

·c

I..-.r:.

U U ..... :::l 0

:::l .....

Q)

I..-

Q)

• • • •

boreholes, trial pits examination of exposures geophysical survey description and mechanical logging of strata for joint data

etc.

CONSIDER PROPERTIES OF WEAK ROCK : • uniaxial compressive strength • stiffness of rock mass • fracture state etc.

Tests on rock cores : • compressive strength • point load strength

CONSIDER LIKELY BEHAVIOUR OF WEAK ROCK : • time dependency • moisture sensitivity • degradability • excavatability etc.

Further tests on cores • plasticity • chemical composition Establish rock mass classification e.g. RMR

Tests in-situ : • pressuremeter/plate • SPT • geophysical

1;)01

~ CONSIDER RELATED GEOTECHNICAL FACTORS • overlying soils • groundwater • mining • seismicity etc.

Tests on overlying soil Piezometers and related observations inspection of abandonment plans, evaluation of seismic records and risk

GEOTECHNICAL MODEL

Figure 2.1 Flowchart for the process of assembling the geotechnical model

16

CIRIA Report 181

2

The geotechnical model

Early recognition of the presence of weak rock and the need for a piled foundation solution are essential for an effective site investigation to be planned and executed. For piling in weak rock the investigation involves three broad considerations: nature, properties and behaviour. The geotechnical model is the outcome of the overall process of investigation and interpretation of the ground conditions. It is the ordering of geotechnical data into a form to be used by the designer of the piles and supported structure and by the constructor. The process and its components are illustrated in Figure 2.1. The geotechnical model should address all of the ground and groundwater conditions for the intended piling site. The identification, characterisation and investigation of weak rock present problems:

•

weak rock is generally not investigated in its own right within a ground investigation (usually the soil borehole is extended into the weak rock, changing to coring techniques when the rock gets harder)

•

weak rocks are rarely tested except by the standard penetration test (SPT)

•

weak rock properties are difficult to measure and will be altered by the piling process

•

weak rock deformability is rarely considered and assessed the distinction between geological rock head and engineering rockhead is illunderstood and rarely made weak rock is heterogeneous, with many inhomogeneities, particularly over horizontal and vertical distances comparable to the length, breadth, and separation of pile sockets.

The following options are likely to help in addressing these problems:

2.1

•

overlapping the soil and rock investigation techniques (in adjacent boreholes) as a means of investigating the weak rock in its own right

•

considering the use of in-situ techniques such as pressuremeter tests and geophysical methods

•

considering the use of rock mass rating (RMR) as a classification system

•

developing a geotechnical model defining the geology, overburden, geometry, groundwater regime, properties of the rock and predicted behaviour. (Note that there is likely to be the need to consider a range of conditions on anyone site.)

DEFINITION OF WEAK ROCK Various terms have been used to describe weak rock: weak weathered and broken rock (BS 8004), indurated soil and soft rock (Oliveira, 1993), intermediate geomaterial (IGM) (FHWA, 1995), and soft rock (Johnston, 1989). Johnston points out that weak (soft) rock is part of the continuous spectrum of materials between rock and soil. In relation to soils, weak rocks are harder, more brittle, more dilatant and discontinuous. In relation to other rocks, they are softer, less brittle, more compressible and more susceptible to changes induced by variations in effective stress. Rocks are weak either

CIRIA Report 181

17

because the rock material is itself weak or because the mass is fractured. Thus the definition of weak rock has to account for material strength and mass structure. For the purpose of this report, the term "weak" is partly defined by a lower-bound uniaxial compressive strength of intact specimens of 0.6 MPa. Figure 2.2, adapted from Kulhawy and Phoon (1993) illustrates the historical perspective of the definition of the strength term "weak". "Weak" is defined by BS 5930: 1981 as a uniaxial compressive strength (oel of 1.25-5 MPa, measured on intact cylindrical rock specimens tested either dry or saturated. Following an approach suggested by the Geological Society Engineering Group Working Party (1995) redrafting of BS 5930 is likely to introduce a lower-bound uniaxial compressive strength for rock of 0.6 MPa and an upper-bound shear strength (su) for soils of 0.3 MPa. With the convention that Su = oc/2, there would thereby be continuity in field strength description between soil and rock. Very Weak Soil

-+-

Geological Society (1970)

Weak Rock

Brach & Franklin (1972)

Extremely Low

Soil

Very Low

Very High

Low \ Med.\ High

Medium

Weak

Bieniawski (1973) Kulhawy (1991 )

Strong

Proposed ReVision to BS5930

0.5

I

2

5

10

20

I

50

100

200

500

Uniaxial compressive strength Dc, MPa

Figure 2.2 Classifications of rock material strength (after Kulhawy and Phoon, 1993)

The other part of this report's definition of weak rock is a mass stiffness value not exceeding 30 GPa, which is less than the typical short-term Young's modulus of concrete. Where mass modulus equals or exceeds the above limiting modulus value of 30 GPa, piles are generally end bearing and pile performance is limited by the properties and behaviour of the pile material rather than the rock. For piles carrying tension loads, pile-socket shear strength is the overriding design consideration. In practice, weak rocks will commonly display uniaxial compressive strengths in the range 0.6-12.5 MPa and mass stiffness values of 100-1000 MPa. Some typical strength and stiffness values for weak rocks in the UK are given in Figure 2.3 (after Hobbs, 1974) with the solid diagonal lines representing modulus ratio, (M, = Em/oc).

18

CIRIA Report 181

.----------r---------.--r--r--~~~

o

__

I__

limiting upper modulus for weak rock 30000 MPa

1W~~~-~-~-~-~~-,~~~~~~~

0.... :::::?!

E

w

10

0.1

0.2

0.4

0.8 1

2

Cl:.

4

6 8 10

20

40 60 80100

MPa

Lower limit of uniaxial compressive strength. 0.6 MPa

Figure 2.3 Modulus ratio ranges (values shown above as diagonal lines) for some Trias rock (after Hobbs, 1974)

2.2

NATURE OF WEAK ROCK Weak rocks are either intrinsically weak (they have undergone a limited amount of gravitational compaction and cementation), or they are products of the disintegration of previously stronger rocks in a process of retrogression from being fully lithified and becoming weak through degradation, weathering and alteration. The various processes by which weak rocks originate are: •

deposition followed by gravitational compaction or consolidation (physical process associated with reduction in water content)

•

mineral changes and cementation (chemical change)

•

disintegration (physical and/or chemical degradation, which is commonly retrograde, for example tectonism, weathering and alteration).

Many of these rock-forming processes can be termed constructive, in contrast to disintegration. This dual origin is discussed by de Freitas (1993) and Knill (1993). The constructive processes apply mainly to sedimentary rocks whereas destructive processes apply to all rock types. Clayton (I993) suggests that the diagenetic processes (which include compaction, consolidation and cementation) influence the behaviour of rocks more than gravitational processes (compaction alone). Cementation (bonding) of the constituent particles of a rock is the major influence on material strength. Table 2.1 gives examples of the above processes with reference to some UK weak rocks. The interplay of these processes controls the nature of the rock. The object of an investigation should therefore be concerned with determining the nature expressed by:

CIRIA Report 181

19

•

past geological history, including depth of burial, degree of tectonism, etc

•

geometry of the rock mass

•

variation in rock type, vertically and laterally

•

rock structure Qoints, faults, fractures, bedding planes and voids)

•

physical, chemical and mineralogical properties.

Many piling problems arise out of the marked heterogeneity of weak rocks, particularly when they are derived from weathering. Thus, alternating sequences of relatively weaker and stronger rock, such as Carboniferous mudstone and sandstone or Mercia Mudstone with variable weathering profiles, present difficult piling conditions.

2.3

PROPERTIES OF WEAK ROCK The properties of weak rock provide quantitative support to engineering geologists' descriptions of rock cores and exposures. They may be used as indices of behaviour or inputs to a classification system and provide a link to key engineering parameters required for the design of pile sockets (ie compressive strength and mass stiffness). Typical engineering properties of weak rock are determined by the methods set out in Section 2.4.

2.3.1

Index properties Depending on rock type, some or all of the following properties are used to characterise the physical attributes of the rock material: •

whether it disaggregates in water

•

clay fraction (% < 2 !-im)

•

moisture content

•

density

•

Atterberg limits.

Other characteristics that may be relevant are swelling index, durability, soundness, abrasivity and solubility. In general, extreme values of properties, such as very low relative density or high porosity, give an indication of the sensitivity of the rock and its susceptibility to reductions in its strength and stiffness brought about by changes in water content and effective stress. These changes inevitably occur in excavations for replacement piles and where displacement piles are installed. For rock materials composed largely of clay minerals (clayey mudstone and shales and weathered volcanic rocks such as tuff and ash), behaviour is highly dependent on these external influences. Smearing of argillaceous materials on the sides of bored pile sockets can significantly reduce shaft friction. Information on the properties of mudrocks can be found in Cripps and Taylor (1981). Table 2.2, assembled from information contained in that paper, shows the link between extremes of the weathering state and several key properties of some British weak rocks. The arenaceous class of rock, particularly Sherwood Sandstone (formerly known as Bunter Sandstone), possesses a relatively high degree of porosity and correspondingly low dry density. This degree of porosity is reflected in other properties, notably uniaxial compressive strength (see Table 2.3). Siltstone, although treated in BS 5930 as an argillaceous rock, has close affinities to sandstone.

20

CIRIA Report 181

Table 2.1 Examples of processes in the formation of some UK weak rocks

..... ..

Geological formation

Rock type

Formational processes and characteristics

Potential engineering significance

Cretaceous Weald Clay

Mudstone with minor limestone and sandstone

Deposited in fresh water. Compacted but weakly cemented. Not well indurated. Can be highly weathered and softened

Low intact strength. Discontinuous. Low mass stiffness. May behave as a stiff overconsolidated clay. Susceptible to swelling and softening

Jurassic Lias Clay

Mudstone with both discontinuous and bedded limestone

Deposited in sea-water. Compacted but weakly cemented. Not well indurated. Contains shear planes and fissures. Surface may be softened and weathered

Low intact strength. Discontinuous. Low mass stiffness. May behave as a stiff overconsolidated clay. Susceptible to swelling and softening

Triassic Mercia Mudstone (Keuper Marl)

Mudstone with siltstone (skerries)

Deposited terrestrially. Some mineral change. Compacted but weakly cemented. Cements replaced by water-soluble salts. Salts removed to leave voids or loose material

Low intact strength and mass stiffness. Susceptible to collapse over voids, loosening and softening. Hard siltstone bands. Spatially frequent changes in degree of weathering and strength

Triassic Sherwood Sandstone (Bunter Sandstone)

Sandstone

As above but salts generally absent. Iron oxide cements weak and scarce - high porosity. Pebbly and mudstone beds

Generally low to medium intact and mass strength and stiffness. Strength dependent on moisture content

Penni an Magnesian Limestone

Limestone (dolomitic)

Deposited in seawater. Can be weakly cemented. May contain voids where gypsum, etc dissolved. Can be flaggy, reeflike or brecciated

Strength and stiffness low to medium. Collapse over voids possible

Carboniferous Coal Measures

Mudstone, sandstone, siltstone, seatearths, coals

Deltaic brackish water or marine deposition. Frequently alternating sequences of hard and soft beds. Mudstone compacted and cemented but still susceptible to weathering

Low to high intact and mass strength and stiffness depending on lithology. Coal workings. Mudstone susceptible to loosening, softening, remoulding. Seatearths fonn lateral shear zones

Carboniferous lavas

Lavas, volcanic ash, tuff and agglomerate, fossil soils, etc

Deposited subaerially. Density variable. May be unconsolidated (rubbly) or "welded"

Variable intact and mass strength and stiffness. Spatially frequent alternations between hard and soft beds possible. Can be expansive. Very susceptible to alteration and weathering

("Shillett")

Slaty mudstone/slate

Deposited in seawater. Compacted, tectonically compressed, indurated and cemented. Very closely fractured and fragmented

High intact strength. Jointed, faulted and closely discontinuous. Potentially low mass strength and stiffness. Susceptible to loosening

A link between weathering grade of Mercia Mudstone (formerly known as Keuper Marl), percentage clay fraction and strength and stiffness was determined by Seedhouse and Sanders (1993) for a Nottinghamshire piling site (see Table 2.4). Young's modulus values were obtained from measured shear modulus values using an assumed value of Poisson's ratio of 0.15-0.2.

CIRIA Report 181

21

Table 2.2 Engineering properties of some British mudrocks (after Cripps and Taylor, 1981) Formation

Water content

Liquid limit

Plasticit y index

W

WI

(%)

(%)

Ip (%)

Clay fraction < 2f.l

n

(%)

(%)

W

U

W

U

32-34

18-30

70-82

50-120

27-80

38-62

Jurassic Lower Oxford Clay

20-33

15-25

45-75

28-50

Triassic Mercia Mudstone

12-40

5-15

25-60

25-35

17-24 6-8 9-14 11

9-22 8 8 9

43-79 39-49 42-45 33-34

35-52 42 44-51 30-35

Cretaceous Gault Clay

Carboniferous Etruria Marl

Coal Measures Mudstone Shale

Porosity

Undrained shear strength Su (kPa) W

U

31-48

17-26

56-1280

30-70

30-54

52-93

96-1300

10-35*

10-50

10-50

70-200

130-800

8-32 9-19 12-19 13-41

12-25 24-53 37-87 33-77

21-35 2-25 3-30 3-28

40-240 15-335 15-335 15-335

120-620 9-103 MPa 29 MPa

Notes: * may be non-plastic W = weathered, U = unweathered

Table 2.3 Typical engineering properties of Sherwood Sandstone (Nottingham Castle formation) (after Bell and Culshaw, 1993) Dry density

Saturated density

'Ydry

'Ysat

(Mg/m.l)

(Mg/m.l)

(%)

1.83

2,09

14,2

2.3.2

Saturation moisture content

Effective porosity

Uniaxial compressive strength (dry)

Uniaxial compressive strength (sat)

neff

Oed

Oes

(%)

(MPa)

(MPa)

26.2

11.8

7.0

Strength and deformability Uniaxial compressive strength is the main index property by which weak rocks are classified and from which design values of ultimate shaft friction and ultimate end resistance for pile socket design are obtained. The influences of specimen size, moisture content (see Figure 2.4) and specimen orientation relative to rock fabric on observed test results are important, see Johnston (I991) and Dobereiner and de Freitas (I 986a,b). The point load strength test is frequently used to determine crushing strength through established empirical relationships. Johnston (I991) suggests that typical multiplication factors to convert from point load strength (Is(SO)) to uniaxial compressive strength are in the order of 7-10 for weaker rocks. These are somewhat lower than values of around 24 usually associated with this correlation (see Broch and Franklin, 1972). Therefore, sitespecific or formation-specific correlations based on testing are essential. For weak rock formations where driven samplers or penetrometers are used, as in highly weathered mudstone of a clay consistency or partially cemented sandstone, the measurement of shear strength can be expressed as effective cohesion and friction angle.

22

CIRIA Report 181

Table 2.4 Engineering properties of Mercia Mudstone (after Seedhouse and Sanders, 1993) Weathering grade

Clay content %

V & VI IV III

10 15 20 25 30

I & II

Undrained cohesion

Shear modulus

Young's modulus

Cu

G (MPa)

E

(kPa)

250 850 1330 1270 1230 1150 1090 1450

50 100 350 265 210 175 150 1270

115 230 820 615 490 405 350 2830

(MPa)

Effective stress approaches to pile design are being used for the design of piles in stiff overconsolidated soils that are almost classifiable as weak rocks. Research is confirming that such an approach is fundamentally more correct for weak rocks (Johnston, 1991). Numerical models can account for the influence of discontinuities on pile-ground performance by simulating the three-component behaviour of inter-joint friction: overriding, shearing and deformation (Johnston, 1991; Leong and Randolph, 1991, 1992 and 1994). The numerical models require that the rock stiffness be known. This information is a prerequisite to any rigorous approach to pile design.

c

8

0...

:::!:

•• x

J>

....c> ~

c

Q) I-

....

6

••

~

Ul

• Shearing mechanism x Splitting mechanism

•

Xx

Q)

> 'iii

)II(

Ul

~

~

~x x

a. 4 E

x)()()(

0

x

u

c ')( c

'c:J

, X

•

••

.,.

2

-c Q)

I-

X

:J Ul

c

X

Q)

:::!:

0 10

12

14

16

18

20

Saturation water content, w, %

Figure 2.4 Influence of water content on measured uniaxial compressive strength of synthetic rock specimens (after Johnston, 1995)

Rowe and Armitage (1987a) suggest that Young's modulus for static loading of the rock mass can be derived from the expression Em =215 G c for mudstone and sandstone without open discontinuities. For first estimates, this expression will be suited to many UK weak rock types. Whitworth and Turner (1989) show that a value of 275 G c is better

CIRIA Report 181

23

suited to the back analysis of bored piles in Sherwood Sandstone. Alternatively the range of values of modulus ratio (Mr =Em/ac) of 75-600 given in Appendix A of BS 8004:1986 may be used in conjunction with an appropriate value of Hobbs' rock mass factor (j). Intact rock modulus (E;) can be determined from strain measurement during uniaxial compression strength tests. Correlations attempt to link modulus of deformability and either RMR value (Boyd, 1993) or SPT N value (Clayton, 1995). It is preferable however for in-situ determinations of deformability to be made. These may be carried out using pressuremeters or by plate-bearing tests. The best method is to conduct preliminary pile loading tests and to back-calculate moduli, although boundary conditions and lithological diversity along the pile shaft can make the analysis difficult. Whichever investigative test method is employed to determine deformability, due regard should be paid to the strain increment over which the modulus values are derived. For working piles under static loading it is usual for mobilised stresses to be between onethird and one-half of ultimate stresses and for stress-strain curves to be distinctly nonlinear up to failure. Strains corresponding to these stresses are therefore appropriate for deduction of modulus values. In determining strength and deformability values for use in design, the concept of representative elemental volume (see Hudson, 1989) should also be considered. This concept recognises that the measured property of a rock is influenced by the volume of rock tested so that account is taken of the influences of discontinuities as well as material characteristics in the zone of ground stressed by the pile. In piling, it is desirable for design parameters to reflect the zones of ground affected by the single pile and the pile group. For end-bearing piles, the stress bulb beneath the pile toe may typically extend to a depth of about two pile diameters, therefore discontinuities within this element of the ground are important. For pile groups, much greater depths may be applicable where equivalent rafts are thought to apply (see Figure 2.5). Distributions of stresses around pile shafts are less certain, but it is suggested that the zone of ground within three pile diameters of the pile is influential in determining behaviour. In practical terms, the concept should be applied to measurements of rock mass modulus. For example, seismic methods will account for the influence of all discontinuities (at constant moisture content conditions) through which source vibrations are induced to pass, whereas the volume of ground tested by the pressuremeter is a function of the dimensions of its expansion mechanism.

2.3.3

Fracture state At a site where piling is expected to be required, rock will generally not be exposed. Thus considerable reliance may have to be placed on assessment of fracture state from borehole cores only, using values of solid core recovery (SCR), fracture index, and rock quality designation (RQD). RQD represents the proportion of the borehole rock core comprising sticks of solid core in excess of 100 mm length, each stick being separated by a natural discontinuity (see BS 5930: 1981). Section 44 of BS 5930 deals with fracture state and rock mass structure and how these may be described and measured. Expressed as a percentage of total coring length, RQD can be used to indicate rock quality where intact strength is relatively high. Weak rocks are, on average, of poor or very poor quality, although this can be markedly influenced by drilling method. Comparative values of RQD, fracture frequency, velocity index and Hobbs' rock mass factor (j) are given in Table 2.5. Whitworth and Turner (1989) note that the values of) for Sherwood Sandstone in Table 2.5 are generally lower than values that may be back-calculated from the results of pile loading tests.

24

CIRIA Report 181

Table 2.5 ROD and its relationship to other rock mass measurements (after Farmer, 1983)

Relation between RQD andj Quality classification

RQD

Fracture frequency

Velocity index

(%)

(per metre)

VF2/VL

Mass factor

2

j

Very poor Poor Fair Good Excellent

0-25 25-50 50-75 75-90 90-100

> 15

15-8 8-5 5-1

0-0.2 0.2-0.4 0.4-0.6 0.6-0.8 0.8-1.0

0.2 0.2 0.2-0.5 0.5-0.8 0.8-1.0

Note: fracture frequency is the number of natural discontinuities per metre length of rock core/rock mass Hudson (1989) gives an approximate relationship between RQD and fracture frequency (A) as: RQD

=

100e·o.lA (O.lA + 1)

(2.1)

This formula should be used with caution, particularly with respect to rocks with high fracture frequency, as it tends to overestimate RQD. Direct measurement is preferred. RQD and fracture frequency give an indication of the intensity of discontinuities within a rock mass. The character of discontinuities - in terms of their size, orientation, roughness, persistence, openness and infill, for example - is also of significance if rock mass modulus is to be derived from an RMR scheme. This can only be obtained by detailed examination and engineering description of rock exposures and core samples, although the latter, on their own, can only provide some of that information.

2.4

INVESTIGATION OF WEAK ROCK The data required to determine the nature of weak rock may be acquired by conventional site investigation processes and techniques.

2.4.1

Desk studies Desk studies are an effective and relatively inexpensive means of collecting a variety of data on regional and site-specific ground conditions. Geological descriptions of the rock succession from published maps or previous ground investigations can give an indication of the probable nature of the site.

2.4.2

Walk-over surveys In addition to providing information on surface conditions that may affect suitability of piling plant, a walk-over survey can identify the surface expression of deep-seated features. For weak rocks, examination of nearby exposures of the same rock formation thought to underlie the site can be useful if the formation is of the same structural domain. It is usually only through mapping of rock masses at exposure that a threedimensional image of rock structure, particularly discontinuity spacings and characteristics, can be assembled.

CIRIA Report 181

25

2.4.3

Exploratory boreholes For piled foundations,.weak rock will be at some depth, requiring the use of boreholes to sample and test the material. The overall depth of exploration will depend on likely socket lengths and pile group spacings plus several other factors. Fleming et al. (1992) recommend that the depth of exploration should be at least one-and-a-haJf times the width of the loaded area (1.5B) below the pile toes (see Figure 2.5). It is essential that exploration does not stop on reaching rockhead so that the boreholes clearly identify this upper surface of the rock. This surface and the underlying bedrock may, for contractual purposes, require precise definition, for failure to define and identify rockhead could lead to contractual disputes and claims. It is also important that water entries and standing water levels in open boreholes are consistently recorded. At least one piezometer should be installed in the weak rock at anticipated pile base level and readings from it should be taken regularly. In the UK, the routine method of sinking boreholes through overlying soils into weak rock is to use small-diameter cable-tool percussion. Boreholes are generally advanced into weak rock by use of weighted clay-cutter and chisels until the method becomes uneconomical or ineffective. This "soft ground" approach usually results in sketchy descriptions of the rock, based on whatever is recovered by way of disturbed samples. In very weak mudstone and sandstone and in the highly or completely weathered zones of other rocks, descriptions will usually follow the methods set out in BS 5930: 1981 for soil, not rock. Cable-tool percussion does not provide a completely satisfactory method of investigation of weak rock. For continuous sampling of the ground, rotary core drilling is required. Frequently, cable boreholes will be extended into rock using small-diameter pendant attachments. The total recovery percentage and the quality of the rock core produced are often poor although with good technique satisfactory results can be achieved in some formations. Rotary drilling with a rock bit and rotary-percussive drilling to form open holes without sample recovery are not suitable. The preferred method is to construct a separate adjacent borehole, which is sunk to a point above the upper surface of the weak rock using an open-hole drilling technique. From this level the hole is advanced by rotary core drilling employing: •

double- or triple-tube core barrels, Mazier barrels or similar

•

large-size core barrels and purpose-selected drilling bits

•

modest lengths of core runs

•

suitable flushing media

•

transparent semi-rigid core barrel liners.

The selection of equipment and method should follow the Specification for Ground Investigation (Site Investigation Steering Group, 1993) and should be based on the contractual requirement to provide SCR of at least 90 per cent and to give explanations for losses. Flush water and sediment returns should be recorded. The success of the drilling operation is critically linked to the skill and experience of the drilling team. A 2-m overlap between base of cable-tool boreholes and top of the cored section of drillholes should be targeted as suggested by Findlay and Brooks (1995).

2.4.4

Samples Good drilling technique makes full recovery of core samples readily possible in very poorly cemented, very weak and friable formations as demonstrated by Scarrow and Gosling (1986) and Reeves et al. (1993). Without near-full recovery of weak rock by good quality core drilling, pile design will be based on incomplete information.

26

CIRIA Report 181

I--B/21

Level of base of raft Level of base of piles

depth below raft 8

Depth

depth below raft 1.58 Depth below of piles

a) Vertical stress envelope for piled raft.

f---B/2

Depths below foundation Depths below 0.50 base of piles

L

8

~ 1.58

~

2.0B ~

~

b) Vertical stress envelope for isolated pile group.

Figure 2.5 Depths of exploration (after Fleming et al., 1992)

CIRIA Report 181

27

For weak rocks, it is important that the core is photographed and logged as soon as possible before it degrades, dries out or is damaged by handling. On-site tests such as point load tests can then be carried out, and the cores immediately selected and preserved for future laboratory index and compression testing. Cores should be stored carefully after logging so that they are available for examination by piling contractors tendering for construction. Close attention should be paid to the full engineering geological description of weak rocks explored by cable-tool and rotary coring methods. In the case of the latter, a full mechanical log including all the measurements recommended by BS 5930:1981 is essential so that the fracture state and quality of the rock mass can be ascertained. Emphasis should also be placed on assigning the correct unconfined compressive strength to the rock material. Sample descriptions which employ a wide strength range, for instance very weak to moderately strong (which implies Gc between 1.25 MPa and 50 MPa according to BS 5930) are unhelpful. The description of weathering grade as recommended by BS 5930:1981 should be followed unless there is a clear opportunity to apply an existing weathering classification for the particular formation being investigated. The following in-situ tests are appropriate for the investigation of the strength, deformability, permeability and other characteristics of weak rocks:

• • • •

standard penetration tests (SPTs) pressuremeter tests permeability tests geophysical measurements.

Plate-bearing tests are likely only to be used for very important structures because of the practicalities and high cost of carrying out the test at depth. Useful reviews of these insitu tests are given by Johnston (1991), Robertson (1986), and De and Saha (1993).

2.4.5

SPT The SPT has a variety of applications in weak rock investigation. It is generally used at the weaker end of the weak rock strength scale in conjunction with cable-tool boring and, occasionally, rotary core drilling. Adoption of the standard test procedure is advisable although Findlay and Brooks (1995), for example, suggest that the test drive should be terminated after 100 blows with a seating drive of 25 blows or 150 mm. In chalk, mudstone, siltstone and sandstone, SPT has been used as a classification tool where actual or extrapolated N values of up to 250 have been used. Although extrapolation of the N value should be avoided where possible, in practice it is often done and correlations with compressive strength have been deduced. Weak rocks can be classified by SPT as given by Clayton (1995), where N60 is the SPT Nvalue corrected to 60 per cent of theoretical free-fall hammer energy.

o .

RIPPING

o

BLASTING

I-

0 0

10 I-

a 5

0

-

0

00 0 0

I>.

)(

Q)

0

I>.

""C

.!: >.

~ 0

:l

0

1

-

0.1

I-

0

0

0 0

0

0

0

0.01 0

I

I

I

I

20

40

60

80

100

Rock moss rating value

Figure 3.3 Applicability of excavation method as a function of rock mass rating value or rock mass quality index (after Abdul/atif and Cruden, 1983)

In broad terms, digging can be considered a similar action to excavation by grab or shell whereas ripping is similar to the action of an auger or coring bucket. Digging is indicated to be possible for rock with an RMR value of less than about 40 (Q < 0.2 approx.), but ripping is appropriate for RMR values of about 40-60 (Q about 1-4). Of particular significance in assessing RMR values for an excavatability assessment will be the relationship between thc size and orientation of the excavation tool and the intensity, orientation and scale of discontinuities. This is accounted for by parameter B in the RMR system (see Appendix 1). Where there are insufficient data to assign RMR values, indications of economical excavation methods may be established from a consideration of uniaxial compression strength and fracture spacing or bed separation (see Figure 3.4). On this figure, point load strength values and uniaxial compressive strength values are not related in the way normally expected (Gc == 24 I s (50»)' As the relationships are based on formation of open excavations in rock, caution should be used in their application to pile excavation.

CIRIA Report 181

43

Uniaxial compressive strength. 0c. MPa

1.25

5

12.5

50

100

200

6 Very Thick

en

GJ ....

+'

BLAST TO FRACTURE

2

GJ

E 0.

Thick

0.6

en

'u en

Medium BLAST TO LOOSEN

0.2

::J 0

~

Thin

Cl I:

'C

0.06

RIP

lL.

0.02

GJ I:

.Q

a.

....GJ

+'

a a.... a a. GJ

I:

a a.

I:

:;;

Very Thin

"'C GJ

CD

DIG Laminated

0.006 0.1

0.3

3

10

Point load strength. IS(50) MPa

Figure 3.4 Excavatability of rock related to compressive strength and fracture spacing (after Franklin et al., 1971)

Bored piling is particularly suited for use in low permeability materials or dry ground where little influx of water is expected. Inflow of groundwater can be controlled by use of casings or bentonite support fluids. To minimise loosening of the ground and to counteract stress relief, careful attention should be paid to maintaining support fluids at the correct density and elevation in the borehole. Rapid excavation of pile sockets in weak rock is desirable so that rocks liable to deterioration in the pile bore wall are not allowed to swell, soften or become remoulded by the boring and casing process. Once excavated, the pile bore should be concreted immediately, preferably on the same day, Holden (1984). Dauncey and Woodland (1984) demonstrated that shaft adhesion values for bored piles in Mercia Mudstone could reduce during delays between excavation and concreting. Mini-diameter piles may be excavated by many of the methods used for replacement bored piles, and wall smoothness varies accordingly. Tremied or pressure-injected cementitious grouts are used; a degree of intrusion of grout into the ground can thus be achieved, and bond between grout and ground improved as a result. In these respects, certain types of mini pile are similar to ground anchorages and are often used as tension elements. Further details can be found in BS 8081:1989.

3.1.2

Small-displacement driven piles Steel universal columns and open-ended tubes and boxes are ideally suited to piling in weak rock when the relatively high cost of the pile material can be offset by other design and constructional advantages. The piles can be driven to virtual refusal provided the pile is not damaged. The piles are well suited to variable ground conditions. Hard points can be added to steel piles to aid both initial penetration into rock and subsequent straight driving, particularly if there is a sloping rock surface.

44

CIRIA Report 181

Small-displacement driven piles generally penetrate weak rock by fracturing and crushing it. Open or clay-infilled joints may be closed up by driving stresses. Thus, while horizontal ground stresses adjacent to the pile may be increased by compaction and dilation on shearing, the rock adjacent to the pile may also be destructured and remoulded. These two effects serve to oppose each other in terms of their influence on socket shear resistance. The mass stiffness of the broken rock around the pile, however, is expected to be lower than that of the undisturbed rock. The piles may also exploit steep discontinuities such as joints and faults so that some piles may be out of alignment or unexpectedly longer than those adjacent. If damage can be avoided, piles should be driven to designed penetrations into the weak

rock and not to satisfy a driving formula alone. As-constructed penetrations should be related to the geology and the driving records and the piles subjected to load tests to verify capacity. Correlations between the results of load tests and records of driving and penetration should be attempted where possible. Accurate predictions of the likely penetration of small-displacement driven piles into weak rock are difficult to obtain and rely largely on experience and judgement. Behaviour is to some extent governed by the ability of material - whether overlying soil or rock - to develop a plug within an open-ended tube or between the flanges of a steel column. Depending on whether a plug develops, the effective base area and therefore the load capacity of the pile in end bearing will vary. Where reliance is to be placed on endbearing resistance, plugging can be encouraged by the following measures: •

welding circular plates across the whole section of a steel tube above the end of the pile

•

welding cruciform or winged plates to the inside of a steel tube or to the web or flanges of a steel column at the toe of the pile.

These measures can also strengthen the pile toe so that high driving stresses can be accepted. As they also increase base area, piles need harder driving to reach designed penetrations. Jeffers (1995) noted that back-calculation of shaft stresses indicated that the effective shaft area for steel H-piles driven into Mercia Mudstone is given by 2BL, not the full perimetral area, where B = flange width and L = length of pile within the rock. Groundwater presents no problems to the piling process in itself although false sets caused by the temporary fall in pore pressures can occur during driving in mudstone and siltstone. Careful consideration should be given to the time period between driving and load testing or redriving.

3.1.3

Large-displacement, driven cast-in-place piles The structural characteristics of this pile type are broadly similar to those of replacement piles. Concrete is placed within a plugged and fuIiy cased hole in dry conditions and sound durable concrete shafts are achieved. Provision of ground support and exclusion of groundwater is inherent in the piling technique, and variability in ground conditions is catered for by the ready variation in pile lengths. In weak rock, the ability to drive the temporary steel casing and the charge of concrete or gravel which forms the driving plug may be restricted. In consequence, limited base enlargement and penetration into weak rock is achieved and this tends to result in a pile that is substantially end bearing within the weak rock layer. An example of the use of this type of pile is given by Jorden and Dobie (1977) for a Franki pile in Mercia

CIRIA Report 181

45

Mudstone at Redcar. Here, penetration of the grade II-grade IV weak rock was as much as 3 m but generally 1-2 m. Where penetration into weak rock is made, beneficial increases in horizontal stresses are achieved and softening of the weak rock is avoided by use of the charge of concrete. In common with driven piles, stress relaxation in weak rock can occur so that timedependent effects should be considered. Unfortunately, there are no methods of reliably estimating these stress changes during and after installation. Healy and Weltman (1980) review some of the potential problems associated with this type of pile. For noise and vibration associated with piling operations, reference may be made to Weltman (1980) and Head and Jardine (1992).

3.1.4

Large-displacement, driven pre-formed piles Of all the generic pile types, large-displacement piles require the greatest amount of driving energy to effect their installation. Noise, vibration and heave can cause problems. Unless the weak rock is towards the lower strength limit or closely fractured (eg Mercia Mudstone where precast piles are sometimes used), deep penetration is unlikely without overstressing the pile material. Without significant penetration these piles are predominantly end bearing, and therefore significant settlements are required to mobilise working loads. The structural capacity of the pile is thus under-utilised, and these pile types should seldom be used in weak rock. If they are used in weak rock, however, hard points should be added to aid penetration and straight driving.

3.2

INSTALLATION AND BEHAVIOUR Factors particular to weak rock and which influence construction and behaviour are examined separately for replacement and displacement piles.

3.2.1

Replacement piles Changes in ground stress during construction involve stress relief on excavation followed by partial recovery of original in-situ stresses under the weight of concrete placed. As the normal stress acting on the concrete/rock interface critically affects socket shaft resistance, quickly formed rough sockets with a minimum of ground relaxation are preferred. Leach and Thompson (1979) took account of stress relief in boring a pile shaft before back-analysing observed test settlements. Methods of counteracting the adverse effects of bore relaxation have been devised recently by use of new construction techniques, and thereby load capacity has been significantly improved. Haberfield et al. (1994), Hassan et al. (1993) and van Bijstervand (1993) describe the experimental and in-service use of chemical expanding grouts and additives to concrete. Robson and Wahby (1994) and Fleming (1993) describe the technique of base and shaft grouting in uncemented sands as a means of improving end and shaft load capacity and stiffness, thereby allowing use of shorter piles. The technique is also likely to be effective in weak rocks where grout can be injected into discontinuities, particularly those around the pile shaft. It should be recognised that improvement in base performance may be a consequence of the development of compressive stresses in the pile and rock immediately beneath from the pressure of grout introduced. Pile uplift may also cause reversal of shaft shear stresses and therefore an improvement in ultimate shaft capacity.

46

CIRIA Report 181

As the major part of the working load of a bored pile socketed into weak rock is usually carried in side shear, methods of excavation which achieve rough socket walls or techniques to roughen the walls are important. Kanai and Yabuuchi (1989) show how nodular insert piles provide superior frictional contact between steel pile material and surrounding concrete. This type of pile consists of a collared steel tube that is driven into the ground at the base of a prebored hole. The hole is then backfilled with concrete to surround the steel tube. Horvath et al. (1983) describe trials on piles bored into mudstone in which socket walls were artificially roughened with a grooving tool. Grooving, in conjunction with preloading the sockets by base grouting at pressures of 100-200 bar, increased ultimate loads and extended the stress range over which there is a quasi elastic response to load. Figure 3.5 illustrates the comparison between load capacity of grooved and non-grooved sockets. The degree of grooving is denoted by the term RF which represents roughness factor. This factor is defined in Equation 4.5 in Section 4. The walls of rock sockets for large-diameter bored piles for the strait crossing between New Brunswick and Prince Edward Island, Canada, were grooved to minimise socket lengths. A series of equally spaced circumferential grooves were cut using retractable groove cutters in the weak sandstone and mudstone (New Civil Engineer, September 1995). Total applied load, MN

0

0

2

3

4

5

6

7

8

4 ~

'0..

8

0

c..

....0

....c

12

,.... E 16

...

E ......

....c

Q)

20

E Q)

u ..Q c.. en

24

non grooved

socket----~~

0

28

.

32

l1li

Figure 3.5 Comparison of total applied load versus displacement behaviour for piles with grooved sockets and non-grooved sockets (after Horvath et al., 1983)

The load resistance of the sides and base of a pile socket also depends on the degree of slurrying of the wall and the amount of boring debris that falls to the pile base. Tomlinson (1995) noted that the amount of slurrying depends on the socket rock type, excavation method and equipment, and the nature of the overlying soil. Base grouting of bored piles appears to be particularly suited to increasing base performance where cleaning of base debris by air-lifting is ineffective or impossible by hand. Williams (1980) has shown that fresh concrete can force base debris a considerable distance up socket walls, and although this can result in a clean pile base, smearing of the socket walls might result. Where the socket is formed under bentonite, side shear resistance may be significantly reduced particularly if the socket is very smooth. A filter cake of bentonite can form along the sides of a socket during construction, perhaps

CIRIA Report 181

47

reaching a thickness of up to 100 mm in mudstone and sandstone (Rowe and Armitage, 1987b). For these reasons, rapid excavation of sockets is necessary and concrete mixes have to be carefully designed and pumped so that the filter cake is displaced and removed. The diameter of the pile may have an effect in this regard with smaller bores being easier to clean by virtue of correspondingly higher concrete flow velocities. Holden (1984) undertook an examination of construction-related design assumptions for bored piles in Melbourne Mudstone, and his conclusions are summarised in Table 3.2.

3.2.2

Displacement piles The success of a displacement pile depends on the ability to penetrate the weak rock without undue damage to the pile or reduction in the strength of the rock. Some fracturing and fragmentation is inevitable and should be considered when rock mass modulus values are computed using} or RMR values. Overdriving the pile should be avoided as it may result in excessive reductions in the strength and stiffness of the rock. Table 3.2 Construction-related design assumptions for bored piles in Melbourne Mudstone (after Holden, 1984) Phenomenon

Possible causes

Possible remedy

Comment

1. Socket diameter incorrect

Excessive overbreak or drilling tool smaller than casings

Stabilise using support fluids

Inspection by calliper. man entry or other means, eg socket inspection device (SID)

2. Unstable socket walls

Adverse pattern of fractures in rock, fast inflow of water. no or insufficient fluid or incorrect support fluid type

Use bentonite. not water. keep in good condition. maintain level at least 1 m above water-table, excavate quickly

Pregrouting to stabilise rock is uneconomic

3. Soft or smeared socket walls/ low wall roughness

Slow construction. equipment in poor condition

Speed up excavation, replace missing reamer teeth on drilling buckets, use grooving tools if necessary

4. Excessive base debris

Unstable socket walls. poor control on bentonite condition. low concrete density and flow rate

Condition bentonite and agitate, in dry bores. clean by hand if possible, monitor concrete parameters

Air-lifting may not be effective with large particles, conventional cleaning buckets can be ineffective

Penetration, however, is essential if a displacement pile is to perform satisfactorily under service loads - a pile in weak rock is never wholly end bearing. Therefore estimates of rock penetration should be made for displacement piles in much the same way as socket lengths are calculated for replacement piles. Good penetration also avoids lifting of the piles by heave in upper soil layers. The load capacity is influenced by the orientation and openness of the rock discontinuities as shown in Figure 3.6. With brittle rocks, overdriving may result in fracturing, which reduces resistance to load particularly at the pile shaft/rock interface. In this case, reinforcement of the pile point may be necessary, as the point carries a greater proportion of load. Shallowly inclined rock surfaces and joints generally aid load capacity and keep the required penetration small compared with steeply inclined discontinuities that may be opened and followed by the pile. Rehnman and Broms (1971) demonstrated this inclination effect for a range of rock types. Where there has been much crushing and fragmentation, pile capacity can be enhanced later by socket grouting. Joer et al (1994) describe the technique for trials in weak calcareous sediments where unit shaft friction values were enhanced by a factor of 10-20.

48

CIRIA Report 181

Open or clay-filled joints

Closed joints remain closed after pile driving

Joints closed by compression due to pile driving

(a)

(b)

il~~~i~ .

~

...

....... -':

~

_ _ ~,'-x·..::v.~.:

Joints closed by compression due to pile driving

Joints open

(c)

Rock strata dragged down by driving pile



c '0..

Check group stability and settlement ego by elastic continuum analysis

Rellised pile sizes. socket dimensions and group arrangement for construction

Figure 4.1 Flowchart for the pile design process

52

CIRIA Report 181

4

Design of piles in weak rock

The particular problems of designing piles in weak rock stem from:

• •

poor understanding of the real behaviour of piles in weak rock difficulty in modelling the interaction between end-bearing and shaft-load transfer mechanisms

•

pile design being generally based on the ultimate limit state (ULS), which gives no indication of performance under serviceability conditions

•

rock properties being modified by the installation of the pile, so that pile performance can be different from that predicted in design

•

pile design being generally based on the weakest parameters of the strata, whereas properties may improve laterally and with depth, leading to needlessly long sockets and over-conservative designs.

The solutions to these problems are: •

an improved understanding of the behaviour of piles in weak rock

•

a move towards the use of improved models of pile/rock-socket behaviour which take account of the interaction between base and shaft responses

•

the use of models which consider the pile performance at the serviceability limit state (SLS)

•

the adoption of flexible design procedures which permit the design to be modified in response to as-found conditions during construction.

Flowchart Figure 4.1 presents the design process in which different approaches are suggested for routine and non-routine designs.

4.1

LOAD TRANSFER The design of piles in weak rock requires performance checks to satisfy both ULS and SLS criteria. As in other areas of geotechnical design, a load-factor approach can be adopted to satisfy serviceability requirements implicitly through ULS calculations. However, the design procedures that have been developed over the past two decades attempt to perform more realistic calculations of behaviour under serviceability conditions. Such calculations require assumptions to be made about the ways in which load is transferred from the shaft of the pile to the surrounding rock: this load transfer mechanism is complex and, while some numerical and analytical procedures have been proposed on the basis of recent research, these are not yet sufficiently established for general design applications. For serviceability calculations, values have to be assumed for the stiffness of the rock around and beneath the socket. It is important that the ground investigation should produce realistic values of these stiffnesses. It also has to be recognised that the relevant values of stiffness and strength properties of the rock will depend on the procedures adopted for the installation of the pile and may be modified by the installation process. Existing design methods do not account for anisotropy of rock stiffness. Therefore if the rock is suspected to be significantly anisotropic, suitable account may be taken of this in the choice of stiffness values, see Section 2.4.

CIRIA Report 181

53

For piles in which much of the working load is carried by the rock at the base, for example, vertical stiffness may be particularly important. In the case of a pile in which the majority of the working load is generated in side shear, normal or horizontal stiffness (for a vertical pile) may be of greater significance. Weak rocks may also display significant heterogeneity, particularly sedimentary rocks with alternating layers of comparatively hard and soft materials, such as Coal Measures mudstone and sandstone. The various design methods can accommodate, in varying degrees, differing rock stiffness around and beneath the pile socket. Where there is heterogeneity within the socket length, a weighted average stiffness value may be appropriate: modulus values are assigned to each rock type and a weighted average value found according to the relative proportions of the rock types. Piles socketed into weak rock transmit their load to the rock through a combination of side shear and base loading. The nature of the load transfer will be controlled by the relative stiffnesses of the pile and rock and by the properties of the pile/rock interface. When measurements are made of the development of load with relative movement between pile and rock, it is typically found that the mechanism of shaft load transfer is initially stiffer than the corresponding mechanism of base load transfer (for example, see Williams, Johnston and Donald, 1980) but that a maximum shear stress is reached at a rather low relative movement. The distribution of load down the length of the pile socket will be strongly influenced by the pile stiffness: a very stiff pile will mobilise rather uniform shaft loads simultaneously with base loads, whereas a compressible pile will show a concentration of load near the top of the socket. Leong and Randolph (1994) and Seidel and Haberfield (1995) have reported numerical analyses in which the shaft resistance is calculated according to an interface model that introduces geometrical properties of the interface. While these models help to show the importance of shaft roughness in influencing the nature of the shaft load transfer, they are not yet fully developed into a form that can be applied in design. Although computer programmes that will predict individual and group pile response are under development, current design procedures are based on empirical assumptions of interface behaviour, usually combined with elastic analysis of the behaviour of the rock socket under load while making allowance for the stiffness of the pile. It will usually be necessary to design piles in weak rock for both ULSs and SLSs. An examination of ultimate capacity is often required as a starting point for serviceability calculations. It is normal to adopt a simplified assumption of separation of the components of pile capacity that come from shaft resistance and from base resistance in estimating the overall ultimate capacity. The ultimate axial load capacity of a pile is a function of many factors (not all of which are independent) which include: •

pile geometry (length and diameter)

•

shear strength and compressive strength of the rock

•

shear strength of the pile/rock interface

•

rock mass modulus (allowing for effects of discontinuities)

•

normal stresses acting on the pile shaft

•

pile socket roughness pile socket cleanliness (shaft and base).

Some of these factors are introduced explicitly in design calculations. Others. such as socket cleanliness, may influence the choice of values of material properties.

54

CIRIA Report 181

4.1.1

Shaft load transfer

When a pile is loaded, the initial transfer of shaft load through shear stresses on the interface may be described as an elastic process and can be analysed for a rock socket in the same way as for any other pile (Randolph and Wroth, 1978). However, for most real rock sockets - especially sockets that are drilled or bored - the interface between the cast concrete and the surrounding rock will be significantly rough, and this roughness will playa major part in the shear load transfer (Figure 4.2). Pile end I" socket die. D.. I

I

Socket dia. D+AD

Verticel displacement of pile

;.

Pile die. D

., Rough wall of rock socket



·iii Cll

"C

"-

Cll

C/l C

-+-'

-+-'

0

U

"C C

"-

::J U ::J

"-

+'

C/l

0 "Cll

c

c> ·iii Cll

"C

c>

c

a:::

• proof load tests • integrity tests • material tests

ACCEPTANCE

Figure 5.1 Flowchart for the process of procurement

84

CIRIA Report 181

5.2

SPECIFICATION Often, the only information available to the pile designer is a value for the working load in a pile and an instruction that a number of piles be load-tested to 1.5 times this value. An additional performance requirement of which the pile designer needs to be aware is in-service displacement criteria. For instance, if a piled foundation is designed to support a crane rail, the tight line and level tolerance requirements applied to the installation and use of the rail may dictate the tolerable pile displacement. The full consideration of group effects may also be an important aspect of the design process. The structure designer's main concem may be the performance of a building or bridge, say, supported by the piled foundation, and in many cases these concerns will not specifically include a preference for a particular type of pile or method of installation. In such a case the structure designer may be relying on a pile designer to provide a piled foundation which allows the superstructure to perform within specific criteria. The pile designer should therefore be fully aware of all in-service performance requirements. While load-carrying capacity is important, it is not the sole requirement of the design. Load-displacement criteria ought to be realistic and allow for compression of the pile material plus displacement of the pile. It should be remembered that notional pile failure involves displacements of, say. 5-10 per cent of the pile diameter. (Vertical displacements of less than, say, 10 mm at the design verification load (DVL) should only therefore be specified where absolutely essential.) As loading of rock may involve non-recoverable strain because of joint closure, significant residual settlement is to be expected for some rock types and should be considered acceptable. If the provision of a clean and rough socket is a fundamental requirement of the design, the constructor must be made aware of this requirement. Although it may be difficult to assess at the tender stage the full range and diversity of conditions which may be encountered on site, the tender documents should include some general guidance with regard to action to be taken in the event of different materials or behaviour being met. This would help to create collaborative attitudes for the work.

5.3

CONTROL DURING INSTALLATION Actual conditions on site rarely mirror those assumed in the design and it is therefore important that the conditions encountered are accurately recorded, interpreted, and responded to in an appropriate manner. The means by which conditions are recorded and interpreted should be agreed before commencement of pile construction. Records may include those items given in Section 5.5.

Specification for Piling (ICE, 1988) and Specification for Highway Works (HMSO, 1994) set out in detail other measurements that should be taken and recorded. The extent to which records are made will depend on, for example, the size of the piling contract and the familiarity with local conditions. Where responsibility for pile design and construction coincide, the need for good records is not diminished. The response to as-found conditions could be a change of plant to cope with obstructions, or perhaps a discussion with the designer to assess whether a higher-thananticipated termination is acceptable. Measures which maintain the quality of supervision, monitoring and reporting will give the designer confidence that the design assumptions are relevant to the conditions encountered.

CIRIA Report 181

85

If there are any aspects of the piling process which the designer considers to have invalidated the design, the constructor and designer should review options available for changing the pile layout, type or installation technique. The designer should also prepare for the possibility that the design assumptions prove unrealistic in practice, perhaps because of modification to the ground caused by the piling process. The designer should be prepared to adapt the design, perhaps at relatively short notice, as the installation progresses. Where unforeseen ground conditions are revealed, alterations to design and/or construction, if necessary, can be made quickly. Speed of decision is important so that piling conditions are not allowed to deteriorate, particularly in replacement piling. Entering the construction stage with a geotechnical model that encompasses an appropriate range of conditions and having associated designs already in place is recommended. tn addition, a reasonable degree of back-up in terms of plant and equipment is desirable, thus providing a means of dealing with the unforeseen without radically altering the type of piling adopted.

5.4

ASSESSING COMPLIANCE WITH PERFORMANCE CRITERIA Load testing of installed piles is the conventional means of confirming that the installation satisfies the specified requirements. A successful pile test confirms the constructor's compliance with the specification, and provides a lower-bound confirmation that the pile will perform satisfactorily in service. It gives only an indication of any excessive conservatism in the pile design unless the pile is tested to failure or near to failure. Pile tests are generally carried out on very few piles on any particular project. It is likely that ground conditions on site will vary from pile to pile, and the relevance of the pile tests undertaken is dependent upon how representative are those piles which are tested. Identification of testing requirements should be expected to evolve as the installation progresses and the full range of ground conditions, and any construction anomalies, are better identified. The number of tests called for should be influenced by the variability in the ground conditions and the types of piles adopted, not merely the number of piles required. It is preferable that load tests are performed on piles installed where the ground conditions are particularly adverse. Where time and cost permit, preliminary pile tests are recommended. These tests ideally load prototype piles to or near notional failure so that design values of ultimate shaft adhesion and base resistance and the load deformation response can be determined. The results may lead to the design of contract piles being altered up or down. As the loads to fail a pile can be very large, scaled-down model piles may be required. Model piles should be chosen such that: •

their installation method is the same as expected for the contract piles

•

socket length-to-diameter ratios are similar to those envisaged for the contract piles they strain a sufficient volume of the rock mass (particularly its discontinuities) to be comparable with the volume that will be affected by loads from the contract piles.

To achieve notional failure or to limit pile material stresses to acceptable values, stubby sockets (lower LID ratios) may be needed. In preliminary and proof load testing, the distinction between the DVL and the specified working load (SWL) should be recognised. In addition to unfactored structural loads, the DVL will also account for:

86

CIRIA Report 181

•

differences between test pile-head elevation and contract pile-head elevation

•

loads arising out of down drag that will operate only when the complete structure is erected.

Thus the DVL is never less than the SWL. The loading sequences adopted for maintained load tests and constant rate of penetration test are given in Specification/or Piling and Specification/or Highway Works. Proof load testing of piles tends to be cumbersome and time consuming, and is often disruptive to progress of the works. However, it is impossible to dispense with proof load testing except on the smallest of contracts where well-proven pile-ground combinations and, perhaps, very conservative designs, have been employed. The decision as to which piles should be tested should be made on site on the basis of as-found ground conditions or driving records - rarely will there be enough tests for a statistical approach to load testing to be adopted. Non-destructive tests can be performed to supplement load tests and to assess the integrity of installed piles (Turner, 1997). In the case of concrete in piles, direct visual inspection is rarely possible and therefore indirect measures of assessing concrete quality are necessary. The use of high-strength concrete grades is not always practical because the increased heat of hydration can be problematical in underground conditions. The use of lower grades of concrete, allied to the inability to inspect, suggests that integrity testing has a role to play. It is a relatively inexpensive procedure which helps in the detection of discontinuities, voids, necking, etc. Such tests are not a panacea, however, and do not provide direct data on pile capacity but will serve to enhance confidence in the installation process. A flexible approach to pile design and installation should be adopted, sympathetically reacting to changing site conditions as a project progresses. On larger projects this iterative process may be pre-empted by the pile design being tuned as a consequence of findings in a trial piling exercise.

5.5

MONITORING AND OBSERVATION DURING INSTALLATION Efficient design and construction of any type of structure will allow for details to change to reflect variation in site conditions. Piling should be no different in this regard. A range of ground and loading conditions should dictate a range of pile types, sizes, or installation techniques. In practice this may not be feasible because of the cost and time involved in, for example, changing rigs and casings. However, it may be possible to adapt certain aspects through observations made on site . In the case of piles in weak rock this flexibility could be attained by maintaining a universal pile size but varying the socket length to reflect changes in loading and site conditions. In order for such a responsive approach to be implemented, close liaison between the designer and the constructor is essential.

... ...

A management control system geared toward critical examination of the piling process is necessary in order that a rapid response can be given: modern piling rigs work quickly, and standing time awaiting decisions on refinements to the design to match ground conditions can outweigh the cost benefits of minor refinements. With more details of the design assumptions and performance requirements included in the contract documentation, together with advance consideration of proposals for different ground conditions, modifications could be dealt with more timeously as their need arises.

CIRIA Report 181

87

For this control system to be effective, observations and measurements should be identified in advance and monitored and reviewed continuously. Examples include: •

description of rock arisings

•

measurable properties of rock excavated from bored piles (eg using point load tests)

•

description of rock from cores taken to prove strata of the socket and beneath the pile base

•

penetration rate for driven piles

•

dynamic pile analysis during driving

•

rig rotation speed, torque, thrust, concrete flow and pressure, etc.

During the piling operation on site, the geotechnical model assumed by the pile designer, and which has been made available to the constructor, will therefore be critically reviewed and updated. In many cases changes will be incidental and neither require nor justify any reappraisal of the design. However, having the system in place will facilitate a speedy response to significant changes from the assumed model where modifications in the design would indeed be appropriate, whether for reasons of safety or economy. The measurement of key parameters during the piling process is not common in current practice, as there is little incentive for recording more than the minimal information required by most specifications. If, however, there was an acknowledgement by all parties that the design would be continuously reappraised on the basis of good-quality feedback during construction, there would be an economic benefit in improving the extent and quality of monitoring and testing during pile installation. The nature of piling rigs and the piling process suggests that, given this incentive, manufacturers would be encouraged to develop more reliable monitoring systems which might permit data to be gathered automatically during such operations as pile driving, sinking of bore casings, augering, and other related activities. For safety considerations in general and when man entry into water- or bentonite-filled pile bores is impossible, remote forms of examination should be considered. Various devices exist for pile bore inspection and these can detect the nature of the socket base and walls. Holden (1988) describes the use of a socket inspection device (SID) which is mechanically lowered into a pile bore to both inspect and, if necessary, clean the base. On large piling contracts, such equipment can be economically employed. From the above it is apparent that some form of observational approach to the design and construction of piles in weak rock offers potential advantages to all parties concerned. Taken out of a contractual context, such an approach is described by Sherwood et al. (1992) for bored piles designed and constructed on a Coal Measures site and where in-construction investigations and preliminary pile tests were performed. The observational approach was strongly recommended by the authors for this site and overall reductions in pile lengths claimed as a result. There exists a need for the development of a contractual form with a framework for establishing an observational approach to piling in weak rock. Further exploration of this theme, however, is outside the scope of this report.

88

CIRIA Report 181

6

Case studies

The five case studies illustrate some of the main points raised in the text of this report. The studies are concise and should not be considered as offering either a definitive or an exhaustive description of how to investigate, select or design piles. The studies present a reasonable blend of realism and detail, with current and recommended practice invoked. Case Study 1 shows how the principal design parameters were established for bored piles founded in mudstone. The study demonstrates the wide-ranging values attributable to a variety of tests and empirical correlations and therefore the need for careful judgement. The use of SPT has been omitted, but examples may be found in the literature (such as Thompson, Newman and Davis, 1993). Case Study 2 examines how rock sockets for bored piles were designed to cope with variably weathered Mercia Mudstone. The design recognised that actual rock conditions differed for each pile and therefore a mechanism for identifying the rock conditions and applying a design rule was used. The design methods chosen are not necessarily those suited only to the particular conditions encountered on that site, but they do consider load-displacement behaviour and the likelihood of achieving a clean pile base. Case Study 3 demonstrates the unreliability of set formulae in predicting the load capacity of driven piles in weak mudstone. A design analysis based somewhat arbitrarily on Rowe and Armitage is used to predict load-displacement behaviour and required penetration depths in the mudstone. The results of the analysis are compared with actual pile performance. Case Study 4 shows a procedure for designing bored pile rock sockets in layered Coal Measure sequence. The design analysis of Rowe and Armitage is employed and socket lengths deduced according to the relative proportions of rock type found in each pile bore. The socket lengths calculated for three typical cases are compared with those obtained by Fleming's method. This study, like Case Study 2, introduces an observational approach to pile design and construction. Case Study 5 focuses on the procurement of a piled foundation in weak rock in which close co-operation and ready exchange of information between the structure designer and pile contractor allows for an economic foundation solution to be achieved. Although the use of artificial roughening of the rock socket by grooving is described, this technique is not being proposed as necessarily advantageous in either a technical or commercial sense, and grooving is in fact rarely used in the UK. However, the effect of socket roughness on the available load capacity of the pile is amply demonstrated.

CIRIA Report 181

89

6.1

CASE STUDY 1: DETERMINING DESIGN PARAMETERS FOR PILES IN MUDSTONE

6.1.1

Background A bored pile is to be socketed into a Carboniferous mudstone described as dark grey thinly horizontally bedded slightly weathered weak with occasional tight planar 45° and subvertical fractures. Borehole logs record total core recovery of 90-100 per cent with RQD values of 0-30 per cent. Bedding plane spacing is generally 75-125 mm with slight discoloration noted on surfaces. Natural moisture content varies from 5 to 11 per cent. Boreholes were noted as damp during drilling with air. Standpipes subsequently recorded water levels some metres above rockhead. Values of Gc and Em are required for design.

6.1.2

Compressive strength (ae) A limited number of uniaxial compressive strength (DCS) tests and a greater number of point load tests on preserved cylinders and lumps of rock core were carried out, and a site-specific DCS-Is(SOl correlation established. Is (50) (MPa)

0.0

0.1

0.2

15

0.3

Site-specific correlation

20 IB (50)

•

12 moisture 9 content %

=

•

Is(50)

o

UCS

UCS

6

3

o

o

2

4

6

UCS, (MPa)

A design value of

6.1.3

Gc

= 4 MPa was selected.

Rock mass stiffness (Em) Five methods were used to establish Em values for design:

(a) Rock Mass Rating (RMR) The RMR was assessed from its components (refer to Appendix 1):

Rl R2

R3 R4

= 1 (1 < G c < 5) = 3 (RQD < 25) = 6 (5-8, thinly bedded, 75-125 mm) = 20-25 (weathered rough surface)

= 4-10 (presence of water) from which RMR = 34-45.

R5

90

CIRIA Report 181

r

Using Equation (2.3):

Em

= 3.5(RMR)3. 75 /1000 MPa = 1937-5541 MPa

Note: if discontinuities were open or infilled with gouge, R4 could be taken as 10 and RMR = 24 with a corresponding Em value of 525 MPa. (b) Rowe and Armitage (1987)

= 430 MPa

Em (c) Modulus ratio (BS 8004)

j is taken as 0.2 (see Table 2.5 of this report). Mr is taken as 250 for group 2/3 rock (see BS 8004 Table 4).

:. Em

= 0.2 X 250 X 4 = 200 MPa

(d) Pressuremeter test results G, (MPc)

o

o

400

BOO

1200 1600 2000

r---~---.----.----.---,

•

•

x Gur

2

depth m below rockhead

Gj

Design value G taken as 1DOOM Po

3

4

x

with U = 0.25 E 2G(1+U) E = 2.5G = 2500MPa

5 6

7 (e) Cross-hole geophysics Three in-line boreholes were used to carry out cross-hole measurements of seismic velocity. The resulting vertically polarised shear wave velocities and density values were used to define a small strain-shear modulus (Go): vs Q

=

1200-1750 mls

= 2-2.35 Mg/m3 (laboratory measurements)

Go = Q (v s)2

= 2880-7197 MPa

for v = 0.25, equivalent values of E are 7560 and 17 992 MPa.

CIRIA Report 181

91

6.1.4

92

Conclusions from Case Study 1 1.

1. There is confidence in assigning a value to Oc because of consistency between point load and ues test results and descriptions.

2.

There is wide variation in the estimation of mass modulus. Geophysics gives the highest result as expected (at the very small strains involved) whereas RMR estimates are considered to be dominated by a somewhat subjective assessment of fracture conditions and the dominant role this plays in arriving at the rating value. The modulus ratio method of BS 8004 gives a lower result and is considered overconservative because of use of cautious values of j and M, and perhaps a degree of inapplicability - as this method relates to weak and broken rock. The pressuremeter test results are consistent, but are likely to represent the horizontal stiffness which could be greater than the vertical value.

3.

Accordingly, a design value Em Em= 500 MPa.

= 1000 MPa is taken with checks using

CIRIA Report 181

6.2

CASE STUDY 2: APPLICATION OF THE GEOTECHNICAL MODEL - VARIABILITY

6.2.1

Background A bored pile foundation was selected for a new bridge. Rock sockets were anticipated in strata of Mercia Mudstone, which is characterised by lateral and vertical variability. The study shows how the geotechnical model was assembled and applied to the design and construction.

6.2.2

Geological sections (vertical scale approx. 1 :500)

BH1 Oc

...

RQD

Q)

BHS

>

il::

~

BH3

RQD

Dc . ~ .~': : .~~ ~

2 III 1.2 2.6 II

o

. :t., :'t.~"

"

~:

~

.,........

:

...

-,"

'

..•.

2 1.8 1.9 III 89 1.7 12

6 1.4 9 2.2 III 7 37 4.9 42

BHS Dc

RQD

Dc

BH6

RQD

RQD

Dc

.~

....

':',. , . :'to'

12 1.8 1.9 III 4 ~2 1.7

.'.

10 2.1 1.6 III 7 9 1.8 2.7 II 30 16 S.O 39

Notes: loose-made ground and sand and gravel overlies Mercia Mudstone, whose RQD and Oc (MPa) values are shown. Hard bands (skerries) are occasionally present. Rock mass factor Ii) values are generally about 0.2-0.3. Water entries are observed in made ground and sand and gravel. Water levels in piezometers installed in overburden and in rock are close to or above surface level of borehole 3 (ie river level).

CIRIA Report 181

93

6.2.3

Rock properties (from site investigation report and literature) Index property I and II

Weathering zone III

IV

2.5-2.3

2.3-2.1

2.2-1.8

2.4-1.9

2.1-1.8

1.8-1.4

Natural moisture content (%)

5-15

12-20

18-35

Liquid limit (%)

25-35

25-40

35-60

Plastic limit (%)

17-25

17-27

17-33

Plasticity index

10-15

10-18

17-35

% clay size (BS 1377)

10-35

10-35

30-50

Bulk density (Mg/m

3

)

Dry density (Mg/m3)

6.2.4

Design parameters These are based on rock characteristics by means of weathering zones and typical published values of Oc and Em supported by a limited amount of testing data.

Weathering zone

RQD

OC

(%)

(MPa)

Em (MPa)

IV

nla

1.1 *

70

III

10

1.9

135

II

25

2.8

360

40

5.0

600

* based on C u = 550 kPa and

6.2.5

Oc

= 2c u

Design methodology Design prepared for Q = 5 MN and 900 mm-diameter pile, Ec = 30 GPa, Qd = 15 mm. Methods of Williams, Johnston and Donald (WJD), Rowe and Armitage (RA) and a conventional method of employing a global factor of 1.5 on shaft resistance (calculated from RA expression for Lsu) are used. The following socket lengths were calculated on the basis of a single weathering zone with the final column showing the design socket lengths adopted. For weathering zones IV-II, by far the greater part of the load is carried in side shear, and little reliance can be placed on base resistance. In zone I material, excavation by normal bucket auger is considered difficult and, therefore, minimal penetration and full reliance on base resistance is assumed.

(WJD) Length (m)

(RA)

Qsu 11.5

L

Length (m)

(m)

(m)

IV

8.5

8.5

5

8

III

5

4

4

4

II

3

2.5

3.5

3

2*

1

3

1**

Weathering zone

* settlement criterion also satisfied with L = 0 m ** or 3 if thickness not> 2D 94

CIRIA Report 181

6.2.6

Construction design review mechanism A review mechanism was instituted which allowed for socket acceptance on the basis of the relative proportions of differing weathering zone material encountered and the associated design socket lengths. Four examples are given below: Succession

Load contribution (MN)

Socket length required (m)

(a) 2 m of zone III

2/4 X 5 = 2.5

2

then 3 m of zone II

3/3

+ (5 - 2.5)

then 2 m of zone III

Not required

(b) 4 m of zone IV

4/8

X

5 = 2.5

4

then 1 m of zone II

113

X

5 = 1.67

+1

then 3 m of zone III

3/4x5=3.75

+ (5 - 2.5 - 1.67) say 6

(c) 2 m of zone III

2/4 X 5 = 2.5

2.5

then 3 m of zone I

311 x 5 = 15.0

Pile terminated on zone I after rock coring proved thickness >2D

(d) 1 m of zone II

113 x 5 = 1.67

then 1 m of zone I then 5 m of zone II

113 x 5 = 1.67 5/3 x 5 = 8.33

X

5 = 5.0

X

3/5 = 3.5, say 4

X

4/5 = 5.66,

+1 + (5 - 1.67 - 1.67) x 3/5 = 3

Recognition of weathering zones was made on the basis of a geologist's description of pile bore arisings supplemented by point load tests on irregular lumps of rock and a record of piling rig rates of penetration.

6.2.7

Discussion and conclusions from Case Study 2 A geotechnical model recognised that the following factors would playa crucial role in pile design and construction: •

variable rockhead levels

•

lateral and vertical variation in rock grade

•

unpredictability of the rock grade at pile base level.

These and other factors, such as the presence of skerries and overlying sand and gravel and made ground, influenced the choice of a partially cased auger-bored pile. More importantly, a design rule was formulated for the on-site acceptance of sockets which vary in length as a function of the rock types encountered in the pile bore .

.... ....

CIRIA Report 181

95

6.3

CASE STUDY 3: DESIGN, CONSTRUCTION AND PERFORMANCE OF DRIVEN H-PILES IN WEAK MUDSTONE

6.3.1

Background The end spans of a new motorway bridge-deck were supported by single rows of 10 no. 305 mm x 305 mm x 186 kg/m universal bearing piles spaced at 1.8-m centres. The piles, each about 10 m long, were driven from the base of 3 m-deep 600 mm-diameter boreholes through 4 m of made ground and 2 m of glacial till to mudstone rock. Information about the rock was sparse, being based on two nearby boreholes that described the material as weak to moderately weak with occasional bands of sandstone.

Brid

SECTION (NTS)

deck

Legend

Compressible void filler

-~--------

- -

-77h..~-

--- -

----

-US pile Mudstone

6.3.2

Design The pile was chosen to form an integral abutment in which resistance to lateral forces at the pile head would be small, to give minimal horizontal restraint to expansion and construction of the bridge-deck. Sufficient lateral restraint in the ground was expected from fixity at depth in the glacial till. Stringent performance requirements were imposed for axial load behaviour:

= 1300 kN

•

working load Qw

•

settlement at 1.5 Qw

•

residual settlement ~ 5 mm.

~

elastic pile compression plus 5 mm

These reflect the transverse continuity between piles and deck, and allow for potential effects of creep in the mudstone. The adoption of universal bearing piles was strongly influenced by the structural requirements of the foundation, namely the need for bending strength and flexibility. Environmental and cost factors were of secondary importance. Relatively little regard was paid to the character of the weak rock, and the pile was considered to derive resistance to axial load primarily from end bearing. No static design analysis was carried out. The designer selected the pile lengths on the basis of nominal penetration of the pile into the weak rock.

96

CIRIA Report 181

6.3.3

Construction and performance The contract for the pile construction required the piling contractor to propose a pile set. One pile in each abutment selected by the designer was to be load-tested to 1.5 Qw. The initial pile set was calculated from the modified Hiley formula with the desired resistance R being taken as 2 Qw giving: S (set per blow) with E =

Yj

= EIR -

CI2

Wd and C taken as 10 mm.

For a W = 5-tonne BSP HH5 hydraulic hammer operating through a drop of d = 600 mm and an assumed hammer efficiency of Yj =0.8, this led to a rounded set of ten blows for 43 mm or less penetration. The result of an incremental loading test on a pile that had been driven to a set of 10 mm for the final three increments of ten blows (ie 30 blows, 30 mm) is given below. It was not known if the pile had penetrated rock to any significant extent.

2000

1600

1200 Load kN

BOO

400

10

20

30

40

Settlement (mm)

6.3.4

Design and construction review The review of the design and installation focused on the following:

CIRIA Report 181

1.

Re-examination of set calculations.

2.

Site investigation to determine position of rockhead and depth of pile penetration into rock.

3.

Dynamic analysis of re-striking undertaken to a revised set, particularly to determine efficiency of energy transfer (Yj).

4.

Calculation of pile performance based on static analysis.

5.

Consideration of delay periods between driving and load testing.

6.

Load testing after re-striking.

97

The results of the review were as follows: Review parts (1) and (3). The assumptions of 11 = 0.8 and C = 10 mm were challenged. Dynamic analysis showed the maximum efficiency to be equivalent to 11 = 0.7. Taking a value of 15 mm for C and 11 = 0.7 was to change the calculated value of S to ten blows for 6 mm penetration. Re-striking with a revised hammer drop of 1.2 m and a minimum penetration into rock of 1.5 m was recommended and subsequently carried out successfull y. Review part (2). Investigation showed initial pile penetration into rock to be negligible and after re-striking to be 1.5-3 m. However, the slope of rockhead was shown to be locally as much as 30° from horizontal. The pile load test result is indicative of a shaft friction failure which also points to a lack of rock penetration or even possible heave caused by adjacent piling. Review part (4). The method of Rowe and Armitage was employed (as given in Appendix A.3.2 for a regular clean socket) with G c = 6 MPa, Ab = D2 and As = 2DL where D = 0.305 m. Allowing for 200 kN of shaft resistance developed over 4 m of till, the ultimate capacity of a pile with a 2 m-long rock socket was calculated to be 2900 kN. For base stresses to be limited to G e , a socket length of 1.5 m was required which indicated settlement of 3.5 mm. Review part (5). Periods of one month between initial driving and static testing elapsed. This period was considered sufficient for false set or set-up phenomena to be unlikely. Review part (6). Load testing of a re-struck pile with a 2 m-long socket showed a linear response to a load of 1950 kN with settlement of only 5 mm (including elastic compression) and a residual settlement of less than 1 mm. CAPW AP analysis indicated a possible ultimate load of 3300 kN.

6.3.5

Conclusions from Case Study 3 The Case Study demonstrates:

98

1.

Driving formulae are only a rough guide to ultimate pile capacity and are not a substitute for socket length design. Assumptions regarding key parameters in a set calculation critically affect the value from which a set criterion is derived.

2.

Piles driven into weak rock should be designed using principles of static equilibrium. For steel universal bearing piles, fully plugged into a mudstone, pile base area Ab and shaft areaA s taken as D2 and 2DL respectively proved to be appropriate. Stringent performance limits can be estimated only by an analysis method based on rock mass stiffness.

3.

Steel piles driven into weak rock should penetrate several metres but in so doing may disintegrate the rock. Variability of the level of the rock surface should be ascertained when selecting pile lengths.

4.

It is unlikely that concrete piles could have been driven as hard as the steel piles selected.

CIRIA Report 181

6.4

CASE STUDY 4: DESIGN OF PILES IN LAYERED COAL MEASURES SEQUENCE

6.4.1

Background A bored pile foundation is required for a building structure. Ground conditions comprise: made ground, peat and soft compressible clay, overlying Coal Measures bedrock. These strata are lithologically variable both laterally and vertically, and engineering rockhead appears in a range of materials as the result of deep glacial action. A commonly occurring bedrock sequence on the site and the design parameters of the various layers are given below. The coal is intact.

6.4.2

Strata sequence and design parameters Nome

Depth (m)

Oc (MPo)

Eb

(MPa)

Er (MPa)

0 Sandstone

15

1000

1000

3 4

200 150

400 200

10

500

700

3

200

400

2 Mudstone

4

Cool Siltstone

6

8

Mudstone

10

The rock mass modulus values apply to the pile base (E b ) and shaft (E r ) and are selected to take account of rock structure and the designer's assessment of likely socket cleanliness and roughness. In this example, stiffness values are not factored further to get to design values (E md ) as suggested by Rowe and Armitage.

6.4.3

Design approach The design approach of Rowe and Armitage (1987a,b) is used (see Appendix A.3.2 of this report). Reference should also be made to their paper, which describes the simple method of dealing with soft seams. Other methods that take account of differing base and shaft stiffness (eg Fleming) could also be used. Design input parameters are:

D

= 6MN = 0.9m

Qd

=

Q

Ec L

= 100E, = 3D (estimate at start)

10 mm (not including elastic compression)

A single bored pile socketed into the rock is selected as the foundation for each of a series of columns. Three possible layer configurations are examined as follows:

CIRIA Report 181

99

a

b

c

Engineering rockhead occurs at top of upper mudstone in sequence shown

Engineering rockhead occurs within lower mudstone. It is suspected that a coal seam may be present beneath

Engineering rockhead occurs in a mudstone with a sandstone at 1.5-m depth. The thickness of the sandstone bed is not certain

Average design parameters for a 3 m-long socket deduced: Oc :::

(1.5x3+0.5x4+ lxl0)/3 = 5.5

Oc :::

3 MPa

As case b but modified by presence of sandstone

MPa E r ::: (1.5x400+0.5x200+ lx700)/3

Eb::: 1000 MPa

E r ::: 400 MPa

::: 467 MPa Eb = 200 MPa

Eb ::: 500 MPa

ts ::: 0.450c 0.5

:::

1.06 MPa

ts ::: 0.78 MPa

tsd::: 0.7ts::: 0.74 MPa

tsd = 0.55 MPa

Figure 14(d) of Rowe and Armitage (1987a) (not in this report)

Lmax::: 6/(Jt.xO.9xO.74)

Lmax::: 3.86 m

Id = (0.0Ix400xO.9)/6 = 0.6 which is out of range of curve

:::2.87 m Lma/D::: 3.19 m

Lma/D::: 4.29

Id ::: (0.09x467)/(6xl000)

Id::: 0.6

"" 0.7 Figure 13(d) of Rowe and Armitage (1987a) (not in this report)

Figure 12(d) of Rowe and Armitage (1987a) (not in this report)

by inspection

Id is too large, ie out of range. It implies that dimensions of socket are too large for geometry/loadsl stiffnesses assumed:

Q'Q::: 0.41, UD

Q'Q

Options:

L::: 2.7xO.9

a - reduce settlement

= 2.7

~

100% as UD ~ 0

Hence no penetration of the sandstone is required

::: 2.43, say 2.5 m

:. a 1.5 m-long socket is satisfactory, subject to acceptance checks

b - reduce stiffness c - reduce diameter d - increase load Reduce settlement to 5 mm

Qb::: 0.41x6

::: 2.46 MN

Id::: (0.005x467xO.9)/6 ::: 0.35

Qb::: 2.71A b

::: 3.87 MPa ie> 3 MPa (oc)

Lma/D::: 2.87/0.9::: 3.19

Qba ::: qbaAb

::: 3xO.636 ::: 1.9 MN For line on Figure 13(d) and

Hence,

Id::: 0.35,

Qb/Qt::: 1.9116 ::: 0.32,

Qb/Qt::: 0.29 and

Id = 0.52, and

UD=2.2

UD=3

100

CIRIA Report 181

a

b

Hence

c

:. L=3xO.9=2.7m

Qb = 0.29x6 = 1.74 MN qb = 1.741Ab = 2.74 MPa L

= 2.2xO.9 = 1.98 m,

say 2 m, ie on top of siltstone qba

=

Oc

= 10 MPa > qb OK

g

=8.7mm qba

qb

length of socket has to be increased to bring base load down to allowable value:

qb > qba so base area is limiting and revised socket geometry required

= 2.8/(rr x 0.75 2/4) = 6.34 MPa

Qba = 4.0 x (rr x 0.75 2/4) = 1.77 MN

The design line drawn on the design chart (Figure 4.16) in Step 6 is then used with Qb/Q, = Qba/Qd To find required socket geometry UD (and hence socket length) and corresponding settlement influence factor Id (and hence actual settlement).

and from line in Figure 4.16, LID = 2.0 implying L = 2.0 x 0.75 = 1.5 m and Id =0.42 implying settlement Qd

= (0.42 x 4.0)/(301 =7.44mm

x 0.75) 5

8 Maximum base stress: qbu ::; qbm = 2.5oc

where qbu calculated from design load using pessimistic estimate of shaft resistance tu = Ysts with Ys = 0.3.

tu

= 0.3 x 0.90 = 0.27 MPa

Then qbu = (4.0 - 0.27 x rr x 1.01 x 0.75 X 1.5)/(rr x 0.75 2/4)

= 6.89 MN allowable maximum base stress qbm = 2.5 x 4 = 10 MPa hence design socket is satisfactory Notes The design method of Rowe and Armitage is entirely based on finite element analysis of the behaviour of piles socketed into weak rock. In summarising the results of these analyses in design charts, the rock and the pile are both treated as isotropic elastic materials and the interface between pile and rock is modelled with a purely cohesive description of the mobilisation of shaft resistance, although Rowe and Armitage (1987a) discuss the effects of describing this interface using a frictional model with dilation. The underlying analysis is perhaps the most thorough of all the methods presented here, and it is left for engineers to introduce their experience through various partial factors for which Rowe and Armitage suggest recommended values. Although intended to give as output a pile socket geometry satisfying the input constraints, the method can be repeated with a series of input settlements to give a complete relationship between load and settlement. Such a load:settlement relationship, calculated for a pile socket 1.5 m long and of 0.75 m diameter, is shown in Figure 4.20.

126

CIRIA Report 181

Rowe and Armitage suggest values of 0.7 for the partial factors 'YT and 'YE on the basis of statistical studies which have shown that the probability of exceeding a specified design settlement is then less than 30 per cent. They suggest that these reduction factors should be used even when the design of the pile socket is based on parameters back-analysed from field tests on prototype piles.

1.

The coefficients 1.421 and 1.90 in the expression for shaft resistance correspond to values of X equal to 2.01 and 2.68 respectively in Table 4.1 and Equation (4.4). 2.

The entire output of the analyses that Rowe and Armitage (1987a) have performed is encapsulated in a series of design charts of which Figure 4.16 is a typical example. Separate charts are provided for values of the modulus ratio EJEmd equal to 10,25,50 and 100 and for ratios of rock stiffness beneath the base and around the shaft of the pile equal to 0.5, 1.0 and 2.0.

3.

The curves drawn in Figure 4.16 for different values of Id assume that slip is occurring at the interface between the pile shaft and the rock. If the specified settlement implies a high stiffness of the pile, it may not be possible to find an intersection between the curve for the appropriate value of Id and the line drawn in Step 6. For example, the design can be repeated with design settlements Qd equal to 5 and 2.5 mm as shown below.

4.

Rowe and Armitage state that this selection of allowable base stress qba is intended to ensure that the rock beneath the base of the pile behaves elastically under design conditions. It is appropriate where (i) the base of the socket is at least one pile diameter below the rock surface, (ii) the rock to a depth of at least one pile diameter beneath the base of the pile is either intact or tightly jointed with average uniaxial compressive strength OC' (iii) there are no solution cavities or voids beneath the pile.

5.

Rowe and Armitage state that statistical studies show that use of 'Ys = 0.3 implies that the probability of the shear resistance being less than Lu is less than 0.5 per cent. For lightly loaded piles or piles with long sockets, the value of qbu may be very small- as demonstrated in Note 6b.

3a For design settlement Qd

=5 mm

Steps in design process

Trial pile design

6a Discover from chart in Figure 4.16 that design settlement lies outside area covered by curves for different settlement influence factor. Use chart of: Id (LlD,EcIEmd) From elastic analysis (Figure 4.17a) to deduce required socket geometry to support load with required settlement but without slip along pile shaft

Id = (0.005 x 300 x 0.75)/4.0 = 0.282 seek intersection with curve EciEm = 100 LID = 3.2 L = 3.2 x 0.75 = 2.4 m

Note

6b Use chart of: Qb/Qt(LlD,EcIEmd) From elastic analysis (Figure 4.17b) to deduce proportion of load carried by base of pile

Qb = 0.2 x 4.0 = 0.8 MN

7 Allowable base stress qba =

qba

CIRIA Report 181

Oc

= 4 MPa 2

qb

= 0.8/(rr x 0.75 /4) = 1.81

qb

< qba so socket geometry satisfactory

127

Steps in design process

Trial pile design

8 Maximum base stress qbm ~ 2.50c where qbm calculated from design load using pessimistic estimate of shaft resistance LU = YSLS with Ys = 0.3.

LU -

Then qba = (Qd - LuJrDL)/(JrD2/4)

Note

0.3 x 0.90 - 0.27 MPa

qba = (4.0 - 0.27 x Jt x 0.75 X 2.4)/(JtxO.75 2/4) = 5.6MPa allowable maximum base stress qbu = 2.5x4 = 10 MPa hence design socket is satisfactory

3b For design settlement gd = 2.5 mm Steps in design process

Trial pile design

6a Use chart of: Id (LlD.EcIE md ) From elastic analysis (Figure 4.17a) to deduce required socket geometry to support load with required settlement but without slip along pile shaft

Id = (0.0025 x 301 x 0.75)/4.0 = 0.141 no intersection with curve EclEm = 100

6b Select a larger pile diameter and repeat Step 6a

Note

select pile diameter D = 1.5 m Id = (0.0025 x 301 x 1.5)/4.0 = 0.28 intersection with curve for EclEm = 100 LID = 3.2 L = 3.2x1.5 = 4.8 m

6c Use chart of: Qb/QJLlD.EcIEmd) From elastic analysis (Figure 4.17b) to deduce proportion of load carried by base of pile 7 Allowable base stress qba

=Oc

8 Maximum base stress qbm ~ 2.5oc where qbm calculated from design load using pessimistic estimate of shaft resistance Lu = Ys LS with Ys = 0.3. Then 2 qba = (Qd - LuJrDL )/(JrD /4)

128

Qb

= 0.2 x 4.0 = 0.8 MN

qba = 4 MPa 2 qb = 0.8/(Jt x 1.5 /4) = 0.45 qb < qba so socket geometry satisfactory

LU

= 0.3 x 0.90

=0.27 MPa

pessimistic estimate of ultimate shaft load Qsu = LuJtDL = 0.27xJtx1.5x4.8 =6.1 MN and no load is required to be transferred to the base hence design socket is satisfactory

CIRIA Report 181

CALCULATION USING METHOD OF KULHAWY AND CARTER (1992)

A3.3

Trial pile design

Steps in design process 1 Assume trial socket length L

Note

AssumeL = 3 m

LID

= 3/0.75 = 4.0 1

2 Estimate relative rigidity of pile socket

IR = (30x 1031300)x(0.75/2x3.0)2

IR = (EoIE m ) (D12L) 2 Pile regarded as rigid for IR ;::: 1

= 1.56

IR> 1, hence pile can be regarded as rigid 2

3 Estimate ultimate shaft resistance ts = 0.63(4.0xO.1)°·5 = 0.398 MPa

where Pa = 100 kPa is atmospheric pressure Ultimate shaft capacity Qsu

= tsrrDL

Qsu = 0.398 X Jt x 0.75 x 3.0 = 2.82 MN 3

4 Estimate ultimate base resistance Obu

= 5 x 4.0

Qbu

= 20.0 x

= 20.0 MPa Jt X

2 0.75 14

=8.84 MN

5 Check ultimate factor of safety F > 2.5

F = (2.82 + 8.84)/4.0

6 Initial elastic response of the pile given by

Q/(gEmD) = 11(1 - v/) + (nA;,) (UD)(111 + vr) where

Vr

is Poisson's ratio for rock and

S = In[5(1 -

vr)(UD)]

Proportion of load carried at base of pile:

= 2.91 > 2.5

Assume Vr = 0.25 then S = In[5 x (1 - 0.25) x 4.0] = 2.708 and 2 Qt/(gEmD) = 11(1 - 0.25 ) + (rr 12.708) x (4.0) x (111 + 0.25) 4.779 implying pile stiffness Q/g = 4.779 x 300 x 0.75 = 1.075 MN/mm Qb/Qt 11[1 + (1 -0.25)(JtI2.708)(4.0)] = 0.223

4

Oc = 4 MPa, Pa = 0.1 MPa hence from 4, 17

5

=

=

7 Full slip along pile shaft: settlement given by

QlD = R3.Q/(nEmD where

2

) -

R4

RJ = (1 + vr)[S + 1I(2tan.tamjJ)](DIL)

c = 0.1 x 0.1 x (4.010.1 )2/3 = 0.117 MPa and tan tan1jJ = 0.001 x (4.010.1)2/3 = 0.0117

R 1 = (1.25) x [2.708 + 1I(2xO.0117) ](V4.0) = 14.205 R2

= [1.25/(2xO.0117)].(0.1l7/300) = 0.0208

CIRIA Report 181

129

Steps in design process

Trial pile design

Rs = (:n:/2) (1 -

0.25 2)

Note = 1.473

R3 = 2 x 14.205 x 1.473/(14.205 + 2 x 1.473) = 2.439 Rs = (JTl2) (1 - v/)

R4 = 2 x 0.0208 x 1.473/(14.205 + 2 x 1.473) = 0.00358

Shaft cohesion c correlated with rock compressive strength Oc (Figure 4.19a) c/pa = 0.1(0c/Pa)213

And the link between settlement and load when full slip is occurring is Q = 3.451Q - 2.683 mm

shaft friction and dilation (tan tan 1jJ) correlated with Oc (Figure 4.19b)

Intersection of elastic and full-slip relationships occurs for Qt = 1.075Q ie Q = 2.683/(3.451 x 1.075 - 1) = 0.99mm and Qt = 1.075 x 0.99 = 1.06 MN

When full slip is occurring, the proportion of load taken at the tip of the pile is

Qb/Qt = Rj/(R j + 2Rs) - [R 6 /(R j + 2Rs)].(:n:D 2c1Qt) where R6 = (1 + v r)/2tan.tan1jJ

R6 = (1.25)/2 x 0.0117 = 53.4118 and Qb/Qt = [14.205(53.418x:n:xO. 75 2xO.117)/Qt]/(14.205 + 2x1.473) = 0.828 - 0.644/Qt

= 4 MN, Qb/Qt = 0.67 ForQt = 4MN,

for Qt

8 Check settlement satisfactory for design load

7,8

Q = 3.451x4 - 2.683 = 11.1 mm It may be appropriate to repeat calculations

with a longer socket to reduce settlement under design load to 10 mm. (Required socket length = 3.75 m). Design is controlled by settlement, not end-bearing stress. Notes The design method of Kulhawy and Carter is based on a simplified approach to the analysis of pile response in which the load transfers along the pile shaft and at the base of the pile are considered separately. It may therefore be regarded as slightly less rigorous in its theoretical approach than the method of Rowe and Armitage. However, in analysing shaft load transfer the method does allow for a frictionalldilatant development of shear stress, with suggested values of the friction/dilatant characteristic obtained from correlations with compressive strength of the rock deduced from observations of field performance of piles in weak rock. As an approach to design of piles it is necessary to iterate, adjusting the assumed pile geometry to produce a pile load:settlement relationship that satisfies the design requirements. The load:settlement relationship that emerges for the assumed pile geometry is shown in Figure 4.20.

130

CIRIA Report 181

The example presented here has a pile geometry that implies that the pile can be considered rigid in comparison with the rock (Step 2 and Note 1). If the condition of pile rigidity is not satisfied, the equations for elastic and for full slip response of the pile become more complicated (although still tractable), and reference should be made to Kulhawy and Carter. ~

1.

Kulhawy and Carter suggest that if IR cent of the theoretical value.

2.

The expression for the socket shear stress is a lower bound to values of "peak average side shear stress" deduced from field tests on piles socketed into weak rock. The coefficient 0.63 corresponds to a value of X = 0.89 in Table 4.1 and Equation

1 the elastic settlement will be within 10 per

(4.4) .

3.

Kulhawy and Carter do not give any suggestions about the calculation of ultimate base capacity. The proposal here follows the recommendation of Williams, Johnston and Donald, mentioned in Section 4.1.2 and in Appendix A3.1. The ultimate base capacity merely sets an upper limit to the load:settlement relationship shown in Figure 4.20, and is necessary for the ULS check, but does not play any other part in the calculations of pile load transfer and pile stiffness.

4.

The analysis presented by Kulhawy and Carter acknowledges the possibility of the rock around and beneath the pile having different elastic properties. Where the pile cannot be treated as rigid (Step 2), a slightly more complex analysis is required: this is presented by Kulhawy and Carter.

5.

Kulhawy and Carter also present the analysis for a pile that cannot be regarded as rigid.

6.

Although the proportion of load carried by the base of the pile alters as the pile is loaded and because the pile is assumed rigid, the base load always increases linearly with pile settlement whether or not shaft slippage is occurring:

Qt!Q = E.."D/(l - v/)

CIRIA Report 181

7.

The criterion of rigidity of piles (Step 2) indicates that, with the concrete and rock stiffnesses and pile diameter assumed here, the limit for rigidity is L = 3.55 m and a redesign to achieve a smaller settlement may be more readily achieved using a larger pile diameter rather than using the full equations for the flexible pile. For example, if pile diameter is increased to D = 1 m, then with L = 3 m a settlement Q = 10.1 mm is calculated for a pile load Qt = 5 MN.

8.

Maintaining the assumption of rigidity of the pile, the socket length required to satisfy the design criteria is 3.75 m. The slope and intercept of the load:settlement relationship are only weakly dependent on UD: a large change in length L is required to give a significant reduction in settlement at a given load. The design of the rock socket is strongly controlled by the cohesion and dilatancy terms: the value used in Step 7 (Equations (4.17) and (4.18)) is a lower bound to the data shown in Figure 4.19. If average values are used instead (a = 0.3, b = 0.003), the design socket length is reduced to 1.35 m although the proportion of load carried at the base remains 60 per cent (see Note 6)

131

FLEMING (1992)

A3.4

Steps in design process

Pile design

1 Assume trial socket length L

Assume L

2 Estimate ultimate shaft resistance ts

ts = 0.84 MPa

Ultimate shaft capacity Qsu = tsrrDL

Qsu = 0.84 X rr x 0.75 x 2.2 = 4.35 MN

Note

=2.2 m 1

3 Estimate ultimate base resistance

2 Obu = 5 x 4.0 = 20.0 MPa

Ultimate base capacity Qbu = oburrD2/4 .;'

Qbu = 20.0xJTxO.75 2/4 == 8.84 MN

4 Shaft load:settlement relationship

Assume Ms

Qs == QsuQ/(M,D + Q)

Qs = 4.35Q/(0.001xO.75 Q: m Qs: MN

=0.001

3,4

+ Q)

5 Base load settlement relationship

5 Qb = (0.75x300x8.84Q)/(0.6x8.84 + 0.75x300Q)Qb = 1989Q/(5.304 + 225Q) Q: m

6a Total pile load for settlement Q (neglecting elastic shortening of pile) Q, == Qs + Qb

Qb: MN

For Q == 2.5 mm = 0.0025 m Qs = 3.35 MN Qb == 0.85 MN Q, = 4.20 MN

6b

For Q = 10 mm = 0.01 m Qs = 4.05 MN Qb = 2.63 MN Q, = 6.68 MN

7a Elastic shortening of pile for Q, < Qsu

Assume KE = 0.5 for Q = 2.5 mm Q, = 4.20 < Qsu = 4.35 MN

QE ==

2 K E.4Q/.J(rrD E c)

QE =

(assumes Lo = 0)

6, 7

3 2 0.5X4x4.20x2.2/(rrxO.75 x30x10 ) = 0.35 mm

Hence total settlement 2.85 mm 7b Elastic shortening of pile for Q, > Qsu

Assume KE = 0.5 for Q = 10 mm Q, = 6.68 > Qsu = 4.35 MN QE

=

{4x2.2.[6.68 - 4.35(1- 0.5)]} l(rrxO.75 2x30x10 3) =0

Hence total settlement 10.75 mm 8 For any input settlement Q calculate shaft load, base load, total load from Steps 4, 5, 6; elastic shortening from Step 7 and hence generate points on complete load:settlement relationship

The full load-settlement relationship is plotted in Figure 4.20

8, 9

J'

132

CIRIA Report 181

Notes To achieve an overallload:settlement relationship for the pile, Fleming proposes the superimposition of separate hyperbolic relationships between load and settlement for pile shaft and pile base. This worked example shows the use of Fleming's method for generating the load:settlement response of a pile of assumed geometry. If assumptions are made concerning the ultimate shaft resistance and ultimate base resistance of a pile socket, Fleming's method can be used to design the socket. The method is entirely empirical, but it is based on extensive experience of back-analysis of pile loading tests which shows that the hyperbolic relationships give a reasonable representation of observed response. It may be useful where piles are to be designed for loading conditions and ground conditions that are not out of the ordinary, but will be most useful for back-analysis of trial piles that may have been designed using other methods. The procedure is readily computerised and is used in reverse by Fleming to deduce values of ground stiffness parameters from results of pile loading tests.

CIRIA Report 181

1.

Fleming does not give recommendations for the calculation of shaft resistance; the value assumed here is the same as that used in the Williams, Johnston and Donald analysis, Appendix A3.I, Step 2. The length L of pile socket assumed in the present example is the value that emerged from the Williams, Johnston and Donald method, Appendix A3.I, Step 3.

2.

Fleming does not give any recommendations for the calculation of base capacity; the value assumed here is the same as that suggested by Williams, Johnston and Donald, Appendix A3.I, Step 10.

3.

The response of the pile is first considered as though the pile were rigid, neglecting elastic shortening of the pile shaft.

4.

The coefficient Ms is the initial slope of the shaft response function and can, in principle, be linked with the elastic properties of the surrounding rock using the same analysis used by Kulhawy and Carter. Fleming suggests that Ms takes values between 0.001 and 0.004 but, for all the examples that he quotes, a value around 0.001 seems to apply. This implies that the elastic pile stiffnesses that would be predicted from standard theoretical approaches are somewhat lower than the stiffnesses deduced from field observations. In this example a value of 0.002 gives results comparable to those of other methods.

5.

The factor 0.6 in the denominator links the elastic response of the base of the pile (calculated using a secant elastic stiffness) with the hyperbolic relationship at a load Qb= QbJ4.

6.

The coefficient KE defines the centroid of the shaft load transfer between pile and rock. For uniform transfer KE = 0.5 but shaft shear stress usually develops more rapidly near the top of the socket and, based on the elastic analysis, a value KE = 0.4 may be more appropriate. However, values of KE between 0.45 and 0.53 are deduced by Fleming from back-analysis of field performance of piles, and for weak rocks a value of KE = 0.5 appears to give best results.

7.

These expressions may be expected to give a slightly conservative estimate of pile shortening since, even at total loads Qt < Qsu, some of the load is taken by the base of the pile.

8.

Having constructed the complete load:settlement relationship, the settlement can be read for any required load, and vice versa. Note that this method suggests that the load carried by this socket for a total settlement of 10 mm will be 6.35 MN, which is somewhat greater than the design load of 4 MN.

133

9.

134

With the equations for Fleming's method set up in a spreadsheet, it is easy to find the socket length that will give a desired settlement (10 mm) at a specified load (4 MN). The result is very dependent on the values of shaft shear stress and ultimate base stress that are assumed. With values chosen to match those used in the other three design methods, the socket lengths that emerge from Fleming's method are:

Ultimate shaft stress

Ultimate base stress

Socket length

MPa

MPa

m

0.84

20

0.65 0.50 0.65

0.63

10

0.77 1.45

0.40

20

1.65

Proportion of load at base at design load

CIRIA Report 181

A4

Load test data

A4.1

SUMMARY OF BORED PILE LOAD TESTS ON CATEGORY 1 AND 2 INTERMEDIATE GEOMATERIALS The following table is a summary of load tests on bored piles in Category 1 and 2 Intermediate Geomaterials (see Section 4.1.3 for definitions of IGM Category and rank) as compiled in the FHW A draft report Load transfer for drilled shafts in intennediate geomaterial (FHW A, 1995). Notes

* Failure not achieved according to the Davisson failure criterion. Resistances were calculated on the basis of the maximum applied loads.

** End-bearing test with shaft length not reported. Table A4.1 Summary of bored pile load tests on Category 1 and 2 IGMs File no

Location

Rock

Date

Dia

Length

Socket length

ac

(mm)

(m)

(m)

(MPa)

type

Voided base

Failure load, Q,

Settlement at 50% ofQ,

(MN)

(mm)

Rank

1001

S Africa

Diabase

1976

610

12.2

10.1

0.77-1.03

No

3.47

3.05

B

1002

Ditto

Ditto

1976

610

12.2

10.1

0.77-1.03

No

2.67

2.54

B

1003

Ditto

Ditto

1976

610

12.2

10.1

0.77-1.03

Yes

2.22

2.03

A

1004

Dallas. Tex

Shale

1991

610

9.1

6.1

0.83

Yes

1.65

1.24

A

1005

Oklahoma

Ditto

1976

762

8.8

6.2

0.39--D.84

No

4.45

2.79

C

1006

Ditto

Ditto

1976

762

11.1

8.5

1.45

No

6.94

3.56

C

1007

Pennsylvania

Ditto

1978

610

5.2

0.0

1.45

No

1.42

2.79

B

1008

Ditto

Ditto

1978

762

6.7

1.5

1.45

Yes

2.22

4.06

A

1009

Ditto

Ditto

1978

762

6.1

0.9

1.45

Yes

2.22

3.30

A

1010

Ditto

Ditto

1978

610

6.1

0.9

1.45

Yes

1.07

2.29

A

1011

Ditto

Ditto

1978

457

6.1

0.9

1.45

Yes

0.89

2.03

A

A

1012

Ditto

Ditto

1978

457

6.1

0.9

1.45

Yes

0.89

1.78

1978

457

5.2

2.5

1.45

No

0.89

1.78

B

1013

Ditto

Ditto

1014

Ditto

Ditto

1978

610

6.1

0.9

1.45

Yes

1.25

2.03

A

1015

Ditto

Ditto

1978

762

5.2

0.0

1.45

No

1.25

3.30

B

1016

Ditto

Ditto

1978

457

7.2

2.0

1.45

No

1.78

4.32

C C

1017

Colorado

Claystone 1970

305

3.5

2.9

1.08

Yes

0.27

2.54

1018

Ditto

Ditto

1970

305

3.5

2.9

1.08

Yes

0.31

1.27

C

1023

N Ireland

Marl

1976

762

13.3

8.3

0.61--D.74

Yes

3.11*

10.16

B

1024

Ditto

Ditto

1976

762

16.0

10.0

0.41--D.81

No

5.12*

1.27

B

CIRIA Report 181

135

Continued from previous page

File no

Location

Rock type

Date

Dia

Length

Socket length

(Jc

(mm)

(m)

(m)

(MPa)

1.9

0.33

Voided base

Failure load, Q,

Settlement at 50% ofQ,

(MN)

(mm)

No

1.60

0.63

C

Rank

1029

Roanoke. Texas

Shale

1988

914

3.0

1030

Ditto

Ditto

1988

914

3.0

1.9

0,41-1.63

No

8.90

2.16

C

5.39

Yes

3.56

2.79

A

1035

Ontario. Canada

Ditto

1983

711

2,4

2,4

1036

Ditto

Ditto

1983

711

2,4

2,4

11.10

No

4,45

4.57

A

1037

Ditto

Ditto

1983

711

2,4

2,4

5.60

Yes

4,45

5.08

A

1038

Ditto

Ditto

1983

711

2,4

2,4

5.50

No

6,49

4.83

A

1039

Ditto

Ditto

1983

711

2,4

2,4

5.39

Yes

6,41

4.32

A

1040

Irving. Texas

Ditto

1989

762

14.1

14.1

2,41-3.45

Yes

3.11

1.27

A

1041

Ditto

Ditto

1989

762

14.1

14.1

Ditto

Yes

4.80

2.79

A

1045

New Mexico

Ditto

1981

610

15.8

15.8

4.18-4.78

No

8.90*

6.35

B

1046

Ditto

Ditto

1981

610

16.9

16.9

Ditto

No

8.90*

6.35

B

1052

Montopolis. Ditto Texas

1975

737

7.3

1.5

1.42

No

4.71

2.54

B

1053

Ditto

Ditto

1975

737

7.3

1.5

1.42

No

3.38

2.54

B

737

7.3

No

3.74

2.54

B

1054

Ditto

Ditto

1975

1.5

1.42

1055

Dallas. Tex

Ditto

1975

889

7.6

1.8

0.62

No

3.11

3.05

B

1056

Melbourne. Australia

Basalt

1984

1524

1.9

1.9

2.80

No

9.79

3.05

B

1057

Ditto

Ditto

1984

1600

2.1

2.1

2.80

No

13.34

1.78

B

1058

Ditto

Ditto

1984

1600

15.8

15.8

3.60

No

13.79

3.81

B

1059

Ditto

Ditto

1984

1524

11.0

11.0

2.80

No

6.67

2.29

B

1060

Ditto

Ditto

1984

1499

18.0

18.0

3.60

No

11.96*

1.02

B

1067

8.5

3.7

1.20-4.00

No

4,45

6.60

B

1061

Coventry.

Siltstone

1975

1062

S Africa

Mudstone 1976

660

6.1

3.0

1.10

No

1.16

3.30

B

1063

Ditto

Ditto

1976

889

6.1

3.0

1.10

No

0.36

1.52

B

1064

Ditto

Ditto

1976

889

6.1

3.0

1.10

No

0.52

2.29

B

1.16

3.56

B

10.5

6,4

0.23-0.69

No

762

11.0

6.9

0.23-0.69

No

1.78

2.29

B

6.2

0.23-0.69

No

0.71

1.27

C

1065

S Antonio. Tex.

Shale

1982

457

1066

Ditto

Ditto

1982

1067

Ditto

Ditto

1968

457

10,4

1068

Ditto

Ditto

1968

762

8.2

2.7

0.57-0.63

No

6.23

3.56

C

1071

Oahu, Hawaii

Basalt

1991

813

16.8

16.8

0.97-1.13

No

11.79*

7.87

B

1072

Ditto

Ditto

1991

838

12.0

12.0

0.97-1.13

No

7.12

6.60

B

1073

Ditto

Ditto

1991

813

17.1

17.1

0.97-1.13

No

10.85*

4.57

B

1078

Dallas, Texas

Shale

1960

457

14.2

6.1

1.57-1.79

No

2.94

4.57

C

136

CIRIA Report 181

Continued from previous page

File no

Location

Rock type

Date

Dia

Length (m)

Socket length (m)

(mm)

(MPa)

1990

457

15,7

13.2

1.54

Oc

Voided base

Failure load, Qt

Rank

(MN)

Settlement at 50% ofQt (mm)

No

1.42

3.05

B

Denver,

Denver

Colorado

Blue (Shale)

1080

Ditto

Ditto

1990

457

10,6

8.2

1.63

No

2.22

5.84

B

1081

Ditto

Ditto

1990

457

6,2

3.7

0.52

No

0.20

1.78

B

1079

1082

Ditto

Ditto

1990

762

16.3

13.9

1.54

Yes

0.76

2.29

B

1083

Ditto

Ditto

1990

762

11.4

9.0

1.63

Yes

3.25*

3.30

B

1084

Ditto

Ditto

1990

762

6,9

4.5

0.52

Yes

0.40

2.03

B

1085

Ditto

Ditto

1990

762

7,9

5.5

0.52

Yes

0.67

1.78

B

1086

Ditto

Ditto

1990

610

3,0

3.0

0.95

No

2.85

4.83

B

1113

Ft Collins, Colorado

Shale

1983

457

13,7

3.0

3.75

No

1.78

2.29

B

1120

Melbourne, Australia

Mudstone

1977

660

2,5

2.5

0.83

Yes

1.78*

2.03

A

1121

Ditto

Ditto

1977

1118

3,6

3.6

0.55

Yes

4.45

1.78

A

1122

Ditto

Ditto

1977

1168

3,5

3.5

0.61

Yes

4.45

1.27

A

1123

Ditto

Ditto

1978

1219

13.5

13.5

2.50

Yes

4.27

1.78

A

1124

Toronto, Canada

Shale

1968

635

6,8

1.3

14.74

No

4.45

4.80

C

1125

St P'burg, Florida

Limestone Undated

711

27,7

10.7

0.34--0.62

No

8.90*

5.33

C

1127

Jacksonville Limestone UnFlorida dated

914

17.4

9.1

3.28-7.20

No

4.00

2.29

C

1128

Dade Co., Florida

Limestone 1986

1067

7.7

5.3

1.93-5.12

No

8.90*

14.48

C

1129

Tempe. Arizona

Hard clay

1990

914

36.6

3.0

0.97

No

12.45*

3.81

A

1131

S Antonio, Texas

Shale

1989

610

15.2

7.7

0.77-1.35

No

0.36

2.29

B

1132

Ditto

Ditto

1989

610

15.2

7.7

0.77-1.35

No

2.67

0.76

B

1134

S Carolina

Limestone 1988

610

13.7

3.8

13.51

No

3.56*

2.03

C

1135

Ditto

Ditto

1988

610

12.4

2.3

10.23

No

1.07

1.27

C

CIRIA Report 181

137

A.4.2

SUMMARY OF SHAFT AND END RESISTANCE VALUES USED IN DESIGN ANDIOR VERIFIED BY LOAD TESTING OF PILES Table A4.2 Shaft and end resistance values for some pile load tests

Rock type

ac

Pile type

(MPa) Upper Carboniferous sandstone. siltstone and mudstone

Varies

Bored 560-1220-mmdiameter

Mercia Mudstone

Varies

Bored. driven castin-place 600-mm-diameter

Pile length (m)

Socket length (m)

Up t09

Comments and reference

'tsd

qbd

'tsu

qbu

(kPa)

(kPa)

(kPa)

(kPa)

2D--4D

127201

31803950

Cole and Stroud (1976)

3.3

209 5300

Values for Q = 5 mm Values for Q = 5 mm

2.25

Jorden and Dobie (1977) Varies Highly weathered mudstone. sandstone. conglomerate of Jurong formation. Singapore Mercia Mudstone grade III

Varies

(CU range 75-725 kPa)

Bored 600-200-mmdiameter typical

15-20 typical

Bored 750-mm diameter compression 600-mm-diameter tension

13.6

8.6

18.8

15.3

2N

N: 150-180 Chang and Broms (1991)

2.4N2.7N (105)

Values apply to concreting immediately after boring Daunceyand Woodland (1984)

Bored 615-mm diameter

122

Shale

Bored 900-mm diameter

250

Sandstone

Bored 600-mm diameter

1200

Diabase

0.3-0.6

0

2650

Thorburn (1967) Thinly to medium bedded 15000 Osterberg (1989) Whitworth and Turner (1989)

1091

Bored 600-900-mmdiameter

5.4-17

2.8-D.9

Mercia Mudstone (varying grades)

Bored 790-1050-mmdiameter

21-34

4-19

1.5N

0

Mercia Mudstone grade I and II

23 max. Bored 1200-mm diameter

varies

138

0 Assumed

Mercia Mudstone grades: IV III II I

305x305-mm steel columns

Mercia Mudstone grades: III/IV

Bored 406-mm diameter

150-180

1500

II

508-mm diameter

250-280

4900

Sherwood Sandstone

138

6.7

150 165 250 250

2.25N

Webb (1977)

0

N: 30-50 Q = 25 mm @ 80% of Qu Kilborn et al. (1989) Q = 5mm Ground Engineering September 1995

A, = 2BL Ab = 4B Ground Engineering

0 0 0 2700

June 1995

Chandler and Davis (1973)

CIRIA Report 181

Rock type

Pile type

Oc

(MPa) Cretaceous mudstone

1.1

Devonian slaty mudstone

Pile length (m)

Socket length (m)

= 420

9

275-mm square precast driven, rotary bored or cfa 450-900-mm diameter

Varies

Varies

Mercia Mudstone grade III and IV

Bored 600-1500 mm diameter

10.6 (dia. = 740mm)

Carboniferous shales

Bored 1065-1220- 12 mm diameter

Mercia Mudstone grade III

Co = 50600 kPa

kPa

Co

CIRIA Report 181

(kPa) (kPa)

tsu

120-184

9

Mercia Mudstone (siltstone)

(kPa)

Bored 670-mm diameter

Bored 610-mm diameter

Co

Florida limestone

qbd

315x315-mm steel columns

Shale with thin limestone beds

and II

tsd

2: 240

kPa

1.7-9.8

Bored various Sizes Driven precast concrete

Various

60-150

150

3000

250

5000

qbu

Comments and reference

(kPa) 6800

Wilson (1977)

25 000 58 000

George, Sherrell and Tomlinson (1977)

0

Expansive concrete used Hassan et al. (1993) Houston (1995)

7.6

119

45N

CO

taken as 5N Leach and Thompson (1979)

3

100

4050

Co taken as SN, N: 80-120 Leach and Thompson (1979)

Various

3931571

Mc Vay et al. (I 992) 130

Leach and Mallard (I 980)

139

CIRIA Core Programme members October 1999 Alfred McAlpine Construction Ltd

IMC Consulting Engineers Ltd

AMEC Pic

Institution of Civil Engineers

Aspinwall & Co Limited

John Laing Construction Ltd

BAA pic

Keller Ground Engineering

Bachy Soletanche Limited

Kennedy and Donkin Environmental

Balfour Beatty Ltd

Kvaerner Technology Ltd

BGP Reid Crowther

London Underground Limited

Binnie Black & Veatch

Maunsell Ltd

Buro Happold Engineers Limited

Miller Civil Engineering Ltd

carillion

MJ Gleeson Group pic

Casella London Ltd

Montgomery Watson Ltd

Cementitious Slag Makers Association

Mott MacDonald Group Ltd

Charles Haswell and Partners Ltd

National Power PLC

Curtins Consulting Engineers pic

Northumbrian Water Limited

Dames & Moore

North West Water Ltd

Davis Langdon & Everest

Ove Arup Partnership

Department of the Environment,

Owen Williams Group

Transport and the Regions

Posford Duvivier

Dudley Engineering Consultancy

Scottish and Southern Energy pic

Edmund Nuttall Limited

Scott Wilson

Entec UK Limited

Sheffield Hallam University

Environment Agency

Shepherd Construction Limited

Galliford pic

Sir Robert McAlpine Ltd

GIBB Ltd

South Bank University

Golder Associates (UK) Ltd

Southern Water Services Ltd

Halcrow Group Limited

Taylor Woodrow Construction Holdings Ltd

Health & Safety Executive

Thames Water Utilities Ltd

Henry Boot Construction (UK) Ltd

Thorburn Colquhoun

High-Point Rendel

United Kingdom Quality Ash Association

Highways Agency, DETR

University of Salford

HJT Consulting Engineers

Wardell Armstrong

HR Wallingford Ltd

WS Atkins Consultants Limited

Hyder Consulting Limited

Yorkshire Water Services Limited

Piling in weak rocks presents particular problems to the foundation designer and piling specialist, partly because of the difficulty of investigating and testing these intermediate geomaterials. This has resulted in a lack of understanding of their behaviour, especially of their deformability and changes in their properties during pile installation. Design methods have thus tended to be over-conservative and assessments of design parameters too pessimistic. This book summarises current knowledge and practice. It looks at the nature of the material and considers the best ways of investigating, characterising and classifying weak rock for the purpose of pile design. Four published design methods are examined, together with worked examples. Five case studies illustrate the main points of the report and there is an extensive bibliography. The work gives recommendations for improving practice and enhancing economy through appropriate geotechnical modelling, pile selection, pile design and procurement. Piled foundations in weak rock will prove to be a valuable reference for geotechnical and structural engineers working in investigation, design and construction, and research.

1;11:11